TRANSPORT RESEARCH LABORATORY Department of Transport

RESEARCH REPORT 377

LOAD TESTS ON A REINFORCED BEAM AND SLAB BRIDGE AT DORNIE

by N J Ricketts and A McC Low

Crown Copyright 1993. The views expressed in this publication are not necessarily those of the Department Of Transport or the Scottish Office Roads Department. Extracts from the text may be reproduced, except for commercial purposes, provided the source is acknowledged. The work described in this paper forms part of a Bridges Engineering Division, DOT funded research programme conducted by the Transport Research Laboratory.

Bridges Resource Centre Transport Research Laboratory Crowthorne, Berkshire, RG11 6AU 1993

ISSN 0266-5247

The Transport Research Laboratory is no longer an Executive Agency of the Department of Transport as ownership was transferred to a subsidiary of the Transport Research Foundation on I st April 1996.

This report has been reproduced by permission of the Controller of HMSO. The views expressed in this publication are not necessarily those of the Department of Transport.

CONTENTS

Page

Abstract 1

1. Introduction 1

2. Condition of structure 1

2.1 Initial site visit 1

2.2 Independent inspection and assessment 1

3. Material properties 3

3.1 Historical data 3

3.2 TRL results 3

4. Opportunities for testing 3

5. Test 1: Crossbeam 4

5.1 Loading arrangements 4

5.2 Instrumentation 4

5.3 Test procedure 4

5.4 The Test 5

5.5 Results 5

6. Test 2: Deck Slab 8

6.1 Loading arrangements 8

6.2 Instrumentation 8

6.3 The Test 8

6.4 Results 9

7. Tests 3 and 4: Main Beams 12

7.1 Loading arrangements 12

7.2 Instrumentation 12

7.3 Test 3 (North Main Beam) 12

7.4 Results of Test 3 14

7.5 Test 4 (South Main Beam) 14

7.6 Results of Test 4 15

8. Analysis and assessment 19

8.1 Test 1: Crossbeam 19

8.2 Test 2: Deck slab

8.3 Tests 3 and 4: Main Beams

9. Discussion

10. Conclusions

11. Future work

12. Acknowledgments

13. References

14. Appendix A: Reinforcement details

Page

21

21

23

23

23

24

24

25

~i~i~,i~ ~

~i!!! ̧!~iiii~ii~! i̧i~/

i~ ! ~!~i~ ~

~ . . . . i I

r / :

._o E o ~3

c- O

J t - O 0

. J

2 0 t~

" 0 .~-

. Q

.m

.:-

LOAD TESTS ON A REINFORCED CONCRETE BEAM AND SLAB BRIDGE AT DORNIE

A B S T R A C T

One span of a concrete bridge across Loch Long at Dornie in the West Highlands was subjected to four separate insitu load tests on its main structural compo- nents. In each case the components tested were able to support a load far in excess of their assessed capacity. In this report the reasons for this are examined with a view to producing less conservative assessment methods.

1. I N T R O D U C T I O N

The bridge at Dornie was constructed in 1939 and carried the A87 trunk road over Loch Long near the Kyle of Lochalsh in the West Highlands. It comprised a central steel bascule span and fifteen spans of insitu reinforced concrete beam and slab construction. The concrete deck consisted of slabs spanning onto crossbeams which were supported by two longitudinal main beams. The main beams were cast into the piers and continuous over three spans. At every forth span a half joint at the point of contraflexure allowed for expansion. The piers supporting the deck were of twin column design cast onto a piled foundation. Details of the bridge are shown in figure 1.

Between 1973 and 1985 W.A.Fairhurst & Partners, Consulting Engineers, completed several reports and assessments on this bridge showing that marked deterio- ration had occurred. Given the results of the assess- ments, the Scottish Office Roads Directorate decided that the vehicle loading on the bridge would have to be reduced. This culminated in the restriction of traffic to one lane, controlled by traffic lights, together with strengthen- ing of the half joints in the deck.

