Embed Size (px)
Citation preview
Bengt H. Fellenius, Dr.Tech., P.Eng.2475 Rothesay Avenue, Sidney, British Columbia, V8L 2B9
TEL: (778) 426-0775 e-address: <[email protected]>Web site: [www.Fellenius.net]
Basics of Design of Piled FoundationsA Course and Seminar
Santa Cruz, BoliviaApril 25, 2013
The primary intent of the course is to demonstrate that deep foundation design is a good deal more thanfinding some value of capacity. The course aims to show what data one must pull together and presentprocesses of analysis and calculations necessary for a design of a specific project. Aspects of negativeskin friction and associated drag load and downdrag are emphasized.
The presentation includes both broad generalities and in-depth details. Aspects of where to installinstrumentation, perform a test, and analyze the test data are addressed. Settlement analysis is of vitalimportance to the design of piled foundations, and the course addresses principles of settlement analysisand provides some of the mechanics of calculating settlement. A few aspects are included ofconstruction aspects as well as of Limit States Design, LSD (Ultimate Limit States, ULS, andServiceability Limit States, SLS, by Canadian terminology and Load and Resistance Factor Design,LRFD, by US terminology).
To simplify following along the flow of the presentation and taking notes, hand-out course notes areprovided, consisting of black-and-white copies of all Power Points slides, six to a page. Full-size colorcopies of the slides are also available on my web site [www.Fellenius.net]. These can be downloadedfrom the link [/Bolivia]. Note, the link is hidden and has to be typed into the command line ("commandribbon").
The slides contain only a minimum of text. For a background and explanation to much of thepresentations, I refer you to my text book "Basics of Foundation Design" also available for downloadingfrom my web site (the file is called “313 The Red Book_Basics of Foundation Design.pdf”. Afterdownloading, the book can be viewed and read on-screen or be printed (color or black & white) withoutany restriction. The book contains a list of references pertinent to the material presented in the course.Copies of the referenced papers where I am the author or co-author are available for downloading at myweb site (click on the link "Download Papers").
I will be glad to respond to any e-mail with a question you might wish to put to me.
Sidney April 2013
Bengt H. Fellenius
Basics of Design of Piled FoundationsA Course and Seminar
Bengt H. Fellenius, Dr.Tech., P.Eng.
The course comprises four main lectures leading up to and presenting the essentials of the Unified Method of deepfoundations design for capacity, drag load, settlement, and downdrag for single piles, pile groups, and piledfoundations. The presentations are illustrated with case histories of testing and design analysis including how toevaluate strain-gage measurements from instrumented pile loading tests and to assess residual load. Settlementanalysis is of vital importance to the design of piled foundations, and the course addresses principles of settlementanalysis and how to calculate settlement of piles and piled foundations. Pertinent aspects of constructionprocedures and Load and Resistance Factor Design, LRFD are discussed.
08:00h Brief Background to Basic Principles Applicable to Piled Foundations
Stress distribution and interaction between adjacent foundations; Settlement analysis; Applications ofwick drains to piled foundations.
09:30h Coffee Break
09:45h Analysis of Load Transfer, Capacity, and Response to LoadLoad-movement response of foundations; Bearing capacity and load-transfer by beta, alpha, and lambdamethods, and by CPT and CPTu methods; Set-up and relaxation; Residual load; Results of predictionevents.
11:30h The Static Loading Test: Performance, Analysis, and Instrumentation
Methods of testing and basic interpretation of the results. How to analyze results from strain-gageinstrumented piles to arrive at resistance distribution along the pile shaft and the pile toe response.
12.00h LUNCH
13:00h The Static Loading Test: Resumed
Determining pile elastic modulus. The importance of residual load and how to include its effect in theanalysis. Principles of the bi-directional test (the O-cell test) and how to analyze the results of an O-celltest. Case histories of analyses on results of static loading tests on driven and bored piles.
14:30h Coffee Break
14:50h 4. Piles and Pile Groups — Long-Term Behavior and how we know what we know; The Unified Design Method.
Important case histories presenting studies that demonstrated the actual long-term response of piles toload and observed settlement of piles and pile groups. The lessons learnt will be referenced to aspects ofdesign applying the Unified Method for Design of Piled Foundations considering Capacity, Drag Load,Settlement, and Downdrag for single piles, pile groups, and piled foundations.
1. Capacity (choice of factor of safety, and rules of LRFD and Limit States Design) and design forstructural strength (including drag load)
2. Settlement of single piles and pile groups due to load directly on the piles and due to influence fromadjacent activity (downdrag)
3. How to combine the various aspects for the design of an actual case with emphasis on foundationsettlement illustrated with examples
17:00h Questions and Discussions; End of Day
1
BASICS OF DESIGN OF PILED
FOUNDATIONSFOUNDATIONS
Bengt H. Fellenius
1
A short course
Santa Cruz, Bolivia, April 25, 2013
08:00h Brief Background to Basic PrinciplesApplicable to Piled Foundations
SCHEDULE
09:30h Break
09:45h Analysis of Load Transfer, Capacity and Response to Load
11.30h The Static Loading Test: Head-down and O-cell Tests
12.00h LUNCH
13.00h The Static Loading Test: Continued
14 00h Case Histories on Pile Analysis Drag Load Downdrag
2
14.00h Case Histories on Pile Analysis, Drag Load, Downdrag,Pile Groups, Piled Raft, Piled Pad
14.30h Break
14.50h The Unified Method of Design
17:00h Questions and Discussions and End of Day
2
www.Fellenius.net
Bolivia
To Download All COURSE SLIDES
Power Point Slides1 - Background Lecture 1.pdf2 - Analysis Methods Lecture 2.pdf3 - Static Loading Test Lectures 3a and 3b.pdf4 - Case Histories and Lectures 4a and 4b.pdf
Design Methods
4
3/24/2013
1
BASICS OF DESIGN OF PILED
FOUNDATIONS
Bengt H FelleniusBengt H. Fellenius
Background and Basic Principles
Bolivia, April 25, 2013
Some Fundamental Principles
22
Determining the effective stress is the key to geotechnical analysis
• The not-so-good method:
hΔ=Δ '' γσ γ’ = buoyant unit weight
33
)'(' hz Δ∑= γσ
)1(' iwt −−= γγγ
It is much better to determine, separately, the total stress and the pore pressure. The effective stress is then the total stress minus the pore pressure.
)( hΔ∑
44
)( hz Δ∑= γσ
uz −= σσ '
Determining pore pressure
u = γw hThe height of the column of water (h; the head representing the water pressure)is usually not the distance to the ground surface nor, even, the distance to thegroundwater table. For this reason, the height is usually referred to as the“phreatic height” or the “piezometric height” to separate it from the depth below
PRESSURE
55
the groundwater table or depth below the ground surface.
The pore pressure distribution is determined by applying the facts that
(1) in stationary conditions, the pore pressure distribution can be assumed to be linear in each individual soil layer
(2) in pervious soil layers that are “sandwiched” between less pervious layers, the pore pressure is hydrostatic (that is, the vertical gradient is unity)
SAND Hydrostatic distribution
CLAY Non-hydrostatic distribution, but linear
SAND Hydrostatic distribution Artesian phreatic head
GW
DEPTH
Distribution of stress below a a small load area
0LBqqz
××=
The 2:1 method
66
)()(0 zLzBqqz +×+
The 2:1-method can only be used for distributions directly under the centerof the footprint of the loaded area. It cannot be used to combine (add)stresses from adjacent load areas unless they all have the same center. it isthen only applicable under the area with the smallest footprint.
3/24/2013
2
The Boussinesq Method Derived from calculation of stress from
a point load on the surface of an elastic medium
33z
77
2/522 )(23
zrzQqz +
=π
Newmark’s method for stress from a loaded area
Newmark (1935) integrated the Boussinesq equation over a finite area and obtained a relation for the stress under the corner of a uniformly loaded rectangular area, for example, a footing
CBAI +×
88
π40CIqqz =×=
2222
22
112nmnm
nmmnA+++
++=
12
22
22
++++
=nmnmB
⎥⎥⎦
⎤
⎢⎢⎣
⎡
−++++
= 2222
22
112arctannmnm
nmmnC
m = x/zn = y/zx = length of the loaded areay = width of the loaded areaz = depth to the point under the corner
where the stress is calculated
(1)
• Eq. 1 does not result in correct stress values near the ground surface. To represent the stress near the ground surface, Newmark’s integration applies an additional equation:
π CBA −+×
99
ππ
40CBAIqqz
+×=×=
For where: m2 + n2 + 1 ≤ m2 n2
(2)
Stress distribution below the center of a square 3 m wide footing
-2
0
) 0 15
0.20
0.25
CTO
R,
IEq. (1)
Eq. (2) Eq. (2)
1010
0 20 40 60 80 100-6
-4
STRESS (KPa)
DE
PTH
(m
0.01 0.10 1.00 10.000.00
0.05
0.10
0.15
m and n (m = n)
INFL
UE
NC
E F
AC
Eq. (1)
0
1
2
0 25 50 75 100
STRESS (%)
met
ers)
Boussinesq
Westergaard
0
1
2
0 25 50 75 100
SETTLEMENT (%)
met
ers)
Boussinesq
Westergaard
1111Comparison between Boussinesq, Westergaard, and 2:1 distributions
3
4
5
DEP
TH (
dia
2:13
4
5
DEP
TH (
dia
2:1
0
1
2
0 25 50 75 100
STRESS (%)
eter
s)
Westergaard
Boussinesq
0
1
2
0 25 50 75 100
SETTLEMENT (%)
met
ers)
Boussinesq
Westergaard
1212
2
3
4
5
DEP
TH (
diam
2:1
2
3
4
5
DEP
TH (
diam
2:1
3/24/2013
3
0
1
2
0 25 50 75 100
STRESS (%)
amet
ers)
Westergaard
Boussinesq
0
1
2
0 25 50 75 100
SETTLEMENT (%)
amet
ers)
Boussinesq
Westergaard
1313
3
4
5
DE
PTH
(di
a
2:1 Characteristic Point; 0.37b from center
3
4
5
DEP
TH (
dia
2:1 Characteristic Point; 0.37b from center
Below the characteristic point, stresses for flexible and stiff footings are equal
Now, if the settlement distributions are so similar, why would we persist in using
Boussinesq stress distribution instead of the much simpler 2:1 distribution?
1414
Because a footing is not alone in this world; near by, there are other footings, and fills,
and excavation, etc., for example:
The settlement imposed outside the loaded
foundation is often critical
0
1
2
0 25 50 75 100
SETTLEMENT (%)
met
ers)
BoussinesqOutside Point Boussinesq
Center Point
1515
2
3
4
5
DEP
TH (
diam
Loaded area
The end result of a geotechnical design analysis
is
1616
Settlement
Stress-Strain
σ' (
KPa
)
Δσ
εσΔΔ
=tM
1717
STRAIN (%)
STR
ESS,
σ
Δσ Δε
Δε
Plotted as Strain-Stress
N (
%)
N (
%)
TIO
, e
Plotted as Void Ratio vs. Stress
1818
STRESS, σ' (KPa)
STR
AIN
STRESS, σ' (KPa)
STR
AIN
STRESS (KPa)
VOID
RA
T
3/24/2013
4
Stress-strain behavior is non-linear for most soils. The non-linearity cannot be disregarded when analyzing compressible soils, such as silts and clays, that is, the elastic modulus approach is not appropriate for these soils.
Non-linear stress-strain behavior of compressible soils, is conventionally modeled as follows.
11 'l'l σσC
1919
where ε = strain induced by increase of effective stress from σ‘0 to σ‘1Cc = compression indexe0 = void ratioσ‘0 = original (or initial) effective stressσ‘1 = final effective stress
CR = Compression Ratio = (MIT)
0
1
0
1
0 'lg
'lg
1 σσ
σσ
ε CRe
Cc =+
=
01 eC
CR c
+=
Some use the term "Ccε" for the "CR", creating quite a bit of confusion thereby
In overconsolidated soils (most soils are)
)''lg
''
lg(1
1 1
00 pc
pcr CC
e σσ
σσ
ε ++
=
2020
where σ‘p = preconsolidation stressCcr = re-compression index
The Janbu Method
The Janbu tangent modulus approach, proposed by Janbu (1963; 1965; 1967; 1998), and referenced by the Canadian Foundation Engineering Manual, CFEM (1985; 1992), applies the same basic principles of linear and non-linear stress-strain behavior. The method applies to all soils, clays as well as sand. By this method, the relation between stress and strain is a function of two non-dimensional parameters which are unique for a soil: a stress exponent, j, and a modulus number, m.
2121
Janbu’s general relation is
])''()
''[(1 01 j
r
j
rmj σσ
σσε −=
where: σ‘r = a “reference stress = 100 KPaj = a stress exponent
m = the modulus number
The Janbu Method
Dense Coarse-Grained Soil j = 1
Cohesive Soil j = 0 1'ln1 σε =
'1)''(101 σσσε Δ=−=
mm
'21)''(
21
01 σσσε Δ=−=mm
σ’ in KPa
σ’ in ksf
2222
Cohesive Soil j = 0
Sandy or Silty Soils j = 0.5
0'lnσ
εm
=
)''(51
01 σσε −=m
pm''(2
1 σσε −=
σ’ in KPa
σ’ in ksf
There are direct mathematical conversions
between m and the E and Cc-e0
For E given in units of KPa (and ksf), the relation between the modulus number and the E-modulus is
2323
m = E/100 (KPa)m = E/2 (ksf)
For Cc-e0, the relation to the modulus number is
cc Ce
Cem 00 13.2110ln +
=+
= And m = 2.3/CR
Typical and Normally Conservative Modulus Numbers
SOIL TYPE MODULUS NUMBER STRESS EXP.
Till, very dense to dense 1,000 — 300 (j=1)
Gravel 400 — 40 (j=0.5)
Sand dense 400 — 250 (j=0.5compact 250 — 150 _ " _loose 150 — 100 _ " _
Silt dense 200 — 80 (j=0.5)compact 80 — 60 _ " _loose 60 — 40 _ " _
This is where the greater value of the Janbu approach versus the MIT CR-approach comes in.
ClaysSilty clay hard, stiff 60 — 20 (j=0)
stiff, firm 20 — 10 _ " _Clayey silt soft 10 — 5 _ " _
Soft marine claysand organic clays 20 — 5 (j=0)
Peat 5 — 1 (j=0)
For clays and silts, the recompression modulus, mr, is often five to twelve times greater than the virgin modulus, m.
This is where the Janbu approach and the MIT CR-approach are equal in practicality.
Reference: Fellenius, B.H., 2012. Basics of foundation design, a text book.Revised Electronic Edition, [www.Fellenius.net], 385 p.
3/24/2013
5
0.80
1.00
1.20
Voi
d R
atio
(- -
)
m = 12(CR = 0 18)
p'c
10
15
20
25
Stra
in (
%)
C
1/m
Slope = m = 12
Evaluation of compressibility from oedometer results
2525
0.40
0.60
10 100 1,000 10,000
Stress (KPa) log scale
V (CR = 0.18)
0
5
10 100 1,000 10,000
Stress (KPa) log scale
p 10p
Cc
Cc = 0.37
e0 = 1.01 p'c
p 2.718p
Void-Ratio vs. Stress and Strain vs. Stress — Same test data
Note, if the "zero"-value -- the e0 -- is off, the Cc-e0 is off (and so is the CR) even ifthe Cc is correctly determined. Not so the "m" (if determined from the test results).
Comparison between the Cc/e0 approachand the Janbu method
0 10
0.15
0.20
0.25
0.30
0.35
PRES
SIO
N IN
DEX
, Cc
Do these values indicate a
compressible soil, a medium compressible
soil, a moderately ibl il
15
20
25
30
35
MO
DU
LUS
NU
MB
ER, m
2626
Data from a 20 m thick sedimentary deposit
0.00
0.05
0.10
0.40 0.60 0.80 1.00 1.20
VOID RATIO, e0
CO
MP compressible soil, or a
non-compressible soil?0
5
10
0.400.600.801.001.20
VOID RATIO, e0
VIR
GIN
The Cc-e0 approach (based on Cc) implies that the compressibility varies by 30± %.
However, the Janbu methods shows it to vary only by 10± %. The modulus number, m, ranges from 18 through 22; It would be unusual to find a clay with less variation.
Conventional Cc/e0
How many of these oedometer results indicate
(o) highly compressible clay
(o) compressible clay
( ) di ibl l20
30
40
50
OD
ULUS
NU
MB
ER, m
0 20 40 60 80 100WATER CONTENT, wn (%)
Janbu Modulus Number m
The Cc-values converted via the associated e0-values to modulus
numbers.
2
3
4
5
MPR
ESSI
ON
IND
EX, C
c
2727
(o) medium compressible clay
(o) non-compressible clay?
0
10
0.00 0.50 1.00 1.50 2.00 2.50 3.00
VOID RATIO, e0
VIRG
IN M
m < 10 ==> Highly compressible Oedometer test data from Leroueil et al., 1983
0
1
0.00 1.00 2.00 3.00
VOID RATIO, e0
CO
M
Stress produces strainLinear Elastic Deformation (Hooke’s Law)
ε = induced strain in a soil layer
= imposed change of effective stress in the soil layer 'σΔ
E'σε Δ
=
2828
p g y
E = elastic modulus of the soil layer (Young’s Modulus)
Young’s modulus is the modulus for when lateral expansion is allowed, which may be the case for soil loaded by a small footing, but not when the load is applied over a large area. In the latter case, the lateral expansion is constrained (or confined). The constrained modulus, D, is larger than the E-modulus. The constrained modulus is also called the “oedometer modulus”. For ideally elastic soils, the ratio between D and E is:
ν = Poisson’s ratio
)21()1()1(νν
ν−+
−=
ED
Settlement is due to Immediate Compression, Consolidation Settlement, and Secondary Compression
Immediate Compression is the compression of the soil grains (soil skeleton) and of any free gaspresent in the voids. It is usually assumed to be linearly proportional to the change of stress Theimmediate compression is therefore often called 'elastic' compression. It occurs quickly and isnormally small (it is not associated with expulsion of water).
Consolidation (also Primary Consolidation) is volume reduction during the increase ineffective stress occurring from the dissipation of pore pressures (expelling water from the soilbody). In the process, the imposed stress, initially carried by the pore water, is transferred to the
il t t C lid ti i kl i i d il b t l l i fi i d
2929
soil structure. Consolidation occurs quickly in coarse-grained soils, but slowly in fine-grainedsoils.
Secondary Compression is a term for compression occurring without an increase of effectivestress. It is triggered by changes of effective stress. It does not usually involve expulsion ofwater, but is mainly associated with slow long-term compression of the soil skeleton. Somecompression of the soil structure occurs and it is then associated with some expulsion of water,but this is so gradual and small that pore pressure change is too small to be noticed. Sometimes,the term "creep" is used to mean secondary compression, but "creep" should be restricted toconditions of shear. Secondary compression is usually small, approximately similar in magnitudeto the immediate compression, but may over time add significantly to the total deformation of thesoil over time. Secondary compression can be very large in highly organic soils, such as peat.Theoretically, seconday compression occurs from the start of the consolidation (effective stresschange), but in practice, it is considered as starting at the end of the consolidation.
On applying load, the soil skeleton compresses and the soil grainsare forced closer to each other reducing the void ratio. Thecompression of the soil skeleton occurs more or less immediately incontrast to the reduction of the void volume which requiresexpulsion of water ("consolidation") and can take a long time.However, in soils containing gas bubbles, the load applicationcauses the bubbles to compress (and partially to go into solution in
Immediate Compression and Consolidation Settlement
3030
the pore water), which also occurs immediately. Then, as the porepressure dissipates during the consolidation process, the gasbubbles expand which slows down the settlement process. The"slow-down" is often mistaken for approaching the end ofconsolidation. The thereafter observed settlement is theninterpreted as a large secondary compression (addressed later on).
3/24/2013
6
2H
Drainage Layer
Clay Layer (consolidating)
Drainage Layer
0
1uu
SS
U t
f
tAVG −==
where UAVG = average degree of consolidation (U)St = settlement at Time tSf = final settlement at full consolidationut = average pore pressure at Time tu0 = initial average pore pressure (on application of the load at Time t = 0)
Basic Relations
UAVG
Consolidation Settlement
3131
vv c
HTt2
=
where t = time to obtain a certain degree of consolidationTv = a dimensionless time coefficient: cv = coefficient of consolidationH = length of the longest drainage path
UAVG (%) 25 50 70 80 90 “100”
Tv 0.05 0.20 0.40 0.57 0.85 ≤1.00
)1(lg1.0 UTv −−−=
HOW TO HANDLE A MULTILAYERED PROFILE?
c/c
d
"Square" spacing: D = √4/π c/c = 1.13 c/c
"Triangular" spacing: D = √(2√3)/π c/c = 1.05 c/c
Vertical Drains
3232
c/cBasic principle of consolidation process in the presence of vertical drains
hh Ud
DT−
−=1
1ln]75.0[ln81
hh UdD
cDt
−−=
11ln]75.0[ln
8
2
and
hh c
DTt2
=
The Kjellman-Barron Formula
Walter Kjellman, inventor of the very first wick drain, the Kjellman Wick, a 100 mm wide, 3 mm thick, cardboard drain that became the prototype for
33
p ypall subsequent wick drains.
Walter Kjellman, 1950
Important Points
Build-up of Back Pressure
The consolidation process can be halted if back-pressure is let to build-up below the embankment, falsely implying that the process is completed
3434
Flow in a soil containing pervious lenses, bands, or layers Theoretically, vertical drains operate by facilitating horizontal drainage. However, where pervious lenses and/or horizontal seams or bands exist, the water will drain vertically to the pervious soil and then to the drain. When this is at hand, the drain spacing can be increased significantly.
Pervious seams (silt or sand) will dry faster than the main body of clay. The pervious seams can be observed in a Shelby sample during the drying process, as indicated in the photos.
3535
p
CPTU soundings with readings every 10 mm can also disclose the presence of sand and silt seams (if they are thicker than about 10 mm; which the illustrated small seams are not).
How deep do the wick drains have to be installed?
In theory, the drains do not need to go deeper than to where the applied stress is equal to the preconsolidation stress.
However in practice it is a good rule to always go down to a
3636
However, in practice, it is a good rule to always go down to a pervious soil layer (aquifer) to ensure downward drainage. But, if the surcharge is by vacuum treatment or combined with vacuum treatment, it is better to avoid having the drains in an aquifer, or they would "suck".
3/24/2013
7
3737
The Kjellman wick, 1942 The Geodrain, 1972
3838
The Geodrain, 1976
Wick drain types
The Burcan Drain, 1978
The Mebra Drain 1984 (a development of the
Castleboard Drain 1979)
3939
0
5
10
15
20
25
30
35
40
0 100 200 300Pore Pressure (KPa)
Dep
th (
m)
Wick Drains Installed
m)
Settlement at center of a 3.6 m high embankment BangkokAirport. Wick drains at c/c 1.5 m were installed to 10 m depth.
PORE PRESSUREEnlarged
40
AVERAGE MEASURED SETTLEMENT
DESIGN CURVE FOR THISSURCHARGE (75 KPa)
1.0 m
FINAL HEIGHTOF FILL
SET
TLEM
ENT
(mm
)
≈200 days
FILL
HEI
GH
T (m
CalculatedTotalSettlements
Settlement and Horizontal Displacement for the 3.6 m Embankment
WICK DRAINS TO 10 m DEPTHWICK DRAINS TO 10 m DEPTH
Settlement was monitored in center and at embankment sides and horizontal displacement was monitored near sides of embankment
Note the steep slopes
4141
Time from start to end of surcharge placement = 9 monthsObservation time after end of surcharge placement = 11 months
1.0 m
2.0 m
WICK DRAIN
Moh and Lin 2006
Horizontal Displacement versus Settlement at Different Test Locations
OVE
MEN
T (c
m)
4242
HOR
IZO
NTA
L M
O
SETTLEMENT (cm)
3/24/2013
8
Secondary Compression
1000
log1 t
te
C ααε+
=
The value of the Coefficient of Secondary Compression, Cα, is usually expressed as aratio to the consolidation coefficient, Cc, ranging from 0.02 through 0.07 with an averageof about 0.05 (Holtz and Kovacs 1981). For example, in a soft clay with Cc of about 0.3
d f b t it (i d l b f 15) C ld b b t 0 01
4343
and e0 of about unity (i.e., a modulus number of 15), Cα, would be about 0.01.
The key parameter, however, it the t100 value, the time it takes for 100 % of consolidation (or 90 %, more realistically) to develop. Also when using wick drains, the 100-% should be the time for vertical drainage, not horizontal.
It is commonly assumed that secondary compression does not start until primary consolidation is completed; U = 100 %. However, the consensus amongst the experts is that secondary compression starts as soon as a change of effective stress has been triggered, i.e., it starts at at 0 % consolidation.
The purpose of calculating stresses is to calculate settlement. The following showssettlements calculated from the Boussinesq distribution. how stress applied to thesoil from one building affect the settlement of an adjacent existing 'identical'building loaded the same constructed about 5 years before.
EXISTING ADJACENT BUILDING
NEW BUILDING
WITH SAME LOAD OVER FOOTPRINT
AREA
The 2nd building was constructed five years after the 1st building. The 1st building had then settled about 80 mm (≈3 inches), which was OK albeit close to what was felt to be
4444The soils consist of preconsolidated (OCR = 2) compressible silt and clay
6.5 m6.5 m 4 m m
1st Building
2nd Building
was OK, albeit close to what was felt to be acceptable. Did the construction of the 2nd building add settlement to the 1st, and what was the settlement of the 2nd building?
(Because the buildings are on raft foundation, the characteristic point is the most representative point for the settlement calculations).
The settlement of the first building calculated using UniSettle Version 4
0 2 4 6 8 10YEARS
SETTLEMENT OVER TIME
4545
020406080
100120
0 2 4 6 8 10
SETT
LEM
ENT
(mm
)
2nd Building constructed
Calculations using Boussinesq distribution can be used to determine how stressapplied to the soil from one building may affect an adjacent existing building(having the same loading as the new building).
0
5
0 20 40 60 80 100
STRESS (%)
STRESSES UNDER AREA
BETWEEN THE TWO BUILDINGS
EXISTING ADJACENT BUILDING
NEW BUILDING
WITH SAME LOAD OVER FOOTPRINT
AREA
4646
10
15
20
25
30
DEP
TH (
m)
STRESSES ADDED TO THOSE UNDER THE FOOTPRINT OF THE ADJACENT BUILDING
STRESSES UNDER THE FOOTPRINT OT THE LOADED BUILDING
TWO BUILDINGS
Calculations by means of UniSettleThe soils consist of preconsolidated
moderately compressible silt and clay
6.5 m6.5 m 4 m m
Calculations using Boussinesq stress distribution can be used to determine howstress applied to the soil from one building may affect an adjacent existing building(having the same loading as the new building).
