DRAFT REPORT GEOTECHNICAL STUDY

MISSION BAY SHORELINE PROTECTION PROJECT PORT OF SAN FRANCISCO, CALIFORNIA

Prepared for:

COAST & HARBOR ENGINEERING, INC.

Prepared By: AGS, INC.

JULY 2009

111 New Montgomery Street, Suite 500, San Francisco, CA 94105 Phone (415) 777-2166 Fax (415) 777-2167

DRAFT REPORT GEOTECHNICAL STUDY

MISSION BAY SHORELINE PROTECTION PROJECT PORT OF SAN FRANCISCO, CALIFORNIA

AGS Job No. KI0301

Prepared for:

COAST & HARBOR ENGINEERING, INC.

Prepared By: AGS, INC.

JULY 2009

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TABLE OF CONTENTS

1. INTRODUCTION ...................................................................................................... 3

1.1. PROJECT DESCRIPTION ............................................................................ 3

1.2. WORK PERFORMED ................................................................................... 3 1.2.1. Review of Available Data ............................................................................... 4 1.2.2. Field Exploration ............................................................................................ 4 1.2.3. Geotechnical Laboratory Testing ................................................................... 5 1.2.4. Engineering Analyses and Report Preparation .............................................. 5

2. FINDINGS ................................................................................................................ 6

2.1 SITE CONDITIONS AND BACKGROUND .................................................... 6

2.2 GEOLOGY .................................................................................................... 7

2.3 FAULTS AND SEISMICITY ........................................................................... 8

2.4 SUBSURFACE CONDITIONS .................................................................... 11 2.4.1. Previous Exploration .................................................................................... 11 2.4.2. AGS Investigation ........................................................................................ 12

2.5 GROUNDWATER ....................................................................................... 13

3. RECOMMENDATIONS .......................................................................................... 15

3.1 GENERAL ................................................................................................... 15

3.2 SEISMIC DESIGN CONSIDERATIONS ...................................................... 16 3.2.1. Fault Rupture ............................................................................................... 16 3.2.2. Maximum Earthquake.................................................................................. 16 3.2.3. Estimated Earthquake Ground Motions ....................................................... 16

3.2.3.1.Deterministic Methods ......................................................................... 16 3.2.3.2.Probabilistic Methods ........................................................................... 17

3.2.4. Liquefaction Hazard .................................................................................... 17 3.2.5. Consequences of liquefaction ..................................................................... 18

3.2.5.1.Seismically-Induced Settlement ........................................................... 19 3.2.5.2.Lateral Deformation ............................................................................. 20

3.2.6. Liquefaction Mitigation ................................................................................. 21 3.2.7. Seismically-Induced Lateral Earth Pressures .............................................. 24

3.3 EXCAVATION AND EARTHWORK ............................................................. 25 3.3.1. Site Preparation ........................................................................................... 25 3.3.2. Fills and Backfills ......................................................................................... 26 3.3.3. Temporary Excavations ............................................................................... 26

3.4 FOUNDATIONS .......................................................................................... 27 3.4.1. General ........................................................................................................ 27 3.4.2. Renovation Scheme .................................................................................... 28

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3.5 SETTLEMENT ............................................................................................. 29

3.6 PERMANENT RETAINING STRUCTURES ................................................ 29

3.7 RESISTANCE TO LATERAL LOADS .......................................................... 30

3.8 EXISTING STRUCTURS ............................................................................. 30

3.9 CONSTRUCTION CONSIDERATIONS ...................................................... 31 3.9.1. General ........................................................................................................ 31 3.9.2. Geotechnical Services During Construction ................................................ 31

4. CLOSURE.............................................................................................................. 33

5. REFERENCES ....................................................................................................... 34 TABLES Table 1 - Historical Earthquakes ..................................................................................... 9 Table 2 - Active Fault Seismicity ................................................................................... 10 Table 3 - Summary of Seismically-Induced Settlements ............................................... 19 Table 4 - Summary of Liquefaction-Induced Lateral Deformations ............................... 21 Table 5 - Summary of Liquefaction Mitigation Techniques ............................................ 24 Table 6 - Estimated Settlements ................................................................................... 29 Table 7 - Allowable Passive Pressures ......................................................................... 30 PLATES Plate 1 - Site Location Map Plate 2 - Subsurface Exploration Location Map Plate 3 - Earthquake Epicenters and Fault Map Plate 4 - Preliminary Design Pressure Data Near Borings B-1 and B-3 Plate 4 - Preliminary Design Pressure Data Near Boring B-2 APPENDICES Appendix A - FIELD EXPLORATION Appendix B - GEOTECHNICAL FIELD AND LABORATORY TESTING Appendix C - LIQUEFACTION POTENTIAL

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1. INTRODUCTION

1.1. PROJECT DESCRIPTION

This report presents the results of our geotechnical study for the Shoreline Protection

project at Mission Bay, San Francisco, California. The shoreline location east of Terry

Francois Boulevard and south of Pier 54 is shown on Plate 1 – Site Location Map.

This report presents findings of the subsurface exploration and gives our geotechnical

engineering conclusions and recommendations regarding the proposed project.

Conclusions and recommendations were developed with regard to geoseismic hazards,

seismic design parameters, site improvements, earthwork recommendations, foundation

types, lateral earth pressures, resistance to lateral loads, and construction monitoring. Of

particular importance to the design of the project is an assessment of the liquefaction

potential of the fills behind the seawall and possible lateral deformations and seismically

induced settlements induced from liquefaction.

The conclusions and recommendations presented in this report are based on the

subsurface conditions encountered at the locations of the three borings drilled during field

exploration programs, available geotechnical information obtained by others at the site and

its vicinity, and available geologic and seismologic information for the site vicinity. These

conclusions and recommendations should not be extrapolated to other areas, or used for

other facilities without our prior review.

1.2. WORK PERFORMED

As stated in our April 2009 proposal, the purpose of our study was to explore and evaluate

subsurface conditions. The work performed for this project included the following:

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1.2.1. Review of Available Data

Available published geotechnical, geologic, and seismologic data, as well as the existing

unpublished data contained in our files pertinent to the project site and its vicinity were

reviewed. We reviewed the 1996 Trans Pacific Geotechnical Report for Improvements at

Pier 52, the Port of San Francisco Boring Profiles completed from China Basin to 25th

Street (1961), and also the Port of San Francisco Seawall Details (Revised 1957) and

Alternate A Wharf Alteration Pier 54 Drawings (1954). Complete citations are listed in the

References section of this report.

1.2.2. Field Exploration

Following marking of the planned boring locations by AGS, Underground Service Alert

(USA) was notified in order that utility companies mark their buried utilities. Based on the

ground markings, none of the originally planned boring locations required adjustment to

avoid utilities. V&W Drilling Inc. of Lodi, California completed the drilling on June 9 and 10,

2009. The field exploration program was performed under technical supervision of a

qualified geologist from our firm who completed logs of each test boring which document

drilling progress and record the encountered subsurface conditions.

Modified California (MC) samples, Standard Penetration test (SPT) samples, and Shelby

tubes were obtained from each boring, as appropriate for the various soils encountered.

Details of the subsurface exploration, including the Logs of Test Borings, are presented in

Appendix A. Boring locations are shown on Plate 2 – Subsurface Exploration Location

Map.

All three borings were located landward (west) of the seawall and were extended to depths

ranging from 26.5 to 36.5 feet below the existing ground surface (bgs). Each borehole was

backfilled with cement grout. Following backfilling and allowance of settlement the ground

surface at each boring location was restored to grade.

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1.2.3. Geotechnical Laboratory Testing

AGS performed a laboratory testing program on selected soil samples obtained during the

field exploration program. The laboratory tests included moisture content, dry density,

sieve analyses, and Atterberg limits as appropriate for the various soils encountered.

Details of the geotechnical laboratory testing program are included in Appendix B -

Geotechnical Laboratory Testing.

1.2.4. Engineering Analyses and Report Preparation

Engineering analyses were performed based on the field and laboratory data to develop

geotechnical conclusions and recommendations for the design and construction of the

proposed project. Our geotechnical findings, conclusions, and recommendations, along

with the supporting field and laboratory data, are presented in this engineering report. The

report addresses the following:

Subsurface soil conditions;

Groundwater elevations;

Local geologic conditions;

Faults and seismicity;

Peak ground surface accelerations for the controlling maximum credible

earthquake;

Geoseismic hazardous including liquefaction potential, seismically-induced

settlements, and seismically-induced lateral deformations;

Liquefaction mitigation and soil improvement recommendations;

Existing foundation capacities;

Lateral loads on the seawall;

Resistance of foundation to lateral loads;

Design criteria for new foundation; and

Construction considerations.

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2. FINDINGS

2.1 SITE CONDITIONS AND BACKGROUND

The project shoreline is east of the Mission Bay Redevelopment area, south of China Basin

and north of Central Basin. The project shoreline extends approximately 1,500 feet south

of Pier 54, next to Terry Francois Boulevard north of the Mariposa Street Intersection.

The northern half of the project site shoreline is protected by a concrete seawall which was

originally designed around 1912 and was partially reconstructed around 1954 (FEMA,

2007). According to our discussion with you, the existing seawall is founded on shallow

spread footings. Rip rap, concrete rubble, and debris are used as a wave break in front of

the seawall. The northern reach of the shoreline behind the seawall is paved with concrete

or asphalt and includes a pedestrian / bicycle trail next to Pier 54, abandoned construction

staging area, construction yard and a one-story wood-framed office, and a makeshift

residence. The fill materials were undercut causing large voids beneath the seawall. These

voids were up to 6 feet in length and up to 2 feet in height. The central and southern portion

of the shoreline is dominated by the remains of Pier 64, including the remains of some large

concrete abutments and walls which would have supported the pier. This area is not paved

and is fenced in by an 8-foot high chain link security fence. The south end of the project

shoreline is protected by a wooden seawall and is bounded by Agua Vista Park, where a

fishing pier extends approximately 80 feet offshore.

Mission Bay was once an embayment at the mouth of Mission Creek, at which time the

shoreline was located near the present day alignment of Interstate 280 (U.S. Coast Survey,

1853). Today the Mission Bay area represents one of the largest areas of historical landfill

in San Francisco. During the 19th century a land bridge was constructed to connect the

north and south sides of the bay by railroad in the vicinity of the present day 3rd Street

corridor. The unfilled portion of the former bay west of 3rd Street formed a lagoon centered

approximately ¼ mile west of the project shoreline (1895 and 1899 United States

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Geological Survey ,USGS, topographic maps). Following the 1906 earthquake the lagoon

was filled, and by 1915 the USGS topographic map shows the shoreline extended as far

east as it exits at the present time. At that time Pier 54 and the original seawall were

already constructed.

2.2 GEOLOGY

The northern San Francisco peninsula, including the project area, is part of the Coast

Ranges geomorphic province. The province is a seismically active region characterized by

northwest-trending mountains, valleys, and faults. The peninsula is bordered on the east

by San Francisco Bay, a drowned, northwest-trending structural depression. The bay and

much of the peninsula are underlain by the late Mesozoic age rocks of the Franciscan

Complex.