The deterioration of the structure was continuing and as the road to Kyle of Lochalsh was programmed for major improvements, the Scottish Office Roads Directorate decided to commission a new bridge as part of the scheme. The replacement bridge was designed to a new alignment which allowed the main road to bypass Dornie village. TRL were offered the existing bridge for testing, a proviso being that the work was to be done during a three week period between the opening of the new bridge and commencement of demolition of the old structure.

2. C O N D I T I O N OF THE S T R U C T U R E

2.1 INITIAL SITE VISIT

The site was visited during October 1989 in order to examine the possibilities for testing and to gain some initial impression of the condition of the structure.

Dornie bridge was in an exposed location at the mouth of Loch Long. The structure had to withstand a mean tidal range of 3.2 metres with very fast flowing current due to the narrowing of the Loch at this point. Extremes of weather that occur in the area meant that the bridge had to withstand much worse conditions affecting durability than most UK bridges. This was reflected in its condition and the chloride levels in the concrete which were probably a result of wind blown salt from the Ioch.

The bridge was found to be in poor condition, particularly on the north side, with much spalling in evidence. A great deal of repair work had been done, large amounts of which were also deteriorated and it was difficult to establish in some cases which areas had been repaired and which had not. The soffits of the beams and slabs were badly spalled, showing in places severely corroded main and shear reinforcement. There were suspicions of shear cracks in one main beam and at one position the anchorage of a bent up bar was exposed.

2.2 INDEPENDENT INSPECTION AND ASSESSMENT

In order to establish the current condition of the structure tenders were sought for an independent Engineers inspection and assessment. W.A.Fairhurst & Partners were commissioned in June 1990 to produce an inspec- tion report, an assessment to BD 21/84 (Department of Transport 1984) and recommendations for testing. They were also asked subsequently to repeat their assessment calculations to the new assessment code for concrete bridges BD44/90 (Department of Transport 1990) which replaces BS 5400:pt4 (BS1,1984) for assessment.

The inspection found as expected, that the deterioration had increased since the previous report in 1985 (also by W.A.Fairhurst & Partners). The north side of the bridge had always suffered a greater deterioration than the south side and this was again found to be the case as the north side main beams showed severe spalling and reinforcement corrosion. All the crossbeams were noted as cracked at the soffit and some were found to have exposed badly corroded reinforcement. The deck slab was noted as basically sound other than Span 7 where there was severe spalling, rust staining and salt leeching out of the soffit of the slab. The half joints, which were strengthened in 1986, had continued to deteriorate and were described as in very poor condition. When as- sessed to BD 21/84, the condition of the main beams was such that a restriction of gross vehicle weight to 17 tonnes was advised.

• C - - - 1

i

z _o l--

l.U

...,.j LLI

I I I I I I

I I I I I I I I I r- I I I I

I I I I - . . . . I I I I I I ~ I I i ] I L I I I I I I I I

I I

~LZ_I7

z

o9 i.- o') U.l I-- i.i_ o z o I ' - -

(...) 0 ._ ]

o I

LI " ~I "6

l i

I I " c~ i i

I I I I I I I I I I I I I I I I I I I I El I I LI

rl i i i i i i

3. MATERIAL PROPERTIES

3.1 HISTORICAL DATA

An inspection by consultants in August 1977 highlighted the high levels of chlorides that were discovered (0.25% by weight of cement). Calculations indicated that the level would have only been 0.14% if the bridge had been constructed using seawater and sea dredged sand and aggregate. The difference was attributed to wind borne salt and de-icing operations. The concrete was also found to be of a porous nature and this it was felt contrib- uted to the take up of chlorides.

Tests on cores taken from the bridge in 1965 gave an "equivalent cube strength" of 59 MPa (sample age 27 years). This high result has never been confirmed in subsequent tests and whilst noted was viewed with suspicion. The inspection of January 1991 determined concrete strengths by taking cores from elements typical of those to be tested, the results are listed below.