EXISTING ADJACENT BUILDING
NEW BUILDING
WITH SAME LOAD OVER FOOTPRINT
AREA
0
5
10
0 20 40 60 80 100
STRESS (%)
STRESSES UNDER THE AREA
BETWEEN THE TWO BUILDINGS
PRECONSOLIDATION MARGIN (Reducingwith depth)
4747The soils consist of preconsolidated moderately compressible silt and clay. The preconsolidation margin reduces with depth.
6.5 m6.5 m 4 m m
10
15
20
25
30
DEP
TH (
m)
CENTER STRESSES COMBINED
STRESSES UNDER THE FOOTPRINT OF THE LOADED BUILDING
STRESSES FROM LOADED BUILDING CALCULATED UNDER THE FOOTPRINT OF THE ADJACENT BUILDING
Settlement distributions (UniSettle Version 4)
0
5
10
0 20 40 60 80 100 120
SETTLEMENT (mm)
1st ONLY
Increase due to 2nd Bldng BOTHSand &
Gravel
Silty Clay
0
5
10
0 20 40 60 80 100 120
SETTLEMENT (mm)
Of ground due to 1st Bldng only
Due to 2nd Bldng
4848
15
20
25
30
35
DEP
TH (
m)
1st BUILDING
Soft Clay 15
20
25
30
35
DEP
TH (
m)
2nd BUILDING
3/24/2013
9
-83 KPa
105 KPa
34 KPa
85 KPa
105 + 34 + 85 = 224 - 83 141 KPa
110 m
38 m74 m
MORE ON SETTLEMENT
YEARYEAR
49
Briaud et al. 2007; Fellenius and Ochoa 2008
0
50
100
150
200
250
300
350
4001936 1946 1956 1966 1976 1986 1996 2006
YEAR
SETT
LEM
ENT
(m
m)
.
0
50
100
150
200
250
300
350
400
1 10 100
SETT
LEM
ENT
(mm
)
1936 1937 1940 1945 1950 1960 1975 2000
LINEAR PLOTLOWER SCALE
LOGARITHMIC PLOTUPPER SCALE
1936 1946 1956 1966 1976 1986 1996 2006
0
20
40
60
80
100
120
140
YEAR
WA
TER
DEP
TH (
m)
132a- 14m217 - 26m216a- 39m115 -153m209 -159m111 -161m501a-180m912 -206m114a-261m618 -267m606 -301m501b-365m132b-442m114b-480m
1925 1935 1945 1955 1965 1975 1985 1995 2005 2015
SHALLOW WELLS
DEEP WELLS
Water Depths Measured in Deep Wells
50
Monument and Well Locations
Well head at Burnett School, Baytown, Texas
YEAR
51
0
50
100
150
200
250
300
350
400
1 10 100
YEAR
SE
TTLE
ME
NT
(mm
)
1936 1937 1940 1945 1950 1960 1975 2000
DEPTH TO WATER TABLE
SETTLEMENT
0
25
50
75
100
125
DEP
TH T
O W
ATE
R T
AB
LE (
m)
San Jacinto MonumentSettlement and Measured Depths to Water in the Wells Plotted Together
1925
The lowering of the pore pressures due to mining of water and subsequent regionalsettlement is not unique for Texas. Another such area is Mexico City, for example.Here is a spectacular 1977 photo from San Joaquin, California.
52
1977
1955
Subsidence at San Joaqu in Valley, California
0.0
0.5
1.0
1920 1930 1940 1950 1960 1970 1980 1990 2000 2010
YEAR
ent
(m)
I II III IV
5353
1.5
2.0
2.5
Settl
eme
NEW ORLEANS 1924 - 1978
I. Initial Period of Pumping II. Increased Pumping III. Further Increased IV. Reduced Pumping
Data from Kolb, C.R. and Saucier, RT., 1982
Site Investigation Techniques
The SPT and the CPT/CPTu
3/24/2013
10
The SPTExample from Atlantic coast of
South USA
0
5
10
15
0 20 40 60 80 100
SPT N-Indices (bl/0.3m)
0
5
0 10 20 30 40 50
SPT N-Indices (bl/0.3m)
5555
20
25
30
35
40
45
50
DEP
TH (
m)
East Abutment
10
15
20
25
DEP
TH (
m)
DETAIL
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H (
m)
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H (
m)
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H (
m)
Example from Atlantic coast of
Canada
5656
40
50
60
70
80
DE
PT 40
50
60
70
80
DE
PT 40
50
60
70
80
DE
PT
SPT for design After problems arose
Forensics
0
10
20
30
0 20 40 60 80 100
N-Index (bl/0.3m)
m)
With all data points
5757
30
40
50
60
70
80
DE
PTH
(m
0.010
0.100
1.000
mv (
1/M
Pa)
30405060708090
100
Mod
ulus
(M
Pa)
Direct numerical use of the SPT N-index
5858
0.0011 10 100
N60-Index (bl/0.3m)
01020
0 10 20 30 40 50N60-Index (bl/0.3m)
(after Terzaghi, Peck, and Mesri 1996 from Burland and Burbidge 1985)
Determining pile Capacity from SPT-indices
0
5
10
15
0 10 20 30 40
SPT N-Index (bl/0.3m)
(m)
0
5
10
15
0 10 20 30 40
SPT N-Index (bl/0.3m)
(m)
0
5
10
15
0 10 20 30 40Cone Stress, qt (MPa)
(m)
5959
20
25
30
35
DEP
TH (
Estimated required depth
20
25
30
35
DEP
TH (
Potentially possible depth
Estimated required depth1
2
Pile 1 had a much smaller capacity than Pile 2!
20
25
30
35
DEP
TH (
N (bl/ft)
Pile 1 had a much smaller capacity than Pile 2!
2
1
Principles of the CPT and CPTU
The Cone Penetrometer
606060
Sleeve friction, fs
Pore PressureU2 position
Cone Stress, qc
“U2 Position” = pore pressure measured on the cone “shoulder”cone shoulder
3/24/2013
11
616161 626262
6363 6464
Continuous cores samples obtained by pushing down a pipe having an inside plastic tube. On withdrawal and cutting the tube open, the soil sample is available in a better condition than a sample in a SPT-spoon.
Courtesy of Pinter and Associates, Saskatoon, SK.
0
10
0 10 20 30
INCLINATION ANGLE (°)
(m)
0
10
0 2 4 6 8
RADIAL DEVIATION (m)
(m)
0
10
0.0 0.3 0.5 0.8 1.0
DEPTH DEVIATION (m)
(m)
The CPT sounding rod is never truly vertical, of course.
How much can it be off?
6565
20
30
40
50
AC
TUA
L D
EPTH
20
30
40
50
AC
TUA
L D
EPTH
20
30
40
50
AC
TUA
L D
EPTH
5
10
15
20
25
Y-D
irect
ion
(m)
20.6 m
PLAN VIEW
"Unfolded"
0
10
20
30
40
50
0 1 2 3 4
DEPTH DEVIATION (m)
EPTH
(m
)
0
10
20
30
40
50
0 5 10 15 20 25
RADIAL DEVIATION (m)
EPTH
(m
)
6666
-5
0
-5 0 5 10 15 20 25
X-Direction (m)
Example 2
60
70
80
90
100
DE
60
70
80
90
100
DE
Inclination plane
X-plane Y-plane
3/24/2013
12
0
5
10
15
0 10 20 30Cone Stress, qt (MPa)
TH (
m)
0
5
10
15
0 100 200
Sleeve Friction (KPa)TH
(m
)
0
5
10
15
0 100 200 300 400
Pore Pressure (KPa)
TH (
m)
0
5
10
15
0.0 1.0 2.0 3.0 4.0
Friction Ratio (%)
TH (
m)
CLAY CLAYCLAY
6767
20
25
30
DEP
T 15
20
25
30
DEP
T 15
20
25
30
DEP
T
20
25
30
DEP
T
SILT SILT SILT
SAND SAND SAND
Results of a CPTU sounding
Soil profilingApplications
6868The Begemann original profiling chart (Begemann, 1965)
1
10
100
Con
e St
ress
, qt
(MPa
)
4
56
7
8
9
10
11
12
Friction Ratio from 0.1 % through 25 %
6969
Profiling chart per Robertson et al. (1986)
01 10 100 1,000
Sleeve Friction (KPa)
C
12
3 25 %
7070Profiling chart per Robertson (1990)
1
10
100
Con
e St
ress
, qE
(MPa
) 5 1 = Very Soft Clays, or Sensitive or Collapsible Soils 2 = Clay and/or Silt 3 = Clayey Silt and/or Silty Clay 4a = Sandy Silt 4b = Silty Sand 5 = Sand to Sandy Gravel
3
4
7171
0.11 10 100 1,000
Sleeve Friction (KPa)
1 2
The Eslami-Fellenius profiling chart (Eslami 1996; Eslami and Fellenius, 1997)
Example of a CPTU sounding from a river estuary delta (Nakdong River, Pusan, Korea)
0
10
20
30
0 10 20 30Cone Stress, qt (MPa)
DEP
TH (
m)
0
10
20
30
0 200 400
Sleeve Friction (KPa)
DEP
TH (
m)
0
10
20
30
0 250 500 750 1,000
Pore Pressure (KPa)
DEP
TH (
m)
0
10
20
30
0 1 2 3 4 5
Friction Ratio (%)
DEP
TH (
m)
Profile
Mixed
CLAY
7272
The sand layer between 6 m and 8 m depth is potentially liquefiable.
The clay layer is very soft.
The sand below 34 m depth is very dense and dilative, i.e., overconsolidated and providing sudden large penetration resistance to driven piles and relaxation problems.
30
40
50
30
40
50
30
40
50
30
40
50
SAN
Reduced pore pressure (“dilation”)
SAND
3/24/2013
13
1
10
100
one
Stre
ss, q
E (M
Pa)
5
3
4
7373
0.11 10 100 1,000
Sleeve Friction (KPa)
Co
1 2
The CPTU data of the Preceding Slide plotted in an Eslami-Fellenius Chart
The CPTU is an excellent and reliable tool for soil identification, but there is more to geotechnical site
investigation than just establishing the soil type.
And, the CPTU can deliver much more than soil profiling
7474
Liquefaction7.4 Magnitude Earthquake of August 17, 1999
Kocaeli, Adapazari, Turkey
7575
Photos courtesy of Noel J. Gardner, Ottawa
7676
Photo courtesy of Noel J. Gardner, Ottawa
dv
v rg
aCSR 'max65.0
σσ
=
CSR = Cyclic Stress Ratio
For earthquakemagnitude of 7.5
An earthquake generates a Cyclic Stress Ratio, CSR
Assessment of liquefaction risk fromresults of a CPTU sounding
7777
amax = maximum horizontal acceleration at ground surface (m/s2)
g = gravity constant (m/s2)
σv = total overburden stress (Pa)
σ'v = effective overburden stress (Pa)
rd = stress reduction coefficient for depth (dimensionless)
z = depth below ground surface (m)
CRR
The safety against liquefaction depends on the Cyclic Resistance Ratio, CRR, determined from the CPTU data
7878
CSRCRRFs = For earthquake magnitude of 7.5
3/24/2013
14
KPaqforqCRR cc 5005.0
100833.0 1
1 <+⎟⎠⎞
⎜⎝⎛=
)(045.0 114.0 cqeCRR =
The following fitted equation represents both equations above
The Cyclic Resistance Ratio, CRR, is expressed in two equations
KPaqKPaforq
CRR cc 1605008.0
10093 1
31 <<+⎟⎠
⎞⎜⎝
⎛=
7979
qc1 = cone stress normalized to depth (i.e., overburden stress)
CNc1 = normalization factor
σr = reference stress = 100 KPa (= atmospheric pressure)
σ'v = effective overburden stress at the depth of the conestress considered (KPa)
'11v
rccNcc qCqq
σσ
==where
CSRCRRFs =
Determining seismic risk from CPTU soundingEvery plotted point represents an earthquake observation (CSR)
with either no liquefaction of with liquefaction observed
Correlations between CRR-valuescalculated from actual earthquakesversus qc1 values for cases ofliquefaction (solid symbols) and noliquefaction (open symbols), andboundary curve (solid line) accordingto Robertson and Wride (1998) andYoud et al. (2001).
The boundary line is the CyclicR i t R ti C CRR hi h
All Data; 0 m through 16.0 m
0.4
0.5
0.6
0.7
SR
Robertson and Wride (1998)
Fines: 35 % 15 %
8080
Resistance Ratio Curve, CRR, whichis also shown as a linear regressioncurve for the boundary values. Thetwo dashed curves show theboundary curves for sand with finescontents of 15% and 35%,respectively (Stark and Olsen 1995).The original diagram has the conestress, qc, divided by atmosphericpressure to make the number non-dimensional.
Note, the effect of fines contents haslately become challenged.
0.0
0.1
0.2
0.3
0 5 10 15 20
Adjusted and Normalized Cone Stress, qc1 (MPa)
CS
0 m through 6.0 m
0.4
0.5
0.6
0.7
max
/g
All Data; 0 m through 16.0 m
0.4
0.5
0.6
0.7
CSR
Separating on two depths and looking at relative seismic force versus not-normalized cone stress.
Re-analysis of data from Moss et al. (2006)
8181
0.0
0.1
0.2
0.3
0 5 10 15 20
Not Normalized Cone Stress, qc (MPa)
a m
A
0.0
0.1
0.2
0.3
0 5 10 15 20
Not Normalized Cone Stress, qc (MPa)
C
BThe 'old' rule that liquefaction risk is small for shallow depth wherethe cone stress is ≥5 MPa appears to hold for quake ratio < 0.25.
In the past, liquefaction risk was based on values of the SPTN-index. Correlations between the CPTU, qc, and the N-indexindicate a ratio between qc and N of about 5. However, thatratio has a very large range between low and high. It isquestionable how relevant and useful a conversion from an
8282
qSPT Index value to a cone stress would be for an actual site.One would be better served pushing a cone in the first place.
Example of determining liquefaction susceptibility before and after vibratory compaction
0
1
2
3
4
0 5 10 15 20Cone Stress (MPa)
(m)
0
1
2
3
4
0 50 100 150 200Pore Pressure (KPa)
H (
m)
0
1
2
3
4
0 20 40 60 80Sleeve Friction (KPa)
H (
m)
0
1
2
3
4
0.0 0.5 1.0Friction Ratio (%)
H (
m)
Sand
PROFILE
Fine sand to Silty Sand
8383
5
6
7
8
9
10
DEP
TH
5
6
7
8
9
10
DEP
TH5
6
7
8
9
10
DEP
TH 5
6
7
8
9
10
DEP
TH
Sand
Silty Clayand Clay
Four CPTU initial (before compaction) soundings at Chek Lap Kok Airport. The heavy lines in the cone stress, sleeve friction, and friction ratio diagrams are the geometric averages for each depth of the four soundings.
10
15
ss, q
E (M
Pa)
1 = Very Soft Clays, Sensitive and/or Collapsible Soils 2 = Clay and/or Silt 3 = Clayey Silt and/or Silty Clay 4a = Sandy Silt and/or Silt
5
Soil chart
8484
0
5
0 20 40 60 80 100
Sleeve Friction (KPa)
Con
e St
res 4b = Fine Sand and/or
Silty Sand 5 = Sand to Sandy Gravel
4b
4a
3
21
3/24/2013
15
0
1
2
3
4
5
0 5 10 15Cone Stress (MPa)
TH (
m)
0
1
2
3
4
5
0 10 20 30 40 50Sleeve Friction (KPa)
TH (
m)
0
1
2
3
4
5
0 50 100 150 200Pore Pressure (KPa)
TH (
m)
0
1
2
3
4
5
0.0 0.1 0.2 0.3 0.4 0.5Friction Ratio (%)
TH (
m)
7 Days7 Days
Before
8585
6
7
8
9
10
DE
PT
6
7
8
9
10
DE
P
6
7
8
9
10
DE
PT
6
7
8
9
10
DE
PT
7 DaysBeforeBefore
Geometric average values of cone stress, sleeve friction, and friction ratios andmeasured pore pressures from CPTU soundings at Chek Lap Kok Airport beforeand seven days after the vibratory compaction.
Fs versus depth0
1
2
3
4
5
0.00 1.00 2.00 3.00 4.00 5.00
Factor of Safety, Fs (--)
PTH
(m
)
Before Compaction
7 Days after
CSRCRRFs =
8686
Factor of safety against liquefaction before and after vibratory compaction
6
7
8
9
10
DEP compaction
CPT and CPTU Methods
for Calculating the Ultimate
Resistance (Capacity) of a Pile
Schmertmann and Nottingham (1975 and 1978)
8787
Meyerhof (1976)
deRuiter and Beringen (1979)
LCPC, Bustamante and Gianeselli (1982 )
Eslami and Fellenius (1997 )
ICP, Jardine, Chow, Overy, and Standing (2005)
But we will save those methods for later
Vibrations from Pile Driving
v = 433 Eh
Z P
M g hr
= 433 Eh
Z P
M g hx2 + z2
V = vertical component of the ground vibration, m/sEh = hammer efficiency coefficientZP il i d N /
88
ZP = pile impedance, Ns/mM = hammer (ram) mass, NG = acceleration, m/s2
H = hammer height-of-fall, m, taken as the equivalentheight-of-fall that corresponds to the kinetic energyat impact
z = pile penetration depth, mx = horizontal distance at the ground surface from pile
to observation point, m
Most ground vibrations are generated from the pile toe
6
8
10
12
14
16
18
20
bration Velocity, v
0 (m
m/s)
89
0
2
4
0 5 10 15 20 25 30 35 40 45 50
Distance to pile toe, r (m)
Vi
Vibrations from driving a long toe bearing pile: measured compared to calculated
3/24/2013
1
BASICS OF DESIGN OF PILED
FOUNDATIONS
Bengt H Fellenius
1
Bengt H. Fellenius
Load Transfer and Capacity of Piles
Bolivia, April 25, 2013 22Driving closed-toe pipe piles into fine sand about 2.5 m above the groundwater table
33Driving 12-inch precast concrete pile into clay for Sidbec in 1974
Head measured in aquifer below the clay layer
44Svärta River 1969
GW
What really is Capacity?
For piles, capacity is
what we determine in
55
— define from —
a loading test
?
e.g.: The Offset Limit Load (Davisson, 1972)
Do you agree that this pointon the curve represents thecapacity of the pile?
Qu
Qu
66
Rs
Rt
3/24/2013
2
γγ NbNqNcr qcu '5.0'' ++=
and for Footings?The Bearing Capacity Formula
where ru = ultimate unit resistance of the footingc’ = effective cohesion interceptB = footing width’ b d ff ti t t th f d ti l l
77
q’ = overburden effective stress at the foundation levelγ‘ = average effective unit weight of the soil below the foundation
Nc, Nq, Nγ = non-dimensional bearing capacity factors
The main factor is the
“Nq”
Nq
88
Nq
But what is the reality?
φ
Results of static loading tests on 0.25 m to 0.75 m square footings in well graded sand (Data from Ismael, 1985)
400
500
600
700
D
( KN
)
1.00 m
0.75 m
0.50 m
0.25 m
1,000
1,200
1,400
1,600
1,800
2,000
S S
( K
Pa
)
Normalized
99
0
100
200
300
0 10 20 30 40 50
SETTLEMENT (mm)
L O
A D
MOVEMENT
0
200
400
600
800
,
0 5 10 15 20
MOVEMENT/WIDTH (%)
S T
R E
S
1.00 m
0.75 m
0.50 m
0.25 m
Normalized
0
2
4
0 5 10 15 20Cone Stress, qt (MPa)
0
2
4
0 100 200 300 400
Sleeve Friction, fs (KPa)
0
2
4
0 20 40 60 80
Pore Pressure (KPa)
0
2
4
0 1 2 3 4 5
Friction Ratio, fR (%)
SAND
CPTU PROFILE
Load-Movement for Five Footings on Sandat Texas A&M University Experimental Site.
J-L. Briaud and R.M. Gibbens, 1994, ASCE GSP 41,
10
6
8
10
12
14
16
DEPT
H (
m)
6
8
10
12
14
16
DEPT
H (m
)
6
8
10
12
14
16
DEP
TH (
m)
6
8
10
12
14
16
DEPT
H (m
)
SANDY CLAYEY SILT
Eslami- RobertsonFellenius
As before the data will tell usmore, if we divide the load withthe footing area (to get stress)and divide the movement withthe footing width, as follows.
Load-Movement of Four Footings on SandTexas A&M University Experimental Site
ASCE GSP 41, J-L Briaud and R.M. Gibbens 1994
8,000
10,000
12,000
N )
3.0 m
3.0 m 1,400
1,600
1,800
2,000
KPa
)
Texas A&MSettlement Prediction Seminar
11
0
2,000
4,000
6,000
,
0 50 100 150 200
MOVEMENT ( mm )
L O
A D
(
KN
1.5 m
1.0 m
2.5 m
0
200
400
600
800
1,000
1,200
0 5 10 15 20
MOVEMENT / WIDTH (%)
S T
R E
S S
(
Load-Movement of Four Footings on SandTexas A&M University Experimental Site
ASCE GSP 41, J-L Briaud and R.M. Gibbens 1994
8,000
10,000
12,000
N )
3.0 m
3.0 m1,600
2,000
)
e
⎟⎟⎠
⎞⎜⎜⎝
⎛=
2
1
2
1
δδ
e = 0.4
q-z curve:
We can also borrow from pileanalysis (Pile toe response) andapply a q-z function to the stress-movement data. The "Ratio" functionis applied here.
Texas A&MSettlement Prediction Seminar
12
0
2,000
4,000
6,000
,
0 50 100 150 200
MOVEMENT ( mm )
L O
A D
(
KN
1.5 m
1.0 m
2.5 m
0
400
800
1,200
0 5 10 15 20MOVEMENT/WIDTH, δ (%)
STR
ESS,
σ
(KPa
)
3/24/2013
3
Lehane et al. 2008Settlement Prediction Seminar
200
250
300
350
400
450
500
OA
D (
KN
)
1.0 m 1.5 m
1.0 m
200
300
400
500
RES
S (K
Pa)
1.0 m
13
Lehane, B.M., Doherty, J.P., and Schneider, J.A., 2008. Settlement prediction for footings on sand. Conference on Deformational Characteristics of Geomaterials. S.E. Burns, P.W. Mayne, and J.C. Santamarina (Editors), Atlanta, September 22-24, 2008, Vol. 1, pp.133-150.
0
50
100
150
0 10 20 30 40 50
MOVEMENT (mm)
L
0
100
0 1 2 3 4 5 6 7 8
MOVEMENT / WIDTH (%)
STR
Footing, 1.5 mFooting 1.0 mFooting 1.0 m
Six footings on gravel
Caisson under air pressure to control water level.
GW//\\//\\//\//\\//\\ //\\//\\//\//\\//\\
14 m16 m
6,000
8,000
10,000
12,000
14,000
TRES
S (K
Pa)
0.3 x 0.3
14Kusakabe, O., Maeda, Y., and Ohuchi, M., 1992. Large-scale loading tests of shallow footings in pneumatic caisson. ASCE Journal of Geotechnical Engineering, 118(11) 1681-1695.
"SCORIA" = Sandy GRAVEL, trace fines. An "interlocked" and highly overconsolidated volcanic soil.
e0 = 1.2, wn = 40 %, ρ = 1,800 kg/m3
``W
Footing test
!?
0
2,000
4,000
0 5 10 15 20 25 30 35 40
NORMALIZED MOVEMENT (%)
ST
0.3 x 0.30.4 x 0.40.7 X 0.71.3 X 1.30.4 X 1.20.4 X 2.0
8,000
10,000
12,000
14,000
ESS
(KPa
)
Considering the "Preloading"/"Reloading"/"Prestress" Effect
15
0
2,000
4,000
6,000
0 5 10 15 20 25 30 35 40
NORMALIZED MOVEMENT (%)
STR
E
0.3 x 0.30.4 x 0.40.7 X 0.71.3 X 1.30.4 X 1.20.4 X 2.0
Data from Kusabe et al. 1992
Plate loading tests on 0.55 m x 0.65 m and 1.10 m x 1.30 m rectangular slabs in silty sand at Kolbyttemon, Sweden
1,500
2,000
(KPa
)
TREND1 1m x 1 3m
16Fellenius (2011). Data from Bergdahl, U., Hult, G., and Ottosson, E. (1984)
0
500
1,000
0 1 2 3 4 5 6 7 8 9 10MOVEMENT (%)
STR
ESS
0.55m x 0.65m
1.1m x 1.3m
Ultimate Shaft Resistance
rs, RsUltimate Shaft Resistance
is a reality
1717
Ultimate Toe Resistance does not exist other than as a definition of load at a certain movement
rt, Rt
Ultimate Toe Resistance does not exist other than as a definition of load at a certain movement
Ultimate Toe Resistance is not
50
100
150
200
AG
E S
HA
FT S
HEA
R(K
Pa)
O-cell to GL3
GL3 to GL1Pile D2000
2,000
3,000
4,000
RA
GE
STR
ESS
AN
DSH
EAR
(K
Pa)
Toe Resistance
Pile D2000
Shaft and toe resistances from full-scale static loading tests on a 2,000 m diameter, 85 m long bored pile in silty clay
Shaft Resistance Toe Resistance
1818
0
50
0 20 40 60 80 100
MOVEMENT (mm)
AVE
R
0
1,000
0 20 40 60 80 100MOVEMENT (mm)
AVE
R S
Shaft resistances(repeated for reference)
The above curve shows the shape of theload-movement every toe resistance."Ultimate Toe Resistance" does not exist!
A pile toe reacts to load by a stiffnessresponse and failure does not occurhowever much the pile toe is moveddown.
3/24/2013
4
• Pile capacity is the combined effect of shaft resistance and toe resistance.
• Shaft resistance is governed by shear strength, which has an ultimate value. That is, shaft capacity is reality.
• In contrast, toe resistance is governed by
1919
In contrast, toe resistance is governed by compression, which does not have an ultimate value. As the load is increased, a larger and larger soil volume is stressed to a level that produces significant compression, but no specific failure or peak value: Toe capacity is a delusion.
Analysis Methods for Determining Shaft Resistance, rs
The Total Stress Method
The Lambda Method
Th SPT M th d
2020
The SPT Method
The CPT and CPTU Methods
The Pressuremeter Method
The Beta Method
where rs = unit shaft resistance
τu = undrained shear strength
α = reduction coefficient for τu > ≈100 KPa
[ ]uusr αττ ==
Piles in Clay
Total Stress Method
"Alpha analysis"
2121
The undrained shear strength can be obtained from unconfined compression tests, field vane shear tests, or, to be fancy, from consolidated, undrained triaxial tests. Or, better, back-calculated from the results of instrumented static loading tests. However, if those tests indicate that the unit shaft resistance is constant with depth in a homogeneous soil, don’t trust the analysis!
