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Chapter 2
Review of literature
2.1 INTRODUCTION
Expansive soils are highly problematic as they have a tendency to increase in
volume on absorption of water and to shrink on evaporation of water. On absorption of
water the density of these soils decreases and they become slushy; they become hard on
evaporation of water due to increase in density. The volumetric change in these soils is
attributed to seasonal variations in the ground water profile resulting in changes in the
moisture content (Rees and Thomas, 1993).
The entire stratum of expansive soil in the field may not be active, as most soils
do not respond quickly to climatic changes. During summer, polygonal shrinkage cracks
appear on the surface, extending to a depth of about 2m, indicating a high potential for
expansion and contraction (Mohan, 1977). The depth of cracking indicates the depth of
active zone in which significant volume changes occur because of high moisture
deficiency (Snethen, 1980).
Katti (1978) observed that, in Indian expansive soils, the depth of active zone is
confined only to the top 1.0 to 1.2m. According to Mohan (1977), however, significant
ground movements occur even up to a depth of 3.5m below the ground level.
Ramaswamy (1990) observed that, in certain areas, the depth of active zone might extend
beyond 3.5m also.
2.2 IDENTIFICATION OF EXPANSIVE SOILS Expansive soils can be identified by their mineralogical composition, index
properties, suction movements or direct measurement of swelling characteristics.
2.2.1 MINERALOGICAL COMPOSITION
The amount of volume change by an expansive soil depends greatly on its
mineralogical composition. Expansive soils are usually composed of montmorillonite
clay minerals (Omari and Hamodi, 1991). The presence of montmorillonite, which has
4
an expanding lattice, contributes most to swelling, clearly indicating the expansiveness
of a clay soil (Shreiner, 1987).
The various techniques employed to determine the mineralogical composition are
X-ray diffraction, differential thermal analysis (DTA) and electron microscope resolution
(EMR). The X-ray diffraction technique gives the proportions of various minerals
present in a colloidal clay fraction. The DTA method is based on the fact that certain
characteristic reactions take place at specific temperatures for different minerals when
these minerals are heated to high temperatures, resulting in a loss or gain in heat. EMR
helps to determine the texture and internal structure of the mineral (Chen, 1988).
2.2.2 IDENTIFICATION FROM INDEX PROPERTIES
The degree of swelling of an expansive soil can be estimated through index
properties also. The Atterberg limits, clay content, activity etc. are used as parameters to
categorize the degree of swelling (Chen, 1988).
Skempton (1953) observed that volume change characteristics of a soil depends
on “activity”, which is defined as the ratio of the plasticity index of the soil to the
percentage clay particles i.e., fraction of soil finer than 2 micron (0.002mm).
Subsequently, this definition was modified for natural soils (Seed et al. 1962) as
5% - m2 finer thanclay %
Index µ
PlasticityActivity= (2.1)
A soil with an activity of less than 0.75 is termed as inactive i.e., it indicates a
low potential for volume change and a soil with an activity of more than 1.25 is termed
as active, indicating a high potential for volume change. The values lying between 0.75
and 1.25 indicate medium potential and soils with these values are classified as normal.
Altmeyer (1955) characterized the degree of expansion of a soil based on its
shrinkage limit and linear shrinkage as given in Table 2.1. Linear shrinkage is defined as
the ratio of decrease in linear dimension of a specimen to its initial dimension, expressed
as a percentage and given by the nearest whole number.
5
Table 2.1 Degree of expansion based on linear shrinkage
Shrinkage limit
(%)
Linear shrinkage
(%)
Probable swell
(%)
Degree of expansion
< 10 >8 >1.5 Critical
10 – 12 5 – 8 0.5 – 1.5 Marginal
> 12 < 5 < 0.5 Non – critical
Holtz and Gibbs (1956) developed a simple test called the free swell test for the
determination of swell potential of a soil. The test is performed by pouring 10cc of dry
soil passing 425micron sieve into a graduated cylindrical glass jar of 100ml capacity
filled with water and observing the swollen volume. Free swell index (FSI) is expressed
as a percentage increase in the volume to the original volume of the soil. Soils having a
free swell of more than 100% pose more damage to lightly loaded structures, and those
having an FSI of less than 50% do not pose serious problems to structures.
Mohan and Goel (1959) suggested a more convenient method, which was later
adopted by Indian Standards Institution. In this method, 10g of oven dried soil passing
425 micron sieve is poured separately into two graduated cylindrical glass jars of 100ml
capacity, one containing distilled water and the other kerosene. Allowing both the jars to
stand for 24 hours, the final volumes of the soil in the two cylinders are noted. The
differential free swell, which was later redesignated as free swell index (FSI), expressed
as a percentage, is given as,
(2.2) 100×
−=
kVks VVFSI
where, Vk = final volume of the soil in kerosene, which is a non-polar liquid, and
Vs = final volume of the soil in water which causes swelling of the soil.
According to Mohan and Goel (1959), a soil with an FSI of 50% or more has a
very high swell potential, and that with an FSI of 20% or less has a low swell potential.
Holtz (1959) proposed some identification criteria based on the probable volume change
in expansive soils, which is the percentage change in the thickness of the sample from
6
oedometer swell test under a surcharge of 6.9kPa or 1psi from the air-dry condition to
saturation. To calculate the probable expansion, the three properties of soil given in
Table 2.2, in which the classification is also given, should be considered simultaneously.
Table 2.2 Degree of expansion based on PI and SL
Colloidal
content %
Plasticity index
(%)
Shrinkage limit
(%)
Probable volume
change (%)
Degree of
expansion
>28 >35 <11 >30 Very high
20 – 31 25 – 41 7 – 12 20 – 30 High
13 – 23 15 – 28 10 – 16 10 – 20 Medium
5 < 18 > 15 < 10 Low
Seed et al. (1962) proposed a relationship between swell potential (S.P) and
plasticity index (PI) as,
S.P = 2.16 X10 -3 (PI) 2.44 (2.3)
where swell potential is the ratio of the increase in thickness to the original thickness of
the soil sample in a consolidation ring, compacted at optimum moisture content, soaked
in water under a surcharge of 6.9 kPa (1Psi). This is expressed as a percentage.
Seed et al (1962) gave an expression for swell potential (SP) in terms of activity (A) and
clay content (C) as
S.P = 3.6 X10-5 (A 2.44) (C 3.44) (2.4)
Seed et al (1962) gave another expression also for swell potential (S.P) in terms
of percent clay (C) i.e., soil fraction with size less than 2 micron as
S.P = k C x (2.5)
where ‘x’ is an exponent and ‘k’ is a coefficient, both depending upon the type of clay.