Concrete

Longitudinal beams

Estimated insitu cube strength Worst credible strength E short term

= 26 MPa = 21.8 MPa = 25.9 GPa

Crossbeams

Estimated insitu cube strength Worst credible strength E short term

= 24 MPa = 21.5 MPa = 25.8 GPa

Bridge deck slab

Estimated insitu cube strength No core recovered

Characteristic strength = 15 MPa E short term = 24 GPa

Reinforcement

characteristic strength = 230 MPa E steel = 200 GPa

It should be noted that the area of the bridge deck which had been cored was found during testing to be severely deteriorated locally. During a period of heavy rain it became apparent that this deterioration had been caused by surface water penetrating the surfacing. The water had formed a drainage path along the surface of the slab under the footpath. The deteriorated concrete broke up very easily and for the subsequent collapse test it was difficult to find sound concrete on which to mount vibrat- ing wire gauges. The consultants could only take cores from this area as the bridge was then open and cores were unable to be taken from the carriageway. The deck slab when tested was found to be in much better condi- tion overall and therefore the concrete strength assumed was unrepresentative.

3.2 TRL RESULTS

During the testing and subsequent demolition of the bridge concrete cores and samples of reinforcement were obtained. These were chosen so as to give as full a representation as possible of the material strengths in the parts of the structure tested.

Concrete strength:

Twenty two cores taken from all parts of the structure were tested. The results indicated that the concrete was of similar strength and quality giving a mean strength of 30.4 MPa and a standard deviation of 5.7 MPa

Steel strength:

Specimens of reinforcement obtained from the structure were tested to BS EN 10 002-1:1990 (BSI, 1990) and mean values are tabulated below. As expected with a structure of this age, all the reinforcement was plain mild steel bar

4. O P P O R T U N I T I E S FOR TEST ING

Practical considerations limited the number of tests that were possible. Firstly it was not practical to provide the loading necessary to collapse a complete span. Secondly

TABLE 1

Reinforcement test results

Bar Yield Ultimate Sample Diam stress stress size

48mm 216 MPa 402 MPa 2

38mm 241 MPa 440 MPa 2

16mm 270 MPa 460 MPa 5

9.5mm 340 MPa 522 MPa 6

3

the rockhead, other than under the first two spans of the bridge, was too deep to be economic for the installation of rock anchors. Speed of the tide was also a problem limiting the erection of free standing scaffolding beneath the bridge.

It was decided to restrict the testing to Span 2 and to test the individual elements of the span insitu. This gave opportunities, with careful planning, for four tests with the object of determining the insitu strength of deck compo- nents with real boundary conditions. The tests carried out were as follows:

Test 1: Crossbeam loaded at midspan.

Test 2: Deck slab loaded centrally with a knife edge.

Test 3:

Test 4:

Main beam loaded at third point on deteriorated north side of bridge.

Main beam loaded at third point on south side of bridge.

The locations of these tests are shown in figure 2.

5. TEST 1: C R O S S B E A M

5.1 LOADING ARRANGEMENTS

The crossbeam was loaded with a single central point load. This was generated from two rock anchors using two 3000 kN capacity prestressing jacks acting on a 2m long steel crosshead, originally fabricated for TRL's

masonry arch tests. The rock anchor capacity was 1500 kN and the loading rig was therefore theoretically able to exert a load of 3000 kN on the beam. Load was meas- ured using the load cells in the prestressing jacks and checked against a load cell placed under the rig.

5.2 INSTRUMENTATION

The instrumentation for Test 1 was as shown in fig 3. Strain was measured using 14 surface mounted vibrating wire gauges positioned in a "back to back" arrangement above and below the crossbeam and slab. Small areas of the carriageway surfacing were cut out to allow the top gauges to be fixed directly on the concrete.

In doing this it was discovered that the concrete on the south side of the deck slab was severely deteriorated beneath the tarmacadam surfacing. Gauges were fixed with some difficulty to this concrete the condition of which was found to have been caused by the surface water penetration described earlier.