Beneath San Francisco Bay, and along much of its margin, the Franciscan bedrock is

overlain by a young, geologically unconsolidated sedimentary sequence, which, in places,

exceeds 400 feet in thickness. The sequence is often subdivided (Goldman, 1969) into

three "natural" units - Older Bay Mud (or old bay clay), Bay Side Sand, and Younger Bay

Mud. In the area of Mission Bay artificial fill has been placed along the margins of the bay

to claim marshland and land once covered by shallow water.

Along the project shoreline the bedrock surface beneath the Mission Bay Basin is estimated

to occur from 100 to 150 feet below the existing ground surface (Goldman, 1969).

Shallower bedrock is found to the northeast and southwest, in the respective vicinities of

Mission Rock and Potrero Hill. Serpentine rock of the Franciscan Formation is mapped at

the Mission Rock, located approximately 1,400 feet northeast of the Pier 54 abutment.

Sandstone, shale, siltstone and serpentine of the Franciscan Formation are mapped across

the northeastern slope of Potrero Hill, located more than 1100 feet west and southwest of

the southern end of the project shoreline (Graymer et al, 2006). Mission Bay Basin is

bounded to the west by the historical delta of Mission Creek, where two small creek

systems once flowed into San Francisco Bay near the present day alignment of Interstate

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280 (U.S. Coastal Survey, 1853). Underneath the project shoreline the artificial fill of

Mission Bay is underlain by the previously discussed young, geologically unconsolidated

sequence of alluvium. This includes areas of very soft Younger Bay Mud which are

estimated to form an approximately 20 to 40 foot thick layer beneath the artificial fill

(Goldman, 1969).

2.3 FAULTS AND SEISMICITY

The project area is located in a seismically active region which has been subjected to

several strong earthquakes during historic time. Active faults in the area are shown on

Plate 3 – Earthquake Epicenters and Fault Map.

The San Andreas Fault, which is situated about 12.7 km southwest of the site, dominates

the tectonics, geology, and physiography of the San Francisco Bay region. The Hayward

Fault is situated about 16 km northeast of the site. Other major active faults, which could

cause significant shaking at the project site, are the San Gregario, Mount Diablo Thrust,

Concord, Calaveras, and Rodgers Creek Faults. Active faults that are pertinent to the site

and historic earthquakes attributed to each fault are listed in Table 1 - Historical

Earthquakes.

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TABLE 1 HISTORICAL EARTHQUAKES

Date

Magnitude Fault Epicenter Area

June 10, 1836 6.51, 6.84 San Andreas San Juan Bautista

June 1838 7.51, 7.04 San Andreas San Juan Bautista

Nov. 26, 1858 6.254 Calaveras San Jose Area

October 8, 1865 6.32, 6.54 San Andreas South Santa Cruz Mountains

October 21, 7.02,4 Hayward Berkeley Hills, San Leandro

February 17, 6.04 San Andreas Los Gatos

April 19, 1892 6.54 Uncertain Vacaville

April 21, 1892 6.254 Uncertain Winters

June 20, 1897 6.254 Calaveras Gilroy

March 31, 1898 6.54 Uncertain Mare Island

May 19, 1889 6.254 Uncertain Antioch

April 18, 1906 7.93 San Andreas Golden Gate

July 1, 1911 6.62, 6.54 Calaveras Diablo Range, East of San Jose

October 22, 6.14 San Gregorio? Monterey Bay

April 24, 1984 6.14 Calaveras Morgan Hill

October 17, 7.14 San Andreas Loma Prieta, Santa Cruz

(1) Borchardt & Toppozada (1996) (2) Toppozada et al (2000) (3) Petersen (1996) (4) Ellsworth, W.L. (1989)

The maximum moment magnitude earthquake (Mmax) is defined as the largest earthquake

that a given fault is considered capable of generating. The Mmax on the San Andreas

Fault would be a magnitude 7.9 event occurring approximately 12.7 km from the project site

(USGS, 2008). The Mmax on the Hayward Fault would be a magnitude 7.1 event occurring

approximately 16 km from the project site (USGS, 2008). The Mmax given for the Hayward

Fault is based on a rupture of the entire length of the fault. The seismicity associated with

each pertinent fault, including estimated slip rates, is summarized below in Table 2 - Fault

Seismicity.

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TABLE 2 ACTIVE FAULT SEISMICITY

Fault Name

Distance to Site1

(Km)

Maximum Moment

Magnitude2

Contributing Segments

Slip Rate 2

(mm/year)

San Andreas

12.7 7.9 SAO, SAN, SAP, SAS 24 3

Hayward

16 7.1 HN, HS 9 2

San Gregorio

19 7.3 SGN, SGS 7 3

Mount Diablo Thrust

33 6.6 MTD 2 1

Calaveras

33 6.8 CN, CC, CS 15 3

Rodgers Creek

35 7.0 RC 9 2

Concord-Green Valley

38 6.7 CCD, GV) 4 2

Monte Vista-Shannon

39 6.8 MVS 0.4 0.3

Point Reyes

44 7.0 1 0.5

West Napa

45 6.5 WN 1 1

Greenville

50 7.0 GN, GS 2 1

Great Valley (segment 5)

66 6.5 CVS5 1.5 1

Great Valley (segment 4)

72 6.6 GVS4 1.5 1

Great Valley (segment 7)

76 6.7 GVS7 1.5 1

Hunting Creek -

Berryessa

77 7.1 HCB 6 3

Zayante Vergeles

83 7.0 ZV 0.1 0.1

Monterey Bay -

Tularcitos

96 7.3 MBT 0.5 0.4

1. Jennings (1994), Nearest Approach

2. WGCEP (2008), Working Group on California Earthquake Probabilities.

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2.4 SUBSURFACE CONDITIONS

2.4.1. Previous Exploration

The locations of past subsurface exploration both onshore and offshore relative to the

project shoreline are shown on Plate 2. During the 1960s the Port of San Francisco

completed soundings along the entire shoreline, including in the area south of Pier 54 (Port

of San Francisco, 1961). Soundings involved recording the depths to which a pipe sank

under its own weight and also under a weight of 3000 pounds. In the project vicinity,

soundings were recorded at three locations immediately offshore of the seawall, at

respective distances of approximately 50 feet, 250 feet, and 500 feet south of Pier 54

(Locations 120, 119, and 118). Findings were that the mudline occurred from 24 to 28 feet

beneath mean lower low water (MLLW) and also that there was an approximately 14 to 20

foot thick layer of very soft mud beneath the mudline. Soundings further indicated that the

very soft mud is underlain by stiff clay 500 feet south of Pier 54 but that in the northward

direction there is a thickening sequence of underlying soft clay or loose to medium-dense

sand which was penetrated to a depth of 63.5 feet beneath the MLLW level 50 feet south of

Pier 54.

Onshore conditions reflect the accumulation of a thick fill sequence both placed and

apparently sunk on top of the underlying native unconsolidated alluvial sequence. The

closest nearby available recent boring information that was reviewed was obtained by

Trans Pacific Geotechnical Consultants (1996) in an area next to Pier 52, approximately

300 feet north of the project shoreline. Trans Pacific Geotechnical Consultants (1996)

drilled four borings onshore, Borings B-1 through B-4. In those borings they encountered a

very thick sequence of fill consisting of loose to very dense sandy and clayey gravel with

boulders that continued to as deep as 71 to 74.5 feet below the existing ground surface

(bgs). Native dense to very dense clayey gravel was penetrated beneath the fill to depths

of 110.5 to 114 feet bgs and stiff silty clay (Older Bay Clay) was penetrated beneath. Trans

Pacific Geotechnical Consultants (1996) also completed four offshore borings, Borings B-5

through B-8. In the offshore borings very soft to soft silty clay (Younger Bay Mud) was

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penetrated to depths ranging from 24 to 36.5 feet below the mudline, with concrete rubble

and rock fill penetrated to 56 feet below the mudline. Gravelly fill was encountered beneath

the rubble and rock fill extending to a depth of 75 feet. Deeper than 75 feet to 99 feet

native dense to very dense gravel with sand was encountered. Stiff silty clay (Older Bay

Mud) was encountered deeper than 99 feet.

Past subsurface explorations indicate that the project shoreline occupies an area of thick

fill, where there was once 20 to 30 feet of water underlain by 20 to 40 feet of very soft to

soft mud and loose sand. In the area north of Pier 54 the water and much of the softer

underlying soils appear to have been displaced by the typically coarse overlying fill soils to

as deep as 70 to 75 feet below the existing ground surface. In the area of the project

shoreline further south, near the remains of Pier 64, there is little available subsurface data,

but soundings indicate the bay is shallower and consequently fill is likely to be thinner.

2.4.2. AGS Investigation

AGS drilled three borings east of Terry Francois Boulevard, along the Mission Bay

Shoreline south of Pier 54 and north of Mariposa Street. Boring B-1 was drilled next to the

seawall inside the vacant lot north of the contractor’s yard at 559 Terry Francois Boulevard.

Boring B-2 was drilled next to the seawall inside the contractor’s yard, approximately 150

feet south of Boring B-1. Boring B-3 was drilled along the shoreline approximately 570 feet

south of Boring B-2. Borings B-1, B-2 and B-3 were drilled, sampled, and logged to

respective depths of 26.5, 36.5 and 27.5 feet below the existing ground surface (bgs). Both

Borings B-1 and B-2 penetrated through a surface layer of asphalt from 2 to 3-inches thick

and also indicate that an aggregate base layer consisting of dense reddish-brown silty

gravel with sand underlies the surface asphalt to a depth of approximately 2 feet bgs.

Fill was encountered beneath the aggregate base layer to maximum depths of exploration

in Borings B-1 and B-2, and to approximately 15 feet in Boring B-3. In Boring B-1, fill

consisted of 5 feet of medium-dense silty, clayey gravel with sand and with a few cobbles.

Typically, coarse poorly graded gravel with silt and sand were penetrated from 5 to 12.5

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feet, while apparent cobbles and boulders were penetrated from 12.5 to nearly 15 feet bgs.

Deeper than 15 feet bgs in Boring B-1 older fill consisting of medium-dense well-graded

sand with gravel was encountered from 15 to 21 feet, with very dense poorly graded gravel

and cobble fill penetrated from 21 feet to the bottom of the hole at 26.5 feet, where practical

drilling refusal was met.

In Boring B-2, fill consisted of 2 to nearly 20 feet of dense to medium-dense silty, clayey

sand with gravel and cobbles underlain by dense to medium-dense well-graded sand with

silt and a few gravelly layers to the bottom of the hole at 36.5 feet. These soils appear to

represent older fill similar to that found in Boring B-1, but with fewer coarse gravel or

cobbles.

In Boring B-3 silty, fill consisted of approximately 15 feet of sandy gravel; poorly graded

gravel with silt and sand; and silty, clayey gravel with sand, cobbles, and concrete rubble

underlain by loose to medium-dense silty, clayey sand with gravel. Loose, medium and

coarse-grained silty sand was encountered from 15 to 16.5 feet. Very soft silty fat clay with

a trace to few coarse sand or fine gravel “Younger Bay Mud” was penetrated from 16.5 feet

to the bottom of the hole at 27.5 feet bgs.