Ranganatham and Satyanarayana (1965) proposed a relationship between swell
potential (S.P) and shrinkage index (S.I) as
S.P= β (S.I)z (2.6)
β = 1/6.3 and z = 1.17 for natural soils and
β = 1/256 and z = 2.37 for artificially prepared sand – clay mixtures.
Based on this relationship, the degree of expansion is classified as shown in Table 2.3.
7
Table 2.3 Degree of expansion based on shrinkage index (Ranganatham and Satyanarayana, 1965) Shrinkage index (%) Swell potential (%) under a
surcharge of 6.9kPa
Degree of expansion
< 20 <10 Low
20 -30 10 – 20 Medium
30 – 60 20 – 30 High
> 60 > 30 Very high
Peck et al. (1974) have related plasticity index to the swelling potential, as shown
in Table 2.4.
Table 2.4 Characterisation of swell potential based on PI
Swelling potential Plasticity index (%)
Low 0 – 15
Medium 10 – 35
High 20 – 35
Very high 35 and above
Chen (1975) gave a relationship for the percentage swell (S) of undisturbed soils
in terms of plasticity index (P.I) in the form
S = B. e A (P.I) (2.7)
where A and B are constants equal to 0.0838 and 0.2558 respectively.
The surcharge pressure used was 6.9kPa. The water content varied between 15% and
20 % and the dry density between 1.6 Mg/cu.m and 1.76 Mg/cu.m.
Mohan (1977) categorized degree of expansion based on liquid limit, plasticity
index, shrinkage index and colloid content as shown in Table 2.5.
8
Table 2.5 Degree of expansion based on liquid limit, plasticity index, shrinkage index and colloid content (Mohan, 1977). Liquid limit
PI Shrinkage index
Colloid content
Degree of expansion
Degree of severity
70 -90 >32 >60 >28 Very high Very critical
50 – 70 23 – 32 30 – 60 20 - 32 High Critical
35 – 50 12 – 23 15 – 30 13 - 23 Medium Marginal
20 – 35 12 <15 <15 Low Non- critical
Shrinkage index in the above table is defined as the difference between liquid
limit and shrinkage limit, while colloid content represents the soil fraction finer than
0.001mm.
Murthy and Raman (1977) defined a term “swelling index” (Isw) expressed as a
fraction and given in the form
Isw = SI/(100+ ws) (2.8)
where ‘SI’ is the shrinkage index (%) and ws is the shrinkage limit (%). Based on
shrinkage index, the degree of expansiveness is classified as shown in Table 2.6.
Table 2.6 Characterization of degree of expansiveness based on swelling index
Swelling index (Isw) Degree of expansiveness
0 -0.1 Low
0.1 – 0.2 Medium
0.2 – 0.5 High
0.5 – 0.75 Very high
0.75 – 1.0 Extra high
Williams and Donaldson (1980) classified the degree of expansiveness
considering the plasticity index of the whole sample, PIws, which is given as
sample whole theof mass
sieve same thefiner than soil of mass 425 through passingfraction soil of P.I. P.Iws ×= µ (2.9)
9
The classification is shown in Table 2.7.
Table 2.7 Degree of expansiveness based on plasticity index of whole sample
(Williams and Donaldson, 1980)
P.I of whole sample Degree of expansiveness
>32 Very high
24 – 32 High
12 – 24 Medium
<12 Low
Bandyopadhyay (1981) proposed an equation for swell potential (S) in terms of
colloidal content (C) as,
S =k’ C x (2.10)
where k’ = a.Ab
in which ‘A’ is the activity of the soil and ‘x’ is an exponent equal to 3.44. The values of
‘a’ and ‘b’ depend upon the plasticity index of the soil. The average values of ‘a’ and ‘b’
applicable to all soils are, a = 2.2x10-5 and b = 2.14
The Indian Standard Institute, now called the Bureau of Indian Standards, has
adopted the classification criteria suggested by Mohan (1977) with a modification
according to which the colloidal content is replaced with the free swell index, as shown
in Table 2.8.
Table 2.8 Characterisation of degree of severity
Liquid limit (%)
Plasticity index (%)
Shrinkage index (%)
Free swell index (%)
Degree of expansion
Degree of severity
70 – 90 >32 >60 >200 Very high Severe
50 – 70 23 – 32 30 – 60 100 – 200 High Critical
35 – 50 12 – 23 15 – 30 50 – 100 Medium Marginal
20 – 35 <12 <15 <50 Low Non-critical
10
2.2.3 SOIL SUCTION
Soil suction which indicates the affinity of soil for water is one of the intrinsic
properties that characterizes an expansive soil and indicates the affinity of soil with water
(Snethen, 1979). It is a measure of the deficiency of moisture in a soil and its capacity to
attract water into its pores.
The total soil suction is a sum of the matrix suction or capillary potential and the
osmotic suction. The capillary potential is equivalent to the negative pore pressure and is
often identified in terms of suction in an unsaturated soil. Volumetric changes respond to
changes in capillary potential (Reese and Thomas, 1993).
Because of this negative pore pressure, expansive soils have remarkable affinity
for water. Hence, soil suction is a fundamental parameter to evaluate wetting (Kenneth et
al. 1993). However, in soils compacted at a water content below plastic limit, suction has
no significant effect on the rate of advance of wetting front. At a water content higher
than plastic limit, suction has a control over the rate of wetting (Weisberg et al. 1990).
Several researchers (Richards, 1967; Aitchison, 1973; Brackley, 1980; Johnson,
1981; Mc Keen, 1981; Snethen, 1980; Richards et al. 1983; Yoshida et al. 1983)
evaluated swell based on the relationships between suction and water content. Swelling
is linearly proportional to the gravimetric water content (Dhowian, 1990). Dhowian
(1992) developed suction potential model to estimate suction as a function of time,
swelling and depth.
Suction can be measured by the filter-paper method (Mc Keen, 1988), in which
the soil sample placed in the presence of a filter paper is allowed to equilibrate and the
water content of the filter paper measured. The suction of the soil sample can be obtained
from a calibration curve for the filter paper. Suction depends on the voids ratio in the
range of low values of suction (Keissar et al. 1990).
2.2.4 PREDICTION OF SWELLING CHARACTERISTICS BASED ON
PLACEMENT CONDITIONS
Placement conditions, which include surcharge, water content and dry density of
the soil, have bearing on the percentage swell and the swelling pressure. The effect of
different placement conditions is explained below:
11
2.2.4.1 Surcharge load and stress history: The increase in surcharge load on an
expansive soil reduces swell. For in situ conditions, any form of overburden on the soil
will reduce swell. Swelling pressure measured depends on the initial surcharge applied
on the specimen (Satyanarayana, 1966). But, according to Chen (1988), it is independent
of the surcharge.
2.2.4.2 Initial dry density and water content: Both swell potential and swelling
pressure increase with increase in dry density (Ranganatham and Satyanarayana, 1965).