Deflection was measured at centre span and supports of the crossbeam using three rotary potentiometric type displacement transducers. These were fixed to the scaffolding and attached to the underside of the beam with a lightly tensioned wire.

5.3 TEST PROCEDURE

Load was applied in approximately 100 kN increments using the prestressing jacks. The instrumentation was scanned and logged after each increment was applied and a load/central displacement diagram updated on a computer screen. This enabled the behaviour of the

Infilled span 1

ierl t2

I L J

/

I .

"3 "1 - - -~ I I I I I

/ I I I

Test 3

F i g . 2 P l a n o f s p a n 2 s h o w i n g p o s i t i o n s o f T e s t s

4600

P i e r ~

I- I - - I I I I

/ I I

L___ L_

/ V Test 1

4

2 VW gauges placed above and below deck

O Displacement gauge

I I I

I I L__L

® 1/8

rl~ 5/12 I I I

¢,

I:• 2/9

I :~ 4/11

1 ~ 6 / 1 3

I I I I

/ I / f, i

I i _j

,~', 3/10,

I 3 I L I

sly,,

F i g . 3 P l a n o f c r o s s b e a m T e s t 1 s h o w i n g i n s t r u m e n t a t i o n l a y o u t

structure to be observed during the test and provided some indication of the approach of failure. The test was continued in increments until a load of 2240 kN was reached. After this the load on the structure was in- creased slowly to failure whilst continuous reading of the instrumentation was taking place.

5.4 THE TEST

For the first nine loading increments there was no discernable damage to the beam. A noise which sounded typical of concrete cracking was heard at increment 6 at a load of 533 kN but on inspection no cracks could be seen.

At increment 10, a load of 932 kN, small flexural cracks were found to have formed at the bottom edge of the centre section of the beam and also a small amount of cracking had developed at the supports which may have been associated with an earlier repair in this area. Also during this increment gauge No 8 at one of the supports was observed to cease functioning.

Over the next few increments the flexural cracking increased slightly until at a load of 1412 kN (increment 15) large 45 degree shear cracks developed in the beam, at either side of the load position. Between increments 15 and 21 these shear cracks became wider and increased in number.

At a load of 1995 kN (increment 21), the jacks had reached the limit of their stroke and had to be locked off and retracted. This took a few minutes to accomplish and the combination of time and the locking off procedure resulted in a reduction of the load. After locking off and retraction the jacks were reactivated and increased the load on the structure to 2240 kN (increment22).

From this point in the test it was decided to load the structure continuously whilst constantly scanning the instrumentation. The reasoning for this is to ensure that the failure achieved is not time dependent, which can happen with prolonged intervals between load incre- ments. The load was increased steadily and at 2530 kN the beam and slab failed simultaneously as a combina- tion of the beam shearing and the slab punching through on either side of it (figure 4). After failure the load cells indicated that there was still a load of 700 kN applied to the failed area which was being resisted by the deformed main reinforcement. Most of the links covering the area of the shear cracks had fractured in tension. There was very little indication at the road surface of the damage that had been produced below it.

5.5 RESULTS

From the results logged during the test, load/displace- ment curves and strain profiles were plotted. A graph of load against central displacement for the beam is shown in figure 5. It exhibits the characteristics that would be expected of shear failure - the rate of displacement with load showing little increase up to the failure point which was sudden and with little warning.

The strain profiles ( see figures 6-8) show a predomi- nantly elastic response at the centre of the beam with high compressive strains at the bottom of the section at the supports. This was indicative of considerable restraint developing at the supports causing axial compression of a large area of the section. It should be noted that no strains were obtained from gauge no 3 because it was mounted on unbonded footpath concrete. This is further described in section 7.4.