Generally, the Borings indicate that 15 to 20 feet of gravelly and cobbly soils, including

areas of concrete rubble and small boulders (rip rap) occur beneath the surface behind the

seawall. In Borings B-1 and B-2 the layer of coarse fill soils is underlain by a relatively thick

well-graded sandy and gravelly layer of older fill, the bottom of which was not penetrated.

In Boring B-3 a thin layer of native sand and then a thick layer of very soft Bay Mud occur

beneath the layer of coarse fill soil.

2.5 GROUNDWATER

Water was observed in each test boring drilled for this study. Water was observed at

depths ranging from approximately 6.2 to 9.8 feet below the existing ground surface and

appears to coincide with the tidal water levels east of the seawall. Generally, the ground

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surface elevation and top of seawall elevation of approximately 10 feet (San Francisco City

Datum) gives approximately 3 feet of freeboard above the maximum high tide elevation of

nearly 7 feet. When groundwater elevations were measured the tides were at intermediate

or low levels. Groundwater is expected to rise closer to the ground surface during periods

of very high tide, and could occur as shallow as 2 feet.

Variations in the water level at the site are likely to occur due to the variations of the water

level in the Bay, changes in precipitation, temperature, and other factors not evident at the

time of this study. The quality of the water was not evaluated as part of this study.

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3. RECOMMENDATIONS

3.1 GENERAL

Based on the results of our field exploration and laboratory testing programs it is our

opinion that the proposed shoreline protection repair is feasible from a geotechnical point of

view, provided the recommendations presented in this report are incorporated in the design

and construction of the project. The existing loose to medium dense sandy, gravelly fill and

loose to medium dense native soils extending to top of the Younger Bay Mud at the site

have liquefaction potential when subjected to significant earthquake shaking. The

consequences of liquefaction include seismically induced settlements, additional lateral

loads on the seawall, localized lateral deformation of fill materials, and floatation of buried

structures.

The major geotechnical concern for this project is the stability of the existing seawall under

a major earthquake and localized lateral deformation of fill materials. It is our opinion that

the existing seawall is founded in the liquefiable material and it will undergo deformation

(both translational and rotational) due to loss of passive resistance. Therefore a new

seawall with or without a ground improvement program is needed to resists lateral forces

induced by a liquefaction event. The ground improvement program may consist of

compaction grouting and/or vibro-replacement methods. We understand that the Port of

San Francisco has instructed the design team to improve the existing seawall to resist

active and earthquake pressures knowing that the improved seawall will fail during a

liquefaction event.

The results of our liquefaction potential evaluations indicate that the seismically-induced

settlements at the site will range from 4 to 6 inches from a 475-year return period

earthquake. The results of our lateral spread analyses indicates that liquefaction-induced

lateral deformation at the site will be about 1 to 3 feet and will cause serious damage to the

existing seawall during an earthquake with moment magnitude of M7.9 on San Andreas

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Fault. The existing buildings at the site may experience damage due to seismically induced

settlements and lateral deformation.

3.2 SEISMIC DESIGN CONSIDERATIONS

3.2.1. Fault Rupture

The site is not located within an Alquist-Priolo Special Study zone. No known or suspected

faults occur at the Mission Bay Shoreline Protection project site, therefore surface fault

rupture is considered to be very low.

3.2.2. Maximum Earthquake

The Maximum Moment Magnitude (Mmax) earthquake is the largest reasonable

earthquake that a given fault appears capable of generating in the current tectonic setting.

The controlling Mmax earthquake that could affect the project site would be a magnitude

7.9 seismic event occurring on the San Andreas fault, with a seismogenic source (focus of

seismic energy at depth) located about 8 miles (12.7kilometers) southwest the site. Values

of Mmax earthquakes on other faults in the region are shown in Table 2. The locations of

active faults, as adopted by California Geological Survey (2002), and the epicenters of

historical earthquakes within 50 kilometers of the site are shown on Plate 3.

3.2.3. Estimated Earthquake Ground Motions

Ground surface accelerations were estimated using both deterministic methods and

probabilistic methods.

3.2.3.1. Deterministic Methods

Correlations between distance from a causative fault and mean values of the peak

horizontal acceleration, developed by Abrahamson and Silva (1997), Campbell (1997),

Boore et al. (1997), and Sadigh et al (1997), were used to estimate the mean value of the

peak horizontal acceleration resulting from the Mmax earthquake on the San Andreas fault.

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Based on the maximum result from these relationships, the Mmax 7.9 earthquake occurring

on the San Andreas fault is estimated to generate a mean peak horizontal acceleration of

0.42g at the site.

3.2.3.2. Probabilistic Methods

The probabilistic seismic hazard analyses were performed using the FRISKSPWIN

computer software package Version 4. Three levels of peak horizontal ground

accelerations were developed at the site for average earthquake return periods of 72, 475,

and 950 years, using attenuation relationships developed by Abrahamson and Silva (1997),

Campbell (1997), Boore et al. (1997), and Sadigh et al (1997.) These earthquake return

periods correspond to approximately 50 percent probability of being exceeded in 50 years

(72-year earthquake), 10 percent probability of being exceeded in 50 years (475-year

earthquake), and 10 percent probability of being exceeded in 100 years (950-year

earthquake). The average peak horizontal accelerations at the site calculated from the

above-referenced attenuation relationships for the three earthquake return periods are as

follows.

72-year earthquake, peak horizontal acceleration is 0.27g

475-year earthquake, peak horizontal acceleration is 0.50g

950-year earthquake, peak horizontal acceleration is 0.60g

3.2.4. Liquefaction Hazard

Soil liquefaction is a phenomenon in which saturated (submerged) cohesionless soils lose

their strength due to the build-up of excess pore water pressure, especially during cyclic

loadings such as those induced by earthquakes. In the process, the soil acquires mobility

sufficient to permit both horizontal and vertical movements, if not confined. Soils most

susceptible to liquefaction are loose, clean, uniformly graded, fine-grained sands. Silty and

clayey sands may also liquefy during strong ground shaking.

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The nature of liquefaction depends greatly on the characteristics of the soils. In loose soils,

liquefaction results in significant loss of soil strength, which can lead to large deformations.

In dense soils, although a condition of liquefaction can be initiated, the tendencies for loss

of strength and deformations are resisted by dilation of the soils. Deformations in dense

soils result in a tendency for soil volume increase (dilation), which in turn results in

reduction of pore water pressures, increase in effective stresses, and increased resistance

to further deformations.

The liquefaction potential of soils at the site was evaluated using a simplified, analytical,

and empirical procedure that is correlated with the liquefaction behavior of saturated sands

during historic earthquakes (Youd, 2001; and Idriss and Boullanger, 2008). The primary

data utilized in the analysis consisted of standard penetration test (SPT) and modified

California (MC) sampler blow counts, which were obtained from the three borings drilled at

the site. The SPT and MC blow counts recorded in the field were corrected for various

factors to obtain corrected N-values, which were used in the liquefaction analysis. The

factors used to obtain corrected N-values, included the effects of overburden pressure, rod

length, sampler type and size, and fines content.

The liquefaction analysis was conducted using the following parameters.

• Magnitude 7.9 earthquake

• Peak horizontal acceleration of 0.50g (475-year return period)

• Depth to groundwater at 2 feet.

Based on the results of the liquefaction analysis, the Mission Bay site is considered to have

a high potential for liquefaction.

3.2.5. Consequences of liquefaction

The main effects of liquefaction at the site include settlement of the ground surface and

utilities, lateral deformation, development of excess pore water pressure, buoyancy effects

on the below groundwater structures, loss of allowable bearing pressure, and increased

lateral pressures on utilities and foundations extending below the groundwater table.

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3.2.5.1. Seismically-Induced Settlement

Seismically-induced settlements were estimated for two scenarios. On the first scenario, we

estimated the seismically-induced settlements using maximum depth of explorations at

each boring. In the second scenario, we assumed that medium dense fill extends to

maximum depth of 50 feet in Boring B-1 and B-2. This assumption is based on the Trans

Pacific Geotechnical Consultants (1966) report which stated that the fill extends to depth of

71 to 74.5 feet. Typically, medium dense soils below 50 feet do not Liquefy. In Boring B-3,

we assumed that either Younger Bay Mud extends to depth of 50 feet or native soils are

dense and will not liquefy. The results of the liquefaction analysis conducted for each boring

is included in Appendix E and summarized in Table 3.

The estimated seismically-induced settlements presented in Table 3 are absolute values. It

is the opinion of AGS that differential seismically-induced settlements along 100 feet of

length are about half of the absolute values.

TABLE 3 SUMMARY OF SEISMICALLY-INDUCED SETTLEMENTS

Boring ID Seismically Induced Settlement at the Existing

Ground Surface (to the bottom of the boring)

(inches)

Seismically Induced Settlement at the Existing

Ground Surface (to maximum depth of 50 feet)

(inches)

B-1 2.5 3.5

B-2 2.0 5.0

B-3 6.0 6.0

AGS estimates that settlements up to 6 inches may occur during seismic events. If the

anticipated seismically-induced settlements are not acceptable to the designer, AGS

recommends that if a seawall design is desired to resist liquefaction effect, either the

proposed improvements should be supported on a deep foundation system, such as pile

foundation, or a soil improvement method, as discussed in Section 3.2.7, be used to reduce

liquefaction consequences to an acceptable level.

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3.2.5.2. Lateral Deformation

Seismically-induced lateral deformation is also another phenomenon which could occur

during a seismic event. The continuity/discontinuity of potentially liquefiable soil layers is a

key consideration in evaluating the potential for lateral deformation. We evaluated the

potential for lateral spreading of the soil behind the seawall using an empirical relationship

developed by Youd et al. (2002) and Zhang et al. (2004). The relationship by Youd et al.

(2002) incorporates the thickness of the liquefiable layer, the fines content and mean grain-

size diameter of the liquefiable soil, the magnitude and distance of the earthquake from the

site, the slope of the ground surface, and boundary conditions, such as a free face, to

estimate the horizontal ground movement.

Based on the predictive relationship for lateral deformation by Youd et al. (2002), liquefiable

soil layer with blow count of 15 and less may exhibit lateral deformation. The test borings

indicates that the potentially liquefiable soil layers with blow count of 15 and less are

present between depths of 2 to 12.5 feet in Boring B-1 and 2 to 16.5 feet in Boring B-3. For

significant lateral deformation to occur, a continuous layer of potentially liquefiable soil

extending for a considerable distance (on the order of several hundred feet) would be

required. Since the test borings and recorded lateral deformation events within the vicinity

indicate the presence of such a layer, it is the opinion of AGS that the potential for lateral

deformation at this site would be moderate to high.

During lateral spreading, surficial soil displaces along a shear zone that has formed within

an underlying liquefied layer. The surficial soil is transported downslope or in the direction

of a free face, such as a channel slope, by earthquake and gravitational forces.

For our analysis, we used two different boundary conditions; one with free face of and one

with gentle sloping ground. For free face boundary condition, we assumed free face height

of 25 feet based on data from offshore sounding. For sloping ground condition, we

assumed a sloping ground gradient of 0.8 percent. Mean grain sizes of 0.6 millimeters

(mm) were used for the fills; the. fines content of the fill was varied from 6 to 12 percent.