2222
Clay adhering to extracted piles
Photo courtesy of K.R. Massarsch
The Lambda MethodVijayvergia and Focht (1972)
)2'( mms cr += σλ
where rm = mean shaft resistance along the pileλ = the ‘lambda’ correlation coefficientσ’m = mean overburden effective stresscm = mean undrained shear strength
Piles in Clay
2323
Approximate Values of λ
Embedment λ(Feet) (m) (-)
0 0 0.5010 3 0.3625 7 0.2750 15 0.2275 23 0.17
100 30 0.15200 60 0.12
The Lambda method was developed for long piles in clay deposits (offshore conditions)
{ } 'tan')/2()()(lg87.0)(016.02.28.0 2.042.0 δσ zts bhOCRSOCRr −−+=
where rs = unit shaft resistance
OCR = overconsolidation ratioSt = sensitivity
Piles in Clay
A method from fitting a variety of parameters to results from static loading tests
2424ICP (Imperial College Pile method)
Jardine, Chow, Overy, and Standing (2005 )
h = height of point above pile toe ; h ≤ 4bb = pile diameterδ’ = interface friction angle
3/24/2013
5
The SPT MethodMeyerhof (1976)
Rs = n N As D
where Rs = ultimate shaft resistance
n = a coefficient
N = average N-index along the pile shaft (taken as a pure number)
Piles in Sand
2525
g g p ( p )
As = unit shaft area; circumferential area
D = embedment depth
n = 2·103 for driven piles and 1·103 for bored piles (N/m3)[English units: 0.02 for driven piles and 0.01 for bored piles (t/ft3)]
For unit toe resistance, Meyerhof's method applies the N-index at the pile toe times a toe coefficient = 400·103 for driven piles and 120·103 for bored piles (N/m3)
[English units: 4 for driven piles and 1 for bored piles (t/ft3)]
CPT and CPTU Methods
for Calculating the Ultimate
Resistance (Capacity) of a Pile
Schmertmann and Nottingham (1975 and 1978)
2626
deRuiter and Beringen (1979)
Meyerhof (1976)
LCPC, Bustamante and Gianeselli (1982 )
ICP, Jardine, Chow, Overy, and Standing (2005)
Eslami and Fellenius (1997 )
caOCRt qCr =The CPT and CPTU Methods
where rt = pile unit toe resistance (<15 MPa)COCR = correlation coefficient governed by the
Schmertmann and Nottingham(1975 and 1978)
CLAY and SAND
SAND (alternative)ccs qKr =sfs fKr =
2727
overconsolidation ratio, OCR, of the soil qca = arithmetic average of qc in an influence zone*)
Kf = a coefficient depends on pile shape and material, cone type, and embedment ratio. In sand, the coefficient ranges from 0.8 through 2.0, and, in clay, it ranges from 0.2 through 1.25.
Kc = a dimensionless coefficient; a function of the pile type, ranging from 0.8 % through 1.8 %
qc = cone resistance (total; uncorrected for pore pressure on cone shoulder)
*) The Influence zone is 8b above and 4b below pile toe2828
Filtering of qc-values and determining pile toe resistance (Schmertmann method)
deRuiter and Beringen(1979)
uct SNr =
us Sr α=Means turning the CPT-
method into the Total St th d
2929
where rt = pile unit toe resistanceNc = conventional bearing capacity factor
Su = undrained shear strength — — — — —>
NK = a dimensionless coefficient, ranging from 15 through 20, reflecting local experience
α = adhesion factor equal to 1.0 and 0.5 for normally consolidated and overconsolidatedclays, respectively
An upper limit of 15 MPa is imposed for rt
k
cu N
qS =
Stress method
LCPC Bustamante and Gianeselli (1982 )
cs qKr =
cat qCr =
3030
C = toe coefficient ranging from 0.40 through 0.55qca = cone stress averaged in a zone 1.5 b above and
1.5 b below the pile toe plus filtering
rt = pile unit toe resistance < 15 KPa, <35 KPa, or <120 KPa, depending on soil type, pile type, and pile installation method
K = a dimensionless coefficient; a function of pile type, rangingfrom 0.5 % through 3.0 % (Compare: Schmertmann proposes 0.8 %
through 1.8 %)
3/24/2013
6
Soil Type Cone Stress Bored Piles Driven Piles Maximum rt
CLCPC CLCPC
(MPa) (- - -) (- - -) (MPa)
CLAY - - qc < 1 0.04 0.50 15
Coefficients and Limits of Unit Toe Resistance in the LCPC Method Quoted from the CFEM (1992)
3131
c
1 < qc < 5 0.35 0.45 15
5 < qc - - - 0.45 0.55 15
SAND - - - qc < 15 0.40 0.50 15
12 < qc - - - 0.30 0.40 15
Soil Type Cone Stress Concrete Piles Steel Piles Maximum rs(MPa) & Bored Piles
KLCPC KLCPC J
(- - -) (- - -) (KPa)
CLAY - - qc < 1 0.011 (1/90) 0.033 (=1/30) 15
1 5 0 025 (1/40) 0 011 ( 1/80) 35
Coefficients and Limits of Unit Shaft Resistance in the LCPC Method Quoted from the CFEM (1992)
3232
1 < qc < 5 0.025 (1/40) 0.011 (=1/80) 35
5 < qc - - - 0.017 (1/60) 0.008 (=1/120) 35
SAND - - - qc < 5 0.017 (1/60) 0.008 (=1/120) 35
5 < qc < 12 0.010 (1/100) 0.005 (=1/200) 80
12 < qc - - - 0.007 (1/150) 0.005 (=1/200) 120
The values in the parentheses are the inverse of the KLCPC-coefficient
cac
t qdbr )5.01( −=
cJs qKr =
σ ' b
ICP (Imperial College Pile method)Jardine, Chow, Overy, and Standing (2005 )
3333
δσσσ tan)')()'(0145.0( 38.013.0
mtr
zcJ h
bqK Δ+=
bq
qqrz
rzccmc 01.0)]
'(10216.1)'(00125.00203.0(2[' 1
265.0 −−− ∗−+=Δ
σσσσσ
Egtt qCr =Eslami and Fellenius
(1997 )
Ess qCr =
rt = pile unit toe resistance
Ct = toe correlation coefficient (toe adjustment factor)—equal to unity in most cases
Shaft Correlation Coefficient
Soil Type*) Cs
Soft sensitive soils 8 0 %
bCt 3
1=
bCt
12=
b in metre
b in inch
3434
qEg = geometric average of the cone point resistance over the influence*) zone after correction for pore
pressure on shoulder and adjustment to “effective” stress rs = pile unit shaft resistanceCs = shaft correlation coefficient, which is a function of soil
type determined from the soil profiling chartqE = cone point resistance after correction for pore pressure
on the cone shoulder and adjustment to “effective” stress
*) The Influence zone is 8b above and 4b below pile toe
Soft sensitive soils 8.0 %Clay 5.0 %Stiff clay andClay and silt mixture 2.5 %Sandy silt and silt 1.5 %Fine Sand and silty Sand 1.0 %Sand to sandy gravel 0.4 %
*) determined directly from the CPTU soil profiling
Unit shaft resistance as a function of cone stress, qc in Sandaccording to the LCPC method and compared to the Eslami-Fellenius method
100
120
140
ce, r
s (K
Pa)
Sandy Silt to silty Sand to sandy Gravel
Concrete
Range for the Eslami Fellenius method
3535
0
20
40
60
80
0 5 10 15 20 25 30 35 40
Cone Stress, qc (MPa)
Uni
t Sha
ft R
esis
tan piles
Steel piles
PILES IN SAND
Cone Stress, qc and qt (MPa)
Pile Capacity or, rather, Load-Transfer follows
principles of effective stress
3636
principles of effective stress and is best analyzed using the
Beta method
3/24/2013
7
the Beta method
Unit Shaft Resistance, rs
zsr 'βσ=
where c‘ = effective cohesion interceptβ = Bjerrum-Burland coefficientσ'z = effective overburden stress
Effective Stress Analysis (Beta-analysis as opposed to Alpha analysis)
3737
dzcAdzrAR zssss )''( βσ+∫=∫=Total Shaft Resistance, Rs
where As = circumferential area of the pile at Depth z(surface area over a unit length of the pile)
Shaft Resistance — in Sand and in Clay
KMr ''tan σφ=
vsr 'σβ=
3838
where rs = unit shaft resistance
M = tan δ’ / tan φ’
Ks = earth stress ratio = σ’h / σ’vσ‘v = effective overburden stress
vss KMr tan σφ=
Approximate Range of Beta-coefficients
SOIL TYPE Phi Beta
Clay 25 - 30 0.20 - 0.35
Silt 28 - 34 0.25 - 0.50
Sand 32 - 40 0.30 - 0.90
Gravel 35 - 45 0.35 - 0.80
0.05 - 0.80 !
3939
Gravel 35 45 0.35 0.80
These ranges are typical values found in some cases. In any given case,actual values may deviate considerably from those in the table.
Practice is to apply different values to driven as opposed to bored piles, but ....
2.0
3.0
4.0
5.0
6.0
coef
ficie
nt i
n s
and
G
Trend line
4040
0.0
1.0
0 5 10 15 20 25 30
LENGTH IN SOIL (m)
ß-c
HK GEO (2005)CFEM (1992)
Gregersen et al. 1973
Beta-coefficient versus embedment length for piles in sand (Data from Rollins et al. 2005). Ranges suggested by CFEM (1993), Gregersen et al 1973, and Hong Kong Geo (2005) have been added.
1.00
1.50
2.00
2.50
OEF
FIC
IEN
T IN
SA
ND
Concrete piles
Open-toe pipe piles
Closed-toe pipe piles
Gregersen
4141
0.00
0.50
0 50 100 150 200 250 300 350
AVERAGE EFFECTIVE STRESS, σ'z (KPa)
ß-C
O et al. 1973
Beta-coefficient versus average σ’ for piles in sand. (Data from Clausen et al. 2005).
1.00
1.50
2.00
2.50
FIC
IEN
T IN
SA
ND
Concrete piles
Open-toe pipe piles
Closed-toe pipe piles
4242
0.00
0.50
0.0 0.2 0.4 0.6 0.8 1.0 1.2
AVERAGE DENSITY INDEX, I D
ß-C
OEF
Beta-coefficient versus average ID for piles in sand. (Data from Karlsrud et al. 2005).
3/24/2013
8
0.20
0.30
0.40
0.50
0.60co
effic
ient
in
cla
y
Norway Japan Thailand Vancouver Alberta
4343
0.00
0.10
0 20 40 60 80
PLASTICITY INDEX, I P
ß-c
Beta-coefficient versus average IP for piles in clay. (Data from Karlsrud et al. 2005 with values added from five case histories).
c
CCI
CKr
vD
C eCe
K φβ σσ
tan'
ln100
10
24
302
⎟⎟⎠
⎞⎜⎜⎝
⎛−
−=
Where K = coefficient of earth stress at restI = density index (“relative density”)
The Beta-coefficient has a certain appeal to the academia it seems. This is what is proposed in a recent issue of the ASCE Journal.
44
ID = density index ( relative density )σ’v = effective overburden stressσr = reference stress = 100 KPaΦ = triaxial-compression critical-state
friction angleC1 = a coefficient: 0.6< C1 <0.7C2 = a constant = 0.2C3 = a constant = 0.4C4 = a constant = 1.3
Unit Toe Resistance, rt
where Nt = toe “bearing capacity” coefficient
D = depth to pile toeσ'z=D = effective overburden stress at the pile toe
Dztt Nr == 'σ
Toe Resistance
4545
Total Toe Resistance, Rt
where At = toe area (normally, the cross sectional area of the pile)
Dzttttt NArAR === 'σ
Approximate Range of Nt-coefficients
SOIL TYPE Phi Nt
Clay 25 - 30 3 - 30Silt 28 - 34 20 - 40Sand 32 - 40 30 - 150Gravel 35 - 45 60 - 300
4646
The Toe Resistance, Rt, while not really an “ultimate” resistance, isusually considered as such in design. However, toe resistance should bethought of as that mobilized in a static loading test at the maximumacceptable movement usually considered applicable to a piled foundation.
Also the toe resistance appears to have certain qualitiesintriguing to the academia. This is what is proposed inthe same recent issue of the ASCE Journal.
DDccD ICC
r
hICCCr
ICtoeu eCeCr 876542 )'()(
21,−−+−=
σσσ φφ
Where ru, toe = ultimate toe resistance for a pile head movementequal to 10 % of the pile diameter
ID = density index (“relative density”)
!!!
47
D y ( y )σ’h = effective horizontal stress (= σ’v/K0?)Φ = triaxial-compression critical-state friction angleC1 = a constant = 0.23C2 = a constant = 1.64C3 = a constant = 0.0066C4 = a constant = 0.10414C5 = a constant = 0.0264C6 = a constant = 0.0002C7 = a constant = 0.841C8 = a constant = 0.0047
Total Resistance (“Capacity”)
tsult RRQ +=
suzsuz RQdzAQQ −=∫−= 'σβ
0
5
10
0 500 1000 1500 2000
LOAD
H
Qult/ Rult
4848
15
20
25
DEP
TH
Rt Rs
Effective stress — Beta — analysis is the method closest to the real response of a pile to an imposed load
3/24/2013
9
0
1
2
0 50 100 150
UNIT SHAFT RESISTANCE (KPa)
0
1
2
0 100 200 300 400 500 600 700 800
TOTAL SHAFT RESISTANCE (KN)
Pile CCPT-3
Calculations of unit and total shaft resistances for a pile driven into asaprolite (residual soil) in Porto, Portugal. The soil can be classified bothas a clay type and sand type.
Shaft resistance by CPT-methods
4949
3
4
5
6
DEP
TH (
m)
DutchSand
MeyerhofSand
LCPCSand
LCPCClay Schmertmann
Clay
Eslami-Fellenius
SchmertmannSand
DutchClay
TumayClaya
3
4
5
6
DEPT
H (
m)
Effective StressBeta = 1.00
DutchSand
MeyerhofSand
LCPCClay &Sand
SchmertmannClay
Eslami-Fellenius
SchmertmannSand
DutchClay
TumayClayb
0
1
2
0 200 400 600 800 1,000 1,200 1,400 1,600 1,800 2,000
CALCULATED PILE RESISTANCE (KN)
TumayClay
Eslami-Fellenius
SchmertmannClayDutch
Clay
LCPC
DutchSand
MeyerhofSand
Pile CCPT-3
Total resistance by CPT-methods
5050
3
4
5
6
DEP
TH (
m) Schmertmann
Sand
LCPCSand
LCPCClay
a
Let’s look at a few case studies
Annacis/Lulu Island Tests, Vancouver,
BC
by UBC 1985
5151
Static loading tests on three 324 mm diameter pipe piles driven to depths of 14 m, 17 m, and 31 m into the Fraser River deltaic soils
0
5
10
15
20
0 5 10 15Cone Stress, qt (MPa)
PTH
(m
)
0
5
10
15
20
0 100 200
Sleeve Friction (KPa)
TH (
m)
0
5
10
15
20
0 500 1,000
Pore Pressure (KPa)
PTH
(m
)
0
5
10
15
20
0 1 2 3 4 5
Friction Ratio (%)
PTH
(m
)
PILES1 2 3 4
PROFILE
Eslami-Fellenius Robertson
CLAY CLAY
SANDSAND
SANDGRAVEL & SAND
CPT and CPTU analysis for capacity
5252
25
30
35
40
DEP
25
30
35
40
DEP
25
30
35
40D
EP
25
30
35
40
DEP
CLAY andSilty
CLAY
CLAY andSilty
CLAY
Annacis/Lulu Island Tests by UBC 1985
The results of the load-movement curves from all three tests combined in
600
800
1,000
1,200
OA
D (K
N)
Depth 16.8 mSet-up Time
85 days
Depth 31.1 mSet-up Time
38 days
5353Data from Lulu Island Tests
by UBC 1985
tests combined in one graph. (With offset limit lines and maximum load in the tests).
0
200
400
0 10 20 30 40
MOVEMENT (mm)
LO
Depth 13.7 mSet-up Time
197 days
Results of CPT and CPTU analysis compared tocapacity from the static loading tests
0
5
10
0 500 1,000 1,500 2,000
SHAFT RESISTANCE (KN)
Eslami-Fellenius
DutchLCPC
SchmertmannUniPile eff.stress
ß = 0 15
0
5
10
0 500 1,000 1,500 2,000
SHAFT and TOE RESISTANCEs (KN)
Eslami-FelleniusDutchLCPCSchmertmannUniPile eff. stressPile static tests
ß = 0.15
5454“UniPile eff.stress” is effective stress analysis matched to results of static tests
15
20
25
30
35
DEP
TH (
m)
ß = 0.15
ß = 0.20
ß = 0.15
15
20
25
30
35
DEP
TH (
m)
ß 0.15 Nt = 7
ß = 0.20 Nt = 25
ß = 0.15 Nt = 3
Test too soon after EOID
3/24/2013
10
150
a)
O-cell to GL3 GL3 to GL2 GL2 to GL1O-cell to GL2 O-cell to GL1
Sunrise City Project, HoChiMinh City, Vietnam1,800 mm diameter bored piles constructed to 70 m depthUnit shaft resistances versus measured downward movement at depths of ≈50 m
150
Pa)
O-cell to GL4 GL4 to GL3 GL3 to GL2O-cell to GL3 O-cell to GL2 O-cell to GL1
SHAFT RESISTANCE
HoChiMinh
Ha Noi
Cai Mep Port
55
0
25
50
75
100
125
0 1 2 3 4 5 6 7 8 9 10
MOVEMENT (mm)
UN
IT S
HA
FT R
ESIS
TAN
CE
(KPa
TBP-1
Next reading was at 56 mm
ß = 0.14
0
25
50
75
100
125
0 1 2 3 4 5 6 7 8 9 10
MOVEMENT (mm)
UN
IT S
HA
FT R
ESI
STA
NC
E (K
P
TBP-2
ß = 0.13
Next reading was at 35 mm
No records were obtained during the sudden movement occurring at about 5 mm
0
500
1,000
1,500
2,000
2,500
0 50 100 150 200
MOVEMENT (mm)
UN
IT R
ESIS
TAN
CE
(KPa
)
TBP-1
Unit Toe Resistance
Unit Shaft Resistances
10% of diameter
TOE RESISTANCE
56
0
500
1,000
1,500
2,000
2,500
0 50 100 150 200
MOVEMENT (mm)
UN
IT R
ESIS
TAN
CE
(K
Pa)
TBP-2
TBP-1Unit Toe Resistance
Unit Shaft Resistances
The stiffness of the toe stress-movement is unusually soft for adense sand and typical of a pilehaving a layer of debris at the bottomof the shaft when the concrete wasplaced. A pile a few metre to the sidewas constructed using the samemethod and equipped with a coringtube. Coring through this pile toe intothe soil two weeks after constructionrevealed presence of about 30 mm ofsoft material between the pile and thesoil.
Core from the pile toe and into the soil below
57
Bridge over Panama Canal, Paraiso Reach, Republic of PanamaO-cell test on a 2.0 m (80 inches) diameter, 30 m (100 ft) deep shaft
drilled into the Pedro Miguel and Cucaracha formations, February 2003.
0
5
0 5,000 10,000 15,000 20,000 25,000 30,000
LOAD (KN) ß
0.30
0.45
5858
10
15
20
25
30
DEP
TH (
m) 0.30
___
1.20
O-cell Tests on an 11 m long, 460 mm square precast concrete pile driven in silica sand in
North-East Florida(Data from McVay et al 1999)
0
2
4
6
8
0 500 1,000 1,500 2,000 2,500 3,000
Shaft Resistance, Rs (KN)
(m
)
E-FLCPCSchmertmannDutchMeyerhofBetaTests
5959
(Data from McVay et al. 1999)
A study of Toe and Shaft Resistance
Response to Loading
10
12
14
16
18
20
DEP
TH
The foregoing analysis results are quite good predictions
They were performed after the test results were known
Such “predictions” are always the best!
So, what about true predictions?
6060
Let’s see the results of a couple ofPrediction Events
p
3/24/2013
11
ULTIMATE R
Prediction Event at Deep Foundations Institute Conference in Raleigh, 1988
6161
44 ft embedment, 12.5 inch square precast concrete driven through compact silt and into dense sand
Capacity in Static Loading Test = 200 tonsRESISTANCE
TonsPREDICTORS (60 individuals)
1,500
2,000
2,500
ity (
KN
)
Orlando 2002 Predictions
Max LoadAvailable
6262
0
500
1,000
Predictors
Cap
ac
500
600
700
KN
)
0 20 40 60 80MOVEMENT (mm)
FHWA Washington, DC, 1986
273 mm diam. closed-toe pipe pile driven 9.1 m into hydraulic sand fill
6363
0
100
200
300
400
1 2 3 4 5 6 7 8 9 10PREDICTIONS
CA
PAC
ITY
( 800
1,000
1,200K
N)
0 2 4 6 8 10 12 14 16 18MOVEMENT (mm)
FHWA Baltimore, MD, 1980
Two 273 mm diam. closed-toe pipe piles driven 13.1 m into Beaumont clay
6464
0
200
400
600
800
PARTICIPANTS
CA
PAC
ITY
(K
1,500
2,000
2,500
3,000
3,500
OA
D (
KN
)
Singapore 2002
1,400
1,600
1,800
2,000
65
0
500
1,000
1 3 5 7 9 11 13 15 17 19 21 23 25 27 29 31 33
L
400 mm H-Pile (168 kg/m) driven through sandy clay to a 15 m embedment
0
200
400
600
800
1,000
1,200
0 10 20 30 40 50MOVEMENT (mm)
LOA
D (
KN
)
Brazil 2004: Bored pile (Omega screw pile) 23 m long, 310 mm diameter
0
2
4
6
8
0 20 40 60 80
Water Content (%)
(m)
0
5
10
0 5 10 15 20 25N-Index (blows/0.3)
(m)
SPT 18at 23 m Pile
0
2
4
6
8
0 25 50 75 100
Grain Size (%)
(m)
SILT
SAND CLAY
Sandy SiltyCLAY (Laterite)
Sandy SILT
6666
10
12
14
16
18
20
DEP
TH
wnwP wLGW
15
20
25
DEP
TH
10
12
14
16
18
20
DEP
TH Sandy SILT
and CLAY
Sandy ClayeySILTGW
3/24/2013
12
Brazil 2004
Static Loading Test
on a 23 m 310 mm bored pile
Load-Movement Response
1,500
2,000
2,500
KN
)
Prediction Compilation
2,000
2,500PUSH L= 23m
0 5 10 15 20 25 30
MOVEMENT (mm)
6767
0
500
1,000
0 10 20 30 40
MOVEMENT (mm)
LOA
D (
K
0
500
1,000
1,500
PARTICIPANTS
LOA
D (
KN
)
Portugal 2004. Precast 350 mm diameter pile driven to 6 m depthin a saprolite, a residual soil consisting of silty clayey sand.
0
1
2
3
0 10 20Cone Stress, qt (MPa)
) 1 500
2,000
2,500
3,000
PAC
ITY
(KN
)
CAPACITY FROM STATIC LOADING TEST
Pile C1
6868
4
5
6
7
8
DEP
TH (
m
0
500
1,000
1,500
1PREDICTIONS
TOTA
L C
AP
0
OFFSET LIMIT LOAD
1,200
1,400
1,600
1,800
KN
)
Pipe-Pile
0 10 20 30 40MOVEMENT (mm)
Northwestern University, Evanston, Illinois, 1989.15 m embedment, 457 mm diameter closed-toe pipe piles driven in sand on clay.
6969
0
200
400
600
800
1,000
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24
CA
PAC
ITY
(K
PREDICTIONS
Finno 1989
Edmonton, Alberta, 2011
Prediction of load-movement and capacity of a 400-mm diameter, 18 mlong, augercast pile constructed in transported and re-deposited glacial till.
2 000
3,000
4,000D
(KN
)E = 20 GPaE = 35 GPa
70
0
1,000
2,000
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT (mm)
LOA
D
10 capacity predictions are at movements > 50 mm
7 mm (4 mm + b/120 mm)TEST RESULTS
Values
Rea
ltive
Fre
quen
cy
2σ
σ = Standard Deviation, σ = 833 µ = Mean, 1,923 σ/µ = Coefficient of Variation, COV = 0.43
Mean, µ
4σ
NORMAL DISTRIBUTIONEdmonton 2011
71
CAPACITY PREDICTIONS
T he area between -σ and +σ f ro m the mean value is 68% o f to tal areaT he area between -2σ and +2σ f ro m the mean value is 95% o f to tal areaT he area between -3σ and +3σ f ro m the mean value is 99% o f to tal area
0
100
200
0 10 20 30 40 50 60 70 80 90 100
PRED
ICTE
D L
OA
D =
100
MOVEMENT (mm)
NORMALIZED TO LOAD
Forthcoming Prediction Event in Bolivia April 2013
Four bored instrumented piles in sand tested in compression
0
5
2.9 m GW1.0 m
4.5 m
1.0 m
4.5 m
Groundsurface
TP1 TP2 TP3 TP4 "Std" FDP FDP "Std" +EB +O-cell +EB BH1 BH3 BH 4 BH2
0.0 m 400 mm 440 mm 400 mm 400 mm
72
5
10
15
20
25
DE
PTH
(m
)
1.2 m17.5 m 2.5 m
15.0 m
O-cell
EB EB
600 mm
5
7.5 m
10.5 m
13.5 m
16.5 m 15.8 m
4.5 m
7.5 m
10.5 m
13.5 m
Test Pile Configurations and Strain-Gage Levels
440 mm400 mm 600 mm
3/24/2013
13
0
2
4
6
8
10
0 10 20 30 40 50N (blows/0.3m)
PTH
(m
)
SPT1SPT2SPT3
0
2
4
6
8
10
0 5 10 15 20 25 30
WATER CONTENT (%)
TH (
m)
0
2
4
6
8
10
0 20 40 60 80 100
GRAIN SIZE (%)
TH (
m)
Fine to Medium Sand
Medium to Coarse SandFines
BH-1
Soil Profile
73
12
14
16
18
20
DEP
10
12
14
16
18
20
DEP
T 10
12
14
16
18
20
DEP
T Gravel
Zone of Clay and Clayey Sand (no samples)
Deadline for submitting a prediction is April 1I will be glad to email the details for how to submit one.
Pore Pressure Dissipation
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
7474Paddle River, Alberta, Canada (Fellenius 2008)
15
20
25
DEP
TH
Before Driving
EOID
Total Stress
15
20
25
DEP
TH
30 Days after EOID 15 Days
after EOID
Before Driving
EOID
Total Stress
15
20
25
DEP
TH
4 Years after Driving
30 Days after EOID 15 Days
after EOID
Before Driving
EOID
Total Stress
800
1,000
1,200
1,400
1,600
D (
KN
)
Effective Stress Analysis
0
5
10
0 500 1,000 1,500 2,000
LOAD (KN)
(m)
4 Years after EOID
7575
0
200
400
600
0 10 20 30 40 50
MOVEMENT (mm)
LOA
Paddle River, Alberta, Canada
15
20
25
DEP
TH (
15 Days after EOID
30 Days after EOID
All three analyses apply the same coefficients coupled with the actual
pore pressure distribution
If we want to know the load distribution, we can measure it. But, what we measure is the increase of load in the pile due to the load applied to the pile head. What about the load in the pile that was there before
76
pwe started the test?