5

Fig.4 Crossbeam after fai lure: Test 1

2500

2000

1500

c~

3

1000

500

0 0

/ I I

Fig.5

5 10

D~splacement ,mm

Load/central d isp lacement : Test 1

-8OO 0

100

2OO

300 E E

o.. 400

E

m 500 -

600 -

700 -

800

- 7 0 0 f

-60O

1 -5OO

f

Strain (microstrain)

-400 - 3 0 0 - 2 0 0 -1 O0 0 1 O0

I r 1 T

O O 2531 kN

1790 kN

A 1511 kN

1197 kN

[ ] [ ] 740 kN

249 kN

e"

Fig.6 Strain profile through crossbeam at support (gauges 1 and 8): Test 1

-1000 O ~

Strain (microstrain)

- 500 0 500 1000 1500 2000 2500

I I I I

100

200

30O E E

400 13

E

500

O O 2531 kN

1993 kN

~ 1685 kN

932 kN

O O 740 kN

249 kN

600

700

800

Fig.7 Strain profile through crossbeam, midspan (gauges 1 and 9): Test 1 r

r- ~ f

- 9 0 0

0

100 -

200 -

300 - E E

400 --

E

500 --

600

700

800

- 8 0 0 - 7 0 0 - 6 0 0

S t ra in (m ic ros t ra in )

- 5 0 0 - 4 0 0 - 3 0 0 - 2 0 0 - 1 0 0 0

I I

O O 2 5 3 1 kN

H 1790 kN

L ~ 1511 kN

1197 kN

[ ] [ ] 740 kN

H 249 kN

I I I I 1 I

F ig .8 S t r a i n pro f i l e t h r o u g h c r o s s b e a m at s u p p o r t ( g a u g e s 3 a n d 10): T e s t 1

6. TEST 2: DECK SLAB

6.1 LOADING ARRANGEMENTS

The deck slab was loaded with a 1200mm knife edge load in the form of a 200mm stiffened universal column section, beneath a large trapezoidal shaped load spreader. This was positioned across the carriageway, centrally about both axis of the slab and represented the effect of the offside wheels of two passing heavy vehi- cles. The crosshead and jacks used for Test 1 were assembled on the load spreader and used as before to apply load to the structure. Load was measured using the calibrated load cells in the prestressing jacks.

6.2 INSTRUMENTATION

The instrumentation for Test 2 was as shown in fig 9. Strain gauges were again placed back to back on the top and bottom surfaces of the slab. One pair of gauges was arranged to record strain in the longitudinal direction at the centre of the slab with 4 pairs recording strain perpendicular to the centre of each of the four edges. Problems were again experienced fixing gauges to the top surface of the slab on the south side of the bridge for the same reasons as described in section 5.2.

Displacements were measured using six rotary potentiometric transducers fixed to the scaffolding.

These were positioned at the centre of all four sides of the slab and also under each end of the knife edge loading position (figure 9).

6.3 THE TEST

Load was applied in approximately 100 kN increments with instrumentation scanned after each increment up to a load of 2050 kN. Beyond this the slab was loaded continuously and data logged until failure occurred.

The first few load increments, as expected, did not result in cracking of the slab. The first small cracks appeared at a load of 600 kN (increment 7) but it was not possible at that stage to establish whether they were part of a general pattern or some pre-existing cracking that was opening under load. The cracking pattern became more extensive over the next five increments. It became apparent that on the west side of the slab cracks were forming in the areas where yield lines would be expected - from under the end of the knife edge running back to the corners of the slab and also under the knife edge itself. Surprisingly no cracking was found on the other two diagonals.

The loading was continued in increments to 2056 kN (increment 23) with some increase in cracking, as far as could be determined given the safety considerations of approaching such a heavily loaded slab.

8

Pier 1

I I I I I®

I I I I I

1/6 nO

2/9

®1] 3/8

® i,---i

4/9

I:::=! 2 VW gauges placed above and below deck

~) Displacement gauge

Fig.9 P lan of s lab T e s t 2 s h o w i n g i n s t r u m e n t a t i o n l a y o u t

The final failure of the slab was in punching and occurred at a load of 2900 kN. It was complete over the whole length of the knife edge which remained parallel to the road surface. No tilting of the loading rig in either direc- tion occurred during failure.