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The results of the liquefaction-induced lateral deformation analysis based on the predictive

relationship for lateral deformation by Youd et al. (2002) and Zhang (2004conducted for

each boring is summarized in Table 4

TABLE 4

SUMMARY OF LIQUEFACTION-INDUCED LATERAL DEFORMATIONS Boring ID Youd et al.

(feet) Zhang et al.

(feet)

Sloping Ground Free Face Sloping Ground Free Face

B-1 0.5-1.0 1.2-2.3 1.0 2.0

B-2 - - 2.0 5.5

B-3 3.0-6.0 6.0-8.0 1.5 4.5

The presence of rock fill along the alignment of the seawall in Mission Bay site would likely

reduce the potential liquefaction-induced lateral deformation that may occur behind the

seawall. We judge this factor could reduce the amount of liquefaction-induced lateral

deformation to about one to three feet.

Liquefaction of soils underlying the existing seawall may also induce temporary buoyant

uplift pressures. The magnitude of such pressures is difficult to estimate, because of the

variability in materials that may be used and construction techniques. However, given that

potentially liquefiable soils are likely only present in continuous layers between depths of 2

to 16.5 feet, it is opinion of AGS that such buoyant uplift pressures would be relatively low.

3.2.6. Liquefaction Mitigation

Seismically-induced settlements of up to 6 inches and lateral deformation up to 3 feet were

estimated at the site, as discussed previously. The designers should either design for such

settlements and lateral deformations, or where estimated seismically-induced settlements

and lateral deformations cannot be tolerated, they should be mitigated through a program

of ground improvement. This section provides several feasible options for liquefaction

mitigation measures. Due to presence of coarse gravel, cobble and concrete pieces within

the existing fill materials, pre-drilling may be required.

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Ground improvement should be performed in areas where the total calculated seismically-

induced settlement exceeds the structurally acceptable level, and be designed to reduce

total liquefaction-induced deformation to a tolerable level. The soil zone to be improved

includes those soils which are at depth of 2 to 16.5 feet. The total thickness of the zone to

be improved depends both on the actual thickness of the soil layer and the desired

reduction in predicted settlement. AGS anticipates that after improvement of soils between

depths of 2 to 16.5 feet, the seismically-induced settlement will reduce to of 1-inch or less.

If seismically-induced settlement of up to 1 inch is not acceptable to the designer, the

unsaturated soils should also be mitigated. AGS recommends that the transition area

(areas with and without soil improvement) be subexcavated at least 2 feet below subgrade

level of the utilities and recompacted with two layers of geogrid to provide a gentle

transition of settlement and minimize the occurrence of differential settlement.

There are several techniques available for soil improvement which may be applicable to

this site: vibro-replacement stone columns and grouting techniques. Alternatively, the

liquefaction-induced settlement can be minimized by supporting the seawall on driven piles.

A low-vibration piling system (such as Screw-in Piling or Press-in Piling) may be used

where vibration due to pile driving activities can not be tolerated. In these low-vibration

systems, a pile is screwed or pushed into the strata, with the resulting skin friction and end

bearing capacities similar to driven piles.

The vibro-replacement stone column technique of ground treatment has proven successful

in reducing the liquefaction potential of sands and low plasticity silt. Stone columns are

used for loose silty sands having greater than about 15 percent fines. Cohesive, mixed and

layered soils generally do not densify easily when subjected to vibration alone, therefore,

the vibro-replacement stone column technique was developed specifically for these soils,

effectively extending the range of soil types that can be improved with the deep vibratory

process.

Grouting techniques (compaction, permeation, deep mixing, chemical, and jet grouting) of

soil improvement have also proven successful in reducing the liquefaction potential of

sandy material. The grouting techniques become less efficient with increased fines

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content, such as silt and clay. Of these grouting techniques, jet grouting appears to be the

most efficient method for the site. Essentially, in jet grouting, the injection of an ultra high

pressure fluids or binders at high velocity. These binders break up the soil structure

completely and mix the soil particles in-situ to create an homogenuous mass which in turn

solififies. Other grouting techniques, such as deep mixing, involve the use of large augers

both to introduce cement grout and to mix it with the soil, producing a treated soil cement

column.

Dynamic deep compaction can densify and reduce the liquefaction potential of sandy soils.

This method becomes less effective with high groundwater level and increased fines

content in soils, but has relatively low costs compared to other methods. However, due to

the effects of vibrations on the adjacent properties, it is the opinion of AGS that this method

is not applicable for this site.

The soil improvement design will depend on the costs of performing the work as well as the

technical specifics of the work, and is beyond the scope of this study.

The practical applications of many of these measures have been presented in the literature

(Hryciw 1995; Stewart et al. 1997; Boulanger et al. 1997; Mitchell et al. 1998b) and

summarized in Table 5.

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TABLE 5 SUMMARY OF LIQUEFACTION MITIGATION TECHNIQUES

Liquefaction

Mitigation

Technique

Advantages Disadvantages Relative Cost

Vibro-Replacement Stone Column

Effective and economical method in many situations. Able to reach depths unattainable by other methods.

Ineffective for densifying soils with greater than 20% fine contents. The liquefiable soil should have a minimum thickness for this method to be effective. Waste spoils disposal is required.

Moderate

Grouting compaction grouting

Pinpoint treatment, Speed of installation, Wide applications range. Can be performed in very tight access and low headroom conditions, Non-hazardous, no waste spoil disposal. Able to reach depths unattainable by other methods.

Not effective at depths with low confining pressure (<15 feet). Ground surface heave due to grout pressure. Very low reinforcing effects of the compaction grout bulbs/columns.

Low to

moderate

deep mixing grouting

Wide applications range (even with high fine contents), Cost savings over deep foundation designs. Installation methods are customized for the site conditions.

Waste spoils disposal is required. Significant overhead clearance is required. Pinpoint treatment is not applicable. Very low reinforcing effects of the compaction grout columns.

High

permeation grouting

Minimum disturbance of the native soil. Can be performed in very tight access and low headroom conditions. Pinpoint treatment.

Construction process is complex. Very costly. limited to clean sands and ineffective in soils with fines.

High

chemical grouting

Minimum disturbance of the native soil. Can be performed in very tight access and low headroom conditions. Pinpoint treatment.

Construction process is complex. Very costly. limited to clean sands and ineffective in soils with fines.

High

jet grouting Nearly all soil types groutable. Most effective method of direct underpinning of structures and utilities. Safest method of underpinning construction. Ability to work around buried active utilities, can be performed in limited workspace, treatment to specific subsurface locations, no harmful vibrations. Much faster than alternative methods.

Soil erodibility plays a major role in predicting geometry, quality and production. Cohesionless soils are typically more erodible than cohesive soils. Pinpoint treatment is not applicable. Very low reinforcing effects of the compaction grout bulbs/columns.

High

3.2.7. Seismically-Induced Lateral Earth Pressures

Horizontal accelerations during seismic events will momentarily increase lateral earth

pressures on underground structures. The existing seawall will experience seismically-

induced earth pressures from a major earthquake on the regional faults. The seismically-

induced earth pressures are in addition to the static lateral earth pressures and should be

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considered in combination with the static lateral earth pressures. For a simplified analysis,

we recommend using an equivalent seismically-induced earth pressure with a rectangular

pressure distribution of F xH psf, where the coefficient F depends on the magnitude of the

ground acceleration and H is the depth to the bottom of the wall in feet. The resultant

seismic force would act at 0.5H above the base of the wall. The seismic earth pressures

are in addition to the static earth pressures and should be considered in design of the wall.

We recommend that a value of 16 be used for F to calculate the magnitude of the

seismically-induced earth pressure resulting from a 475-year return period earthquake. The

choice of the value of F to be used at the site depends on the level of risk accepted by the

designer. This level of risk involves the probability of occurrence of the design earthquake,

the consequences of distress/damage to the seawall, and other factors.

The magnitude of seismically induced earth pressures was calculated based on the

simplified procedures developed by Seed and Whitman (1970) and Ebeling and Morrison

(1992) and incorporated a reduction factor on the order of 25 percent to judgmentally

account for possible effects of wave scattering or passage, the transient nature of

earthquake ground motions, and possible wall-soil interaction effects.

3.3 EXCAVATION AND EARTHWORK 3.3.1. Site Preparation

Prior to site grading, pavements, existing slabs, piles, and decks should be removed and

debris should be disposed of outside the construction limits. Existing underground or under

the pier deck utilities located within the proposed construction areas, if affected by

construction activities, should be relocated prior to excavation. Debris generated from the

demolition of underground utilities, including abandoned pipes, should be removed from the

site as construction proceeds.

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3.3.2. Fills and Backfills

Utility trench backfills and below pavement granular base course materials will be placed

during construction of this project.

Fills and backfills may either be structural or nonstructural. Structural fills and backfills are

those defined as providing support to foundations, slabs, and pavements; nonstructural fills

and backfills include all other fills such as those placed for landscaping, and not planned for

future structural loads. Structural fills and backfills should be compacted to at least 95

percent of the maximum dry density as determined per ASTM D-1557; nonstructural fills

and backfills should be compacted to at least 90 percent of this criteria.

Structural fill and backfill materials should be placed in lifts not exceeding approximately 8

inches in loose thickness, brought to near-optimum moisture content and compacted using

mechanical compaction equipment. Nonstructural fills and backfills may be placed in lifts

not exceeding 12 inches in loose thickness and compacted in a similar manner.

Import fill may be necessary to achieve the design grades and should be placed and

compacted under the full time inspection and testing of the project geotechnical engineering

firm. Material to be used as compacted fill and backfill should be predominantly granular,

less than 3 inches in any dimension, free of organic and inorganic debris, and contain less

than 20 percent of mostly nonplastic fines passing the No. 200 sieve. The fill and backfill

soils should have a liquid limit less than 35 and plasticity index less than 12. Samples of fill

and backfill materials should be submitted to the geotechnical engineer prior to use for

testing to establish that they meet the above criteria.

3.3.3. Temporary Excavations

Excavations must comply with the current requirements of OSHA or Cal-OSHA, as

applicable. Additionally, all cuts deeper than 5 feet should be sloped or shored. Some of

the excavations may need to be shored; however, shallow excavations above the

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groundwater level may be sloped if space permits. It is our opinion that temporary

excavations to the west of the seawall may be sloped at 1H:1V (horizontal to vertical) or

flatter and 1.5H:1V or flatter above and below the groundwater level, respectively. The

groundwater is estimated to be as shallow as 2 feet below the existing ground surface;

however, it is the responsibility of the contractor to maintain safe and stable slopes or

design and provide shoring during construction. Flatter slopes will be required if clean or

loose sandy soils are encountered along the slope face.

Heavy construction equipment, building materials, excavated soil, and vehicle traffic should

not be allowed within 7 feet of the top of excavations.

Based on the results of our field exploration program and our review of available

subsurface data, it is our opinion that the majority of the proposed excavations can be

made using conventional equipment. However, in some locations obstructions may be

encountered.