According to Satyanarayana, (1966) and Vijayvergiya and Ghazzaly (1973), the initial
moisture content affects swelling pressure whereas, according to Chen (1988) and
Brackley (1973), it does not. The lower the initial water content, the higher will be the
suction and the consequent volume change.
2.2.4.3 Method of compaction: Statically compacted specimens swell more than those
compacted by kneading (Parcher and Liu, 1965). Some more factors that influence the
swelling characteristics are side friction (EI Sayed and Rabbaa, 1986), size of the
specimen (Uppal and Palit, 1969; EI Sayed and Rabbaa, 1986) and deformability of
oedometer (Fredlund, 1982; EI Sohby et al. 1989).
2.2.5 DIRECT MEASUREMENT OF PERCENT SWELL
The most convenient method of determining the swelling characteristics of
expansive clays is by direct measurement in the laboratory.
For the determination of percent swell, the soil specimen is sandwiched between two
porous stones and confined in an oedometer ring. The sample is inundated with water.
Water enters both from the top and the bottom. The final increase in the thickness of the
sample following absorption of water by the soil specimen is recorded and reported as a
percent of the original thickness.
2.2.6 DIRECT MEASUREMENT OF SWELLING PRESSURE
Swelling pressure of an expansive soil is determined by the following three methods in
the laboratory:
i) the method of different surcharges,
ii) the constant volume method and
12
iii) the free swell method ( or, the swell – consolidation method).
Based on the method of determination, swelling pressure is defined (Jennings, 1963) as,
i) the pressure corresponding to zero volume change obtained by interpolation
from the equilibrium swell or compression values of identical specimens
inundated under different surcharges (known as the method of different
surcharges). The swelling pressure thus determined is designated by the
symbol, psds;
ii) the pressure that is necessary to hold the soil sample at its original volume
when it is inundated under no – load condition, or under a surcharge equal to
the field effective overburden pressure (known as the constant volume
method). The swelling pressure thus determined is designated by the symbol,
pscv and
iii) the pressure that is required to recompress a fully swollen soil sample (on
inundation) to its original volume under a nominal surcharge of about 10kPa
(known as the free swell method). The swelling pressure thus determined is
designated by the symbol psfs.
Of these, the free swell method has been found to give the largest value of swell
(Shanker et al. 1982; Sridharan et al. 1986). However, while the former investigators
found the constant volume method to yield the least value, the latter found the method of
different surcharges to yield the least value.
Sullivan and Mc Clelland (1969) and Uppal and Palit (1969) recommend the use of
the constant volume method as it simulates the stress path followed in the field.
However, Holtz and Gibbs (1956) maintain that the entry of water into the sample under
load is difficult as the fissures would be closed. This affects the determined value of the
swelling pressure. Zeitlen (1961) and Justo and Saetersdal (1979) recommend the use of
different surcharges method as it gives the swell expected under any initial surcharge.
Jennings (1965), however, observed that the stress path is not an important factor
in the determination of the percent swell, even though Brackley (1975) and Sullivan and
Mc Clelland (1969) emphasize its importance.
13
2.3. EXISTING FOUNDATION PRACTICES IN EXPANSIVE SOILS The various options in foundation practices adopted to minimize heave in
expansive soils are,
a) avoiding expansive material,
b) mechanical, physical or chemical alteration and
c) adopting special foundation techniques.
2.3.1 AVOIDING EXPANSIVE MATERIAL
Avoiding the expansive soil in favour of a safer foundation soil is not an
economically viable proposition in most of the situations. With the development of
several modern techniques for effectively combating problems posed by expansive soils,
it is seldom adopted these days.
2.3.2. ALTERATIONS FOR MINIMIZING HEAVE
2.3.2.1. MECHANICAL ALTERATIONS
These include excavation of expansive soil and replacement with non-expansive
material, where the depth of active zone is small and where a suitable replacement
material is available. Sand cushion method (Satyanarayana, 1966) and cohesive non-
swelling (CNS) layer method (Katti, 1978) are very popular.
2.3.2.1.1. Sand cushion method: In this method, the entire depth of the expansive clay
stratum if it is thin, or a part thereof, if it is deep enough, is removed and replaced by a
sand cushion compacted to the desired density and thickness. Swelling pressure varies
inversely as the thickness of the sand layer and directly as its density. Hence, sand
cushions are formed in their loosest state without violating the criterion of bearing
capacity. Fig 2.1 shows the concept of sand cushion technique.
Expansive clay
Expansive clay
Sand cushion
Fig 2.1 Sand cushion method
14
The basic philosophy of this method is that, in monsoon, the saturated sand occupies less
volume, accommodating some of the heave of underlying soil, and in summer, partially
saturated sand bulks and occupies the extra space left by the shrinkage of the soil.
This method, however, has some limitations. Firstly, the high permeability of sand
creates accumulation of water, and secondly, the thickness of the sand cushion depends
on the depth of active zone termine. Besides, effects of
tage with this system is, it is not easy to get a material which conforms to the
specifica
pansive soil were arbitrarily
and 150-75 µm,
sed on the grain size. Swell potential and
swelling p
increasing fines content. Coefficient of
which itself is difficult to de
fatigue in swelling have to be considered in designing the thickness of the sand cushion.
2.3.2.1.2. Cohesive non- swelling (CNS) layer method: In this method, about top 1m
to 1.2m of the expansive soil is removed and replaced by a cohesive non-swelling soil
layer.
According to Katti, with saturation of expansive soil, cohesive forces are
developed up to a depth of about 1.0m – 1.2m and counteract heave. The electric charge
of clay particles produces absorbed water bonds and develops this cohesion. A CNS
layer creates an environment similar to that around 1m depth in an expansive soil with
equivalent cohesion to counteract heave. Moorum is an example of CNS material. The
disadvan
tions laid down by Katti for an ideal CNS material.
2.3.2.2. PHYSICAL ALTERATION
In this, granular material is mixed with expansive clay to minimize swelling
properties (Satyanarayana, 1966). Phanikumar et al. (2012) studied swell-consolidation
characteristics of these artificially prepared sand-clay mixes in one dimensional
consolidometer. Fine sand content and fines content in the ex
varied in the investigation. The fines content was varied as 425-300 µm
separated from the same expansive soil ba
ressure decreased with increasing fine sand content but increased with
compressibility, coefficient of volume
compressibility and compression index of the samples decreased initially up to a sand
content of 15% and thereafter increased at higher sand contents. One of the
disadvantages of clay-sand and clay-gravel mixes is the faster ingress of water due to
increased permeability.