That is, the Residual load.
Normalized Applied Load
Load distributions in
static loading tests on
four instrumented
77
D E P T H
piles in clayS d
Example from Gregersen et al., 1973
0
2
4
6
8
0 50 100 150 200 250 300
LOAD (KN)
(m)
0
2
4
6
8
0 100 200 300 400 500 600
LOAD (KN)
(m) True
Residual
True minus Residual
78
B. Load and resistance in DA
for the ultimate load applied
Sand8
10
12
14
16
18
DE
PTH
(
Pile DA
Pile BC, Tapered
8
10
12
14
16
18
DE
PTH
(
A. Distribution of residual load in DA and BC
before start of the loading test
3/24/2013
14
FHWA tests on 0.9 m diameter bored pilesOne in sand and one in clay
(Baker et al., 1990 and Briaud et al., 2000)
0
2
4
0 10 20 30 40
Cone Stress and SPT N-Index(MPa and bl/0.3 m)
Silty Sand
0
2
4
0 10 20 30 40
Cone Stress (MPa)
ClaySilty
Sand Clay
79
6
8
10
12
DEPT
H (m
)
Sand
Pile 4
6
8
10
12
DEPT
H (m
)
Pile 7
N
qc Sand Clay
ANALYSIS RESULTS: Load-transfer curves
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
)
True Distribution
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
Measured Distribution
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
True Distribution
Residual Load
80
6.0
8.0
10.0
12.0
DEP
TH (
m
PILE 4SAND
Measured Distribution6.0
8.0
10.0
12.0
DEP
TH (
m)
PILE 4SAND
Residual Load
Measured Distribution
6.0
8.0
10.0
12.0
DEP
TH (
m
PILE 7CLAY
6.0
8.0
10.0
12.0
DEP
TH (
m
PILE 7CLAY
Results of analysis of a Monotube pile in sand(Fellenius et al., 2000)
0
5
0 1,000 2,000 3,000
LOAD (KN)
Measured Resistance
Residual Load
81
10
15
20
25
DE
PTH
(m
)
True Resistance
Method for evaluating the residual load distribution
0
2
4
0 500 1,000 1,500 2,000
RESISTANCE (KN)
Measured Load
Shaft
82
6
8
10
12
14
16
DE
PTH
(m
)
Measured Shaft ResistanceDivided by 2
Residual Load
True Resistance
ExtrapolatedTrue Resistance
Resistance
0
5
10
15
20
0 500 1,000 1,500 2,000 2,500LOAD (KN)
(m)
Static Loading Testat Pend Oreille, Sandpoint, Idaho, for
the realignment of US95
406 m diameter,45 m long, closed-toe pipe pile
driven in soft clay
Determining True Resistancefrom Measured Resistance (“False Resistance”)
Cl
83
25
30
35
40
45
50
DEP
TH (
Fellenius et al. (2004)
driven in soft clay
200+ m
Clay
0
5
10
15
20
-500 0 500 1,000 1,500 2,000
LOAD (KN)
(m)
ß = 0.60
ß = 0.06
AS MEASURED,i.e. "FALSE RES."
A
ß = 0.09
0
5
10
15
20
-500 0 500 1,000 1,500 2,000
LOAD (KN)
(m)
ß = 0.60
ß = 0.09
ß = 0.09
AS MEASURED,i.e. "FALSE RES."
CPTu Eslami-Fellenius
B
84
Test on a strain-gage instrumented, 406 mm diameter,45 m long pile driven in soft clay in Sandpoint, Idaho
25
30
35
40
45
50
DE
PTH
ß = 0.06
"TRUE RES." RESIDUAL LOAD
AFTER 1st UNLOADING
25
30
35
40
45
50
DE
PTH
ß = 0.10
"TRUE RES." per CPTu
RESIDUAL LOAD
AFTER 1st UNLOADING
ß = 0.10
Extrapolated
3/24/2013
15
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
PTH
(m
)
True Resistance
HEAD-DOWN AND FULL RESIDUAL LOAD
Residual Load
True Resistance
False Resistance
Silty Sand
Silty Clay
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
PTH
(m
)
HEAD-DOWN AND PARTIAL RESIDUAL LOAD
True
False Resistance
Shaft Resistance
Typical Example: Table 7.3 in the Red Book
85
20
25
30
35
DE
P Resistance
Residual and TrueToe Resistance
Transition Zone
Silty Sand
Glacial Till
20
25
30
35
DEP
Residual Load
Resistance
Residual and TrueToe Resistance
Transition Zone
Resistance
The effect of residual load on an uplift test
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
True Resistance
TENSION TESTAND FULL RESIDUAL LOAD
Residual Load
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
Residual Load
True Resistance
TENSION TESTAND PARTIAL RESIDUAL LOAD
8686
15
20
25
30
35
DE
PTH
(m
False Resistance
Toe Resistancein an Uplift Test?!
15
20
25
30
35
DEP
TH (
m
False Resistance
Toe Resistancein an Uplift Test?
Combining the results of a head-down test with those of a tensions test will help determining the true resistance
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
H (
m)
HEAD-DOWN AND PARTIAL RESIDUAL LOAD
FalseHead-down
True Shaft
False TensionTest
8787
20
25
30
35
DEP
TH
Residual Load
True Resistance
Residual and TrueToe Resistance
Transition Zone
True Shaft Resistance
Not directly useful below this level
Now you know why some claim that resistance in tension is smaller than that in compression
400
600
800
1,000
LOA
D (
KN
)
No Residual Load
Residual Load present
No Strain Softening
Presence of residual load is not just of academic interest
400
600
800
1,000
LOA
D (
KN
)
With Strain Softening
Residual Load present
No Residual Load
8888
0
200
400
0 5 10 15 20 25 30
MOVEMENT (mm)
L
OFFSET LIMIT LOAD
0
200
400
0 5 10 15 20 25 30
MOVEMENT (mm)
L
OFFSET LIMIT LOAD
• "Residual Load " follows the same principle and mechanism as "Drag Load". The distinction made is that by residual load we mean the locked-in load present in the pile immediately before we start a static loading test. By drag load we mean the load present in the pile in the long-term.
Additional Comments on Residual load
8989
• Residual load as well as drag load can develop in coarse-grained soil just as it does in clay soil.
• Both residual load and drag load develop at very small movements between the pile and the
soil.
600
800
1,000
1,200
D (
KN
)
HEADTOE TELLTALE
A
Does not this shape of
Residual Load Affects Toe Resistance Response
9090
0
200
400
600
0 5 10 15 20 25
MOVEMENT (mm)
LOAD TOE
Does not this shape of measured toe movement
suggest that there is a distinct toe capacity?
3/24/2013
16
400
600
800
1,000
1,200
LOAD
(K
N)
HEAD
TOE
TOE TELLTALEA
400
600
800
1,000
1,200
LOAD
(K
N)
HEAD
TOE
B
9191
0
200
0 5 10 15 20 25
MOVEMENT (mm)
0
200
0 5 10 15 20 25
MOVEMENT (mm)"Virgin" Toe Curve
No, it only appears that way when we forget to consider the residual toe load (also called the initial, or “virgin” toe movement)
Miscellaneous DetailsOpen vs. Closed Toe
Tapered sectionH section
9292
H-section. . . . . . .
Special Conditions
Step-tapered pile
9393
"Add-on" toe resistance acting on a donut-shaped area
Special Conditions
Step-tapered pileSmooth-tapered pile
Conical pile (wood pile)
Calculate in elements
(increments) at t
9494
"Add-on" toe resistance acting on a donut-shaped area
every metre or so the shaft resistance acting along the pile and toe resistance for the “donut” of
each element
Just because the design assumes that the pile shaft issmooth and straight with parallel sides does not mean it is.
9595
A A
B B
A-A and B-B
The "donut" area A minus B projection acting like an extra Pile Toe
An unintentional effect for many bored piles and intentional for “multi-underreamed” piles
9696
3/24/2013
17
9797 9898
PILES FOR AN EXPANSION OF A LOADING DOCK
9999
CALCULATION OF PILE CAPACITY and
LOAD-TRANSFER CURVES
355 mm diameter closed-toe pipe pile to 32 m embedment
Area, As = 1.115 m2/m Live Load, Ql = 200 KN
Area, At = 0.099 m2 Dead Load, Qd = 800 KN
SILT
CLAY
4 m
W
5 m
100100
, t , d
LAYER 1 Sandy Silt ρ = 2,000 kg/m β = 0.40
LAYER 2 Soft Clay ρ = 1,700 kg/m3 β = 0.30
LAYER 3 Silty sand ρ = 2,100 kg/m3 β = 0.50With artesian head of 5 m
LAYER 4 Ablation Till ρ = 2,200 kg/m3 β = 0.55Nt = 50 TILL
SAND
27 m
21 m
32 m
CALCULATION OF LOAD TRANSFERArea, As = 1.115 m2/m Live Load, Ql = 200 KN Shaft Resistance, Rs = 1,817 KNArea, At = 0.099 m2 Dead Load, Qd = 800 KN Toe Resistance, Rt = 1,205 KN
Total Load, Qa = 1,000 KN Total Resistance, Ru = 3,021 KNF.S. = 3.02 Depth to N. P. = 26.51 m Load at N. P., Qmax = 1,911 KN
DEPTH TOTAL PORE EFFECTIVE INCR. Qd+Qn Qu-RsSTRESS PRES. STRESS Rs
(m) (KPa) (KPa) (KPa) (KN) (KN) (KN)
LAYER 1 Sandy Silt ρ = 2,000 kg/m3 β = 0.400.00 30.00 0.00 30.00 0.0 800 3,0211.00(GWT) 48.40 0.00 48.40 17.5 817 3,0044.00 104.30 30.00 74.30 82.1 900 2,922
LAYER 2 Soft Clay ρ = 1,700 kg/m3 β = 0.304.00 104.30 30.00 74.30 900 2,9226.00 136.04 57.06 78.98 26.0 951 2,8708.00 168.08 84.12 83.96 27.7 1,005 2,816
10 00 200 37 111 18 89 20 29 4 1 063 2 758
101101
10.00 200.37 111.18 89.20 29.4 1,063 2,75812.00 232.88 138.24 94.64 31.2 1,125 2,69714.00 265.55 165.29 100.26 33.1 1,190 2,63116.00 298.38 192.35 106.03 35.0 1,259 2,56218.00 331.33 219.41 111.92 37.0 1,332 2,48920.00 364.40 246.47 117.93 39.0 1,409 2,41321.00 380.97 260.00 120.97 40.0 1,449 2,373
LAYER 3 Silty sand = 2,100 kg/m3 β = 0.5021.00 380.97 260.00 120.97 1,449 2,37323.00 422.17 280.00 142.17 76.3 1,596 2,22625.00 463.45 300.00 163.45 88.2 1,766 2,05527.00 504.80 320.00 184.80 100.1 1,960 1,861
LAYER 4 Ablation Till = 2,200 kg/m3 β = 0.5527.00 504.80 320.00 184.80 1,960 1,86130.00 569.93 350.00 219.93 372.4 2,332 1,48932.00 613.41 370.00 243.41 285.1 2,617 1,205 for Nt = 50
0
5
10
0 1,000 2,000 3,000 4,000
LOAD and RESISTANCE (KN)
Qd + qn
Qd
QallowQlive Qu
Plot of the Calculated Values
zs cr '' βσ+=
Dztt Nr == 'σ
Calculation of shaft and toe resistance per the
effective stress method
102102
15
20
25
30
35
DEP
TH (
m)
Qu - rs
Rt
dzcAdzrAR zssss )''( βσ+∫=∫=
Dzttttt NArAR === 'σ
Mother Nature no like no kinkie stuff
3/24/2013
18
0
5
10
15
0 1,000 2,000 3,000 4,000
LOAD and RESISTANCE (KN)
H (
m)
103103
20
25
30
35
DEP
TH
Transition Zone
Qn
Note, just because we carried the static loading test to acertain toe movement does not mean that Nature willimpose the same toe load and toe movement for thelong-term condition.
0
5
10
0 500 1000 1500 2000
LOADQult/ RultQdead
0
5
10
0 500 1000 1500 2000
LOADQult/ RultQdead
104104
A) Small settlement only in the surrounding soils B) Large settlement in the surrounding soils
15
20
25
DEP
TH
Rs
Qn
(Rt)
15
20
25
DEP
TH
(Rt) Rs
Qn
RESIDUAL LOAD
0
5
0 500 1000 1500 2000
LOADQult/ Rult
A test pile.
Before the start of the test there is no
105105
10
15
20
25
DEP
TH
Residual Toe Load
load on the pile head
A Case history of evaluation of static and dynamic tests on a 300 mm, 12 m long pile driven in sand. Data from Axelsson (2000).
GW
Silty CLAY
SAND with lenses of clay and silty clay
Uniform SAND (80% sand size)
with occasional
9.25"235 mm
0 m
2.5
m
T E S T S
Static loading test 5 days after driving at Depth 12.8 m
Restrike after static test to final depth 13.0 m with PDA/CAPWAP
106106
with occasional lens of Silty CLAY13
.0
Redrive to 13.0 m depth
Static loading test 1 day after redrive
Static loading test 8 days after redrive
Static loading test 120 days (4 months) after redrive
Static loading test 670 days (22 months) after redrive
Total unit weight 0 m - 2.5 m = 18 KN/m3
Total unit weight 2.5 m - 13.0 m = 19 KN/m3
Hydrostatic pore pressure distribution
cnt.
107107
0123
0.00 0.20 0.40 0.60 0.80 1.00
Equivalent ß (- - -)
0123
0 25 50 75 100
Unit Shaft Resistance (KPa)
0123
0 25 50 75 100
Unit Shaft Resistance (KPa)
ß-Method E-F Method
Equivalent ß-coefficient from CPTU sounding and Eslami-Fellenius Method
Unit Shaft Resistance from Equivalent ß-coefficient and CPTU Method plus LCPC-Method
cnt.
108108
456789
10111213
DEP
TH (
m)
E-F Method
456789
10111213
DEP
TH (
m) 4
56789
10111213
DEP
TH (
m)
LCPC Method
3/24/2013
19
250
300
350
400
450
500
(KN
)
Static test 8 days after driving
PDA/CAPWAP after static test
Load-movement curves from a static loading test andthe CAPWAP-determined load-movement curve from asubsequent same-day dynamic test.
cnt.
109109
0
50
100
150
200
250
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT (mm)
LOAD
Load-Movement Curves for static tests after the redrive
cnt.
250
300
350
400
450
500
OAD
(KN
)
1 Day
8 Days
4 Months
4 Months(Reloading)22 Months
110
0
50
100
150
200
0 10 20 30 40 50 60 70
MOVEMENT (mm)
LO
An obvious example of set-up in sand — Right?
300
350
400
450
500
KN)
1 Day
8 Days
cnt.
When plotting the data in sequence as the testsprogressed from unloadings to reloadings, notime-dependent increase can be discerned.
111
0
50
100
150
200
250
0 25 50 75 100 125 150 175 200
MOVEMENT (mm)
LOAD
(K 8 Days
4 Months
4 Months(Reloading)
22 Months
100
150
200
OE
LOA
D (
KN
)
1 Day
8 Days
4 Months
22 Months
Toe load from earth stress cell at pile toe
cnt.
112
0
50
0 50 100 150
MOVEMENT OF PILE HEAD (mm)
TO
This indicates an ultimate toe resistance, i.e., no increase of toe resistance for increasing toe movement — Right?
Toe load from earth stress cell at pile toe
cnt.
100
150
200
INC
REA
SE
(KN
)
1 Day8 Days4 Months22 Months
113
Residual Toe Load
The entire history of the toe response needs to be considered. A plot of entire history does not show an ultimate value. Residual load can be determined from instrumented tests.
0
50
0 25 50 75 100 125 150MOVEMENT (mm)
LOA
D
Redundancy is nothing to look down on
114
3/24/2013
1
BASICS OF DESIGN OF PILED
FOUNDATIONS
B t H F ll iBengt H. Fellenius
The Static Loading TestPerformance, Instrumentation, Interpretation
Bolivia, April 25, 2013
33
Candidates for Darwin Award, First Class
44
! ! !Testing piles is a risky business.
55
! ! ! 2 SPACER
1. SWIVEL PLATE
What do you think could happen to the stack of four pieces on the pile head when
66
4. JACK
3. LOAD CELL
2. SPACER the load is applied? And, therefore, to the three oblivious persons next to the pile?
3/24/2013
2
77This is how experience taught the three, and others, to arrange the units on the pile head 88
99 1010
1111 12
3/24/2013
3
1313 14
Fellenius 1984
250
300
350
d (K
N)
Head-down O-cell Pile
August 2006
The error can be small or it can be large. Here are resultsfrom two tests at the same site using the same equipmenttesting two adjacent piles, one after the other.
1,500
2,000
) 15% Error
Shinho-Pile August 2006
15
0
50
100
150
200
0 2,000 4,000 6,000 8,000 10,000Loadcell (KN)
Erro
r in
Jack
Loa
d
2.5% Error
0
500
1,000
0 5,000 10,000 15,000Loadcell load (KN)
Erro
r (K
N)
2.5% Error
Note, the test on the pile called "O-cell pile" is a head-down test after a preceding O-cell test.
A routine static loading test provides the load-movement of the pile head...
and the pile capacity?
16
The Offset Limit MethodDavisson (1972)
LLEAQ Δ=
Q
17
OFFSET (inches) = 0.15 + b/120
OFFSET (SI-units—mm) = 4 + b/120
b = pile diameter (inch or mm)
LΔ
The Decourt ExtrapolationDecourt (1999)
1,000,000
1,500,000
2,000,000
-- Q
/s (
inch
/kip
s)
1
2
CC
Qu =C1 = Slope
C2 = Y-intercept
δQ
18
0 100 200 300 400 5000
500,000
, ,
LOAD (kips)
LOAD
/MVM
NT
-
Ult.Res = 474 k ips
Linear Regression Line
Q
3/24/2013
4
Other methods are:
The Load at Maximum Curvature
Mazurkiewicz Extrapolation
Chin-Kondner Extrapolation
19
DeBeer double-log intersection
Fuller-Hoy Curve Slope
The Creep Method
Yield limit in a cyclic test
For details, see Fellenius (1975, 1980)
DECOURT 235
20
1,500
2,000
2,500
N)
Definition of capacity (ultimate resistance) is only needed when the actual value is not obvious from the load-
movement curve
21
0
500
1,000
0 5 10 15 20 25 30 35 40
MOVEMENT (mm)
LOAD
(KN
Offset-LimitLine
The capacity is not a constant, but changes with time
1,500
2,000
2,500
3,000A
CIT
Y (
KN
)8 years
BOR
16 h BOR
48 days Static Test
CASE 14 years20 m
22
0
500
1,000
0.01 0.10 1.00 10.00 100.00 1,000.00 10,000.00
DAYS AFTER EOID
CA
PA
1 h BOREOIDs CASE 2
16 m
Pile Toe Movement
3,000
4,000
AD
(K
N) HEAD
HEAD LOAD vs. TOE MOVEMENT 65 ft long, 14 inch pipe pile.
With a telltale to the toe arranged to
23
0 5 10 15 20 25 300
1,000
2,000
MOVEMENT (mm)
LOA
D A
T P
ILE
HE
A
determine pileshortening. Don’t arrange it to measure toe movement directly.
Analysis of toe resistance
An adjacent pull test on a similar pile established that the pile shaft resistance (2,000 KN) was approximately fully
3,000
4,000
D (
KN
)
HEAD
HEAD LOAD vs. TOE MOVEMENT
PILE SHORTENING
24
was approximately fully mobilized just short of a 5-mm upward movement at the pile toe. Therefore the load applied in the push test beyond a toe movement of 5 mm goes to toe resistance, only.0 5 10 15 20 25 30
0
1,000
2,000
MOVEMENT (mm)
LOA
D A
T P
ILE
HE
AD
ESTIMATED T OE LOAD vs.
TOE MOVEMENT(Based on the assumption thatshaf t resistance is 2,000 KN)
+10 %
-10 %
3/24/2013
5
20 inch square diameter, prestressed concrete pile driven to 58 ft embedment, through about 45 ft of soft silt and clay, 5
ft of sand, and to bearing 6 ft into hard clay
PUSH and
PULLTo separate
f
Unloading-reloading once or a couple of times “on the way up”
400
500
600
)
Push test
Offset LimitTOEHEAD
25Data from AATech Scientific Inc.
shaft and toe resistances. The pile is
equipped with a toe telltale.
y pserves no purpose and may result in distorted analysis results
0
100
200
300
0.0 0.2 0.4 0.6 0.8 1.0 1.2
MOVEMENT (in)
LOA
D (
kips
Pull test
Combining the push and pull test results with the telltale measurements to determine the load-movement for the pile toe
400
500
600
ps)
PUSH TEST
TOE
"Toe Telltale "
26Data from AATech Scientific Inc.
0
100
200
300
0.0 0.2 0.4 0.6 0.8 1.0 1.2
MOVEMENT (in)
LOA
D (
ki
PULL TESTSHAFT
From pull test with the head movement adjusted to the toe movement
Instrumentation
a d
27
and
Interpretation
T e l l t a l e s• A telltale measures shortening of a pile and must never be arranged to
measure movement.• Let toe movement be the pile head movement minus the pile shortening.• For a single telltale, the shortening divided by the distance between
the pile head and the telltale toe is the average strain over that length.• For two telltales, the distance to use is that between the telltale tips.
28
, p• The strain times the cross section area of the pile times the pile material
E-modulus is the average load in the pile.
• To plot a load distribution, where should the load value be plotted? Midway of the length or above or below?
Load distribution for constant unit shaft resistance
00 100 LOAD, Q
A1
Average Load 0
PILE HEAD
29
50
100h A2
MidheightLoad
Distribution Q = az
DEPTH, zPILE TOE
ars =
21 AA =
Linearly increasing unit shaft resistanceand its load distribution
00 Unit Shaft
Resistance
az 31
3axA =
)23( 323 xhxha +
00 Average
LoadLOAD, Q0
A1x
30
h DEPTH, z
21 AA =
hhX 58.03==“X” is where the average load should be plotted
6)23(2 xhxhaA +−
=
A2
LoadDistribution
Q = az2/2 h
3/24/2013
6
• Today, telltales are not used for determining strain (load) in a pile because using strain gages is a more assured, more accurate, and cheaper means of instrumentation.
• However, it is good policy to include a toe-telltale to measure toe movement. If arranged to measure shortening of the pile, it can also be used as an approximate back-up for the average load in the pile.
Th f ib ti i t i ( ti l t i l
31
• The use of vibrating-wire strain gages (sometimes, electrical resistance gages) is a well-established, accurate, and reliable means for determining loads imposed in the test pile.
• It is very unwise to cut corners by field-attaching single strain gages to the re-bar cage. Always install factory assembled “sister bar” gages.
32
Rebar Strain Meter — “Sister Bar”
Instrument Cable
Three bars?!
Reinforcing Rebaror Strand
Instrument Cables
33
Reinforcing Rebar
Rebar Strain Meter
Wire Tie
or Strand
Tied to Reinforcing Rebar
Hayes 2002
Wire Tie
Tied to Reinforcing Rings
(2 places)
Rebar Strain Meter(3 places, 120° apart)
8
10
12
14
16
18
20
LOA
D (
MN
)Load-strain of individual gages and of averages
4
6
8
10
12
14
16
18
LOA
D (
MN
)
LEVEL 1 D CA B
A&C B&D
34
0
2
4
6
0 50 100 150 200STRAIN (µε)
Level 1A+1C
Level 1B+1D
Level 1 avg 0
2
4
0 100 200 300 400 500 600 700
STRAIN (µε)
The curves are well together and no bending is discernable
Both pair of curves indicate bending; averages are very close; essentially the same for the two pairs
If one gage “dies”, the data of surviving single gage should be discarded. It must not be combined with the data of another intact pair.Data from two surviving single gages must not be combined.
12
14
16
18
20
N)
A&C+D
Means: A&C, B&D, AND A&B&C&D
A&C+BB+CA+D
LEVEL 1
35
0
2
4
6
8
10
0 100 200 300 400 500 600 700
STRAIN (µε)
LOA
D (
MN
Error when including the single third gage, when either Gage B or Gage D data are discarded due to damage.
Glostrext Retrievable Extensometer (Geokon 1300 & A9)
36
Lee Sieng Kai, 2010. Recent development in pile instrumentationtechnology for driven, jacked-in and bored cast-in-place piles.Lecture notes. [www.glostrext.com.my]
Anchor arrangement display Anchors installed
3/24/2013
7
Gage for measuring displacement, i.e., distance change between upper and lower extensometers. Accuracy is about 0.02mm/5m
37
corresponding to about 5 µε.
That the shape of a pile sometimes can be quite different from the straight-sided cylinder can be noticed in a retaining wall built as a pile-in-pile wall
38
0
5
10
0.00 0.50 1.00 1.50 2.00 2.50
DIAMETER RATIO AND AREA RATIO
(m)
Nominal Ratio
Determining actual shape of the bored hole before concreting
39
15
20
25
DEP
TH
Gage Depth
DiameterRatio
Area Ratio
O-cell
3/24/2013
1
We have got the strain.How do we get the load?
• Load is stress times area
1
• Stress is Modulus (E) times strain
• The modulus is the key
εσ E=
For a concrete pile or a concrete-filled bored pile, the modulus to use is the combined modulus of concrete,
reinforcement, and steel casing
cs
ccsscomb AA
AEAEE
++
=
2
Ecomb = combined modulus Es = modulus for steelAs = area of steelEc = modulus for concreteAc = area of concrete
• The modulus of steel is 200 GPa (207 GPa for those weak at heart)
• The modulus of concrete is. . . . ?
Hard to answer. There is a sort of relation to the cylinder strength and the modulus usually appears as a value around 30 GPa, or perhaps 20 GPa or so, perhaps more.
This is not good enough answer but being vague is not necessary.
The modulus can be determined from the strain measurements.
3
Calculate first the change of strain for a change of load and plot the values against the strain.