3.4 FOUNDATIONS

3.4.1. General

The Port of San Francisco has instructed the project design team to prepare a seawall

design which is stable statically. The proposed renovation shall also make the existing

seawall withstand dynamic earth pressure due to a 475-year earthquake. . The Port does

not seek a renovation scheme which withstands pressures generated during a liquefaction

event. Recommendations provided in this section do not take into consideration

liquefaction-induced pressures. If liquefaction-induced pressure should be taken into

consideration in design of the project, it is recommended that either soil improvement

should be performed or the existing seawall should be founded on deep foundation system.

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3.4.2. Renovation Scheme

We understand that foundation grouting with full height revetment will be used for

renovation of the existing seawall.

Voids below the existing seawall should be filled completely with grout to provide intimate

contact between the entire seawall foundation and the bearing soils beneath the seawall.

The contact between the grout and seawall and between the grout and underlying soils

should be verified by the Geotechnic a Engineer. The Contractor is responsible to make

sure that grout does not leak into the Bay.

The most desirable type of revetment is rock riprap. We recommend using plated or keyed

riprap. Plated riprap is placed on the bank with a skip and then tamped into place using a

steel plate, thus forming a regular, well organized surface. Experience indicates that during

the plating operation, the larger stones are fractured, producing smaller rock sizes to fill

voids in the riprap blanket.

The median riprap particle size should be determined using the average velocity of flow,

rock riprap specific gravity of 2.65, and the riprap material's angle of repose of 40 degrees

and not less than Light Class as described in Section 72 of Caltrans Standard

Specifications.. we recommend using bank slope of 1.5H:1V or gentler.

After grouting beneath the existing seawall, a design allowable bearing pressure of 2,000

pounds per square feet may be used at the base of the seawall. The allowable bearing

pressure is a net value. Therefore, the weight of the foundation and the backfill over the

foundation may be neglected when computing dead loads. The bearing pressure applies to

dead plus live load and includes a calculated factor of safety of three. The allowable value

may be increased by one-third for dead plus live plus seismic loading conditions. If a

liquefaction event occurs, a significant loss of bearing pressure and subsequently bearing

failure is expected.

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3.5 SETTLEMENT

The primary consolidation of the Younger Bay Mud underlying the site is estimated to be

complete, therefore, settlements of the Younger Bay Mud under its own weight and the

weight of the existing fill are expected to be negligible. It is estimated that additional

settlements will occur as a result of placing additional rock riprap in front of the existing

seawall. The majority of the static settlements will be time dependent and will result from

consolidation of the Younger Bay Mud. The magnitude and time rate of the settlements will

depend upon thickness of the existing fill and the Younger Bay Mud. Our estimated

settlements are presented in Table 6.

TABLE 6

ESTIMATED SETTLEMENTS

Boring

No.

Estimated Bay Mud

Thickness (ft)

Settlement (in) Elapsed Time to

complete 90 percent of

primary consolidation

Thickness of Rock Riprap

1 ft 2 ft 3 ft

B-1 32 3 6 8 15 yr

B-2 20 2 5 7 5 yr

B-3 15 1 4 6 2 yr

If the estimated static settlements described above are not acceptable, we recommend that

the proposed grading plan be adjusted to limit the lateral extent and thickness of additional

fill required. Surcharging with or without wick drains before construction may be used to

reduce the amount of time-dependent settlements.

3.6 PERMANENT RETAINING STRUCTURES

Lateral earth pressures for level backfill condition may be calculated using the values

presented on Plates 4 and 5. The pressures should be computed for the height of the wall

bearing against the soil, except that for passive pressures the upper 1 foot should be

ignored.

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The presented values include hydrostatic forces. In addition, surcharge loads should be

considered where appropriate. Where traffic surcharge loads cannot be accurately

anticipated, a nominal uniform lateral surcharge pressure of 150 psf per foot may be used

for the full height of the walls, to a maximum depth of 20 feet.

To prevent excessive lateral forces from being applied to walls, heavy construction

equipment should not be allowed within 7 feet of the tops of the walls.

3.7 RESISTANCE TO LATERAL LOADS

Lateral resistance may be provided by passive pressure against foundations and by

frictional resistance against the bottoms and sides of those elements. Allowable passive

pressure in front of the seawall for different revetment geometry may be taken as

equivalent to the pressure exerted by unit fluids weighing presented in Table 7:

TABLE 7 ALLOWABLE PASSIVE PRESSURES

Revetment Geometry equivalent fluid unit weight (pcf

10 feet level bench changing to 1.5H:1V 400

2H:1V rock slope from the top of the revetment 140

3H:1V rock slope from the top of the revetment 180

At least 5 feet level bench changing to 1.5H:1V 270

Friction along the bottoms and sides of the foundations may be used in combination with

the above allowable passive resistance. The frictional resistance can be estimated by

using a coefficient of friction of 0.40. The net downward load on the foundations should be

used to estimate the base friction.

3.8 EXISTING STRUCTURS

The existing buildings are immediately underlain by uncontrolled fill and potentially

liquefiable soils. Some settlements and cracking may result from differential seismically-

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induced settlements of liquefiable soils during an earthquake and could damage the

building foundations. We recommend that either the existing structures be founded on

driven piles extending below liquefiable soils, or soil improvement be utilized at the

perimeters of the building to improve their performance against a liquefaction event.

3.9 CONSTRUCTION CONSIDERATIONS

3.9.1. General

Although the information in this report is primarily intended for the design engineers, data

from our borings will also be useful to the contractor. However, it is the responsibility of the

bidders and contractor to evaluate soil and groundwater conditions independently and to

develop their own conclusions and designs regarding soil densification, excavation,

grading, shoring, foundation construction, and other construction or safety aspects.

Extreme care should be exercised by the Contractor to avoid excessive deflections of the

existing seawall during repair, due to lateral deformation as the excavations are made in

front of the seawall. Where the existing fill in front of the existing seawall are removed, the

excavated area should not excess 10 feet in width, with 20 feet clear between the

excavated area; thus 3 increments of excavation and replacement would be required.

3.9.2. Geotechnical Services During Construction

AGS, Inc. should review project plans and specifications prior to construction to ascertain

that the geotechnical aspects of the project are consistent with the intent of the

recommendations presented herein. AGS should also be retained during construction to

observe the following items.

Site preparation and earthwork; Excavations for the working pits; Grouting and soil improvement; and Placement and compaction of fills, if any, and backfills.

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Our presence during construction will allow us to provide consultation regarding the

geotechnical aspects of the project. Our representative will observe the soil conditions

encountered during construction, verify the applicability of the recommendations presented

in this report to the soil conditions encountered, and recommend appropriate changes in

design or construction procedures, if the conditions differ from those described herein. In

addition we will take field density tests during the placement and compaction of engineered

fill and backfill.

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4. CLOSURE

This report has been prepared in accordance with generally accepted professional

geotechnical engineering practice for the exclusive use of Port of San Francisco and the

design team for the proposed Mission Bay Renovation project in San Francisco, California.

No other warranty, express or implied, is made.

The analyses and recommendations submitted in this report are based upon the data

obtained from three borings drilled for this study. The nature and extent of variations

between the borings may not become evident until construction. In the event variations

occur it will be necessary to reevaluate the recommendations of this report.

It is the responsibility of the owner or its representative to ensure that the applicable

provisions of this report are incorporated into the plans and specifications and that the

necessary steps are taken to see that the contractor carry out such provisions.

Respectfully submitted,

AGS, Inc.

Bahram Khamenehpour, Ph.D. Geotechnical Engineer 2104

Kamran Ghiassi, Ph.D. Geotechnical Engineer 2792

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for Cohesive Soils, Rep# UCB/EERC 88/15, Earthquake Engineering Research Center, University of California, Berkeley, CA.

Tokimatsu, Hohji, and H. Bolton Seed, 1987, Evaluation of Settlements in Sands Due to

Earthquake Shaking, J. of Geotechnical Engineering, V. 113, No. 8. Toppozada, T.R., 1984, History of Earthquake Damage in Santa Clara County and

Comparison of 1911 and 1984 Earthquakes, in the 1984 Morgan Hill, California Earthquake, California Division of Mines and Geology Special Publication 68.

Toppozada, T.R., Real, C.R., and Park, D.L., 1981, Preparation of Isoseismal Maps and

Summaries of Reported Effects for pre-1900 California Earthquakes; CDMG Open-File Report 81-11 SAC.

Trans Pacific Geotechnical Consultants, Inc., 1996, “Progress Report No. 2, Geotechnical

Engineering Services Proposed Public Boat Ramp Café / Bait Shop and Public Access Pier 52, San Francisco, California”, for the Port of San Francisco, December 3, 1976.

AGS

37

United States Geological Survey, 1989, Lessons Learned from the Loma Prieta, California Earthquake of October 17, 1989, Circular 1045.

Vucetic, M. and Dobry, R. Effect of Soil Plasticity on Cyclic Response, Journal of

Geotechnical Engineering, American Society of Civil Engineers, vol. 117, #1.

AGS

PLATES

AGS, Inc.CONSULTING ENGINEERS

PLATE 1

MISSON BAY SHORELINE PROTECTION PROJECT

SAN FRANCISCO, CALIFORNIA

JOB NO. Ki0301 DATE: 07/09

SITE LOCATION MAPN

PROJECT LOCATION

SITE

AGS, Inc.CONSULTING ENGINEERS

PLATE 2

Mission Bay Shoreline Protection Project

San Francisco, CA

JOB NO. KI0301 DATE: JUN 2009

SUBSURFACE EXPLORATION LOCATION MAP0 150 300 ft

USGS, courtesy of terraserver-usa.com2004, Aerial Photo

N

B-1

B-2

B-3

4

1-3

5-8

x

x

x

x

x

x

x

x

x

118

119

120

121

Trans Pacific Borings (1 through 8; 1996)

xx

x

x

x

x

x

x

122

123124

140

139

138137

136

135

134

133

132

Pier 54

Pier 64

Remains

AGS Borings (B-1, B-2, B-3; 6-9,10-09)

x Port of San Francisco Borings and Soundings (1961)

Mission Bay Shoreline

Protection Area Project

Area

Pier 52

153

Reference USGS Web Site

http://pubs.usgs.gov/sim/2004/2848/:

EARTHQUAKE EPICENTERS

AND FAULT MAPAGS, Inc.

CONSULTING ENGINEERS

MISSION BAY SHORELINE PROTECTION PROJECT

SAN FRANCISCO, CALIFORNIA

DATE: 7/09 PLATE 3PROJECT NO.: KI0301

SCALE

0 5 10 MILES

SITE

PLATE 4

MISSION BAY SHORELINE PROTECTIONS PROJECT

SAN FRANCISCO, CALIFORNIA

JOB NO. KI0309 DATE: 06/09

PRELIMINARY DESIGN PRESSURE DATA

NEAR BORINGS B-1 AND B-3

NOTES:

1. Add surcharge pressure to active pressures. If surcharge is kept 40 feet behind seawall,

lateral surcharge will be negligible.

2. Earth pressures include hydrostatic pressures.

3. Earth Pressures represents liquefaction condition of the existing fill.

4. .Hp is equal to 2 feet for loose rock riprap and 1 foot for plated rock riprap.