15
2.3.2.3. CHEMICAL ALTERATION
This involves addition of chemicals to expansive clay to reduce heave by altering
the nature of clay minerals. Stabilisation of expansive soil with various additives
including lime, cement, calcium chloride and fly ash has shown a promising reduction in
heave and improved strength characteristics. (Shanker and Maruthi, 1989; Desai and
Oza, 1997; Phanikumar et al. 2001; Cokca, 2001).
cur when fly ash is blended with expansive soils
(Cokca
would decrease causing reduction in
free sw
and an increase in maximum dry density
and she
Two mechanisms are likely to oc
, 2001; Phanikumar and Sharma, 2007). They are (i) physico-chemical
interactions and (ii) mechanical changes. Expansive clay particles are replaced by the
non plastic fines of fly ash. Fly ash, which is composed of silica, alumina and iron
oxides, promotes flocculation of clay particles by cation exchange. The surface area of
the flocculated particles and their affinity for water
ell index, swell potential and swelling pressure. The reduction in swelling
pressure on addition of fly ash is due to the reduction in the amount of suction (Sharma,
1998). Swell potential and swelling pressure decreased by nearly 50% at 20% fly ash
content (Phanikumar and Sharma, 2004). It was also observed that swell pressure
reduces with increase in the curing time (Cokca, 1999). Further, Phanikumar et al. (2009)
developed the technique of compacted fly ash columns (FAC) as an innovative
foundation technique for expansive clay beds.
Phanikumar (2009) compared the effect of lime and fly ash on free swell index,
swell potential, swelling pressure, co-efficient of consolidation, compression index,
secondary consolidation characteristics and shear strength. Lime content was varied as
2%, 4% and 6% whereas fly ash content was varied as 0%, 10% and 20%. It was noted
that at 20% fly ash there was a significant reduction in swell potential, swelling pressure,
compression index and secondary consolidation
ar strength. A similar pattern was observed when the soil was stabilized with 4%
lime.
A technique called lime-slurry pressure injection (LSPI) is used wherein lime
slurry is injected into drill- holes under a pressure of 15kg/cm2. Lime columns or lime-
soil columns were also tried to stabilize expansive clays in-situ (Rao, 1984). It has been
16
reported (Venkataratnam et al. 1985; Shanker and Maruthi, 1989) that diffusion of lime
is effective up to a radial distance of about 3 times the diameter of the lime soil column.
Table 2
ansive soil (Rao et al. 2008). In this technique fly ash cushion was stabilized
with 1
% of lime.
ent of 8%.
e in the thickness of the
cushion
.9 shows the summary of the data on stabilizing expansive soils with different
chemicals.
Lime-stabilized fly ash cushion ( Rao et al. 2008), stabilized with 10% lime and
with thickness equal to half that of the active zone, found that heave reduced by about
60% and further reduced to 99.2% in the subsequent swelling and shrinkage cycles.
Cement stabilized fly ash cushion technique was also found to be effective in controlling
heave of exp
0% cement with the thickness equal to that of the expansive clay bed. 75% of
heave was reduced when compared with CNS cushion. Heave reduced to 99.1% at the
end of the fourth swell-shrink cycle.
Effect of lime, cement and artificial pozzolan and the combination of these three
stabilizers at different proportions were studied (Al Rawas et al. 2005). Lime, cement
and sarooj (artificial pozzolan produced by burning calcining clay) were mixed at
dosages of 3%, 6% and 9% by dry weight of soil. It was observed that swell percentage
and swell pressure reduced to zero at 6
Sahoo and Pradhan (2010) studied the effect of lime stabilized soil cushion on
strength behaviour of expansive soil. In this study, lime contents of 2%, 4%, 6%, 8% and
10% by dry weight of cohesive non swelling soil were used. Tests were conducted at
different soaking periods of 7, 14, 28 and 56 days. Maximum increase in strength was
observed at 14 days of curing period at a lime cont
Rice husk ash stabilized with lime or cement was used as cushion between the
expansive soil and the foundation to counteract the effect of heaving (Sivapullaiah et al.
2004; Sharma et al. 2008). Rice husk ash was stabilized with a lime content of 3% to
9% or 10% of cement and cured for a period of 7 days to effectively function as a
cushion. It was observed that heave decreased with increas
. Lime stabilized cushion was found to be more effective when compared to rice
husk stabilized with cement.
17
Table 2.9 Summary of data on stabilizing expansive soils with different chemicals
Type of stabilization
Properties of stabilizers
% of stabilizer
Tests conducted Conclusions References
Lime
Hydrated
lime
Lime-slurry
2% -8%
2500 lbs of
lime:900
gallons of
water
Consistency limits, Proctor
compaction, free swell and
shear strength tests
Lime-slurry pressure
injection (LSPI)
Liquid limit, plasticity index and swell
potential of expansive clays decrease
with increase in lime content up to 4%.
Plastic limit and shear strength increase.
MDD increases and OMC decreases.
Diffusion of lime is effective up to a
radial distance of 3 times the diameter
of the lime soil column
Abduljawad
(1995)
Basma et al.
(1998)
Rao (1984)
18
Fly ash
Class C fly
ash
Class C fly
ash
Class C fly
ash
Class C fly
ash
0%, 3%,
5%, 8%,
10%,15%,
20% and
25%
15% and
25%
0%, 5%,
10%, 15%
and 20%
0%, 5%,
10%, 15%
and 20%
Liquid limit, plastic limit
and swell pressure
Plasticity index, swell
potential and cation
exchange capacity(CEC)
Free swell index (FSI),
swell potential, swelling
pressure, compression
index (Cc) and secondary
consolidation
Plasticity index, hydraulic
conductivity, compaction
and undrained cohesion
Grain size distribution of the soils is
altered by the addition of fly ash.
Liquid limit and plasticity index
decreased with increasing fly ash
content.
CEC reduces due to the formation of
new pazzolanic reactions causes the soil
to become more granular and results in
less water absorption potential.
20% of fly ash content reduced FSI,
swelling pressure and swell potential by
about 50%. Cc and secondary
consolidation decreased by 40% at 20%
fly ash.
Plasticity index reduced by 50% at 20%
fly ash.Hydraulic conductivity and
OMC decreased whereas MDD and
undrained cohesion increased.
Cokca (1999)
Cokca (2001)
Phanikumar and
Sharma (2007)
Phanikumar and
Sharma (2004)
19
Class F fly
ash
0%, 3%,
6%, 9%,
12% and
15%
Swell potential, swelling
pressure and unconfined
compressive strength
(UCS)
Swell potential and swelling pressure
decreased with increase in curing time.
UCS increased with increase in curing
time.
Optimum content of fly ash was found
to be between 9% and 12% at a curing
time of 7 days.
Fesheng Zha et
al. (2008)
Cement Portland
cement
Cement kiln
dust
3% -9%
0%-30%
Consistency limits, swell
potential and shear strength
Plasticity index, free swell
and unconfined
compressive strength
Reduces liquid limit, plasticity index
and swelling potential, and increases the
shrinkage limit and shear strength
Reduces plasticity index and swell %
and increases compressive strength
Nelson and
Miller (1992)
Basma et al.