Values are knownεσΔΔ
=tE
50
60
70
80
90
100
DU
LUS
(G
Pa)
Level 1
Level 2
Level 3
Level 4
Level 5
Example of “Tangent Modulus Plot”
4
0 200 400 600 8000
10
20
30
40
50
MICROSTRAIN
TAN
GE
NT
MO
Best Fit Line
baddEt +=⎟
⎠⎞
⎜⎝⎛= ε
εσ
εεσ ba+⎟
⎞⎜⎛= 2
Which can be integrated to:
B t stress is also a f nction of
In the stress range of the static loading test, modulus of concrete is not constant, but a more or less linear relation to the strain
5
εεσ b+⎟⎠
⎜⎝
=2
εσ sE=
But stress is also a function of secant modulus and strain:
Combined, we get a useful relation:
baEs += ε5.0 and Q = A Es ε
50
60
70
80
90
100
DU
LUS
(G
Pa)
Level 1
Level 2
Level 3
Level 4
Level 5
Example of “Tangent Modulus Plot”
6
0 200 400 600 8000
10
20
30
40
50
MICROSTRAIN
TAN
GE
NT
MO
Best Fit Line
Intercept is
”b”
Slope is “a”
3/24/2013
2
Note, just because a strain-gage has registered some strain values during a test does not guarantee that the data are useful. Strains unrelated to force can develop due to variations in the pile material and temperature and amount to as much as about 50±microstrain. Therefore, the test must be designed to achieve strains due to imposed force of ideally about 500 microstrain and
7
beyond. If the imposed strains are smaller, the relative errors and imprecision will be large, and interpretation of the test data becomes uncertain, causing the investment in instrumentation to be less than meaningful. The test should engage the pile material up to at least half the strength. Preferably, aim for reaching close to the strength.
Unlike steel, the modulus of concrete varies and depends on curing, proportioning,mineral, etc. and it is strain dependent. However, the cross sectional area of steel in aninstrumented steel pile is sometimes not that well known.
y = -0.0013x + 46.79145
50
55
60
STIF
FNES
S, E
A (
GN
)
45
50
55
60
STIF
FNES
S, E
A (
GN
)
EAsecant (GN) = 46.5 from tangent stiffnessEAsecant (GN) = 46.8 - 0.001µε from secant stiffness
8
30
35
40
0 100 200 300 400 500 600
STRAIN, με
SEC
AN
T S
30
35
40
0 100 200 300 400 500 600
STRAIN, με
TAN
GEN
T
y = 0.000x + 46.451
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = Q/ε
(Data from Bradshaw et al. 2012)
Pile stiffness for a 1.83 m diameter steel pile: open-toe pipe pile. Strain-gage pair placed 1.8 m below the pile head.
Field Testing and Foundation Report, Interstate H-1, Keehi Interchange, Hawaii, Project I-H1-1(85), PBHA 1979.
4
5
Q/∆ε
(G
N)
TANGENT STIFFNESS, ∆Q/∆ε
4
5
ε (
GN
)
SECANT STIFFNESS, Q/ε
Strain-gage instrumented, 16.5-inch octagonal prestressedconcrete pile driven to 60 m depth through coral clay andsand. Modulus relations as obtained from uppermost gage(1.5 m below head, i.e., 3.6b).
9Data from PBHA 1979
y = -0.0014x + 4.082
0
1
2
3
0 500 1,000 1,500 2,000
STRAIN (µε)
TAN
GEN
T S
TIFF
NE
SS, ∆Q
y = -0.0007x + 4.0553
0
1
2
3
0 500 1,000 1,500 2,000
STRAIN (µε)
SEC
AN
T S
TIFF
NES
S, Q
/ε
Secant DataSecant from Tangent DataTrend Line
10
15
TIFF
NES
S, E
A (
GN
)
10
15
STIF
FNES
S, E
A (
GN
)
y = -0.003x + 7.41
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = Q/ε
For the "calibrating" uppermost gage level, the secantmethod appears to be the better one to use, right?
10
Pile stiffness for a 600 mm diameter concreted pipepile. The gage level was 1.6 m (3.2b) below pile head
Data from Fellenius et al. 2003
0
5
0 50 100 150 200 250 300
STRAIN, µε
SEC
AN
T S
T
0
5
0 50 100 150 200 250 300
STRAIN, µε
TAN
GEN
T S
y = -0.004x + 7.21
EAsecant (GN) = 7.2 - 0.002µε from tangent stiffnessEAsecant (GN) = 7.4 - 0.003µε from secant stiffness
y = -0.0053x + 11.231
20
30
40
50
NT
STI
FFN
ESS,
EA
(G
N)
20
30
40
50
ENT
STI
FFN
ESS
, EA
(G
N)
EAsecant (GN) = 10.0 - 0.003µε from tangent stiffnessEAsecant (GN) = 11.2 - 0.005µε from secant stiffness
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = ∆Q/∆ε
Or this case? Here, that initial "hyperbolic" trend canbe removed by adding a mere 20 µε to the strain data,"correcting the zero" reading, it seems.
11
0
10
0 100 200 300 400 500
STRAIN (µε)
SEC
AN
y = -0.0055x + 9.9950
10
0 100 200 300 400 500
STRAIN (µε)
TAN
GE
Secant stiffness after adding 20µε to each strain value
Secant stiffness from tangent stiffness
Pile stiffness for a 600-mm diameter prestressed pile.The gage level was 1.5 m (2.5b) below pile the head.
Data from CH2M Hill 1995
Or the adding of a mere8 µε for this case?
40
50
S, E
A (
GN
)
40
50
SS, E
A (
GN
)
Secant for Virgin Loading
Trend Line from Tangent Stiffness
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = ∆Q/∆ε
0
2,000
4,000
6,000
8,000
10,000
12,000
14,000
0 1 2 3 4 5 6MOVEMENT (mm)
LOA
D (
KN
)
12
y = -0.008x + 30.295
10
20
30
0 100 200 300 400 500STRAIN (με)
SEC
AN
T S
TIFF
NES
y = -0.0115x + 29.234
10
20
30
0 100 200 300 400 500
STRAIN (με)
TAN
GE
NT
STI
FFN
ES
EAsecant (GN) = 29.2 - 0.006µε from tangent stiffness
EAsecant (GN) = 30.2 - 0.008µε from secant stiffness
gRelation
Secant Stiffness after adding 8µε to each strain value
Pile stiffness for a 900-mm bored pile constructed in Indonesia. The gage level was 2.0 m (2.2b) below pile the head.
3/24/2013
3
After completion of the test, the pilewas reloaded. Below, the 2nd cycle datahave been added to the first cycle plot.
0
2,000
4,000
6,000
8,000
10,000
12,000
14,000
0 1 2 3 4 5 6MOVEMENT (mm)
LOA
D (
KN
)
40
50
SS, E
A (
GN
)
40
50
ESS,
EA
(G
N)
Secant for Reloading(1st cycle strains removed)
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = ∆Q/∆ε
13Data from Geo Optima Pt. 2011
10
20
30
0 100 200 300 400 500STRAIN (με)
SEC
AN
T S
TIFF
NE
S
10
20
30
0 100 200 300 400 500STRAIN (με)
TAN
GE
NT
STI
FFN
E
Secant for Reloading
Tangent for Reloading
Illustration of the adverse effect of unloading/reloading.
What really do we learn fromunloading/reloading and what
14
unloading/reloading and whatdoes unloading/reloading do tothe gage records?
The Testing Schedule
150
200
250
300
ERC
EN
T"
A much superior test schedule. It presents a large number of values (≈20 increments), has no destructive unloading/reload cycles, and has constant load-hold duration. Such tests can be used in analysis for load distribution and settlement and will provide value to a project, as opposed to the long-duration, unloading/reloading, variable load-hold duration, which is a next to useless test.
Plan for 200 %, but make use of the opportunity to go higher if this becomes feasible
0
50
100
0 6 12 18 24 30 36 42 48 54 60 66 72
TIME (hours)
"PE
The schedule in blue is typical for many standards. However, it is costly, time-consuming,and, most important, it is diminishes or eliminates reliable analysis of the test results.
XXXXX
What about keeping the load on the pile until "zero" movement?
(Long-duration load-holding)
30
40
50
60
D (
MN
)
Pile TP-1 Pile TP 1
30
40
50
60
(MN
)
16
0
10
20
30
0 5 10 15 20 25 30
DAYS
LOA
D
Lakhta Center, St Petersburg, Russia2.0 m diameter, 84 m long, bored pile
0
10
20
30
0 20 40 60 80 100
MOVEMENT (mm)
LOA
D
Pile TP 1
0
10
20
30
40
50
0 50 100 150 200 250 300
TIME (hours)
MO
VEM
ENT
(mm
)
2L-13
2L-112L-10
2L-12
2L-14
Pile TP 1
0
5
10
15
20
25
30
35
0 5,000 10,000 15,000 20,000
TIME (minutes)
LOA
D B
ETEE
N G
Ls (
MN
)
GL6 to GL7 GL5 to GL6 GL5 to GL7
2L-142L-13
2L-122L-11
2L-142L-132L-12
2L-10
2L-122L-112L-10
2L-14
2L-10
2L-11
2L-13
17
Lakhta Center, St Petersburg, Russia2.0 m diameter, 84 m long, bored pile
The long-duration load-holding and variations of load increments have obviously hadconsiderable costs consequence for the project. Yet, nothing was "bought" by thosecosts. On the contrary, the uneven load-holding durations and the differing load incrementmagnitudes messed up the data and reduced the usefulness of the detailed analysis ofthe test records.
( ) TIME (minutes)
The occasional unloading/reloading and varying load-holding durationsprovide no information of any value for assessing pile response to load. Itis nothing but a vestigial practice, i.e., remnant of old, now obsolete, partof the practice, much like our tailbone.
Figuratively speaking, it is strange that so many still appear to believe thatthey have a tail at their rear end to wag despite the fact that the vestigial
On unloading/reloading
18
they have a tail at their rear end to wag, despite the fact that the vestigialtailbone is not connected to the head. Indeed, to schedule a test toinclude unloading/reloading and varying load-holds duration is nothing butakin to a insisting on that there is a tail to wag, disregarding all evidenceto the contrary. Those who argue for the wag seem to be too busycontemplating their navel to realize that nothing useful happens.
3/24/2013
4
The strain-gage measurement is supposed to be the change of strain due to the applied load relative the “no-load” situation (i.e., when no external load acts t th l ti )
Determining load from strain-gage measurements in the pile
1919
But, is the “no-load” situation really the reading taken at the beginning of the test? What is the true “zero-reading” to use?
at the gage location).
• We often assume – somewhat optimistically or naively – that the reading before the start of the test represents the “no-load” condition.
• However, at the time of the start of the loading test, loads do exist in the pile and they are often large.
• For a grouted pipe pile or a concrete cylinder pile,
2020
these loads are to a part the effect of the temperature generated during the curing of the grout.
• Then, the re-consolidation (set-up) of the soil after the driving or construction of the pile will impose additional loads on the pile.
Concrete hydration temperature measured in a grouted concrete cylinder pile
40
50
60
70
RA
TUR
E (°
C)
Temperature at various depths in the grout of a 0.4 m center hole in a 56 m long, 0.6 m diameter, cylinder pile.
2121
0
10
20
30
0 24 48 72 96 120 144 168 192 216 240
HOURS AFTER GROUTING
TEM
PER
Pusan Case
20
30
40
50
60
70
TEM
PER
ATU
RE
(°C
)
Start of Static Loading Test
Temperatures measured in a 74.5 m long, 2.6 m diameter bored pile
20
30
40
50
60
70
TEM
PER
ATU
RE
(°C
)
Temperatures measured in a 74.5 m long, 2.6 m diameter bored pile
2222
100 5 10 15 20 25 30 35
DAY AFTER GROUTING
100 1 2 3 4 5
DAY AFTER GROUTING
Temperature records during curing of grout in the Golden Ears Bridge test pile, Vancouver, BC. Data courtesy of Trow Engineering Inc. and Amec Inc.
-200
-100
0
100
NG
E O
F ST
RA
IN (
µε) Change of strain measured in a 74.5 m
long, 2.6 m diameter bored pile
2323
-400
-300
0 5 10 15 20 25 30 35
DAY AFTER GROUTING
CH
AN
Rebar ShorteningSlight Recovery of Shortening
Change of strain during the hydration of the grout in the Golden Ears Bridge test pile
The strain gages themselves are not are temperature sensitive, but the records may be!
The vibrating wire and the rebar have almost the same temperaturecoefficient. However, the coefficients of steel and concrete are slightlydifferent. This will influence the strains during the cooling of the grout.
2424
g g gMore important, the rise of temperature in the grout could affect the zeroreading of the wire and its strain calibration. It is necessary to “heat-cycle” (anneal) the gage before calibration. (A routine measure ofGeokon, US manufacturer of vibrating wire gages).
3/24/2013
5
0
5
10
15
20
25
-300 -200 -100 0 100 200 300 400
STRAIN (µε)
m)
Zero Reading. Does it mean
l d?
Immediately before the test, all gages must be checked and "Zero Readings" must be taken.
2525
25
30
35
40
45
50
55
60
DEP
TH (
m
Shinho Pile
zero load?
Answer to the question in the graph:
No, there's always residual load in a test pile.
0
5
10
15
20
25
-300 -200 -100 0 100 200 300 400
STRAIN (µε)
H (
m)
9d
15d
23d
30d
39d
49d
Strain measured during the 218-day wait-period between driving (grouting) and testing.
Do these strains really represent
2626
30
35
40
45
50
55
60
DEP
TH
59d
82d
99d
122d
218d Day ofTest
At an E- modulus of 30 GPa, this strain change corresponds to a load change of 3,200 KN
really represent load in the pile, as present before the start of the static loading test?
• Readings should be taken immediately before (and after) every event of the piling work and not just during the actual loading test
• The No-Load Readings will tell what happened to the gage before the start of the test and will be helpful in assessing the possibility of a shift in the reading value representing the no load condition
2727
reading value representing the no-load condition
• If the importance of the No-Load Readings is recognized, and if those readings are reviewed and evaluated, then, we are ready to consider the actual readings during the test
Of course,
we must consider also other aspects:
2828
Also the best field work can get messed up if the analysis and
conclusion effort loses sight of the history of the data
1,000
1,500
2,000
D (
KN) STATIC TEST
DYNAMIC TEST
1,000
1,500
2,000
D (
KN)
DYNAMIC TEST
1,000
1,500
2,000
D (
KN)
DYNAMIC TEST in a series of blows
Repeated STATIC TEST
2929The dynamic test (CAPWAP) was performed after the static test.
The redriving (ten blows) forced the pile down additionally about 45 mm.
0
500
0 100 200 300
MOVEMENT (mm)
LOA
0
500
,
0 100 200 300
MOVEMENT (mm)
LOAD
0
500
,
0 100 200 300
MOVEMENT (mm)
LOAD
Result on a test on a 2.5 m diameter, 80 m long bored pile
Does unloading/reloading add anything of value to a test?
20
25
30
)
Acceptance Criterion
20
25
30
)
Acceptance CriterionRepeat test
30
0
5
10
15
0 25 50 75 100 125 150 175 200
MOVEMENT (mm)
LOA
D (
MN
)
0
5
10
15
0 25 50 75 100 125 150 175 200
MOVEMENT (mm)
LOA
D (
MN
)
3/24/2013
6
Plotting the repeat test in proper sequence
15
20
25
30
AD
(M
N)
Acceptance Criterion Repeat test plotted in sequence of testing
31
0
5
10
0 25 50 75 100 125 150 175 200
MOVEMENT (mm)
LOA 500
1,000
1,500
2,000
LOA
D (
KN
)
3232
The above series of unloading/reloading has added nothing but they have cost the client a lot of money.
0
500
0 1 2 3 4
MOVEMENT (mm)
Good measurements do not guarantee good conclusions!
A good deal of good thinking is necessary, too
Results of static loading tests on a 40 m long, jacked-in,
instrumented steel pile in a saprolite soil
0
5
10
15
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
0
5
10
0 50 100 150 200 250
UNIT SHAFT SHEAR (KPa)
3333
A good deal of good thinking is necessary, too15
20
25
30
35
40
45
DEP
TH (
m) 15
20
25
30
35
40
DEP
TH (
m) ?
0
5
10
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
ß = 0.3
A more thoughtful analysis of the data
0
5
10
0 50 100 150 200 250
UNIT SHAFT SHEAR (KPa)
ß = 0.3
3434
15
20
25
30
35
40
45
DEP
TH (
m)
ß = 0.4
15
20
25
30
35
40
45
DEP
TH (
m)
ß = 0.4
And a second pile:
0
5
10
15
0 50 100 150 200 250 300 350 400 450
UNIT SHAFT SHEAR (KPa)
ß = 0.3
0
5
10
15
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
ß = 0.3
3535
15
20
25
30
35
40
45
DEP
TH (
m)
ß = 0.5
15
20
25
30
35
40
45
DEP
TH (
m)
ß = 0.5
400
600
800
ER P
ILE
(KN
) Single PileAverage of 4 Piles
Average of 9 Piles
Group Effect and Interaction
36
0
200
0 2 4 6 8 10
PILE HEAD MOVEMENT (mm)
LOA
D P
E
O’Neill et al. (1982)
20 m
3/24/2013
7
800
mm
800 mm
60 mm
#1
#2 #3
#4 #5
Loading tests on a single pile and a group of 5 piles in loose, clean sand at Gråby, Sweden.
37Data from Phung, D.L (1993)
2.30
m
Pile numbers indicate order of driving. Pile #1 wasdriven first and tested as a single pile. Piles #2 - #5were then driven and tested in sequence as singlepiles. Finally, the full five-pile group was tested withpile cap not in contact with the ground.
8
10
12
14
16
AD
/PIL
E (K
N) Average
#2
#5
#4
#3
#1
#1 as single
38Data from Phung, D.L (1993)
0
2
4
6
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT OF PILE HEAD (mm)
LOA
#12,3 m
340 mm
#2
#3 #4
#5
680 mmc/c = 5.7 b
sq60 mm
Pile #1: Effect of compaction caused by driving Piles #2 — #5 and re-testing
8
10
12
14
16
PILE
(K
N) Pile #1 reloaded as
part of the group after Piles #2 - #5 were driven
#1
#1
Data from Phung, D.L (1993)39
0
2
4
6
0 10 20 30 40 50 60 70 80
MOVEMENT OF PILE HEAD (mm)
LOA
D/P
#1
#2
#3 #4
#5
8
10
12
14
PILE
(K
N)
Head average
Head single
Toe single
Toe average
The main change occurred along the shaft
The toe resistance showed little change, only
40Data from Phung, D.L (1993)
0
2
4
6
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT (mm)
LOA
D/P
Shaft single
Shaft average
The Bi-Directional Static Loading Test
The Osterberg
“O-cell Test”
4141
Jorj Osterberg
2001
Schematics of the Osterberg O-Cell Test(Meyer and Schade 1995)
Telltales
andPile Head
4242
Upward Load
Downward Load
THE O-CELL
Grout Pipe
3/24/2013
8
4343
Three O-Cells inside the reinforcing cage(My Thuan Bridge, Vietnam) 4444
The O-cell can also be installed in a driven pile (after the driving). Here in a 600 mm cylinder pile with a 400 mm central void.
4545 46O-cell in a pipe pile inserted in a augercast pile after grouting.
4747Inchon, Korea
-60-50-40-30
0 5,000 10,000 15,000 20,00
Load (KN)
Results of an O-cell test on a 2.8 m by 0.8 m, 40 m deep barrette in Manila, Philippines
4848
30-20-10
010203040506070
Mov
emen
t (m
m)
Upward
Downward
Upward
Approximate extrapolation of toe movement back to starting conditions
3/24/2013
9
O-Cell test on a 1,250 mm
diameter, 40 m long, bored pile at US82 Bridge 20
40
60
80
100
120
MEN
T (m
m)
UPPER PLATE UPWARD MVMNT
Shaft
4949
across Mississippi
Riverinstalled into dense sand
-80
-60
-40
-20
0
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
MO
VEM
LOWER PLATE DOWNWARD MVMNT
Weightof
Shaft
Residual Load
Toe
From the O-Cell results, one can produce the equivalenthead-down load-movement curve that one would haveobtained in a routine “Head-Down Test”
“Head –down”
cnt.
5050
cnt.
5151
O-Cell Results Shown Two Ways
40
60
80
100
120
(mm
)
6,000
7,000
8,000
9,000
N)
Shaft Movement
Toe Movement
cnt.
5252
-80
-60
-40
-20
0
20
40
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
MO
VEM
ENT
0
1,000
2,000
3,000
4,000
5,000
0 20 40 60 80 100 120
MOVEMENT (mm)
LOA
D (
KN
Weight of Shaft
Residual Load
38 m
Equivalent Head-down Curve
10
15
20
25
OA
D (
MN
)
Acceptance ReferenceLoad 25 mm Movement
O-cell Load-Movement Curves
-40
-20
0
20
MEN
T (m
m)
UPWARD
DOWNWARD
Evaluation of Proof Test Results from an O-cell Test
10
15
20
25
OA
D (
MN
)
Offset Limit
Acceptance ReferenceLoad 25 mm Movement
53
O-cell0
5
10
0 20 40 60 80
MOVEMENT (mm)
LO
-100
-80
-60
0 5 10 15LOAD (MN)
MO
VEM
Fellenius and Tan, 2010
0
5
10
0 20 40 60 80
MOVEMENT (mm)
LO
Kahuku Bridge across Kamehameha Highway, Hawaii
Test on a 16-inch (600 mm), 75 ft (23 m) long, bored pile in hard clay and
weathered rockDesired allowable load = 340 kips
7
O cell is
5454
-1
0
1
2
3
4
5
6
0 50 100 150 200 250 300 350 400
O-cell Load (kips)
MO
VEM
ENT
(inc
hes)
UPWARD
DOWNWARD
O-cell is placed at
46 ft (14 m) depth
3/24/2013
10
0
10
20
0 100 200 300 400 500 600 700 800
LOAD (kips)
Silty stiff clay
The load distribution after tangent-modulus evaluation of pile stiffness
The equivalent "head-down load distribution"and speculative fully mobilized resistance
0
10
20
0 100 200 300 400 500 600 700 800
LOAD (kips)
SPECULATIVE, FULLY MOBILIZEDDISTRIBUTION
1,100
cnt.
55
30
40
50
60
70
80
DEP
TH (
ft)
DATA FROM O-CELL TEST
O-cell
Shaft and toe resistances are not fully mobilized below the O-cell
Silty hard clay
Weathered rock
Buoyant Weight
Project rules: Qallow. = RULT (=600)/2.0 for test results ===> = 300 kips
Qallow. = RULT(=1,100)/3.0 for calculations ===> = 365 kips
30
40
50
60
70
80
DEP
TH (
ft)
CONVERTED TO HEAD-DOWN "TEST"
APPROXIMATED
O-cell
Orchard Center SingaporeFellenius and Tan (2010)
Combining O-cell and Head-down Tests
56Photo of scale model Photo of building
Head-down Test
Bored Pile, Singapore, 2007
#13
#12
#11
ft)
1,000 mm (39 i )
Stiff
mar
ine
clay
1,030 mm
cnt.
5757
#1
#8
#7
#6
#5
#4
#3
#2
#9
37.5
m (1
23 f
Sapr
olite
Sand
, Silt
, and
cla
y as
mat
rix in
w
eath
ered
gra
nitic
bed
rock
O-cell
Head-down Tangent Modulus
30
40
50∆σ
/∆ε
(GPa
)
SG-13
SG-12
SG-11
SG-9
SG-8
UTP-3, Stage 1b
cnt.
5858
10
20
0 200 400 600 800 1,000 1,200
STRAIN (µε)
∆
SG-7
SG-6
0
5
10
15
20
0 5 10 15 20 25
LOAD (MN)
H (
m)
Now, having determined the
relation between strain and
secant modulus (directly or from
the tangent-modulus), we are
ready to convert measured
strain to ”measured” load in the
Marine clay
cnt.
5959
20
25
30
35
40
DEP
T
UTP-3
Shaft Resistance is not fully mobilized below 20 m depth.
pile: The load distribution. Saprolitic soil
Note
Stage 1 Head-down Test (Stages 1a and 1b)
Stage 2 O-cell test. O-cell is left open
The tests were repeated: Stage 3 head-down and Stage 4: O-cell
5
10
15
20
25-60-50-40-30-20-100
MOVEMENT (mm)
LOA
D (
MN
)
Stage 1
Acceptance Criterion
0 2 4 6 8 10 12 14 16LOAD (MN)
cnt.
60
0
UTP-3 Head-down
-60
-40
-20
0
200 2 4 6 8 10 12 14 16
MO
VEM
ENT
(mm
)
UTP-3 O-cell test
Downward
Stage 4
Stage 2
Max toe movement in Stage 1
Upward
3/24/2013
11
Head-down Test
LOAD DISTRIBUTIONS STAGES 3 AND 4O-cell Test
0
5
10
0 5 10 15 20 25
LOAD (MN)
Max LoadSt 1b
Before Start
After Unload
0
5
10
-5 0 5 10 15 20
LOAD (MN)
cnt.
61
15
20
25
30
35
40
DEP
TH (
m)
UTP-3, Stage 3
Unload
O-cell
C
15
20
25
30
35
40
DEP
TH (
m)
UTP-3, Stage 4
O-cell
Before Start
D
0
5
10
15
0 10 20 30 40 50 60
LOAD (MN)
"Flipped" UTP-3O-cell Curve
Maximum ResistanceHead-down Test (1b)
Head-down curves movedto connect ith
0
5
10
15
0.0 0.5 1.0 1.5 2.0
BETA COEFFICIENT (--)
)
ß
rs, avg. = 100 KPa
Combining the results of the head-down and O-cell tests
cnt.
6262
20
25
30
35
40
DEP
TH (
m) to connect with
load in O-cell test
Effective stress analysis fitted to tests
15
20
25
30
35
40
DEP
TH (
m)
O-cell
Head-downPILE UTP-4
A B
rs, avg. = 450 KPa
rs, avg. = 550 KPa
Head-downPILE UTP-3
The effect of residual load on an uplift test ("Head-up")
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
True Resistance
TENSION TESTAND FULL RESIDUAL LOAD
Residual Load
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
Residual Load
True Resistance
TENSION TESTAND PARTIAL RESIDUAL LOAD
63
15
20
25
30
35
DE
PTH
(m
False Resistance
Toe Resistancein an Uplift Test?!
15
20
25
30
35
DEP
TH (
m
False Resistance
Toe Resistancein an Uplift Test?
Combining the results of a head-down test with those of a tension test will help determining the true resistance
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
H (
m)
HEAD-DOWN AND PARTIAL RESIDUAL LOAD
FalseHead-down
True Shaft
False TensionTest
6464
20
25
30
35
DEP
TH
Residual Load
True Resistance
Residual and TrueToe Resistance
Transition Zone
True Shaft Resistance
Not directly useful below this level
Now you know why some claim that resistance in tension is smaller than that in compression
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
(m)
O-CELLAND FULL RESIDUAL LOAD
Residual Load
The effect of residual load on an O-cell test
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
(m)
O-CELLAND PARTIAL RESIDUAL LOAD
True Shaft
65
20
25
30
35
DEP
TH Load
False Resistance
True Resistance
Residual and TrueToe Resistance
EquivalentHead-down
False Resistance
The O-cell
The O-cell load includes the residual load whereas the load evaluated from the strain-gages does not.