A great deal of the punched material was retained in its approximate position by the deformed lattice of reinforce- ment within the slab and could be examined in more detail than is usual, for this type of failure. Longitudinally, half of the slab reinforcement was curtailed beyond the central portion by bending up the bars and this line of curtailed bars appeared to be the limit of the punching failure in this direction. A photograph of the failed area of the slab is shown in fig 10.

Apart from a roughly rectangular hole where the load was applied there was again little indication at the road surface of the damage to the slab below it. The reinforce- ment across the punched area continued to support a residual load of 350 kN from the rig at the end of the test.

6.4 R E S U L T S

As with Test 1 load/displacement graphs and strain profiles were plotted. Unfortunately strain gauges 2 and 3 did not work successfully which meant in the case of gauge 2 that a strain profile for the centre of the slab could not be plotted.

The load/displacement graph shown in figure 11 is again indicative of an abrupt failure, in this case punching of the slab. As described in section 6.3, yield lines indicated by cracking were forming before the punch occurred on the pier 2 side of the slab. This becomes important when interpreting the strain profiles shown in figures 12-14.

The strain profiles shown in figures 12-14 relate to points at the boundary of the slab. Figures 12-13 are longitudi- nal strains adjacent to the crossbeams. There is no apparent explanation for the significant difference between their compression strains. The high tensile strain shown in figure 13 indicates a crack within the gauge length. No crack was found but the top surface of the slab was uneven because of the removal of surfacing and this could have concealed a crack. Figure 14 shows trans- verse strains adjacent to a main beam close to the position where the nominal top transverse reinforcement is curtailed.

,.bo" ~ " ~ ,,a .

.% , .

Fig.lO Underside of slab after failure: Test 2

3000

2500 -

2000

15oo

S

1000

5OO

I 1 5 10

D~sptacement /mm)

Fig.11 Load/central displacement: Test 2

10

-5O0 0

50,

-~- lOO E

v

o _

7 o

.Jo

150 -

200 -

250

-400 -300

I I

O O 1967kN

H 1727 kN

~ 1324 kN

931 kN

[ ] O 502 kN

108 kN

Strain (microstrain)

-200 -100

I I 0 100 200

Fig.12 Profile of longitudinal strain through slab next to c rossbeam (gauges 4 and 9): Test 2

-1500 0

5O

"~ lOO E

1 5 0 - -

200 --

25O

Fig.13

-1000 -500

I I

O O 1967kN

1727 kN

~ 1324 kN

931 kN

O O 502 kN

108 kN

Strain (microstrain)

0 500 1000 1500 2000

Profile of longitudinal strain through slab next to c rossbeam (gauges 5 and 10): Test 2

11

- 1 0 0 0

0

5 0 -

~ 10o - E

O_ o~

t-,

03 150 -

200 -

250

Fig.14

-500 0 500

St ra in (m ic ros t ra in )

1000 1500 2000 2500 3000

O O 1967 kN

1727 kN

~- Z~ 1324 kN

931 kN

[ ] 13 502 kN

108 kN

P r o f i l e o f transverse strain through slab next to main beam (gauges 1 and 6): Test 2

7. TESTS 3 A N D 4: MAIN BEAMS Two main beam tests were attempted because of the differing condition of the north and south beams. The north side of the bridge had suffered severe spalling. Its repairs had also suffered and it was in a far more deterio- rated condition. It was hoped that by testing both beams the effects of that deterioration might be established. The span tested was generally in better condition than other more exposed spans near the centre of the bridge. However testing was not possible at these locations for the reasons stated in section 4.