(1998)
Zaman et al.
(1992)
Calcium
chloride
Powder form 1% Plasticity index and swell
pressure
Plasticity index and swelling pressure
decreased with increasing CaCl2
content.
Ramanamurty
and Harikrishna
(2006)
20
Expansive soil can also be stabilized with calcium chloride also (Ramanamurthy
and Harikrishna, 2006). One percent of calcium chloride was applied to the expansive
clay bed by ponding and also through boreholes, and plasticity of the clay reduced to
nearly 40% - 60% whereas the plasticity reduced to only 7% -15% when lime slurry was
used. Similarly the swell pressure of the clay decreased by 50% – 60% and 20% - 25%
respectively in cases of CaCl2 application and lime application. It was observed that
calcium chloride was several times better than conventional lime.
2.3.2.4. FIBER REINFORCEMENT OF EXPANSIVE SOILS
Fiber reinforcement of expansive soils also proved to be quite successful in
reducing volume changes and increasing shear strength of expansive soils. Puppala and
Musenda (2000) investigated the effect of discrete and randomly oriented polypropylene
fibre reinforcement on strength and volume change behaviour of expansive soils. Fiber
reinforced clayey samples were prepared by varying fiber percentage as 0%, 0.3%, 0.6%
and 0.9% by dry weight of soil for the both types of fibers and the samples were tested
for unconfined compressive strength, volumetric shrinkage, free swell and swelling
pressure. Test results indicated that fiber reinforcement enhanced strength and reduced
volumetric shrinkage and swelling pressure. Fiber reinforcement also decreased swell
potential of the soil.
Investigations were also carried out with nylon fibers and natural fibers having
different aspect ratios (Al-Akhras et al. 2008) to study the influence of fibers on swelling
properties of clayey soils. Four aspects ratios (l/d) of 25, 50, 75 and 100 and five
different fiber contents of 1%, 2%, 3%, 4% and 5% were used in the study. Results
revealed that the swelling pressure and swell potential reduced with increase in the fiber
content significantly. It was also observed that natural fiber is more efficacious than
nylon fibers. Further, a lower aspect ratio appeared to have a greater effect in reducing
swelling pressure in both the types of fibers.
Viswanatham et al. (2009) studied the effect of polypropylene type fibers on
swelling behaviour of expansive soils. One-dimensional swell tests were conducted on
remoulded expansive soils without reinforcement and with reinforcement in the
percentage range of 0.25% and 0.5% and at fiber lengths of 30mm, 60mm and 90mm.
21
22
Digital imaging technique as well as dial gauge was used to measure the heave of
reinforced and un-reinforced expansive clay beds. Reduction in heave was proportional
to fiber content and maximum heave was observed at low aspect ratio at both the fiber
contents of 0.25% and 0.5%. It was also observed that the length of the fiber was the key
factor that influenced the reinforcing effect of fiber. Discrete and randomly distributed
fibers were found efficacious in reducing heave.
Investigation on mixture of fiber and fly ash was also carried out (Puppala, 2001)
to stabilize expansive soils. This technique was also found effective in reducing plasticity
and free swell characteristics. Kumar et al. (2007) studied the influence of fly ash, lime
and polyester fibers on compaction and strength properties of expansive soils. Different
percentages of randomly oriented fibers were introduced in mixes such as soil- lime,
soil-fly ash, and soil-lime-fly ash and compaction tests, unconfined compression tests
and tensile strength tests were conducted on specimens cured for 7, 14 and 28 days. It
was observed that strength increased with increase in curing period. Table 2.10 shows
the summary of test results.
23
Type of fiber
% of fiber Types of tests conducted on expansive soils
Conclusions References
Polypropylene
fiber
0%, 0.3%, 0.6% and
0.9%
0.25% and 0.5%
aspect ratio: 15, 30
and 45
Unconfined compressive
strength, volumetric
shrinkage, free swell and
swelling pressure
1-D swell test
Unconfined compressive strength
increased whereas volumetric shrinkage
and swelling pressure decreased
Reduction in heave was proportional to
fiber content;
Discrete and randomly distributed fibers
were found efficacious in reducing heave
Puppala and
Musenda (2000)
Viswanatham et al.
(2009)
Nylon fiber
1%, 2%, 3%, 4% and
5%
aspect ratio: 25, 50,
Swelling pressure and
swell potential
Swelling pressure and swell potential
decreased with increase in fiber content;
Low aspect ratio has greater effect in
Al-Akhras et al.
(2008)
Table 2.10 Summary of data on fiber reinforcement of expansive soils
75 and 100
0%, 1% and 2%
aspect ratio: 15
Heave test
reducing swelling pressure
Heave decreased as the number of
wetting-drying cycles increased
Heave decreased in both fiber-reinforced
and unreinforced conditions.
Phanikumar et al.
(2011)
Polyester fiber
0.5%, 1.0%, 1.5%
and 2%
Fiber length : 3mm,
6mm and 12mm
Compaction , unconfined
compressive strength,
tensile strength
Split tensile strength and unconfined
compressive strength increased with
increase in fiber content.
Kumar et al. (2007)
24
2.3.3. ADOPTING SPECIAL FOUNDATION TECHNIQUES
Apart from the above mentioned physical alteration and chemical alteration
techniques, there are some special foundation techniques adopted in expansive soils.
They are tension- resistant foundations, devised to absorb the tensile force generated in
expansive clay beds during swelling. Some of the special foundation techniques
employed to counteract heave in expansive soils are explained below:
2.3.3.1 Drilled piers
Piers of small diameter but of considerable length are drilled (Fig 2.2) into
expansive soils up to a zone where there is no effect of moisture change (Chen, 1988).
These are straight-shafted piers with a uniform diameter. Both the end bearing resistance,
E (psf) and friction resistance of drilled piers increase with depth at the rate of about 10%
per meter. Therefore, the load carrying capacity, Q (kips) of a drilled pier is given (Chen,
1988) as,
Q = [ A’ (E+0.03 EL’) + 0.1(E+0.03 EL) C’L’] (2.11)
where A’ = Area of cross-section of the pier (sq.ft.) and
L’ = Depth of penetration into the bearing stratum (ft.) and
C’ = Perimeter of the pier (ft.).
The uplift force, U (lb) is given by,
U = 2πr’fu(D’-d’) (2.12)
where r’ = radius of the pier (ft.),
d’ = depth of the zone of soil unaffected by wetting (ft.),
D’ = total length of the pier (ft.),
U = Swelling pressure (psf) and
fu = coefficient of uplift between concrete and soil
The resisting force comprises of the dead load of the pier and the skin friction and is
given by
25
W= (π r’2p + 2πr’s’d’) (2.13)
where p = Unit dead load pressure (psf),
s’ = skin friction surrounding the pier (psf),
w = total withholding force (lb),
d’ = depth of the zone of soil unaffected by wetting (ft) and
r’ = radius of the pier (ft).