LEGEND:

H = DEPTH TO GROUNDWATER, IN FEET

Z =DEPTH BELOW EXISTING GROUND SURFACE, IN FEET

w

F

ZF

Existing

Seawall

H

ACTIVE

RESISTING PRESSURES

(psf)

DRIVING PRESSURES

(psf)

1

EQUATIONS (VARIABLES IN FEET, RESULTS IN PSF)

AGS, Inc.CONSULTING ENGINEERS

2

SEISMICALLY -INDUCED

EARTH PRESSURE

FINISHED GRADE

Hw=2 ft

Bay Mud

Fill

16.5 ft5

3

4

1 35Hw

3

2

Active Pressure

4

80Z -20Hwf

80Z -45Hwf

At-Rest Pressure

50Hw

140Z -90Hwf

122Z -72Hwf

Seismically-Induced Earth Pressure

16H

5

Passive Pressure

See Table 7 of the text

Hp

Zp

PLATE 5

MISSION BAY SHORELINE PROTECTION PROJECT

SAN FRANCISCO, CALIFORNIA

JOB NO. KI0309 DATE: 06/09

PRELIMINARY DESIGN PRESSURE DATA

NEAR BORING B-2

NOTES:

1. Add surcharge pressure to active pressures. If surcharge is kept 40 feet behind seawall,

lateral surcharge will be negligible.

2. Earth pressures include hydrostatic pressures.

3. Hp is equal to 2 feet for loose rock riprap and 1 foot for plated rock riprap.

LEGEND:

H = DEPTH TO GROUNDWATER, IN FEET

Z =DEPTH BELOW EXISTING GROUND SURFACE, IN FEET

w

F

ZF

Existing

Seawall

H

ACTIVE

RESISTING PRESSURES

(psf)

DRIVING PRESSURES

(psf)

1

EQUATIONS (VARIABLES IN FEET, RESULTS IN PSF)

AGS, Inc.CONSULTING ENGINEERS

2

SEISMICALLY -INDUCED

EARTH PRESSURE

FINISHED GRADE

Hw=2 ft

Fill

min 35 ft5

3

4

1 35Hw

2

Active Pressure

4

78Z -43Hwf

At-Rest Pressure

50Hw

87Z-37Hwf

Seismically-Induced Earth Pressure

16H

Hp

Zp

Passive Pressure

See Table 7 of the text5

AGS

APPENDIX A

FIELD EXPLORATION AND TESTING

AGS

AppendixAcover.doc July 2009

A-1

A.1 EXPLORATION

Field work was performed by AGS in June of 2009, including reconnaissance mapping and

marking of prospective boring locations, utility clearance, and drilling. Prior to drilling, the

ground surface at each proposed boring location was marked in white paint in order that

Underground Service Alert (USA) members could mark the locations of their respective

subsurface utilities. Locations of the borings are shown on Plate 2 – Subsurface

Exploration Location Map.

Drilling was performed by V&W Drilling, Inc. of Stockton, California on June 9 and 10, 2009

under the supervision of a qualified geologist experienced with area subsurface conditions.

Borings were drilled and sampled to depths ranging from 26.5 to 36.5 feet below the

existing ground surface. Following completion of drilling and sampling, each boring was

backfilled with cement-bentonite grout. The ground surface at the locations of Boring B-1

and B-3 was patched with quick-setting concrete, while the top of Boring B-2 was backfilled

with a 6-sack portland cement and pea gravel concrete mixture to avoid risk of future

subsidence in a sinkhole area. Backfilling was monitored by the San Francisco Department

of Public Health Inspector.

The subsurface conditions encountered in the borings were continuously logged in the field

during drilling operations by a geologist from AGS. Plates A-1.1 through A-1.3 - Log of

Borings B-1 through B-3, respectively, give descriptions and graphic representations of the

materials encountered, the depths at which samples were obtained, and the laboratory

tests performed. The legend to the logs is shown on Plate A-2 - Soil Classification Chart

and Key To Test Data.

AGS

AppendixAcover.doc July 2009

A-2

A.2 SAMPLING

Samples of the soils, as appropriate for the various earth materials encountered, were

obtained by AGS with a modified California (MC) sampler, Standard Penetrometer (SPT)

sampler, or Shelby tube. The Shelby tubes were pushed with the drill rig. The MC and

SPT samplers were driven with a 140-pound hammer, falling 30 inches for an 18-inch

penetration, where possible. The hydraulic pressure or blow counts required for advancing

the samplers were recorded for each 6 inch interval of sample and are shown on the Log of

Borings (Plates A-1.1 through A-1.3). Relatively undisturbed soil samples were collected in

2.5 by 6-inch stainless steel tubes from the MC sampler, plastic Ziploc bags from the SPT

sampler, or inside 3 by 24-inch steel Shelby tubes. The tubes were immediately capped,

sealed with tape, and labeled. The tubes were kept upright and cushioned from shock. All

samples were preserved in a cool, dark area until delivery to the AGS laboratory

3

SA(6)

103

110

105

7

12

14

19

2A B

5

LBG

X

KI0

301.

GP

J 6

/25/

09

NR

7A B

81119

358 4 710

31013 5

50/6"

1011 9

81330

1027

50/6"

1A B

6

ASPHALT PAVEMENT: 2-inches thick, poor condition. [FILL - AC]SILTY GRAVEL WITH SAND (GM), slightly moist, reddish-brown,medium dense, gravel is mainly fine, subangular, few coarse gravel.[FILL - AB]SILTY, CLAYEY GRAVEL WITH SAND (GC-GM) moist, dark brownand gray, medium dense, mostly fine subangular gravel, tracecobbles, some fine- to coarse-grained sand. [FILL]POORLY GRADED GRAVEL WITH SILT AND SAND (GP-GM) wet,brown, medium dense, few cobbles.

SILTY GRAVEL WITH SAND (GM) wet, yellowish-brown,medium-dense, mostly fine and coarse subangular gravel, tracecobbles, some fine- to coarse-grained sand. [FILL]

COBBLES AND BOULDERS (GP), wet, mainly sandstone. [FILL]

WELL-GRADED SAND WITH GRAVEL (SW) wet, medium dense, fineand coarse subrounded to subangular gravel. [FILL]

change to dense

POORLY GRADED GRAVEL (GP) very dense, wet, few cobbles.[FILL]

Boring completed to 26.5 feet below ground surface (bgs).Groundwater measured at depth of 6.20 feet bgs at 2:12 PM, duringdrilling.Boring backfilled with cement grout, pavement patched with asphalt.

TE

ST

S

DRILLING METHOD: 8-Inch Hollow-Stem & Rotary Wash

PROJECT: Mission Bay Shoreline Protection

HAMMER TYPE: 140-lb, falling 30 inches

DE

PT

H (

FE

ET

)

SA

MP

LE T

YP

E

DR

Y D

EN

SIT

Y

(PC

F)

MO

IST

UR

E

SA(12)

PLA

ST

ICIT

Y

4A

CHECKED BY: DH

SA

MP

LE N

O.

IND

EX

(%

)

AD

DIT

ION

AL

5

10

15

20

25

30

35

40

GEOTECHNICAL DESCRIPTION AND CLASSIFICATIONB

LOW

CO

UN

T

GR

AP

HIC

LO

G

LIQ

UID

LIM

IT (

%)

CO

NT

EN

T (

%)

BORING

B-1

PLATE A-1.1

LOG OF

SHEET 1 OF 1

SA(2)

JOB NO. KI0301

LOGGED BY: JF

DATUM:

SURFACE ELEVATION: ftDRILLING DATE: 6/9/09

DRILL RIG TYPE: CME 75

SA(2)

105

117

97

13

9

6

16

1A B

2A

3

4

6

8A B

LBG

X

KI0

301.

GP

J 6

/25/

09

10

92726

2256/6"

6 512

3 323

51217

122335

81527

4 711

12 912

SA(31)

5

9A B

ASPHALT PAVEMENT: 2 to 3" thick, poor condition. [FILL - AC]SILTY GRAVEL WITH SAND (GM) moist, reddish-brown, dense,gravel is mainly fine, subangular, few coarse gravel. [FILL - AB]SILTY, CLAYEY SAND WITH GRAVEL AND COBBLES (SC-SM)moist, yellowish-brown, black and greenish-gray, dense, some fine- tomedium-grained sand, few coarse-grained sand, some fine subangulargravel, trace cobbles. [FILL]

change to medium-dense, wet.

change to dense.

POORLY GRADED GRAVEL AND COBBLES (GP) wet,yellowish-brown and gray, medium dense, mostly coarse subangulargravel to 3-inch diameter, few cobbles, trace fine- to coarse-grainedsand. [FILL]

WELL-GRADED SAND WITH SILT (SW-SM) wet, darkyellowish-brown, dense, mostly medium- to coarse-grained sand, tracefine subrounded gravel. [FILL]cuttings sample indicates sand continues.

change to medium-dense.

Boring completed to 36.5 feet below ground surface (bgs).Groundwater measured at depth of 9.82 feet bgs at 11:30 AM, duringdrilling.Boring backfilled with cement grout to approximately 6 feet bgs.Top backfilled with 6-sack concrete mix poured to grade on 6/11/09.

TE

ST

S

DRILLING METHOD: 8-Inch Hollow-Stem & Rotary Wash

PROJECT: Mission Bay Shoreline Protection

HAMMER TYPE: 140-lb, falling 30 inches

DE

PT

H (

FE

ET

)

SA

MP

LE T

YP

E

DR

Y D

EN

SIT

Y

(PC

F)

MO

IST

UR

E

LIQ

UID

LIM

IT (

%)

SA(9)

IND

EX

(%

)

AD

DIT

ION

AL

PLA

ST

ICIT

Y

GR

AP

HIC

LO

G

BLO

W C

OU

NT

GEOTECHNICAL DESCRIPTION AND CLASSIFICATION

5

10

15

20

25

30

35

40

CHECKED BY: DH

CO

NT

EN

T (

%)

SA

MP

LE N

O.

7

BORING

B-2

SHEET 1 OF 1

LOG OF

PLATE A-1.2

DRILL RIG TYPE: CME 75

DRILLING DATE: 6/10/09

JOB NO. KI0301

LOGGED BY: JF

DATUM:

SURFACE ELEVATION: ft

57

SA(12)

81

4452

60

8

15

38

65

9169

95

1

3

LBG

X

KI0

301.

GP

J 6

/25/

09

5

6

NR

545

102

6 311

315111

SA(8)

80

4A B

SILTY, SANDY GRAVEL (GM), dry, yellowish-brown, loose, gravel isfine and coarse, subangular [FILL].POORLY GRADED GRAVEL WITH SILT AND SAND (GP-GM) moist,black and dark brown, loose, few organics, trace shell fragments,mostly fine subangular gravel, few coarse gravel and pieces ofconcrete rubble, some fine- to coarse-grained sand. [FILL]SILTY, CLAYEY GRAVEL WITH SAND, COBBLES, CONCRETERUBBLE (GC-GM) moist, brown, very loose, fine and coarsesubangular gravel, few concrete blocks from 1 to 2 foot diameter.[FILL]

SILTY, CLAYEY SAND WITH GRAVEL (SC-SM) wet,yellowish-brown, greenish-gray and black, loose to medium dense,some fine gravel. [FILL]

SILTY SAND (SM) wet, loose, fine- to medium-grained, tracecoarse-grained sand, few shell fragments. [Bayside Sand]SILTY FAT CLAY (CH) wet, dark gray, very soft. [Bay Mud]

change to stiff

(No sample recovery with pushed Shelby Tube.)