20
25
30
35
DEP
TH
Residual Load
False Resistance
True Resistance
True Shaft Resistance
EquivalentHead-down
False Resistance
The O-cell
0
5
10
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
True Shaft Resistance
False Head-down Resistance
Combining an O-cell test with a subsequent head-down test to determine the distribution of residual load
66
15
20
25
30
35
DEP
TH (
m) Resistance
False O-cell Resistance
True Resistance
O-cell Equivalent Head-down False
Resistance
Residual Load
3/24/2013
12
Measured load-movements can besimulated (fitting) to t-z and q-z relations
Pile shaft by t-z relation; Pile toe by q-z relation ("Ratio function")
80
100R = MVMNT^Exp
TOE
SHAFT
exp
2
1
2
1 )(δδ
=RR
Ratio function
cnt.
6767
0 20 40 60 80 1000
20
40
60
Movement (%)
Res
ista
nce
(%)
Exp. = 0.75
Exp. = 0.05
Exp. = 0.50
Exp. = 0.33
Exp. = 0.20
Exp. = 0.10
t-z and q-z (and p-y) functionsHypothetical case of load-movement curves modeling a test on a pile with a capacity of 100 % at a pile head movement of 4 mm
b
uurr ⎜⎜⎝
⎛⎟⎟⎠
⎞=
δδ
21 CCr
+=
δδ
RATIO
HYPERBOLIC
80
100
120
140
AR
(%
of r
ult)
Ratio (b = 0.25)
Hyperbolic (C2 = 0.006)
Exponential (b = 0.75)rU
80 % (strain-hardening)
Zhanga = 0.0083r∞/ru = 50 %
δb
6868
EXPONENTIAL
80 %
Zhang0
20
40
60
0 5 10 15 20 25
RELATIVE MOVEMENT BETWEEN PILE AND SOIL ELEMENT (mm)
SH
AFT
SH
EA
80 % (strain-softening)δU
2)()(
δδδ
bacar
++
=
21 CCr
+=
δδ
uu
ar
bδ
−=21
2bc
rru =∝
)1( δbu err −−=
uu
ar
cδ
−=41
The ratio function applies also a toe response and, some of the functions can also be used as p-y curves which is the term applied to horizontally loaded piles.
Pensacola, Florida410 mm diameter, 22 m long, precast concrete pile driven into silty sand
6969
Pensacola, Florida, USA
1
2
3
4
(mm
)
10
20
30
40
50
60
MO
VEM
ENT
(mm
)
O-cell equipped, 16-inch, 72-ft, prestressed pile driven into sand.
cnt.
7070
-4
-3
-2
-1
0
0 500 1,000 1,500 2,000 2,500
LOAD (KN)M
OVE
MEN
T (
-10
0
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
O-cell
After the push test, the pile toe is located higher up than when the test started!
Test at Bangkok Airport
7171
Stage 1Lower Cell activatedUpper cell closed
Stage 2Lower Cell openUpper Cell activated
Stage 2
cnt.
7272
Stage 2Lower Cell closedUpper Cell activated
Data fromFox, I., Du, M. and Buttling,S. (2004)Buttling, S. (2006)
3/24/2013
13
Downward movements during test phases 1, 2, and 3
0
25
50
75
100
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
EMEN
T (m
m) P1 P2
P3
1 2 3
cnt.
7373
Concern was expressed (Buttling 2006) that the toe resistance (Phase 1) was ≈3,000 KN and the shaft resistance for the lower segment was ≈5,000 KN (Phase 2), while in Phase 3 the combined shaft and toe resistances were only ≈6,000 KN. Should not the Phase 3 resistance be ≈8,000 KN rather than ≈6,000 KN (i.e., the sum of the values ≈5,000 KN and ≈3,000)?
125
150
175
MO
VE
Active CellInactive, Open CellInactive, Closed Cell
0
25
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
P3
1 2 30
25
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
(mm
)
1 2 3
The downward toe movements. However, the movements are best plottedper sequence of testing. Particularly when considering the example toeresistance, one must evaluate the load-movement response in comparingPhase 1 + Phase 2 to Phase 3 (i.e., P2 shaft below cell plus P1 toe).
cnt.
7474
50
75
100
125
150
175
MO
VEM
ENT
(mm
)
Active CellInactive, Open CellInactive, Closed Cell
P1 P2P3
50
75
100
125
150
175DO
WN
WA
RD
MO
VEM
ENT
Active CellInactive, Open CellInactive, Closed Cell
P2
P1 and P2 data combined
P3
O-cell Tests on a 1.4 m diameter
bored pile in North-West Calgary
constructed in silty glacial clay till
7575
glacial clay till
A study of Toe and Shaft Resistance Response to Loading and correlation to CPTU
calculation of capacity
0
5
10
0 10 20 30Cone Stress, qt (MPa)
0
5
10
0 200 400 600 8001,00
0
Sleeve Friction, fs (KPa)
0
5
10
-100 0 100 200 300 400 500
Pore Pressure (KPa)
0
5
10
0 1 2 3 4 5
Friction Ratio, fR (%)PROFILE
The upper 8 m will be removed for basement
uneutral
GW
Cone Penetration Test with Location of Test Pile
cnt.
7676
15
20
25
30
DEP
TH (
m)
10
15
20
25
30
DEP
TH (
m)
10
15
20
25
30D
EPTH
(m
)
10
15
20
25
30
DEP
TH (
m)14 m
net pile length
30
40
50
60
70
80
ENT
(mm
)
Pile Profile and O-cell Location O-cell Load-movement Up and Down
ELEV.(m)
cnt.
7777
-30
-20
-10
0
10
20
0 1,000 2,000 3,000 4,000 5,000 6,000
LOAD (KN)
MO
VEM
E
#5
#4
#3
#2 #1
Load Distribution
0
5
0 5 10 15 20LOAD (MN)
0
5
0 5 10 15 20LOAD (MN)
cnt.
7878
10
15
20
25
DEP
TH (
m)
O-cell
10
15
20
25
DEP
TH (
m)
ß = 0.75
ß = 0.35
3/24/2013
14
0
5
0 5 10 15 20LOAD (MN)
After consideration of potential presence of residual load and applying judgment
Load Distribution
cnt.
7979
10
15
20
25
DEP
TH (
m)
ß = 0.75
ß = 0.35
ß = 0.65
0
5
0 5 10 15 20LOAD (MN)
Schmertmann
LCPC withoutmax. limitsTEST
LCPC withmax. limits
Load Measured Distribution Compared to Distributions Calculated from the CPTU Soundings
cnt.
8080
10
15
20
25
DEP
TH (
m)
E-F
8,000
10,000
12,000
14,000
(KN
) Loadtest's Curvefor the Pile Head
Pile Head
Pile Shaft
The Test Data for Shaft and Toe and Evaluation of Head-down Movement Using t-z and q-z evaluations
cnt.
8181
0
2,000
4,000
6,000
0 20 40 60 80 100 120
MOVEMENT (mm)
LOA
D Pile Shaft
Pile Toe
Test range
Los Angeles Coliseum, 1994Case Record
82
The Northridge earthquake in Los Angeles, California,in January 1994 was a "strong" moment magnitudeof 6.7 with one of the highest ground accelerationever recorded in an urban area in North America.
The piles had been designed using the usual design approachwith adequate factors of safety to guard against the unknowns.Moreover, the acceptable maximum movement was morestringent than usual.
The earthquake caused an estimated $20 billion inproperty damage. Amongst the severely damagedbuildings was the Los Angeles Memorial Coliseum,which repairs and reconstruction cost about $93million. The remediation work included constructionof twenty-eight, 1,300 mm diameter, about 30 m long,bored piles, each with a working load of almost9,000 KN (2,000 kips), founded in a sand and graveldeposit.
It was imperative that all construction work was finished in sixmonths (September 1994, the start of the football season).However, after constructing the first two piles, which took sixweeks, it became obvious that constructing the remainingtwenty-six piles would take much longer than six months.Drilling deeper than 20 m was particularly time-consuming.The design was therefore changed to about 18 m length,combined with equipping every pile with an O cell at the piletoe. — Note, the O-cell was now used as a construction tool.
Los Angeles Coliseum, 1994
-100
-90
-80
-70
-60
-50
-40
-30
-20
-10
0
10
20
30
40
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
MO
VEM
ENT
(mm
)
100
-90
-80
-70
-60
-50
-40
-30
-20
-10
0
10
20
30
40
MO
VEM
ENT
(mm
)
20
30
40
ONE OF THE PILES
cnt.
83
-1000 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
Schmertmann (2009, 2012), Fellenius (2011)
-100
-90
-80
-70
-60
-50
-40
-30
-20
-10
0
10
0 2,000 4,000 6,000 8,000 10,000
LOAD (KN)
MO
VEM
ENT
(mm
)
THE O-CELL AS A CONSTRUCTION TOOL
TESTS ON EVERY PILE
ONE OF THE PILES
THREE OF THE PILES
ALL PILES
2,000
4,000
6,000
8,000
10,000
LOA
D (
KN
)
Downward, First Loading
2,000
4,000
6,000
8,000
10,000
LOA
D (
KN
)
Downward, Second Loading
cnt.
84
FIRST STAGE LOADING; "VIRGIN" TESTSWITH LINE SHOWING AVERAGE SLOPE
RELOADING STAGE. DOWNWARD DATA ONLY
The first stage loading is of interest in the context of general evaluation of test results and applying them to design
00 25 50 75 100 125 150
MOVEMENT (mm)A0
0 25 50 75 100 125 150MOVEMENT (mm)B
1
BASICS OF DESIGN OF PILED
FOUNDATIONS
Bengt H Fellenius
1
Bengt H. Fellenius
Case Histories of Results from Instrumented Piles and Soil
Leading up to the Unified MethodBolivia, April 25, 2013 22
• Design of piles requires understanding of how load is transferred from pile to soil and, less obvious but equally important, from soil to pile.
Th t t t f th t h l d f f
33
• The current state-of-the-art has evolved from a few ground-breaking case histories that appeared in the late 1960s and early 1970s.
Bjerrum et. al., (1965; 1969)
presented case histories of 300-mm diameter steel piles
CASE #1
44
driven through marine clay
deposited directly on bedrock in Oslo Harbor, Norway.
Heröya
55
Profile of test site and piles. Heröya site.
(Bjerrum et al., 1969)
0
5
10
15
20
0 100 200 300 400
EFFECTIVE STRESSAND PORE PRESSURE (KPa)
DEP
TH (
m)
σ'z after full dissipation of excess pore
pressure
σ'zΔu
Marine Clay
FILL
0
5
10
15
20
0 300 600 900 1,200 1,500
PILE LOAD (KN)
DEPT
H (m
)
uncoated
Distribution calculated from
ß=0.3 times σ'z for actual excess u
Measured distribution
0
5
10
15
20
0 5 10 15
PILE SHORTENING (mm)
DEP
TH (
m)
Bitumen coated uncoated
66
Distribution of soil stress, excess pore pressure, pile shortening, and load distributions. Heröya site. (Data from Bjerrum et al., 1969).
25
30
35
Start of Bedrock
Gravel
25
30
35
Bitumen coated
Notice the distinct Force Equilibrium, the Neutral Plane
25
30
35
2
Compilation of Norwegian results
cm cm β
Drag load
77
0
Endo et al. 1969, presented a very
ambitious study in Japan on drag load on four
i t t d t l il
CASE #2
88
instrumented steel piles during a period of three
years. The soils consist of silt and clay on sand. The case history is one of the few that actually also
measured settlement.
99
Profile of test site and piles
Closed-toe, Open-toe, Inclined, and Short Pile
(Endo et al., 1969)
Neutral plane = Force Equilibrium = Settlement Equilibrium
0
5
10
15
20
25
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
H (
m)
SandySilt
Clay
0
5
10
15
20
25
0 50 100 150 200
SETTLEMENT (mm)
Soil
Pile
1010
Load distribution in the three long piles together and settlement of soil and piles measured March 1967 672 days after start. (Data from Endo et al., 1969).
25
30
35
40
45
50
DEP
TH
Closed-toePiles: Inclined and Vertical
Open-toePile
Sand
Silt
( ) ?
25
30
35
40
45
50Closed-toe PilesToe
Penetration
- NEUTRAL PLANE -
Toe forces and toe penetrations extracted from the graphs of Endo et al.
1,500
2,000
D (
KN
)
Approximation
11
0
500
1,000
0 10 20 30
TOE PENETRATION (mm)
TOE
LOA
D
From data in paper
0
5
10
15
20
25
0 2,000 4,000 6,000 8,000
LOAD (KN)
PTH
(m
) Measured load
Calculated Curve
ß = 0.40
ß = 0.35
1212
30
35
40
45
50
DE
P
ß = 0.30
ß = 0.25
Measured load distribution and distribution matched to measured values in effective stress analysis. (Data from Endo et al., 1969).
3
Fellenius (1972) presented a study
of two instrumented,
precast concrete
CASE #3
1313
piles driven through marine clay and
into sandat Bäckebol,
Göteborg, Sweden
1414
M# = Pile Force gages
0
5
10
15
20
25
0 25 50 75 100
PL, LL, and wn (%)
DEP
TH (
m)
LAYER 2
LAYER 3 wn LAY
1515
30
35
40
LAYER 4
LAYER 5
SAN
DC
Pile II segments with Gage M4 at pile toe
1616
First load (360 KN) placed on the piles
1717
Piles with loads applied but before fill was placed over the site
1818
4
800
1000
1200
1400
1600
1800
E A
T G
AG
E (K
N)
First load placed on piles
Second load placed on piles
2 m thick fill placed over site M1 & M5
M2 & M6
M3 & M7
M2 & M6
M3 & M7
1919Measured loads in piles versus time after driving
0
200
400
600
0 500 1,000 1,500 2,000 2,500 3,000 3,500
DAYS AFTER END OF DRIVING
FOR
CE
M4
M4
M1 & M5
0
10
20
30
0 500 1000 1500 2000FORCE AT GAGE (KN)
= “LIVE LOADS”
Placing the fill
2020
40
50
60
2,650
19881923
Distribution of load in Piles I and II
Note, the dragload was eliminated by the “live load” Neutral
Plane
That the toe resistance is small is
due to that the movements are not
large enough to mobilize any larger toe
resistance
Distribution of measured and calculated consolidation settlement due to the fill
The settlement measured at depth
0
5
10
15
0 50 100 150 200 250
STRESS (KPa)
(m)
(σ'Z)f
0
5
10
15
0 100 200 300 400 500
SETTLEMENT (mm)
(m
)
MEASURED
CALCULATED FINAL (after 80
years)
2121
pamounted to only a few millimeters,
but this was enough to fully
mobilize the negative skin
friction
20
25
30
35
40
DEP
TH
(σ'Z)i PRECONSOLIDATION
STRESS, σ'c
20
25
30
35
40
DEP
TH
Force gage locations
FHWA Project, Keehi Interchange, Honolulu,Hawaii 1977
CASE #5
2222
Test piles, embankment fill, and soil profile
OLD FILL
EMBANKMENT
SOFT CLAY
27 m
SLEEVE
PLAN VIEW
PILE #8PILE #6
PILE #7
18 m
6 m
CASE #5
2323
Data from Clemente, 1981
SOFT CLAY w = 70 %
LL = 80PL = 45
SILTY CLAY
MIXED STRATACORAL SANDCORAL CLAY
0
10
20
30
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
TH (
m)
450 mm
10 mmEND OF
SETTLEMENT
0
10
20
30
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
TH (
m)
ß = 0.25
450 mm
10 mmEND OF
SETTLEMENT
Total stress evaluation Effective stress evaluation
24243/16 inch = 4 mm
30
40
50
60
DEP
T
0 mm
END OF COAT
ß = 0.25
#7 #6
#8
30
40
50
60
DEP
T
ß = 0.25
0 mm
35 m uncoated pile ==> 1,770 KN35 m bitumen coated pile ==> 375 KN for less than 1/16-inch (1.5 mm) coat = Reduction to 20 % of uncoated load 3/16 coat + stiff outer skin ==> elimination of drag load
END OF COAT
ß = 0.25
#7 #6
#8
5
2525
Bitumen coating of piles
Laboratory tests on bitumen coats at different
2626
at different rates of shear
A study in Australia of two 760 mm, strain-gage instrumented, open-toe pipe piles driven through a 6 m sand layer over a 15.5 m thick overconsolidated silty clay deposited on silt and sand. Ground surface settlement was induced by placing a 3 m high surcharge over 200m x 100m area around the test piles, causing drag load.
CASE #6
2727Data from Walker et al. (1973
0
5
10
15
0 50 100 150 200 250 300
DAYS AFTER START
T (m
m)
1,500
2,000
KN)
Fill completed
Load in pile at 20m depth
CASE #6
0
5
10
15
20
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
DEP
TH (
m)
FILL
FineSand
StiffSiltyClay
Tran
sfer
Zon
e
Bitumen coated
pile
2828
Load distribution on two pipe piles, one bitumen-coated and one uncoated
20
25
30
35
40
SE
TTLE
ME
NT
0
500
1,000
LOAD
(K
Settlement
Ground surface settlement due to a 3 m high surcharge placed over 200m x 100m area around the test piles.
Data from Walker et al. (1973
25
30
35
SandySilt
Sandand
Gravel
Leung, C.F, Radhakrishnan, R., and Tan Siew Ann (1991) presented a case history on
instrumented 280 mm square precast concrete piles driven in marine clay in Singapore
Note, the distribution of negative skin friction is linear (down to the beginning of the transition zone) indicating the proportionality to the effective overburden stress
CASE #7
2929
Neutral PlaneTRANSITION
ZONE
0
5
10
0 200 400 600
LOAD (KN)
(m)
OldSilt&
ClayFill
Marine
ß = 0.5
Two months
after start(57 days)
Two years later
(745 days)
Variable load i.e., Live Load
3030Data from Leung, Radhakrishnan, and Tan (1991)
15
20
25
30
DEP
TH
Clay
Weak Shale
Bedrockand
Residual Soil
Clay
6
• Indraratna et al. (1992) reported results from full-scale field tests on instrumented 400 mm diameter, 25 m long cylinder piles driven into Bangkok clay
CASE #8
3131
piles driven into Bangkok clay. Tests were made in push and pull on bitumen-coated and uncoated piles.
-10
-5
0
5
-15 -10 -5 0 5 10 15
Distance (m)
Dist
ance
(m
)
Coated Uncoated
3232Indraratna et al. 1992
10
-5
0
5
10
15
20
25
30
-15
DEPT
H (
m)
PiezometersLoad Cells
Coa
ted
Leng
th
Weathered
Soft
Stiff
2 m high embankment Fill
Installation was made one 4 m long segment at a time. Beforesplicing on the next segment and continuing, pull tests were carriedout. The diagrams below show the measured load-movementcurves.
300
400
500
AD
(K
N)
8 m Length
12 m Length
16 m Length
20 m Length300
400
500
AD
(K
N)
8 m Length
12 m Length
16 m Length
20 m Length
Uncoated Pile Bitumen Coated Pile
00 200 400 600 800
3333
0
100
200
0 10 20 30 40 50
MOVEMENT (mm)
LOA
0
100
200
0 10 20 30 40 50
MOVEMENT (mm)
LOA
Data from Indraratna et al. 1992
Pull Test Results
5
10
15
20
25
30
0
5
0 100 200 300 400 500 600 700 800
Axial Load (KN)
Ground Surface
Loads from pull tests
The measurements compiled and put into
the context of the basic soil response.
0
5
0 100 200 300
SETTLEMENT (mm)
PILE
0 5 10 15 20
3434Data from Indraratna et al. 1992
10
15
20
25
30
DE
PTH
(m
)
β = 0.2
β = 0.3 β = 0.3
Loads at 265 days
p
Settlement distributionMeasured and calculated load distribution
10
15
20
25
30D
EPTH
(m
)
3 Days
25 Days
53 Days
81 Days
141 Days
209 Days
265 Days
10
15
20
25
35
CASE #9
36
Okabe, T., 1977. Large negative skin friction and friction-free methods.
Proc. 9th ICSMFE, Tokyo, Vol.1, pp. 679 – 683.
7
Strain-gage instrumented, 600mmdiameter, pipe piles driven throughsilty clay and silt with silty sand.
A fill was placed on the ground (no
CASE #9
0
5
10
15
20
0 10 20 30 40 50 60 70 80
wn, wP, and wL (%)
(m)
37Data from Okabe (1977)
A fill was placed on the ground (noinformation of fill height) over a vastarea of the site and pumping ofwater at depth lowered the porepressures at depth (no informationon pore pressure distribution).
25
30
35
40
45
50
DEP
TH (
0
10
0 2,000 4,000 6,000 8,000
LOAD (KN)
83 days
131 days
277 days
944 days
Test Pile #1 — Single Pile
Load distribution with time after driving
Test Pile #2 — Single Pile
Load distributions as loads (700 KN and 1,000 KN have been placed on the pile head
0
10
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
550 days
+1 day
+5 days
+28 days
38
20
30
40
50
DEP
TH (
m)
1,663 days20
30
40
50
DEP
TH (
m) +63 days
++1 day
++5 days
++19 days
++43 days
++59 days
++194 days
?
Data from Okabe 1977
0
10
-2,000 0 2,000 4,000 6,000 8,000
LOAD (KN)
-
Note: Tension
Average of sleeved piles
Group of piles connected to a common cap
Four piles are “sleeved” one pile is not
0
10
-1,000 0 1,000 2,000 3,000 4,000
LOAD (KN)
?
-
Sacrificial
Group of piles connected to a common cap surrounded by “sacrificial” piles supposedly(?)
not connected to cap
39
20
30
40
50
DEP
TH (
m)
Depth of sleeve
No-sleeve pile
20
30
40
50
DEP
TH (
m)
Sac c apiles
Foundation piles
Sacrificial pile
Data from Okabe 1977
Case historieson
Damaging Drag Load
4040
g g gand
Damaging Downdrag
Inoue, Y., Tamaoki, K., and Ogai, T., 1977. Settlement of building due to pile downdrag. Proc. 9th ICSMFE, Tokyo, July 10-15, Vol. 1, pp. 561– 564.
CASE #10
41
A three-storey building with a foot print of 15 m by 100 m founded on500 mm diameter open-toe pipe piles driven through sand and siltyclay to bearing in a sand layer at about 35 m depth. The piles hadmore than adequate capacity to carry the building. Two years afterconstruction, the building was noticed to have settled some 150 mm.Measurements during the following two years showed about 200 mmadditional settlement. The building was demolished at that time.
A Downdrag Case
< 102 m >
SAND FILL
42
FINE SAND
SILTY CLAY: w = wL = 40% - 60%; τu = 40 KP
SILTY CLAY: w = wL = 40% - 60%; τu = 80
FINE SAND
SAND FILL
FINE SAND
SILT & SAND
Pile Toe Depth Inoue 1977
8
FINE SAND
SILTY CLAY: w = wL = 40% - 60%; τu = 40 KP
SILTY CLAY: w = wL = 40% - 60%; τu = 80
FINE SAND
SAND FILL
FINE SAND
SILT & SAND
43Settlement between piles in Row 6 and Row 10 from Sep. 1967 through May 1969 = 150 mm.
Slope ≅ 1 : 100 (Sep 67 Apr 71)
Inoue 1977
LOAD — SETTLEMENT
m)
Soil settlement
Load in Pile
Building
-5
0
5
10
15
0 200 400 600 800
m)
Stress and Pressure (KPa)
SAND
SAND
CLAY
Settling Layer
44
Dep
th (
Speculative distribution
Data from Inoue et al. 1977
20
25
30
35
40
45
Dep
th (m
SAND
CLAY
SANDsilt"Current"
Ef fective Stress
Pore Pressure
30 m long piles driven to bedrock 50 m long
piles driven to shaft bearring
Province A Province B
Marine l
XXXX ≈NEUTRAL PLANE
Highly loaded (max allowed by code)
Lightly loaded
≈NEUTRAL PLANE
4545One Bridge — Two foundations
?
clay on bedrock
A CASE HISTORY OF A STRIP-MALL FOUNDED ON PILES
GROUND SURFACE
4646Limestone bedrock providing good bearing
The soils investigation revealed 54 ft (16m) of no-strength "muck".
Design called for 54 ft long piles. Designer discounted all shaft resistance contribution.
54 ft(16 m)
54 ft (16m) f " k'
GROUND SURFACE
4747
of "muck'
Limestone bedrock providing good bearing
Strip-Mall as designed
A real DOWNDRAG case
ORIGINALGROUND SURFACE
5 ft (1.5m) of fill added before the piles were
4848
X x X x X x X x X x X x X x X
54 ft (16m) of "muck"
Limestone bedrock providing good bearing
pdriven
5 ft (1.5m) of "muck"
9
Office Building placed on longtoe-bearing piles in Brisbane,Australia. A later built, lightextension was placed on shortwood piles.
5 m
0 m
FILL
Soft CLAY
49
10 m
15 m
0 1 2 3 MPaqc :
Stiff CLAY
SANDSTONE
Data Courtesy (2007) of Wagstaff Piling Pty. Ltd., Queensland, Australia.
View toward the roof
5050
Foundations and underpinning
5151
The original piles had a capacityabout three times the applied load,but downdrag got the better of them.Luckily, new piles could be installed(driven to the sandstone; for somecolumns to a“four-for-one” ratio).
$ $ $ $
A downdrag Case
52Courtesy of Joram M. Amir, Amir Geotechnical Engineering Ltd.
The photo shows a building placed on piled foundations made up of 14 m longbored piles through loose non-engineered coarse-grained fill into soft rock.Once the building was occupied and the lawn and garden established andwatered, the fill lost volume and ground settled. The pile capacity was morethan adequate to support the building loads.
Under one wing of the building, the fill layer was thicker, in fact, approachingthe length of the piles. The reduced pile length in the rock plus the pile toeresistance under the wing was not sufficient to prevent a neutral plane —settlement equilibrium — from developing in the fill. Therefore, considerabledowndrag developed and the building wing suffered extensive damage.
Monitoring deformations and settlement in Uppsala, Sweden (1965)
70
53530
10
20
30
40
50
60
70
0 10 20 30 40 50 60
EAST-WEST (m)
NO
RTH
- SO
UTH
(m
)
Stand-pipe Piezometers, ZHouse Bench MarksGround Benchmark, STest Piles, PBorehole, B+
+
N
P88
P52
P2P1
Z1
Z2
Z3
B1
S3 S1
S2
0
5
10
0 10 20 30 40 50UNDRAINED SHEAR STRENGTH (KPa)
m)
0
5
10
0 10 20 30 40 50 60 70 80WATER CONTENT (%)
m)
CLAYCLAY
(mean level) GW GW
Basement Excavation
+11.00 m
Clay crust
Soil Profile After construction, a one metre thick fill was placed over the site around all buildings.
54
15
20
25
DEP
TH (
m
15
20
25
DEP
TH (
m
SILT SILT
SAND & GRAVEL SAND & GRAVELPilePile
Drop-hammer driven, 250 mm diameter, ordinary reinforced, precast concrete pile.