A rational pier formula is obtained by equating the uplift force and the resisting force.
Therefore,
2πr’fu(D’-d’) = (πr’2p+2πr’s’d’)
or, p = 2/r’[fu(D’-d’) –s’d’] (2.14)
Reinforcement for tension Grade beam
Dead load
Air space beneath Grade beam
Uplift pressureD-d Zone of wetting
D
Zone of soil unaffected by wetting
d
Fig. 2.2 Drilled piers
26
2.3.3.2. Belled piers
Belled piers have an enlarged bell-shaped base at the end of a straight pier shaft
for increasing the load carrying capacity. Such piers such as having enlarged diameter at
the bottom are referred to as belled piers (Fig 2.3). The swelling soil tends to grip the
pier shafts and lift them. The uplift force ‘U’ exerted on the belled pier is given as
U = ( P + Fw + Fs) (2.15)
where P = total vertical pressure on the pier,
Fw = total weight of the soil above the bell,
and Fs = total shearing resistance along the circular failure surface.
The most prominent disadvantage of belled piers is the cost and the difficulty of
inspection.
2.3.3.3. Friction piers
Friction piers can be used where the bedrock is very deep and the upper layers of
the soil are expansive (Chen, 1988).
Reinforcement for tensionGrade beam
Dead load
Air space beneath Grade beam
Uplift pressure Zone of wetting (D-d)
D
Zone of soil unaffected by wetting (d)
Assumed circular plane of failure
Fig. 2.3 Belled piers
27
The ult
(2.16)
ned from field load tests.
ould not be encountered within the length of the pier.
India) and designed for Indian conditions. Under-reamed
piles a
ansfer the load of a building
through the piles to a depth beyond the zone of seasonal moisture variation or active
zone.
inactive zone
Fig 2.4 Under-reamed pile foundation
imate shaft resistance (S) for a homogeneous clay stratum is given by
S = πr’2α’Su
where r = radius of the pier,
Su = undrained shear strength, and
Α’ = a reduction factor, obtai
The disadvantage is that soft soil sh
2.3.3.4. Under- reamed piles
Under-reamed piles are similar to that of belled piers. Under-reamed piles were
developed by CBRI, Roorkee (
re most effective in isolating the structure from foundation soil. Fig (2.4) shows
the schematic diagram of an under-reamed pile. They have an enlarged bottom anchored
in the inactive zone as shown. This resists the uplift force on the foundation. Sometimes
double under-reams are also cast.
Grade beam
under-ream with enlarged bottom
active zone of expansive clay bed
Under-reamed piles
The principle involved in under-reamed piles is to tr
Under-reamed piles are usually bored cast in situ piles bored down to at least
0.6m into the inactive zone with the lower portions reamed in the form of a bulb. The
28
spacing of the piles may range from 1.5m to 3.0m. The piles are connected at their top by
plinth beams which support the super structure.
Some of the techniques discussed above have some limitations and some
techniq
.4. GRANULAR PILES
.4.1. INTRODUCTION
ed as a ground improvement technique to improve soft clays
and loo
.4.2. FAILURE MODE
weak soil when loaded vertically on top may fail with zero
lateral
greater than critical length (Madhav et al. 1994).
ues are not economical for lightly loaded structures. Hence, a simple technique in
the form of a granular pile anchor (GPA) was suggested as a new foundation technique
for expansive soils (Phanikumar, 1997; Phanikumar and Rao, 2000; Phanikumar et al.
2004; Rao et al. 2007; Rao et al. 2008) is explained in the following section.
2
2
Granular piles are us
se cohesionless deposits (Thornburn and Mc Vicar, 1968; Greenwood, 1970;
Baumann and Bauer, 1974; Hughes and Withers, 1974; Rathegeb and Kutzner, 1975;
Hughes et al. 1975; Priebe, 1976; Balaam et al. 1977; Aboshi et al. 1979; Goughnour
and Bayuk, 1979; Rao and Bhandari, 1980; Datye and Nagaraju, 1981; Rao, 1982;
Ranjan and Rao, 1983; Rao and Ranjan, 1985; Rao, 1986; Ranjan, 1989). The response
of granular pile reinforced ground under load is assessed in terms of its bearing capacity
and settlement.
2
A granular pile in a
strain (Madhav and Vitkar, 1978), by bulging (Hughes and Withers, 1974; Rao
and Bhandari, 1980; Datye and Nagaraju, 1981; Rao, 1982; Ranjan and Rao, 1983; Rao
and Ranjan, 1988), by general shear failure (Greenwood, 1970; Madhav and Vitkar,
1978) and by sliding (Aboshi et al. 1979). An end- bearing granular pile of length greater
than 3 or 4 diameters fails in bulging (Hughes and Withers, 1974). The length of the
bulge is limited to 4-5 pile diameters. A very short end-bearing granular pile undergoes
either a general or local shear failure at the surface. For the soil conditions encountered
in practice, bulging is the controlling failure mechanism of most granular piles of length
29
Failure in the case of a group of granular piles is by a combination of bulging and
local shear failure. Groups with short column lengths fail in end bearing. Spacing of piles
in a gro
The ultimate bearing capacity of the ground treated with granular piles is
elow:
2.4.3.1
e pressure of which resists these
b
pressure of the granular material.
qual to the depth of the footing from the ground level
up depends on the degree of improvement and settlement tolerance for the design
load (Watt et al. 1967; Engelhardt and Kirsch, 1977). If granular piles are closely spaced
(sp < 5d), settlements are significantly reduced (Balaam et al. 1977) and failure avoided.
Bulging of granular piles can be avoided by replacing the bulged portion of the pile by
concrete plugs or cement grout which is called skirting (Engelhardt and Kirsch, 1977;
Floss, 1979). Provision of all round surcharge is also suggested to increase passive
resistance and consequently the load bearing capacity. Recent research (Alamgir, 1989;
Adayat and Hanna, 1991; Madhav et al. 1994) has shown that granular piles either
reinforced by geogrids or enveloped by a geomembrane perform better as bulging is
prevented and ultimate capacity is increased. The effectiveness of granular piles will be
better if they are combined with other soil-replacement techniques (Bergado et al. 1991).
2.4.3. ULTIMATE BEARING CAPACITY
determined by different approaches explained b
. Passive pressure approach: In this approach, the pile material dilates and exerts
lateral stresses on the surrounding clay, the passiv
stresses (Greenwood, 1970). Hence, the technique of granular piles is feasible only in
soils of undrained strength more than 15 kPa (Barksdale and Bachus, 1983; Juran and
Guermazi, 1988). The ultimate bearing capacity (qult) of a single granular pile is equal to
the ultimate lateral strength of the surrounding soil (Hughes and Withers, 1974).