Boring completed to 27.5 feet below ground surface (bgs).Groundwater measured at depth of 8.32 feet bgs, at 9:35 am, duringdrilling.Boring backfilled with cement grout.

TE

ST

S

DRILLING METHOD: 8-Inch Hollow-Stem & Rotary Wash

PROJECT: Mission Bay Shoreline Protection

HAMMER TYPE: 140-lb, falling 30 inches

DE

PT

H (

FE

ET

)

SA

MP

LE T

YP

E

DR

Y D

EN

SIT

Y

(PC

F)

SA(12)

LIQ

UID

LIM

IT (

%)

2

AD

DIT

ION

AL

MO

IST

UR

E

GR

AP

HIC

LO

G

BLO

W C

OU

NT

GEOTECHNICAL DESCRIPTION AND CLASSIFICATION

5

10

15

20

25

30

35

40

CHECKED BY: DH

CO

NT

EN

T (

%)

SA

MP

LE N

O.

IND

EX

(%

)

PLA

ST

ICIT

Y

UC(.07)WA(99)UC

(.48)

PLATE A-1.3

B-3BORINGLOG OF

SHEET 1 OF 1JOB NO. KI0301

LOGGED BY: JF

DATUM:

SURFACE ELEVATION: ftDRILLING DATE: 6/9/09

DRILL RIG TYPE: CME 75

COARSE FRACTION

ML

WITH LITTLECLEAN SANDS

OVER 12% FINESGRAVELS WITH

NO FINESWITH LITTLE ORCLEAN GRAVELS

LIQUID LIMIT GREATER THAN 50

LIQUID LIMIT LESS THAN 50

SANDS WITHIS SMALLER THAN

OVER 12% FINES

MORE THAN HALF

NO. 4 SIEVEIS LARGER THANCOARSE FRACTIONMORE THAN HALF

Pt

OH

CH

MH

OL

NO. 4 SIEVE

OR NO FINES

SC

INORGANIC SILTS AND VERY FINE SANDS, ROCK FLOUR,

CLAYEY SANDS, POORLY GRADED SAND-CLAY MIXTURES

SILTY SANDS, POOORLY GRADED SAND-SILT MIXTURES

POORLY GRADED SANDS, GRAVELLY SANDS

WELL GRADED SANDS, GRAVELLY SANDS

MIXTURESCLAYEY GRAVELS, POORLY GRADED GRAVEL-SAND-CLAY

MIXTURESSILTY GRAVELS, POORLY GRADED GRAVEL-SAND-SILT

POORLY GRADED GRAVELS, GRAVEL-SAND MIXTURES

WELL GRADED GRAVELS, GRAVEL-SAND

CL

TYPICAL NAMES

SM

SP

SW

GC

GM

GP

GW

HIGHLY ORGANIC SOILS

SILTS AND CLAYS

SILTS AND CLAYS

GRAVELS

MAJOR DIVISIONS

SANDS

R-Value

SOIL CLASSIFICATION CHART AND KEY TO TEST DATA

eveis 002# < flaH naht ero

MSLI

OS

DE

NIA

RG

ENIF

eveis 002# > flaH naht ero

MSLI

OS

DE

NIA

RG

ES

RA

OC

Water Level after Drilling(with date measured)

Water Level at Time of Drilling

(with % Passing No. 200 Sieve)

Wash Analysis

(Shear Strength, ksf)

Unconfined Compression

Torvane Shear

Unconsolidated Undrained Triaxial

Cyclic Triaxial

Swell Test

KI0301 PLATE A-2.1

Terry Francois Boulevard, San Francisco, California

DATEJOB NO. June 2009

PP

INORGANIC CLAYS OF HIGH PLASTICITY, FAT CLAYS

Sieve Analysis

Pitcher Barrel

Standard Penetration Test

Modified California

ADDITIONAL TESTS AND KEY TO TEST DATA

UNIFIED SOIL CLASSIFICATION SYSTEMPEAT AND OTHER HIGHLY ORGANIC SOILS

NX Core Barrel

ORGANIC CLAYS OF MEDIUM TO HIGH PLASTICITY,

Bulk Sample

SANDY OR SILTY SOILS, ELASTIC SILTSINORGANIC SILTS, MICACEOUS OR DIATOMACIOUS FINE

PLASTICITYORGANIC CLAYS AND ORGANIC SILTY CLAYS OF LOW

LEAN CLAYSGRAVELLY CLAYS, SANDY CLAYS, SILTY CLAYS,INORGANIC CLAYS OF LOW TO MEDIUM PLASTICITY,SLIGHT PLASTICITYSILTY OR CLAYEY FINE SANDS, OR CLAYEY SILTS WITH

ORGANIC SILTS

TX

PM

DS

CP

CN

CA

(20)

WA

(1.2)

TV

TC

SW

SA

RV

Pocket Penetrometer

Permeability

Direct Shear

Compaction

Consolidation

Chemical Analysis

Sample Attempt with No Recovery

UC

Mission Bay Shoreline Protection Project Geotechnical Study

AGS

APPENDIX B

GEOTECHNICAL LABORATORY TESTING

AGS

AppendixBcover.doc July 2009

B-1

B.1 GENERAL

Preliminary visual soil classifications were made by AGS in the field in accordance with

ASTM D-2488 -93, Standard Practice for Description and Identification of Soils (Visual-

Manual Procedure). Upon completion of drilling, the samples collected from each boring

were taken to AGS’ laboratory or other testing laboratory for examination and analyses.

The soil classifications were verified by observation of the samples in the laboratory and by

completion of a testing program in accordance with ASTM D-2487 -93, Standard

Classification of Soils for Engineering Purposes (Unified Soil Classification System).

Geotechnical field and laboratory tests were performed on selected soil samples in order to

evaluate the engineering properties of the materials and to corroborate their classification.

Tests were completed for particle size, moisture content and density, Atterberg limits, and

unconfined compressive strength in accordance with the ASTM Standards discussed

below.

B.2 FIELD TESTING

The blows required to drive the MC and SPT samplers, using a 140-pound hammer falling

30 inches for an 18-inch penetration, were used to assist in classifying the relative density

of cohesionless soil deposits, and the stiffness of cohesive soil deposits. Blow counts

recorded by AGS in the field for each 6-inch interval of the 18-inch penetration are shown

on the Log of Borings.

AGS

AppendixBcover.doc July 2009

B-2

B.3 AGS LABORATORY TESTING

The laboratory tests were performed using the techniques and procedures discussed

below.

B.3.1 Moisture and Density Tests

Moisture content and density tests were performed on selected samples to evaluate their

consistencies and the moisture variation throughout the explored profile. The moisture

content was evaluated in accordance with ASTM D-2216 -92, Standard Test Method for

Laboratory Determination of Water (Moisture) Content of Soil and Rock, and was

considered to represent the moisture content of the entire sample for dry density

evaluation. The test results are presented on the Log of Borings at the appropriate sample

depth.

B.3.2 Particle Size

Particle size analyses were conducted on selected samples in accordance with ASTM D-

422-63, Standard Test Method for Particle Size Analysis of Soils or ASTM D-1140 -97,

Standard Test Method for Amount of Material in Soils Finer than the No. 200 (75-µm)

Sieve. The results of the particle size analyses are presented on Plates B-1.1through B-

1.2. The weight percent passing the No. 200 sieve are shown on the Log of Borings.

AGS

AppendixBcover.doc July 2009

B-3

B.3.3 Atterberg Limits Tests

Atterberg limits were evaluated on selected cohesive, fine-grained soil samples to assist in

their classification. Liquid limits, plastic limits, and plasticity indices were evaluated in

accordance with ASTM D-4318, Standard Test Method for Liquid Limit, Plastic Limit, and

Plasticity Index of Soils. The results of the Atterberg limits tests are shown on Plate B-2.

Liquid limits and plasticity indices are also shown on the Logs of Borings.

B.3.4 Unconfined Compressive Strength Tests

Unconfined compressive strength tests were performed on selected cohesive soil samples

to evaluate their strength characteristics. The tests were conducted in accordance with

ASTM D-2166, Standard Test Method for Unconfined Compressive Strength of Cohesive

Soil. The unconfined compressive strength test results are shown on Plates B-3.1 and B-

3.2.

95

0

5

10

15

65

30

35

40

45

50

20

60

100

70

75

80

85

90

55

1 0.0010.110100

25

0.01

267.8

Poorly Graded Gravel (GP)

12

14

19

13

6

3.32

1.72

1.46

PARTICLE SIZE ANALYSIS

25.8

Silty Gravel with Sand (GM)

2.6

3.7

B-1 @

B-1 @

B-1 @

B-2 @

B-2 @

6.5'

12.0'

2.21

B-1 @

Silty, Clayey Sand with Gravel (SC-SM)

B-1 @

Poorly Graded Gravel (GP)B-1 @

B-2 @

B-2 @

6.5'

12.0'

26.5'

3.5'

16.0'

Poorly Graded Gravel with Silt and Sand (GP-GM)

16.0'

Mission Bay Shoreline, Terry Francois Blvd., San Francisco

26.5' 1.4

42.2

5.0

6.2

12.1

2.2

31.4

1.8

34.0

Jun 2009

93.1

JOB NO. DATE

Mission Bay Shoreline Protection

KI0301

2.846

%Clay

19.00

25.00

37.50

19.00

37.50

7.93

13.13

26.43

28.8

26.57

3.5'

1.053

19.815

20.481

0.3072

10.1770

7.1365

59.9

59.1

96.4

26.41.45

103 2 1.5 1 3/4 1/2 3/8 3 4

8

HYDROMETER

14 16646

PERCENT

FINER

BY

WEIGHT

U.S. SIEVE OPENING IN INCHES

GRAIN SIZE IN MILLIMETERS

U.S. SIEVE NUMBERS

PL

SILT OR CLAYcoarse fine coarse medium fine

SAMPLE SOURCE

SAMPLE SOURCE

CLASSIFICATION

GRAVEL

LL

COBBLES

PI Cc Cu

D100 D60 D30 D10 %Gravel %Sand %Silt

MC%

PLATE B-1.1

SAND

20 30 40 50 70 100 140 200

1100

100

60

65

70

75

80

85

95

45

0.0010.010.110

90

10

55

5

50

15

20

25

30

35

40

0

1.37

Poorly Graded Gravel with Silt and Sand (GP-GM)

Silty, Clayey Sand with Gravel (SC-SM)

Silty Sand (SM)

16

8

15

38

1.48

26.5'

1.9110.0'

16.6

94.8

88.7

7.5

B-2 @

B-3 @

B-3 @

0.78

%Gravel %Sand %Silt %Clay

PARTICLE SIZE ANALYSIS

Well-graded Sand with Silt (SW-SM)

16.0'

B-2 @

B-3 @

B-3 @

B-3 @

26.5'

3.0'

3.0'