10
LID
LID
BASEMENTWALL
BASEMENTFLOOR
PILE
CENTERTUBE
3
21
CLAY
+ 11.00 m
+ 10.30 m
ORIGINALGROUNDSURFACE CONSOL
+ 12.00 m
Fill
50 mm 'SETTLEMENT' PIPE
4 7
8
9
10
0 200 400 600 800 1000 1200
DAYS
PHRE
ATI
C E
LEVA
TIO
N (
m)
WINTER -67/68 WINTER -68/69 WINTER -69/70
+
+
+
+
Z1 & Z2
Z3
0 200 400 600 800 1 000 1 200
DAYS
Phreatic elevation measured in Piezometers Z1 - Z3
55
UPPERTELLTALE
LOWERTELLTALE
PILE TOE
19.5 mm SOUNDINGROD
25 mm PROTECTION PIPE
SILTSAND
GRAVEL
- ≈ 5.0 m
- ≈ 10.5 m
0
5
10
15
20
25
30
0 200 400 600 800 1,000 1,200G
RO
UN
D S
UR
FAC
E SE
TTLE
MEN
T (m
m)
S1S2P1P2P52P88
WINTER -67/68 WINTER -68/69 WINTER -69/70
Settlement of ground surface at benchmarks close to building and at test piles
Measurement arrangement at the test pilesfor measuring pile shortening and settlementof building and soil
-3
-2
-1
0
1
0 200 400 600 800 1 000 1 200
SETT
LEM
ENT
(mm
)
From Survey At Pile 1 At Pile 2
At Pile 52 At Pile 88
WINTER -67/68 WINTER -68/69 WINTER -69/70
-2
-1
0
OVE
MEN
T (m
m)
F ll L th Sh t i
56
0 200 400 600 800 1,000 1,200
DAYS
-30 200 400 600 800 1,000 1,200
DAYS
MO Full Length Shortening
Upper Length ShorteningLower Length ShorteningWall SettlementPile Toe Penetration
Pile 1Pile P1 shortening and building settlement measured after completion of building. The shortening of the upper and lower lengths of pile were about 1.5 mm and 0.5 mm, respectively, which corresponds to an average strain of 100 µε for both lengths
Settlement of pile-supported basement wall measured near the test piles and according to survey
0
5
10
0 200 400 600 800 1,000
LOAD (KN)
EPTH
(m
)
Increase of load after three years
0
5
10
0 200 400 600 800 1,000
LOAD (KN)
EPTH
(m
)
Load distribution after three years
Load distribution at end of construction
Drag load after three years
57
15
20
25
DE
Average Load in Upper, Lower,and Full Lengths of Pile
BA. Typical distribution of load in a pile at end of construction and three years later
B. Increase of load in the pile during three years
15
20
25
DE
Neutral plane
Assumed short and long term mobilized toe resistances
A
1
BASICS OF DESIGN OF PILED
FOUNDATIONS
Bengt H Fellenius
1
Bengt H. Fellenius
The Unified Method for Design of Piled FoundationsCapacity, Drag Load, Settlement, and Downdrag
Bolivia, April 25, 2013
A foundation design is carried out by a team of professionals
2
3
Distribution of load at the pile cap
The case histories have made us realize how piles and piled foundations will respond to load and settling soil. Here are a few applications of the knowledge gained.
00 500 1,000 1,500 2,000 2,500 3,000
LOAD and RESISTANCE (KN)
DEAD LOAD LIVE LOAD CAPACITY
44
5
10
15
20
DEP
TH (
m)
CLAY
SAND Neutral PlaneTransition Zone
Static Loading Test Distribution
The effect of different pile length and/or different toe resistance response
0
5
0 500 1,000 1,500 2,000 2,500 3,000
LOAD and RESISTANCE (KN)
m)
CLAY
55
10
15
20
DEP
TH (
m
SAND
This pile happens to have more (about 50%) toe resistance or is longer.
0
5
10
-500 0 500 1,000 1,500 2,000 2,500
LOAD and RESISTANCE (KN)
EPTH
(m
)
CLAY
66
15
20
DE
SAND Neutral PlaneTransition Zone
Now, let’s assume that this pile is damaged at the pile toe, or that debris collected at the toe, eliminating the toe resistance.
So, what is the effect of this?
2
0
5
-500 0 500 1,000 1,500 2,000 2,500
LOAD and RESISTANCE (KN)
(m)
CLAY
77
10
15
20
DEP
TH
SAND Neutral PlaneTransition Zone
AVERAGE"AVERAGE" or Design Pile
The analysis of the capacity and resistance distribution detailed in the foregoing deals withanalysis of a single pile or small groups of piles, where interaction between the piles doesnot occur or is negligible. However, for larger groups, the group effect is substantial. Thereis a difference between a single pile and a pile in a group of piles, as well as between pilesinside a pile group as opposed to at the perimeter of the group.
Most of our knowledge comes from studies of the response of single piles. Yet, piles are rarely single. So, what about the response of a group of piles?
8
A moderate size pile group (36 piles), two small groups (4 and 2 piles), and a single pile.
Pile Group: b = 0.32 m, c/c = 1.0 m (1.4 m) = 3.1b (4.4 b) Pile area to footprint area = 8 %
0
5
10
15
20
25
0 50 100 150
Neg. Skin. Friction (KPa)
DEP
TH (
m)
0
5
10
15
20
25
0 1,000 2,000 3,000
Drag Load (KN)
DEP
TH (
m)
0
5
10
15
20
25
0 50 100 150
Neg. Skin. Friction (KPa)
DEP
TH (
m)
9
30
35
40
45
D
CENTER SINGLE
Distributions of unit negative skin friction and of accumulated drag load for a single pile and for a fully shielded pile in the center of the group.
30
35
40
45
D
30
35
40
45
D
CENTER SINGLE SINGLECENTER
0
5
10
15
20
25
0 500 1,000 1,500 2,000 2,500 3,000
LOAD (KN)
DEP
TH (
m)
SINGLE
CENTER
SIDE
CORNER
10
30
35
40
45
Load distribution in the pile group piles ("Center", "Side", and "Corner") and single pile ("Single), assuming the Center piles fully shielded, and the Center and Corner piles partially shielded from the negative skin friction effect.
Let's assume that conditions at depth > 40 m are such that the neutral plane lies at a depth of 40 m
0
5
10
15
20
25
0 500 1,000 1,500 2,000 2,500 3,000
LOAD (KN)
EPTH
(m
)
SINGLE
CENTER
OUTER
11
Load distribution in the pile group piles ("Outer" and "Corner") and single pile("Single"), assuming that the center piles ("Center") are fully shielded, andthat the outer piles are not shielded from the negative skin friction effect.Note, the effect of pile cap stiffness is not included in the foregoing.
30
35
40
45
DE
We can use our understanding to critically review design recommendations in current text books and standards. For example:
12
text books and standards. For example:
3
A quote from a textbook *)
“The net effect negative skin friction is that the pile load capacity is reducedand pile settlement increases. The 'allowable load capacity' (sic!) is given as:”
negnegult
allow QF
QQQ −
−=
It could have been worse. Logically, the drag load should here been increased by a factor of safety. But so what there is little logic in
13
If you think this ghastly recommendation is correct, you have not been paying attention!
*) Compassion—perhaps misdirected—compels me not to identify the author
SF But so what, there is little logic in the approach anyway.
Do not include the drag load when determining the allowable load!
0
5
10
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
TH (
m)
ALLOWABLE
LOAD - (Fs = 2.5)CAPACITY
Drag load must neither subtracted from the pile capacity nor from the allowable load
0
5
10
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
TH (
m)
ALLOWABLE LOAD minus DRAGLOAD*1.0
CAPACITY
Effect of subtracting the drag load
14
10
15
20
DEP
T
DRAG LOAD
10
15
20
DEP
T
DRAG LOAD
INCREASE!If the pile capacity had been reduced with the amountof the drag load before subtracting the drag load, therewould have been no room left for the working load!
Similarly for the LRFD:
Do not include the drag load when determining the factored resistance!
Drag load not subtracted from the factored resistance Drag load factored and subtracted!
0
5
0 500 1,000 1,500 2,000 2,500
LOAD (KN)FACTORED RESISTANCE
CAPACITY
0
5
0 500 1,000 1,500 2,000 2,500LOAD (KN)
FACTORED RESISTANCEminus FACTORED DRAGLOADFactors = 0.6 and 1.5, respectively
FACTORED RESISTANCE CAPACITY
1515
10
15
20
DEP
TH (
m)
DRAG LOAD
10
15
20
DEPT
H (
m)
DRAG LOAD
If a factor of safety of 2.0 is applied and the drag load is subtracted from the allowable load . . . , then ?
The allowable load becomes zero!
Imagine a shaft-bearing pile (no toe resistance) with a certain capacity and an allowable load for a factor of safety of 2.0. In a settling soil, the drag load amounts to half the capacity value.
1616
Imagine that same pile designed for uplift: Logically, if one subtracts the drag load for the push case, should one not add it for the pull case ??!!??
Do you think that there is a difference in bearing capacity between an
ordinary precast and a prestressed pile? — Be the pile prestressed or not
prestressed, the prestress has nothing to do with the pile bearing capacity.
Load placed on a pile causes downward movements of the pile head due to:
1. 'Elastic' compression of the pile.
2. Load transfer movement -- the movement response of the soil.
3. Settlement below the pile toe due to the increase of stress in the soil. This isonly of importance for large pile groups, and where the soil layers below the pilesare compressible.
SETTLEMENT
1717
A drag load will only directly cause movement due to Point 1, the'elastic' compression. While it could be argued that Point 2 also is atplay, because the stiffness of the soil at the pile toe is an importantfactor here, it is mostly the downdrag that governs (a) the pile toemovement, (b) the pile toe load, and (c) the location of the neutralplane in an interactive — "unified" — process.The drag load cannot cause settlement due to Point 3, because there has been no stress change in the soil below the pile toe.
Negative-skin-friction/drag-load does not diminish
capacity. Drag load (and dead load) is a matter
for the pile structural strength, and the main
question is "will settlement that can cause
1818
downdrag occur around the pile(s)"? The
approach is expressed in “The Unified Design
Method”, which is a method based on the
interaction between forces and movements.
4
The Unified Design Method is a three-step approach
1. The dead plus live load must be smaller than the pile capacitydivided by an appropriate factor of safety. The drag load is not included when designing against the bearing capacity.
1919
2. The dead load plus the drag load must be smaller than the structural strength divided with a appropriate factor of safety. The live load is not included because live load and drag load cannot coexist.
3. The settlement of the pile (pile group) must be smaller than a limiting value. The live load and drag load are not included in this analysis. (The load from the structure supported by the piles does not normally cause much settlement, but the settlement due to other causes can be large. The latter is called downdrag).
"
2020
Construing the Neutral Plane and Determining the Allowable Load
The distribution of load at the pile cap is governed by the load-transfer behavior of the piles. The “design pile” can be said to be the average pile. However, the loads can differ considerably between the piles depending on toe resistance, length of piles, spacing, etc.
2121
The location of the neutral plane is the result of Nature’s iterations to find the force equilibrium. If the end result — by design or by mistake — is that the neutral plane lies in or above a compressible soil layer, the pile(s) will settle even if the total factor of safety appears to be acceptable.
The principles of the mechanism are illustratedin the following three diagrams
2222
The mobilized toe resistance, Rt, is a function of the Net Pile Toe Movement
Pile toe response for where the settlement is small (1) and where it is large (2)
00 1,500LOAD and RESISTANCE
00
SETTLEMENT
21
NEUTRAL PLANE 1
Utimate Resistance
2323
DEP
TH
1 2
NEUTRAL PLANE 2
Toe Penetrations
Note, the magnitude of settlement affects not only the magnitude of toe resistance but also the length of the Transition Zone
= Movement into the soil
Pile toe response for where the settlement is small (1) and where it is large (2)
00 1,500LOAD and RESISTANCE
00
SETTLEMENT
21
NEUTRAL PLANE 1
Utimate Resistance
2424
DEP
TH
1 2
NEUTRAL PLANE 2
Toe Penetrations
Note, the magnitude of settlement affects not only the magnitude of toe resistance but also the length of the Transition Zone
1 TOE PENETRATION
TOE
RES
ISTA
NC
E 1
2
3
a b c
5
25
0
5
10
0 10 20 30 40 50
FILL
Silty SAND
SAND
N
N (bl/0.3m); qt (MPa); wP, wn, and wL (%);
GW
Example of the Unified Design Approach as Applied to a Refinery Structure
Design for a large refineryexpansion was undertaken at asite reclaimed from a lake inthe 1960s. The natural soilsconsist of sand deposited onnormally consolidated, com-pressible post glacial lacustrineclay followed by silty clay till onlimestone bedrock found atabout 25 m to 30 m depth
Soil Profile
26
15
20
25
30
DEP
TH (
m)
Silty SAND
CLAY TILL
LIMESTONE BEDROCK
wn
CLAY
m ≈ 100
mr ≈ 600 ∆σ' ≈ 9 MPa
m ≈ 20
mr ≈ 200 ∆σ' ≈ 20 KPaw LwP
qt
about 25 m to 30 m depthbelow existing grade. The sitewill be raised an additional1.5 m, which will cause long-term settlement. Some of thenew units are 30 m to 70 m inheight and will be supported onpiles—several thousand in all.
Fellenius and Ochoa (2010)
40
60
80
n)
Results of an O-cell test on a 575-mmdiameter test pile, a 26 m deep boredcylindrical pile. A 1.5 m thick fill willbe placed over the site beforeconstruction. Piles are single or insmall groups.
Results of analysis of test data:Load Distributions
0
5
10
0 1,000 2,000 3,000 4,000 5,000LOAD (KN)
m)
O-cell Test
Strain-gage Value
"FlippedSilt
Sand
27
-60
-40
-20
0
20
40
0 500 1,000 1,500 2,000 2,500LOAD (kips)
MO
VEM
ENT
(in
9 %
15
20
25
30
DEP
TH (m
O-cell
Long-term. (After Completed Consolidation due to the Fill).
Clay
Till
0
100
200
3000 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
DO
WN
WA
RD
M
OVE
MEN
T (m
m)
O-cell Testq-z extrapolation
0
5
0 2,000 4,000 6,000LOAD (KN)
SiltSand
0
5
0 2,000 4,000 6,000LOAD (KN)
SiltSand
Qd
Distribution of residual load in the pile after installation, but before load is applied to the pile.
Distribution of load in the pile immediately after the pile starts to sustain the load from the structure.
28
10
15
20
25
30
DEP
TH (m
)
O-cell
Clay
Till
Residual Load Distribution Before Construction
10
15
20
25
30
DEP
TH (m
)O-cell
Clay
Till
Load Distribution Immediately After Construction
Long-term load distribution
The shaft shear isassumed to be fullymobilized. However, thetoe resistance value touse is a function of the
0
5
10
0 2,000 4,000 6,000
LOAD (KN)
(m)
SiltSand
Cl
Qd
29
toe penetration due todowndrag and can onlybe determined fromassessing the soilsettlement distribution.
15
20
25
30
DEP
TH
O-cell
Clay
Till
Long-term Load Distribution
0
5
10
15
0 2,000 4,000 6,000LOAD (KN)
PTH
(m)
SiltSand
Clay
Qd
0
5
10
15
0 50 100 150 200
SETTLEMENT (mm)
PTH
(m)
0
5
10
15
0 2,000 4,000 6,000LOAD (KN)
PTH
(m)
SiltSand
Clay
Qd Pile Cap Settlement
Soil Settlement
Force and settlement (downdrag) interactive design. The unified pile design for capacity, drag load, settlement, and downdrag
30
20
25
30
DEP
O-cell
Till
Pile toe load in the load distribution diagram must match the toe load induced by the toe movement (penetration), which match is achieved by a trial-and-error procedure.
20
25
30
DEP
20
25
30
DEP
O-cell
Till
0
1,000
2,000
3,000
4,0000 50 100
TOE
LOAD
(KN
)
Pile toe load in the load distribution diagram must match the toe load induced by the toe movement (penetration), which match is achieved by a trial-and-error procedure. PILE TOE PENETRATION (mm)
q-z relation
6
0
5
10
15
0 50 100 150 200
SETTLEMENT (mm)
PTH
(m)
0
5
10
15
0 2,000 4,000 6,000LOAD (KN)
PTH
(m)
SiltSand
Clay
Qd Pile Cap Settlement
Soil Settlement
Force and settlement (downdrag) interactive design. The unified pile design for capacity, drag load, settlement, and downdrag
31
20
25
30
DEP
20
25
30
DEP
O-cell
Till
0
1,000
2,000
3,000
4,0000 50 100
TOE
LOAD
(KN
)
Pile toe load in the load distribution diagram must match the toe load induced by the toe movement (penetration), which match is achieved by a trial-and-error procedure. PILE TOE PENETRATION (mm)
q-z relation
0
5
10
15
0 50 100 150 200
SETTLEMENT (mm)
PTH
(m)
0
5
10
15
0 2,000 4,000 6,000LOAD (KN)
PTH
(m)
SiltSand
Clay
Qd Pile Cap Settlement
Soil Settlement
Force and settlement (downdrag) interactive design. The unified pile design for capacity, drag load, settlement, and downdrag
Summary
32
20
25
30
DE
P
20
25
30
DE
P
O-cell
Till
0
1,000
2,000
3,000
4,0000 50 100
TOE
LO
AD
(K
N)
Pile toe load in the load distribution diagram must match the toe load induced by the toe movement (penetration), which match is achieved by a trial-and-error procedure. PILE TOE PENETRATION (mm)
q-z relation
The final solution is based on three "knowns": The shaft resistance distribution, the toe load-movement response, and the overall settlement distribution. Which all comes from basic site and project knowledge.
The Unified Design Analysis for load-transfer and long-term downdrag involves an iterative procedure applied to single piles and small pile groups
2. Soil settlement due to sustained pile load and changes of effective stress around the pile(s)due to other causes: Calculate and plot the distribution of soil settlement developing after thesustained load has been placed on the pile. For single piles and small pile groups, the settlementbelow the pile toe due to the sustained pile loads will be small, usually negligible. However, settlementdistribution due to increase of effective stress from other causes, such as fills, groundwater (pore
1. Load-transfer response: Calculate and plot the distribution of the shaft resistance. Determine —make an assumption on — the magnitude of toe resistance and toe movement that developed beforethe sustained load was placed on the pile. This requires applying a pile-toe load-movement relation.You need either to have measurement results showing the relation, or assume a q-z relation to use.Maybe you have the results of an O-cell test available. This analysis produces a preliminary location ofthe Neutral Plane, N.P.
33
In a routine case, it is usually sufficient to just make sure that the neutral plane lies below a levelindicating settlement that can be accepted. Moreover, when analyzing not just single piles or a fewpiles clustered together, but pile groups, matters can become more complicated.
distribution due to increase of effective stress from other causes, such as fills, groundwater (porepressure) changes, etc. could be substantial and must be calculated. At the N.P., the settlement of thepile(s) and the soil are equal. Determine now what increase of toe resistance the enforced extra piletoe penetration has resulted in and let an iterative set of calculations determine agreement betweenN.P. as force equilibrium and settlement equilibrium. This is established once the toe load and net toemovement (difference between the total toe movement and the soil settlement at the pile toe) fits thetoe relation the downdrag (i.e., pile settlement).
The Unified Design Analysis for Pile Groups
A large pile group within a footprint made up of one raft or more adjoining rafts can be defined a collection of piles in more or less perpendicularly placed rows of piles or small piled footings where the smallest number of rows is at least three, usually five or more. The load applied to the footprint of such groups could significantly increase the effective stress below the pile toe level and cause the soil to compress (and the pile group to settle).
The transfer of the stress from the load on the raft, rafts, or pile footings over the particular footprint will start at the N.P. First by shaft resistance and, finally, as toe resistance. The settlement between the N.P. and the pile toe level can normally be disregarded because the presence of the piles have created a very stiff unit of "reinforced soil" that will not compress appreciably. The key question is how the pile
34
very stiff unit of reinforced soil that will not compress appreciably. The key question is how the pile load will distribute between the N.P. and the pile toe level. The shaft response can be considered as a series of individual elements between the N.P. and the pile toe, each spreading the load at 2(V):1(H) from the pile raft, rafts, or footings to the pile toe level. Progressively, however, each such element along the pile will have a shorter distance to the pile toe, and, therefore affect a smaller footprint. The pile toe resistance, of course, acts right a the pile toe. This series of stress footprints can be approximated to an equivalent raft at the N.P. with the footprint of the piled raft, rafts, or footings that spreads the total load at 5(V):1(H) to the pile toe level. From the pile toe level, the stress is distributed by either 2(V):1(H) for the average settlement of by Boussinesq distribution for calculation of the settlement at different locations within the pile group. Note, the drag load must not be included in the analysis of settlement.
Settlement Analysis of Large Pile Groups by the Equivalent Footing Method
0 300 600 900 1,200
LOAD (KN)
Qd+Qn alternatives
Qd0
5
10
0 300 600 900 1,200
LOAD (KN)
Qd60
80
AT
TOE
#7
Qd on Equiv. Raft Projected
35
DEP
TH (
m)
Ru-Rs
Rt
Neutral PlaneEquivalent Raft8m x 12 m
15
20
25
30
35
40
DEP
TH (
m)
Rt
Qd#1
#2#3
#4#5
#6#7
The Qd transferred to the soil in several steps along the pile from the neutral plane to the pile toe
0
20
40
-15 -10 -5 0 5 10 15 20
STR
ESS
A
AVERAGE WIDTH (m)
#2
#3
#4
#5
#6
#1
Qd on Equiv. Raft Projected 2(V):1(H)
5(V):1(H)
Pile Group F t i t
Settlement Analysis of Large Pile Groups by the Equivalent Footing Method
40
60
80
AT
TOE
#7
Qd on Equiv. Raft Projected 5(V):1(H)
36
5(V):1(H)
2(V):1(H) or Boussinesq
Neutral Plane
Footprint
Pile Toe Depth
Footprint Projected 5(V):1(H) to form an Equivalent Raft
Projection of Raft for Settlement Analysis
0
20
40
-15 -10 -5 0 5 10 15 20
STR
ESS
AVERAGE WIDTH (m)
#2
#3
#4
#5
#6
#1
Qd on Equiv. Raft Projected 2(V):1(H)
7
Settlement Analysis of Large Pile Groups by the Equivalent Raft Method
soilsoilpilepile EAEA +The compressibility in this
Equivalent Footing placed at the Location of the Neutral Plane
G.W
FILLS, etc.
5:1 5:1
Start by placing an "Equivalent Raft" at the
depth of the Neutral Plane
37
soilpile
soilsoilpilepilecombined AA
E+
=The compressibility in this zone must be of soil and pile combined
2:1 distribution 2:1 distribution
Settlement of the piled foundation is caused by the compression of the soil due to increase of effective stress below the neutral plane from external load applied to the piles and, for example, from fills, embankments, loads on adjacent foundations, and lowering of groundwater table.
5:1 5:1
The approach is far beyond the 1948 Terzaghi and Peck approachof placing The Equivalent Raft at the lower third depth
Liquid storage tank in Tessaloniki, Greece (Savvaidis 2009)
38
-20
-15
-10
-5
0
5
10
15
20
-20 -15 -10 -5 0 5 10 15 20
East-West
Nor
ht-S
outh
Pile 11
Pile 16
Pile 7
112 1.0 m diameter, bored piles installed to 42 m depth
Dense, silty sand to 50+ m depth Settlement
No Settlement
0
10
0 5 10 15 20 25 30 35 40
ACROSS THE DIAMETER (m)
NT
(mm
)
Settlement measured across a diameter during a 30 Hydro Test
39
20
30
40
SETT
LEM
EN
Curve calculated for a flexible footing locatedat the pile toe level with parameters fitted tothe settlement measured at the tank mid-point
0
5
10
0 5 10 15 20 25 30 35 40
CONE STRESS, qc (MPa)
CLAYEY SAND
Ghent Grain Terminal — Settlement of a large pile group(Goossens and VanImpe, 1991)
85 m
34 m41 x 17 = 697 PILES
B
PILE GROUP CASE HISTORYFootprint ratio = 9 %
40
15
20
25
30
35
40
DEO
PTH
(m
)
CLAY
SAND
CLAY
Very dense SAND
0
500
1,000
1,500
2,000
2,500
0 2 4 6 8 10
MOVEMENT (mm)
LOA
D (
KN)
Pile #585
Pile #085
Can we use the results of the two static loading tests to estimate the settlement of the pile group?
0
50
NT
(mm
)
770 days
1 080 days
Footprint of Silo Foundation84 m by 34 m
BM 4 BM 2A BM 2 BM 1 BM 3
0
50
100
T (m
m) Measured and
calculated at the benchmarks
Footprint of Silo Foundation84 m by 34 m
BM 4 BM 2A BM 2 BM 1 BM 3
41Center to corner differential settlement is 0.25÷45 ≈ 1:200
100
150
200
SETT
LEM
EN
1,080 days1,245 days
1,960 days2,637 days3,471 days3,808 days
150
200
250
300
SETT
LEM
EN
benchmarks
Calculated value fitted to measured
Calculated for center line points
Calculated
42
Settlement of a Pile Group Supporting Five Furnaces at QIT Plant, Sorel, Quebec
Golder, H.Q. and Osler J.C., 1968.Settlement of a furnace foundation, Sorel, Quebec.Canadian Geotechnical Journal 5(1) 46-56.
8
0
500
1,000
1,500
2,000
0 1 2 3 4
MOVEMENT (mm)
LOA
D (
KN
)
#1 #2 #3 #4 #5
54 m
16 m
NORTH SOUTH
Static loading test used to predict settlement of the five furnaces: 10 mm
v
v
v
vv v v v v v v v
v
v
10 m
16 m
6 m Pile Embedment Pile Toe at 8.5 m Depth
LAYOUT OF ONE FURNACE
Footprint ratio = 7 % (28 %?)
43
0
20
40
60
80
SETT
LEM
TN (
mm
)
Apr. 1951 Nov. 1951
Aug. 1952
Jan. 1962
Furnace #1 Furnace #2 Furnace #3 Furnace #4 Furnace #5
Measured Settlement24 m of Compact SAND50 m Champlain CLAY
0.60
0.80
1.00
1.20
10 100 1,000 10,000
Stress (KPa)
Voi
d R
atio
(- -
)
CR = 0.026
mr = 90
CR = 0.257 m = 9
DEPTH = 14 mσ'0 = 230 KPaσ'c = 280 KPa
0
10
20
30
40
50
60
70
80
YEARS
SETT
LEM
ENT
(mm
)
1950 1955 1960 1965
North and South Side Furnaces
CenterFurnaces
0
10
20
30
0 25 50 75 100 125
SETTLEMENT (mm)
DEP
TH (
m)
Center of Furnace Row
Inside edge of Furnace #2
SANDCLAY+SAND
SAND
SANDY CLAY
CHAMPLAINCLAY
Outside edge of Furnaces #1 and #5
Between Furnaces#1 & #2 and #4 and #5
44
40
50
60
0
20
40
60
80
SETT
LEM
TN (
mm
)
Furnace #1 Furnace #2 Furnace #3 Furnace #4 Furnace #5
Nov. 1951
Aug. 1952
Jan. 1962
CALCULATED SETTLEMENT
SINCE APRIL 1951
Be the one of the pack to dareto design for settlement rather than "capacity"
45
Piled foundations in current codes
The Canadian Building Code and Highway Design Code (1992), as well as the Hong Kong Code (Geo Guide 2006) apply the Unified Design method. That is, the drag load is only of concern for the structural strength of the pile. Indeed, the Canadian Highway Code even states that for piles with an aspect ratio (embedment depth over diameter, D/b), smaller than 80, the design does not have to check for drag load. However, the design must always check for downdrag.