Therefore (2.17)
pupbpult KcKZPq 2+== γ
where, γ is the bulk density of clay,
Z is the total depth of bulge and
Kp is the coefficient of passive earth
The total depth of bulge (Zb) is e
plus the depth of the bulge of the pile.
30
2.4.3.2. General shear failure approach: Madhav and Vitkar (1978) considered the
plane strain version of a granular pile as a granular trench and proposed the failure
(2.18)
(2.19)
and
Nc1, Nc2 ensionless factors, depending on the properties
ch material. c1 and γ1 are cohesion and density of the trench material, and c2 and
ssible, rigid plastic column contained in a semi-infinite, rigid, plastic
ter test as,
p1 is the lateral limit stress in a pressuremeter.
Unit-cell approach: Priebe (1976) and Goughnour and Bayuk (1979) considered
ar pile is
ribution of
(2.22)
mechanism. The theory is developed for a c-φ soil. The ultimate capacity (qult) is given
as
where
qfcult NDBNNcq 222 21 γγ γ ++=
212
1cCc NN
ccN +=
211
γγγγ NNN +=
2
(2.20)
of soil , Nγ1, Nγ2 and Nq are dim
and tren
γ2, those of soil.
2.4.3.3. Pressuremeter approach: In this approach, the granular pile is considered as a
single, incompre
soft soil (Schlosser and Juran, 1983). The lateral limit stress, σr, is found from triaxial
compression test as,
σr = 2cu +σs (2.21)
or from a pressureme
σr = p1
where σs is the normal stress and
2.4.3.4
a single granular pile and its surrounding tributary soil as a unit cell. The granul
assumed to be rigid, plastic and incompressible and the soft soil to be elastic.
As the vertical settlement of the granular pile and the soil is approximately the same
(Vautrain, 1977), stress concentration occurs in the granular pile. The dist
vertical stress within a unit cell is expressed by a stress concentration factor, n, as
s
p
n =
31
where qp and qs are stresses in the pile and the soil respectively.
For equilibrium, ) at a given depth of the unit cell is given as,
(2.23)
s = (Area of the pile) / (Area of the unit cell)
(2.24)
(2.25)
pproxima itely long cylinder expanding into the soil, and
(2.26)
,
(2.27)
the average stress (q
q = qp as+ qs(1-as)
where ‘as’ is area replacement ratio given as,
a
The stresses in the pile and soil are given as,
s
p anq
)1(1 −+=
qn .
and
ss an
qq)1(1 −+
=
2.4.3.5. Cavity expansion approach: The lateral bulging of the granular pile is
a ted in this approach as an infin
the confining pressure on the pile evaluated (Hughes and Withers, 1974). Based on the
elastic –plastic theory given by Gibson and Anderson (1961) for a frictionless material,
the ultimate lateral stress (σ3) of the surrounding soil is given (Hughes and Withers,
1974) as,
⎥⎤
⎢⎡
++= log1σσ Ec C
where σro is the total in-situ lateral stress,
Ec is the elastic modulus of the soil,
c is the undrained shear strength and
µ is the poisson’s ratio.
The ultimate capacity (qult) is given as
⎦⎣ + )1(203 µCer
( ) ⎟⎟⎠
⎞⎜⎜⎝
⎛−+
⎭⎬⎫
⎩⎨ ⎥
⎦
⎤⎢⎣ +
++=s
sceroult c
cqϕϕ
µσ
sin1sin1
12log1
⎧ ⎡ E
where φs is the angle of internal friction of the pile material.
32
Vesic (1972) extended this theory to soils with both friction and cohesion and gave an
(2.28)
ess (σ1+σ2+σ3) /3 at the equivalent failure depth and
cording to Vesic (1972), is given as
(2.29)
.4.3.6. Expe ithers (1974) conducted a series of
d Withers (1974) is given as,
(2.30)
ere σro is and Kp is the coefficient of passive pressure, which
Bayuk (1979), Rao (1982) and Greenwood and Kirsch (1983)
showed
.4.4. SETTLEMENT OF FOUNDATIONS ON COMPOSITE GROUND
ments of
compo
method (Aboshi et al. 1979; Barksdale, 1981),
[ ]
expression for the ultimate lateral resistance (σ3) of the surrounding soil as,
σ3 = cF’c + qF’q
where ‘c’ is the cohesion
‘q’ is the mean str
F’c and F’q are cavity expansion factors.
The ultimate capacity of stone column, ac
⎟⎟⎞
⎜⎜⎝
⎛−+′+′=
s
sqCult FqFcq
ϕϕ
sin1sin1
⎠
2 rimental approaches: Hughes and W
model tests on single granular piles in normally consolidated clay and found that the
ratio of the applied stress to the undrained shear strength in the case of treated ground
was more than that in the case of untreated clay.
The ultimate pile capacity predicted by Hughes an
uro
ult cq
4=σ
pK+
wh the in-situ radial stress,
varies from 4 to 5.89.
Goughnour and
that the shear stresses at the pile-soil interface would be small as there is no
relative movement of the pile with respect to soil.
2
The analytical and experimental approaches to predict the settle
site ground are explained below:
Analytical approaches: In the equilibrium
the load applied to the unit cell is equally shared by the pile and the soil. The settlement
(st) over a depth ‘H’ of the treated ground is calculated from the conventional equation
from consolidation settlement.
33
The settlement reduction ratio (β´) which is the ratio of the settlements of treated
and untreated grounds ( st and s) is given in terms of average effective stress(σo) and
applied stress (σ) as,
(2.31)
where
(2.32)
Priebe (1976) also considered the unit cell concept and estimated reduction in
settlement. The change in volume within the soil is directly related to vertical shortening
of the pile. Both the pile and soil are assumed to undergo equal settlements.
Goughnour and Bayuk (1979) and Goughnour (1983) assumed the granular pile
to be in plastic equilibrium and the volume changes to be accommodated by the
surrounding clay. The vertical strain are calculated assuming the pile to be linearly
elastic.
Based on unit cell concept, Van Impe and De Beer (1983) predicted ‘β’ considering the
granular piles as
(i) plastic, incompressible columns and
(ii) linearly elastic columns.
2.4.4.1. Experimental approach: Hughes et al. (1975) measured radial strains from the
pressuremeter tests and calculated the vertical strain within the granular pile as twice the
radial strain. The settlement is found by dividing the pile into layers of different
thicknesses.
Floss (1979) added the vertical displacement of the top of the pile to the
settlement of the soil strata below the pile tip.