50.0

87.9

8.7

8.1

12.1

12.1

KI0301 Jun 2009

B-3 @ 0.0

JOB NO. DATE

Mission Bay Shoreline ProtectionMission Bay Shoreline, Terry Francois Blvd., San Francisco

0.551

10.0'

16.0'

9.50

25.00

19.00

4.75

1.84

10.99 34.2

0.45

89.6

1.000

0.624

0.193

0.1112

0.1160

1.7

57.7

37.9

D60

4.25

102 1.5 1 3/4 1/2 3/8 3 4

D10

86 14 16634

PERCENT

FINER

BY

WEIGHT

U.S. SIEVE OPENING IN INCHES

GRAIN SIZE IN MILLIMETERS

U.S. SIEVE NUMBERS HYDROMETER

SAMPLE SOURCE

SAMPLE SOURCE

COBBLESGRAVEL SAND

SILT OR CLAYcoarse fine coarse

140

fine

100

CLASSIFICATION MC% LL PL PI Cc Cu

D100

PLATE B-1.2

medium

D30

20020 30 40 50 70

12040 1008060

30

00

20

20

40

50

60

70

80

10

CH

MH or OH

ML or OL

38 57

PL

AS

TIC

ITY

IND

EX

(P

I)

KI0301 Jun 2009

CL

95

JOB NO. DATE

Mission Bay Shoreline Protection

99

Mission Bay Shoreline, Terry Francois Blvd., San Francisco

LIQUID

PLASTICITY CHART

SAMPLE SOURCE

CL-ML

LIQUID LIMIT (LL)

PLATE B-2

LIMIT (%)

PLASTIC

LIMIT (%)

PLASTICITY

INDEX (%)

% PASSING

#200 SIEVE

B-3 @ 20.5' Silty Fat Clay (CH)

CLASSIFICATION

200

0

50

150

400

250

500

450

300

350

100

UNCONFINED COMPRESSIVE STRENGTH

Strain

(%)

DryDensity

(pcf)

MoistureContent

(%)

B-3 @ 21.0' Silty Fat Clay (CH) 14 52

UC = Unconfined Compression

Ultimate

KI0301 Jun 2009

Mission Bay Shoreline, Terry Francois Blvd., San Francisco

JOB NO. DATE

Mission Bay Shoreline Protection

69.1

0

PLATE B-3.1

4 8 12 16 20

(psf)

UC

Type of TestClassificationSample Source

UN

CO

NF

INE

D C

OM

PR

ES

SIV

E S

TR

EN

GT

H (

psf

)

71

Strength

STRAIN (%)

200

0

50

150

400

500

450

250

300

350

100

UNCONFINED COMPRESSIVE STRENGTH

Strain

(%)

DryDensity

(pcf)

MoistureContent

(%)

B-3 @ 22.5' Silty Fat Clay (CH) 7 60

UC = Unconfined Compression

Ultimate

KI0301 Jun 2009

Mission Bay Shoreline, Terry Francois Blvd., San Francisco

JOB NO. DATE

Mission Bay Shoreline Protection

79.9

0

PLATE B-3.2

4 8 12 16 20

(psf)

UC

Type of TestClassificationSample Source

UN

CO

NF

INE

D C

OM

PR

ES

SIV

E S

TR

EN

GT

H (

psf

)

478

Strength

STRAIN (%)

AGS

APPENDIX C.docx C-1 July/2009

APPENDIX C

LIQUEFACTION POTENTIAL AND

SEISMICALLY-INDUCED SETTLEMENT

AGS

APPENDIX C.docx C-2 July/2009

C.1 LIQUEFACTION POTENTIAL

Appendix C presents the results of our liquefaction potential evaluation for the proposed

The Shoreline Protection Project at Mission Bay in San Francisco. The liquefaction

potential evaluation was based on the results of our field exploration program, whereby

blow counts were recorded by driving the California Modified (MC) and Standard

Penetration Test (SPT) samplers. The measured blow counts were later corrected for

various factors, as discussed below, and used in the liquefaction analyses for two levels

of design earthquake.

Peak horizontal ground surface accelerations were estimated using probabilistic

analyses, for events with return periods of 72 years, 475 years and 975 years as

discussed in Section 3.2.3 of the main text. A moment magnitude of 7.9 was assumed

for a 475-year return period event, corresponding to a horizontal ground surface

acceleration of 0.50g. Magnitude assumptions were made for the purpose of

performing the liquefaction analyses. Our liquefaction evaluations were made using the

procedure developed by National Center for Earthquake Engineering (NCEE),1996,

Youd (2001), AND Idriss and Boulanger (2008).

A comprehensive collection of site conditions at various locations where some evidence

of liquefaction was known to have or to have not taken place was collected by Seed and

others (1984). These data on sandy soils with a fines content less than 5 percent under

magnitude 7.5 earthquake conditions can be presented as relationships between field

values of average cyclic stress ratio, τav/σ'o (where: τav = average horizontal shear stress

induced by an earthquake; and σ'o = initial effective overburden pressure on the soil

element), and the SPT blow counts corrected for certain effects. For an earthquake of

magnitude 7.9, the cyclic shear stress ratio necessary to cause liquefaction in Seed’s

curve was corrected by a factor of 0.94.

For the first step in estimating liquefaction potential, the measured SPT blow counts

should be corrected for various factors using the method proposed by Seed and others

AGS

APPENDIX C.docx C-3 July/2009

(1984). The raw SPT blow count, N, is corrected to obtain the modified penetration

resistance, (N1)60. The modified penetration resistance is computed as follows:

(N1)60 = N x Cm x Cz x Ch x Cs x Cn (C.1)

where:

N: rawSPT or Mod Cal blow count (blows/ft)

Cm : a factor to correct for the larger size of the modified California sampler.

Raw blow counts using a modified California sampler were multiplied by 0.55

Cz: a factor that depends on the length of the drive rods; the following Cz factors may

be used for various depths:

Depth Cz

20 < x <30 ft 1.0

13 < x <20 ft 0.95

10 < x <13 ft 0.85

<10 ft 0.75

Ch: a factor that accounts for the hammer efficiency used in the field, where the blow

count is multiplied by a factor of 0.75;

Cs: a factor that depends on the sampling tube; for a split-spoon sampler without

liner (ID = 1.5" and OD = 2.0"), the following Cs factors may be used:

Raw Blow Count, N CS

< 10 1.0

> 10 1.2

Cn: a factor that depends on the effective overburden pressure at the depth when

the penetration test was conducted.

As presented by NCEE (1996), another correction factor, (N1)60, should be added to

(N1)60 to account for fine contents as follows:

(N1)60 Fine = (N1)60 + (N1)60

AGS

APPENDIX C.docx C-4 July/2009

The average cyclic stress ratio, τav/σ'o, at a specific depth can be estimated from

dynamic site response analyses. It also can be estimated with reasonable accuracy

from the following equation as discussed by Seed and Idriss (1982).

τav/σ'o=0.65 x amax/g x σo/σ'o x rd (C.2)

where:

amax: maximum acceleration at the ground surface

σo : total overburden pressure on sand layer

σo’: effective overburden pressure on sand layer

rd : a stress reduction factor.

Based on the magnitude of the design earthquake, and the peak ground acceleration

generated by that earthquake, the cyclic stress ratio was calculated using Equation C.2.

The cyclic stress ratio was then corrected to account for an earthquake magnitude other

than 7.5.

As Plates C-1 through C-3 show, the results of our analyses indicate a significant

potential for liquefaction of the upper fill and loose to medium dense on-site materials.

The Bay Mud, a silty clay, is not considered liquefiable, however there were thin, loose

sand lenses observed in the samples. These sand lenses should be considered

liquefiable, however they are mostly less than 6 inches thick and likely to be lenticular,

and are confined by the surrounding clay. The thin, lenticular nature of these sands and

the clay confinement should greatly reduce or eliminate potential ground surface

deformation due to liquefaction of these materials. The dense to very dense silty sand

and sand immediately below the Bay Mud would have a very low liquefaction potential

under a significant earthquake. The fill materials are also likely to lead to seismically-

induced settlement, as discussed in the following section.

AGS

APPENDIX C.docx C-5 July/2009

C.2 SEISMICALLY-INDUCED SETTLEMENT

We have estimated volumetric strain of liquefiable and non-liquefiable sandy soils using

different empirical methods based on the performance of liquefied sites in previous

earthquakes. The three methods of analyses include Tokimatsu & Seed (1987),

Ishihara & Yoshimine (1992), Based on the results of our evaluation, it is estimated that

the medium dense to dense sandy soils at the site could experience significant vertical

strain. Based on the estimated thicknesses of the sandy soils, the estimated

seismically-induced areal settlement of the site would range from approximately 2.0 to

6.0 inches.

18 110 35

7 120 517 120 5

14 120 514 120 5100 120 12

20 120 2

26 120 2

47 120 2

20 120 2

Silty Clayey Gravel

Gravel

Gravel With Sand

Sand With Gravel

Gravel

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LIQUEFACTION ANALYSISMISSION BAY SHORELINE RENOVATIONS

SAN FRANCISCO, CA Plate C-1

Hole No.=B-1 Water Depth=2 ft Magnitude=7.9Acceleration=0.5g

Raw Unit FinesSPT Weight %(ft)

0

10

20

30

40

50

60

70

Shear Stress Ratio

CRR CSR fs1Shaded Zone has Liquefaction Potential

0 1Soil Description Factor of Safety

0 51Settlement

SaturatedUnsaturat.

S = 3.12 in.

0 (in.) 10

fs1=1

32 110 31

100 120 3117 120 31

26 120 31

29 120 2

58 120 9

26 120 9

11 120 9

21 120 9

21 120 9

Silty Gravel

Clayey Sand

Gravel

Sand With Silt

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LIQUEFACTION ANALYSISMISSION BAY SHORELINE RENOVATIONS

SAN FRANCISCO, CA Plate C-2

Hole No.=B-2 Water Depth=2 ft Magnitude=7.9Acceleration=0.5g

Raw Unit FinesSPT Weight %(ft)

0

10

20

30

40

50

60

70

Shear Stress Ratio

CRR CSR fs1Shaded Zone has Liquefaction Potential

0 1Soil Description Factor of Safety

0 51Settlement

SaturatedUnsaturat.

S = 4.76 in.

0 (in.) 10

fs1=1

9 110 8

2 110 8

14 110 12

4 110 122 84 NoLq2 84 NoLq

6 84 NoLq

2 84 NoLq

2 84 NoLq

Silty Clayey GravelGravel

Clayey Gravel

Clayey Sand

Silty SandFat Clay

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LIQUEFACTION ANALYSISMISSION BAY SHORELINE RENOVATIONS

SAN FRANCISCO, CA Plate C-3

Hole No.=B-3 Water Depth=2 ft Magnitude=7.9Acceleration=0.5g

Raw Unit FinesSPT Weight %(ft)

0

10

20

30

40

50

60

70

Shear Stress Ratio

CRR CSR fs1Shaded Zone has Liquefaction Potential

0 1Soil Description Factor of Safety

0 51Settlement

SaturatedUnsaturat.

S = 5.43 in.

0 (in.) 10

fs1=1