The Manual of US Corps of Engineers indicate a similar approach (but less explicit), stating that the drag load constitutes a settlement problem (as opposed to a bearing capacity problem).
4646
The ASCE “Practice for the Design and Installation on Pile Foundations (2007)” includes the following definitions:
DOWNDRAG: The settlement due to the pile being dragged down by the settling of surrounding soil;
DRAG LOAD: Load imposed on the pile by the surrounding soil as it tends to move downward relative to the pile shaft, due to soil consolidation, surcharges, or other causes.
and the following statement: . . . , the allowable load, as well as the pile embedment depth, is governed by concerns for settlement and downdrag, and by concern for structural strength for dead load plus drag load, rather than by bearing capacity.
The FHWA has produced one of the most extensive recent guidelines document. The full reference is: Report No. FHWA-NHI-05-042, Design and Construction of Driven Pile Foundations - Volume I and II. National Highway Institute, Federal Highway Administration, U.S. Department of Transportation, Washington, D.C., April 2006. 1,450 pages.
The current issue, drag load and downdrag, is covered in about 20 of the total number of pages. in all essential parts, the FHWA document adheres to the principles of the Unified Design Method.
The FHWA document indicates the following criteria for identifying a drag load and/or downdrag problem. If any one of these criteria is met, drag load or downdrag shall be considered in the design.
The criteria are:
4747
1. The settlement of the ground surface (after the piles are installed) will be larger than 10 mm (0.4 in) *).
2. The piles will be longer than 25 m (82 ft).
3. The compressible soil layer is thicker than 10 m (33 ft).
4. The water table will be lowered more than 4 m (13 ft).
5. The height of the embankment to be placed on the ground surface exceeds 2 m (6.5 ft).
*) This must not be taken to mean that negative skin friction would not develop unless the settlement is larger than 10 mm (0.5 inch)! On the contrary, both positive shaft resistance and negative skin friction are often mobilized at a movement between the pile and the soil as small as ≈2-3 mm.
The trend is toward Load and Resistance Factor Design(LRFD). The Canadian Highway Code has been based on LRFD for about 20 years. With regard to the drag load and downdrag issue, the Canadian Code follows the unified design method.
4848
Since 1995, the Australian Piling Standard is also a Limit States Design Code (LRFD), and, like the Canadian Code, the recommendation for the design of piled foundations is according to the Unified Method, as quoted in the following.
9
The Australian Piling Standard, AS 2159—1995
3.3.2 Load combinations for strength design The load combinations for strengthdesign shall be as follows:
(a) The design load for ultimate strength design of piles shall be the combination of factored loads which produces the most adverse effect on the pile in accordancewith AS 1170.1
(b) If there are loads induced by soil movement (see Clause 3.3.1.2), they shall becomputed as follows:
49
(i) Design structural strength (see Clause 4.3.5)—determined as follows:(A) 1.2 Fnf — negative friction loads (i.e., drag load).(B) 1.5 Fes — compressive and tensile loads(C) 1.5 Fem— bending moments, shear forces, and axial loads.
(ii) Design geotechnical strength—loads induced by soil movement shall not be taken into account. (EDITORIALLY CORRECTED: axial loads
induced by soil movement shall not be included)
4.3.5 Negative friction In the absence of other information, the geotechnical strength in compression or uplift shall be assumed to be unaffected by negative frictionand shall be computed as set out in Clauses 4.3.1 and 4.3.2 for a single pile, andClause 4.3.3 for a pile group.
The additional axial forces induced in a pile by negative friction shall be considered inthe structural design of the pile.
4.5.3 Settlement Consideration shall be given to the settlement of both a pile and a pile group resulting from effects caused by settlement of the surrounding ground NOTE: In the absence of an analysis in
The Australian Piling Standard, AS 2159—1995
50
from effects caused by settlement of the surrounding ground. NOTE: In the absence of an analysis in which pile-soil interaction is allowed for, the settlement of a pile or pile group subjected to negative friction may be approximated as the greater of the following:(a) The settlement of the ground at the ‘neutral plane’ in the ground, that is the depth at whichthe shaft friction on the pile changes from negative (downward) to positive (upward). Applied compressive loading tends to raise the ‘neutral plane’ and increase the settlement of the pile or pile group.(b) The sum of the following three components:
(i) the compression of the pile shaft due to the design action;(ii) the compression of the pile shaft due to the computed forces arising from negative friction; (iii) the settlement of the portion of the pile shaft in the ‘stable’ soil (the part of the soil profile not
subjected to movement) under the sum of the design action and the maximum computed forcein the pile arising from negative friction.
As many other codes and standards, the Australian Standardcan go overboard with some details
TABLE 4.1RANGE OF VALUES FOR GEOTECHNICAL STRENGTH REDUCTION FACTOR
Method of assessment of ultimate geotechnical strength Range of vStatic load testing to failure 0.70Static proof (not to failure) load testing 0.70Dynamic load testing to failure supported by signal matching 0.65Dynamic load testing to failure not supported by signal matching 0.50Dynamic proof (not to failure) load testing supported by signal matching 0.65
Dynamic proof (not to failure) load testing not supported by signal matching (!) 0 50
51
Dynamic proof (not to failure) load testing not supported by signal matching (!) 0.50
Static analysis using CPT data 0.45
Static analysis using SPT data in cohesionless soils (!) 0.40Static analysis using laboratory data for cohesive soils 0.45
Dynamic analysis using wave equation method (!) 0.45
Dynamic analysis using driving formulae for piles in rock (!) 0.50
Dynamic analysis using driving formulae for piles in sand (!) 0.45
Dynamic analysis using driving formulae for piles in clay (!)
Measurement during installation of proprietary displacement piles,using well established in-house formulae 0.50
As many other codes and standards, the Australian Standardcan go overboard with some details
TABLE 4.1RANGE OF VALUES FOR GEOTECHNICAL STRENGTH REDUCTION FACTOR
Method of assessment of ultimate geotechnical strength Range of valuesStatic load testing to failure 0.70–0.90Static proof (not to failure) load testing 0.70–0.90Dynamic load testing to failure supported by signal matching 0.65–0.85Dynamic load testing to failure not supported by signal matching 0.50–0.70Dynamic proof (not to failure) load testing supported by signal matching 0.65–0.85
Dynamic proof (not to failure) load testing not supported by signal matching (!) 0 50 0 70
52
Dynamic proof (not to failure) load testing not supported by signal matching (!) 0.50–0.70Static analysis using CPT data 0.45–0.65
Static analysis using SPT data in cohesionless soils (!) 0.40–0.55Static analysis using laboratory data for cohesive soils 0.45–0.55
Dynamic analysis using wave equation method (!) 0.45–0.55
Dynamic analysis using driving formulae for piles in rock (!) 0.50–0.65
Dynamic analysis using driving formulae for piles in sand (!) 0.45–0.55
Dynamic analysis using driving formulae for piles in clay (!)
Measurement during installation of proprietary displacement piles,using well established in-house formulae 0.50–0.65
The Euro CodeThe European Community has recently completed EuroCode 7, which issupposed to be adopted by all member states. The EuroCode treats the dragload as a load acting similarly to the load from the structure, and requires it to beadded to that load (or subtracted from the pile capacity). Moreover, the shaftresistance in the soil layer that contributes to the drag load is disregarded whendetermining the pile resistance. That is, when a capacity has been determinedto, say, 1,000 and the drag load is expected to be, say, 400, the usable capacityis 1,600 and the usable unfactored resistance is a mere 1,600. This value isthen factored and reduced by the factored drag load. If resistance factor anddrag load factors are say 0 5 and 1 5 respectively the amount left is 200! What
5353
Unfortunately, the recently issued AASHTO LRFD Specs have adopted the EuroCode approach! A few US State DOTs, e.g., Utah, have wisely rejected the AASHTO Specs and apply the Unified Method.
drag load factors are, say, 0.5 and 1.5, respectively, the amount left is 200! What“salvages” the economy of some designs is that the EuroCode clausesadvocate that the designer maintain the faithful approach that “the drag loadcannot really be that large, can it, please?” to determining the magnitude of thedrag load. Incredibly, the EuroCode says little on how to calculate settlement ofpiled foundations and nothing is stated about downdrag!
5.0 m
SOFT CLAY
SILTY CLAY
11.5 m
FILL
Average unit shaft resistance, rs = 20 KPa
Rs = 94.2 KN; Rs = Qn
Average rs = 50 KPa
Rs = 543 KN fq*300 + fn*94 ≤ 543/fr
Q (unfactored) = 300 KN
Eurocode Guide , Example 7.4 (Bored 0.3 m diameter pile)
CALCULATIONS
and AASHTO Specs example
5454
"The settlement due to the fill is sufficient to develop maximum negative skin friction in the soft clay ".�
1.35*300 + 1.35*94 ≤ 543/1.0
532 ≤ 543
(Alternative: If fr = 1.1, the length in the silty clay becomes 12.4 m)
Rt = 0 KN ?!
The Guide states that the neutral plane lies at the interface of the two clay layers, which based on the information given in the example, cannot be correct. But there is a good deal more wrong with this “design” example.
The Guide states that the two rs-values are from effective stress calculation. The values correlate to soil unit weights of 18 KN/m3 and 19.6 KN/m3, ß-coefficients of 0.4 in both layers with groundwater table at ground surface, and a fill stress of 30 KPa.
10
Analysis using the same numerical values for the pile shaft, but including the benefit of a small toe resistance
0
5
10
0 200 400 600 800 1,000LOAD (KN)
EPTH
(m
)
Fs = 2.50
5
10
SETTLEMENT (mm)
EPTH
(m
)
NeutralPlane
THE KEY QUESTION:
?
5555
If the settlement is acceptable, there may be room for shortening the pile or increasing the load. That would raise the location of the neutral plane. Would then the pile settlement still be acceptable?
15
20
DE Maximum
Load = 500 KNQn = 200 KN not 94 KN
Rf = 760/1.35 KN > 1.35*300 KN
Rf = 560 KN > 405 KN
Rt
125 KN
= Factored resistance
15
20
DE
Toe Movement
THE KEY QUESTION:is the settlement acceptable?
5656
TYPICAL EXAMPLEAn 84 m (275 ft) wide LNG tank is founded on 1,200 400 mm (16 in) driven piles at c/c 5.25b (footprint ratioof 3.5 %). The soil profile consists of 25 m of normally consolidated moderately compressible clay on 10 mof dense sand (with artesian pore pressure) followed by 25 m of moderately compressible, slightlypreconsolidated clay on very dense gravel at 60 m depth. A fill is placed under and around the tank to raisethe ground by one metre.
0
5
10
0 500 1,000 1,500 2,000 2,500 3,000 3,500 4,000
LOAD (KN)
CLAYQd
FILL
TestCondition
2,000
2,500
3,000
N)
Pile Head
Offset Limit2,300 KN
Pile Toe Mvmnt vs. Load at Head
Pile Shortening
57
15
20
25
30
35
40
DEP
TH (
m)
SAND
CLAY
Long-termCondition
0
500
1,000
1,500
0 5 10 15 20 25 30
MOVEMENT (mm)
LOA
D (
KN
Pile Shaft
Pile Toe unaffected by residual load
Pile Toe Mvmnt vs. Load at Toe
vs Mvmnt at Head
Load-movement curves from a static loading test Short and long term load distributions
SETTLEMENT OF A SINGLE PILE OR SMALL PILE GROUP OUTSIDE THE MAIN GROUP
0
10
20
0 1,000 2,000 3,000 4,000
LOAD (KN)
Long-termQd
0
10
20
0 25 50 75 100 125 150
SETTLEMENT (mm)
CLAY
FILL
Soil
Pile
Neutral Plane
58
30
40
50
60
70
DEP
TH (
m) 30
40
50
60
70
DEP
TH (
m) SAND
CLAY
GRAVEL
SETTLEMENT OF THE PILED FOUNDATION
05
10152025
0 5 10 15 20 25 30 35
DAYSWA
TER
HEI
GH
T (m
)
HYDRO TEST
0
100
200
0 10 20 30 40 50
TIME (years)
T (m
m)
LONG-TERM SETTLEMENTS OF THE PILED TANK FOUNDATION
Only Fill;Away from the tank
Tank Perimeter
59
DAYSW
0
2040
6080
0 5 10 15 20 25 30 35
SETT
LEM
ENT
(mm
)
200
300
400
500
SETT
LEM
ENT
Tank Center
A recent modern application of a piled pad foundation is the foundations for the Rion-Antirion bridge piers (Pecker 2004). Another is the foundations of the piers supporting the Golden Ears Bridge in Vancouver, BC (Sampaco et al.; Naesgaard et al. 2012), pictured below.
60
11
Piled pad foundation piers supporting the Golden Ears Bridge in Vancouver, BC.
Bored piles (900 mm; 8 m) to provide lateral resistance
and
driven (300 mm; 30 m)piled-pad piles
BRIDGE DECK
FOOTING AND PILE CAP Sh t
61
over an about 100 m thick deposit of soft compressible clay
FOOTING AND PILE CAP
Longslenderpiles
Shortboredpile
Pad
ShinHo and MyeongJi Housing Project,in the estuary of the Nakdong River, Pusan, Korea
Project Managers: Drs. Song Gyo Chung and Sung Ryul Kim, Dong-A University, Busan
62
63 64
65
AIR VIEW(Shinho Site)
66
12
SITE PLAN (SH Site)
67
Silty clay
0
10
20
30
0 10 20 30Cone Stress, qt (MPa)
EPTH
(m
)
0
10
20
30
0 25 50 75 100
Sleeve Friction, fs (KPa)
PTH
(m
)
0
10
20
30
0 500 1,000 1,500
Pore Pressure (KPa)
EPTH
(m
)
0
10
20
30
0 1 2 3 4 5
Friction Ratio, fR (%)
EPTH
(m
)
Profile
FILL
Silty CLAY(marine)
68
40
50
60
DE
40
50
60
DE
40
50
60
DE
40
50
60
DE
SILT&CLAY
Very dense SAND
SAND
CPTU sounding at the location of the O-cell
0
10
20
0 10 20 30Cone Stress, qt (MPa)
EPTH
(m
)
0
10
20
0 200 400
Sleeve Friction (KPa)
PTH
(m
)
0
10
20
0 250 500 750 1,000
Pore Pressure (KPa)
PTH
(m
)
0
10
20
0 1 2 3 4 5
Friction Ratio (%)
EPTH
(m
)
Profile
Mixed
CLAY
6905-08-08 Myeongji Site C-block
30
40
50
DE
30
40
50
DE
30
40
50
DE
30
40
50
DE
SAN
Reduced pore pressure (“dilation”)
SAND
The pile considered is a 600 mm diameter cylinder pile with a 100 mm wall driven closed-toe
The questions to resolve in the design are
1 What is the capacity in the different layers?
70
1. What is the capacity in the different layers?
2. What is the depth to the force equilibrium/settlement equilibrium, i.e., the neutral plane
3. What will be the maximum load in the pile? Is the structural strength adequate?
4. What is the settlement of the pile as a function of the location of the neutral plane.
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000
LOAD and RESISTANCE (KN)
)
0
5
10
15
0 10 20 30 40 50 60 70 80
SETTLEMENT (mm)
)
SETTLEMENT OF PILE HEAD
PILE "CAPACITY"
DEAD LOAD
The Unified Method for Design of Piled Foundations(typical only)
7171
20
25
30
35
40
45
DEP
TH (
m)
20
25
30
35
40
45
DEP
TH (
m)
NEUTRAL PLANE
TOE MOVEMENT THAT MOBILIZES THE TOE
RESISTANCE
TOE RESISTANCE
DRAGLOAD
*) Portion of the toe resistance will have developed from the driving
*)
O-cell TestFor Pile Toe Response and to "Open up" the Pile Toe,i.e., remove pile Toe Resistance for the Subsequent Head-Down Test
20
25
30
T (m
m)
The pile head did not move
2nd Cycle
For Pile Toe Response and to "Open up" the Pile ToeSo that toe resistance is removed
72-90
-60
-30
0
30
0 1,000 2,000 3,000 4,000 5,000 6,000
LOAD (KN)
MO
VEM
ENT
(mm
)
Downward
Upward
1st Cycle
2nd Cycle
0
5
10
15
UPW
AR
D M
OVE
MEN
T did not move. A 16 mm pile compression of the shaft
for decreasing load is not realistic.
1st Cycle
Pile Toe Broke
So that toe resistance is removed from the subsequent head-down Test
13
0
10
20
-5,000 0 5,000
LOAD 2nd TOE-UP (KN)
(m)
0
10
20
-5,000 0 5,000
LOAD 1st TOE-UP (KN)
(m)
73
30
40
50
60
DEP
TH (30
40
50
60
DEP
TH ( 5 000
6,000
7,000
8,000
9,000
10,000
OW
N T
ESTS
(K
N)
Now The Head-down Test
74
0
1,000
2,000
3,000
4,000
5,000
0 10 20 30 40 50 60 70 80 90
MOVEMENT (mm)
LOA
D H
EAD
-DO
First Head-down Test
40
50
60
ANG
E O
F ST
RAIN
, Mt
SG-12 CD
SG-12 AB
SG 11
SG-10
SG-9
SG-8
The Shinho test pile — head-down test
75
0
10
20
30
0 500 1,000 1,500
STRAIN (µε)
CHA
NGE
OF
STRE
SS/C
HA(G
Pa)
Q =A(-0.0035(µε)2 + 29µε)
0
10
20
0 2,000 4,000 6,000 8,000 10,000
LOAD, 2nd HEAD-DOWN (KN)
)
ß = 1.0
ß = 0.4 (0.25
ß = 0.1(0 1)
RESIDUAL (maximum)
76
30
40
50
60
DEP
TH (
m
ZERO LINE IS AT START OF 2ND HEAD-DOWN TEST
(0.1)
ß = 0.7(0.2)
ß = 0.3 (0.1)
TRUE RESISTANCE (for maximum residual load)
After Unloading
The shaded force area corresponds to a shortening of just about 3 mm
PRESUMED RESIDUAL LOAD AT START OF O-CELL TEST
Estimated Residual Load Distribution at Start of the O-cell Test
0
10
20
30
0 2,000 4,000 6,000 8,000 10,000 12,000 14,000
RESISTANCE and LOAD (KN)
TH (
m)
Qd ≈RULT
77
40
50
60
DEP
T
Qn
Shinho Pile
Long-term resistance and load distributions at the Shinho site
This has been a long day with lots of material and a sometimes heavy message I am afraid I just hope that I
78
message, I am afraid. I just hope that I have not overloaded you.
14
79
The New International Airport, Bangkok Thailand
Data from: Fox, I., Due, M. and Buttling,S. (2004) and Buttling, S. (2006)
THAILAND
8080
Current and Future Pore Pressure Distribution
0
5
10
15
20
25
0 100 200 300 400 500
Pore Pressure (KPa)
pth
(m)
0
10
20
1975 1980 1985 1990 1995 2000 2005 2010
YEAR
er T
able
(m
)
DesignPhase
ConstructionPhase
Nearby Observations of Groundwater Table
81
30
35
40
45
50
Dep
Long-Term
Short-Term (Current)
30
40
50
60
70
Dep
th to
Gra
ounw
ate
Pumping (mining) of groundwater has reduced the pore pressures. At the start of the design process, pumping in the area was stopped.
82
The clay is soft and normally consolidated with a modulus number smaller than 10.
All foundations — the trellis roof, terminal buildings, concourse, walkways, etc. —are placed on piles. The stress-bulbs from the various foundations will overlap each other’s areas resulting in a complicated settlement analysis.
Several static loading tests on instrumented piles were performed to establish the load-transfer conditions at the site at the time of the testing, i.e., short-term conditions. Effective stress analysis of the test results for the current pore pressures established the coefficients applicable to the long-term conditions after water tables had stabilized.
83
A total of 25,000+ piles were installed.
The design employed the unified pile design method.
Example of resistance distribution for 600 mm diameter bored pile installed to a 30 m embedment depth.
0
10
LOAD (KN)
(m)
Qd = 1,040 KN RULT = 2,870 KNFs = 2.0
Short-Term
Fs = 2.0 on long-term capacity
0
10
LOAD (KN)
m)
Long-Term
Qd = 1,040 KN RULT = 2,160 KNFs = 2.0
84
The extensive testing and the conservative assumption on future pore pressures allowed an Fs of 2.0. The structural strength of the pile is more than adequate for the load at the neutral plane: Qd + Qn ≈ 1,500 KN.
20
30
DEPT
H (
Qn = 770 KN
20
30
DEPT
H (
m
Qn = 500 KN
Clay
Sand
15
The settlements for the piled foundations were calculated to:
Construction Long-term TotalTrellis Roof Pylons 20 mm 90 mm 110 mm
Terminal Building 30 15 45
85
Concourse 35 20 55
* * *
0
5
10
15
0 20 40 60 80 100
Soft Clay, compressible
Sand and Gravel
Highway Viaduct over
Railroad
(m
)
Milford, Beaver County, Utah
12.75-inch Diameter 0.5-inch Wall
Pipe Piles
Driven closed-toe to52 ft (16 m) embedment
86
20
25
30
35
40
Sand
DEP
TH To be concrete-filled
Load at SLS = 240 kips 1,068 KN
Required Factored Resistance = 540 kips
2,400 KN
0
5
10
15
0 20 40 60 80 100Cone Stress, qt (MPa)
(m)
0
5
10
15
0 500 1,000
Sleeve Friction, fs (KPa)
m)
0
5
10
15
0 500 1,000
Pore Pressure (KPa)
(m)
0
5
10
15
0 1 2 3 4 5
Friction Ratio, fR (%)
(m)
Profile
CPTU Sounding Results
87
20
25
30
35
40
DEP
TH
20
25
30
35
40
DE
PTH
(
20
25
30
35
40
DE
PTH
(
20
25
30
35
40
DE
PTH
(
Profile0
5
10
15
0 2 4 6 8 10Cone Stress, qt (MPa)
Enlarged Cone Stress Scale Soil Profiling Chart
88
15
20
25
30
35
40
DE
PTH
(m
)
“Correlation” CPT - SPT
0
5
10
0 50 100
Cone Stress, qt (MPa)
1.5
2.0
2.5
Blow
s)
Utah case
Florida case
8989
10
15
20
25
30
35
DEP
TH (
m)
0.0
0.5
1.0
0.00 0.50 1.00 1.50 2.00
Mean Particle Size (mm)
q c/N
(M
Pa/B
S A N DFine Medium Coarse
0
2
4
6
0 100 200 300 400
Unit Shaft Resistance (KPa)
m)
ß= 1.2
ß= 0.5
0
2
4
6
0.00 0.50 1.00 1.50 2.00
Equivalent ß (- - -)
m)
90
8
10
12
14
16
18
20
DE
PTH
(m
LCPC and Schmertmann
ß= 1.2
ß= 0.80
E-F andß-Method
8
10
12
14
16
18
20
DEP
TH (
m
16
0
2
4
6
8
0 500 1,000 1,500 2,000
Shaft Resistance (KN)
H (
m)
0
2
4
6
8
0 2,000 4,000 6,000Total Resistance (KN)
H (
m)
Required unfactored capacity
91
10
12
14
16
18
20
DEP
TH
E-F andß-Method
LCPC and Schmertmann
10
12
14
16
18
20
DEP
TH
LCPC E-F
0
5
10
15
0 10 20 30 40 50 60Cone Stress, qt (MPa)
(m)
qt filtered
qt
0
5
10
15
0 100 200 300 400 500 600 700 800Modulus Number, m)
(m)
LAB. TESTS, Oedometer
92
20
25
30
35
DEP
TH (
qt filtered and depth adjusted
20
25
30
35
DEP
TH
Filtered and unfiltered MODULUS NUMBER
0
5
10
15
0 100 200 300 400 500 600
Modulus Number, m Settlement (mm)
(m)
SETTLEMENT
93
20
25
30
35
DEP
TH (
MODULUS NUMBER
A repeat: Distribution of unit shaft shear and of load and resistance
SHAFT SHEAR LOAD
Settling Soil
Qd
9494
DEP
TH (–)
(+)
DEP
TH
Soil
Non-Settling
Soil
The shear force along the pile in a swelling soil is the opposite to that insettling soil, of course — "positive skin friction" as opposed to "negative skinfriction". But the same analysis method applies.
How would the distributions look for a pile in a swelling soil?
SHAFT SHEAR
Swelling Soil
NEGATIVE POSITIVE
PileLOAD
(+)
Qd
TENSION
0
95
Distributions of unit shaft shear
DEP
TH
(+)
(–)
Non-Swelling
Soil
A
and load for a pile in swelling soil
DEP
TH
(–)
(+)
COMPRESSION
BToe Load = 0
DEP
TH
Swelling Soil
Non-
(–)
Qd TENSION 0 COMPRESSION
Pile in swelling soil loaded in uplift Tension load
96
D
Swelling Soil, but Settling
Soil,perhaps
So, what does it mean that a pile loaded in tension in swelling soil can have a neutral plane in settling soil below the swelling soil?
17
Conventional piled foundations with floor supported on the piles or as a ground slab
Piled Raft and Piled Pad Foundations
9797
Piled raft foundation with loads supported by contact stress and piles
Remaining load on raft evenly distributed as contact stress
Evenly distributed load on the raft supported by evenly distributed piles (Fs = 1.0)
Uneven load on raft supported by the piles
(Fs = 1.0)
9898
Piled pad foundation with loads supported by contact stress and piles
Engineered Backfill
Conventional raft or mat Geotextile
9999
The exception to this is in the case of a piled raft, which is a term referring to a piled foundation designed with a factor of safety for the piles of close to unity, or better expressed: The neutral plane is designed to be located close to or at the
At the level of the pile cap, there is no contact stress between the underside of the pile cap and the soil, because the soil will always settle more than the pile cap. Therefore, it is incorrect to allow any contribution from contact stress.
Comments on Contact Stress, Piled Raft, and Piled Pad
100100
better expressed: The neutral plane is designed to be located close to or at the underside of the raft. Only if the external loads on the pile cap are equal to or larger than the combined pile capacities will there be a contact stress.
The emphasis of the design for a piled raft is on ensuring that the contact stress is uniformly distributed across the raft. The piled-raft design intends for the piles to serve both as soil reinforcing (stiffening) elements reducing settlements and as units for receiving unavoidable concentrated loads on the raft. This condition governs the distribution across the raft of the number and spacing of the piles.