The use of plain granular piles, at times with a skirting too, was confined to non-
swelling soils only. However, Rao (1986) suggested the use of these piles in expansive
soil deposits also as an alternative to the existing foundation practices, without
supporting it with any laboratory or field investigations. But, a mere granular pile, which
⎪⎭⎪⎩⎟⎠
⎜⎝ 0
10 σ⎪⎬
⎫
⎪⎨
⎧
⎟⎞
⎜⎛ +
⎞⎛ +
==′0
0
log σσ
σµσ
βs
t ⎪⎪
⎪⎪ ⎟⎟
⎠⎜⎜⎝
0010log
σs
sc
1=µ
an )1(1 −+
34
is an un-cemented particulate medium, cannot counteract heave of a foundation caused
by a swelling soil. Hence, Phanikumar (1997) modified a granular pile as a granular-pile
anchor (GPA) by anchoring the foundation at the bottom of the granular pile to an anchor
plate through an anchor rod and rendered it into a tension-resistant foundation system.
Phanikumar (1997) conducted an extensive laboratory testing on small scale GPAs and
found the GPA system extremely efficacious. .
2.5. GRANULAR PILE-ANCHORS (GPA) Phanikumar (1997) suggested a simple and a new foundation technique in the
form of granular pile anchor (GPA), by modifying a granular pile into a tension-resistant
foundation through an anchor plate and an anchor rod.
A granular pile-anchor (GPA) is an innovative modification of the conventional
granular pile. In a GPA the foundation is anchored to an anchor plate (or base plate) at
the bottom of the granular pile through a mild steel anchor rod. Fig 2.5 shows a
schematic representation of a GPA. Because of the anchor introduced in the granular
pile, the system becomes a tension-resistant foundation. Hence, the tensile uplift force
(Pu) caused on the foundation by the swelling soils is resisted by downward friction,
mobilized over the cylindrical pile-soil interface. Fig 2.6 shows the forces acting on a
GPA.
Fig 2.5 Concept of a granular pile-anchor (GPA)
Footing Anchor rod Granular pile-anchor Granular fill Anchor plate
35
The uplift force (PU) is caused by the swelling pressure (ps) of the soil and the
resistance (PR) to uplift is due to
(i) the weight of the GPA (Wgpa) acting in the downward direction and
(ii) the shear resistance or the frictional resistance mobilized over the cylindrical
pile-soil interface due to the shear parameters of the interface, namely, c´ and
φ´
In the case of a single GPA, the uplift force (Pu) on the foundation due to
swelling pressure (ps) of the soil can be written as,
(2.33)
where, Df is the diameter of the foundation and
Dgpa is the diameter of GPA.
However, if there are ‘n’ number of GPAs under a footing, Pu can be written as
(2.34)
Footing Df
Lgpa
Uplift force (Pu) due to swelling pressure on the footing
Granular pile-anchor
Anchor plate
Anchor rod
Fig 2.6 Forces acting on a GPA
Resistance to uplift (PR)
Lateral swelling pressure (Ksps)
⎟⎠⎝ 44 f⎞
⎜⎛ −= 22
gpasu DDpP ππ
⎟⎠
⎜⎝
−= 22
44 f DnDpP ⎞⎛ × gpasuππ
Dgpa
36
where ‘n’ is the number of GPAs in the group anchoring the foundation,
Similarly, in the case of single GPA, the resisting force (PR) can be written as,
(2.35)
( )( )[ ]φσπ ′++′+= tanKsPsLDP KcW vogpagpagpaR
where,
Wgpa is the weight of the GPA
Dgpa is the diameter of GPA,
Lgpa is the length of the GPA,
K is the lateral earth pressure coefficient,
Ks is the lateral swell pressure coefficient and
is the effective overburden pressure. ′
voσ
If there are ‘n’ numbers of GPAs anchoring a footing, the resisting force (PR) can be
written as
(2.36) ( )( )[ ]nKcWR KsPsLDP vogpagpagpa φσπ ′++′+= tan
Hence the factor of safety (FS) against uplift is
(2.37) u
R
PPFS =
A series of laboratory scale tests was conducted on expansive clay beds
reinforced with GPAs of varying diameter (Dgpa), length (Lgpa) and relative density (Dr)
to study the efficacy of this system (Phanikumar, 1997). Heave tests were conducted, and
rate and amount of heave were studied in the light of the above parameters. Test results
showed that the rate of heave improved with the introduction of GPA in the clay beds,
and the amount of heave of the clay beds also decreased significantly (Phanikumar,
1997; Phanikumar et al. 2004).
Rao et al. (2007), Rao et al. (2008) and Phanikumar et al. (2008) conducted field-
scale test program to study the pullout response of GPAs embedded in expansive clay
beds. Pullout tests were conducted for varying lengths and diameters. The length of the
granular pile anchor was varied as 500 mm, 750 mm and 1000 mm. The diameter of the
37
granular pile anchor was varied as 100mm, 150mm and 200mm for each length of the
GPA. Of the various single granular pile anchors, the GPA with l/d ratio of 5 showed the
best pullout load response when tested alone, resulting in a failure uplift capacity of
14.7kN. One pullout test was also performed to study the group effect on the behaviour
of GPAs. In the group test five GPAs were used with a c/c spacing of two times the
diameter of the GPA. All the GPAs were of diameter 150mm and length 1000mm. The
GPA at the center of the group was loaded. The granular pile anchor under the effect of a
group of GPAs resulted in increased resistance to uplift load for a given upward
movement. The failure load of the GPA was 18kN as against a failure pullout load of
12kN for the GPA when tested single indicating a 50% improvement in the failure load.
Phanikumar et al (2008) conducted an extensive field scale study on GPAs and found
that heave was reduced significantly in the field scale GPAs also. Rao et al. (2008)
subjected field scale GPAs to compressive loading and obtained excellent results. It was
found that the bearing capacity of the clay beds reinforced by field scale GPAs improved
significantly.
The investigations carried out on granular pile anchors (GPAs) reveal that the
GPA system is an effective foundation technique for reducing heave in expansive soils.
It is also a simple and cost-effective technique. The behaviour of GPA during swelling
was thoroughly investigated but its behaviour during shrinkage has not been studied so
far. It is important to understand the behaviour of GPA-reinforced expansive clay beds
subjected to swelling and shrinkage also as they are under the influence of alternate
swelling and shrinking due to the changes in climatic conditions. This thesis presents
research results from a detailed experimental investigation conducted to understand the
behaviour of expansive clay beds subjected to swelling and shrinkage cycles. Effect of
number of GPAs in an expansive clay bed (n = 0, 1, 2 and 3) and the number of swell-
shrink cycles (N = 0, 1, 2 and 3) on the amount of heave and shrinkage was studied.
Useful results were obtained.
The details of experimental investigations carried out are presented in Chapter 3
and the analysis of test results in the subsequent chapters.
38