7.1 LOADING ARRANGEMENTS The main beams were loaded with a single point load in a similar manner to the cross beam test. The test rig was positioned at a point 4.6 m from pier 2 to try to produce a combined flexural and shear failure as suggested in the assessment carried out by W.A.Fairhurst & Partners. Load was applied using a large 600 tonne capacity crosshead designed specifically for the tests. A combina- tion of two 2500 kN capacity ground anchors and two 3000kN prestressing jacks were used to react against the crosshead and apply load to the structure. Load was again measured using the jack load cells.

7.2 INSTRUMENTATION The instrumentation for Tests 3 and 4 was planned to be identical However, due to spall ing on the main beams, some of the gauges had to be repositioned to ensure

they were attached to firm concrete. In addition to this displacement transducers were positioned at the end of the beam haunches on Span 3 so that any rotation at pier 2 could be measured. The layout of the gauges for Tests 3 and 4 are as shown in fig 15. Vibrating wire gauges were again placed above and below the beams in a back to back arrangement, the top gauges being placed in recesses cut in the tarmacadam. The displacement transducers were fixed to the scaffolding and attached to the beams in the same manner as the previous tests.

7.3 TEST 3 (NORTH MAIN BEAM) The test commenced with a few small increments to take the slack out of the rock anchor cables and to make sure the crosshead was positioned correctly and stable. This was particularly important for the main beam tests because of the high loads required. After the initial load was applied to the bridge the test was progressed in approximately 200 kN increments. A photograph of the beam under test is shown figure 16.

The first signs of damage were noticed at a load of 1600 kN (increment 11) when some flexural cracks were found in the soffit of the beam between the load position and the centre of the span. Also cracks were noticed at pavement level between the non structural concrete of the footpath and the back of the parapet. Over the next two increments the flexural cracking extended 100mm above the soffit of the beam but no other damage was noted.

12

® / / I • • • 3 / 6

I I I

Infilled span 1 I I

I I I I

I I

[ ~ . . . . I - ~ 3 / 6 - -

7®

I

_---- ,,~C ~. 'R

/

t - ~ 2 VW gauges placed above and below deck

• O Displacement gauge

1 3 8 1 7

I ®

I iI

.... l i ,,o

2 7 6 4 I _ 4 6 0 0

"1 p ®1. ~est4 0

F . . . .

z I I I i t I I I I I ,

I I i

I L I l ,

T®22= ®

,4

-I I

\

®

I I 1 I I I

I

®

Fig.15 Plan of span 2 s h o w i n g i n s t r u m e n t a t i o n l a y o u t f o r T e s t s 3 and 4

Shear cracks appeared at a load of 3000 kN (increment 18) on either side of the load position. The full extent of these cracks was difficult to determine as the concrete repairs to the lower edges of the beam may have given a false impression of how the cracking was progressing.

At the commencement of increment 20 loading was suspended at 3220kN when a large section of repair material fell from the centre of the main beam in the adjacent Span 3. This was not entirely unexpected as the main beams were continuous. Displacement gauge5, under the haunch of Span 3 had recorded some upward displacement on that side of the pier. This must have been enough to detach some deteriorated repair concrete at mid span.

Following this, a close inspection was made of the pier area which revealed that some of the existing cracks in the main beam haunch (Span 3) had extended and also the cracks in the crossbeam (Test 1) had spread at the point where it joined the pier. When the next loading

increment was applied (3410 kN) the displacement gauge in Span 3 did not record any change. The pier area was again inspected to see if the reason could be determined and a hogging crack was discovered over the pier running through the parapet and into the beam.

The load was increased to 4000 kN (increment 24) and the jacks were locked off and retracted as they were reaching the end of their stroke. Two more hogging cracks over the pier were found and at the other support, a crack had opened between the end of the parapet and the existing stone wing wall.

The load was again taken on the jacks and the test continued. Three more load increments were applied to increase the load to 4590 kN. An inspection revealed a forth hogging crack over the first pier, additional cracking in the beam haunches and a large flexural crack at centre span.

A further three load increments were applied whilst the beam was observed using binoculars. At a load of 4970

13

Fig.16 North main beam during testing: Test 3

kN a bang, later found to be a rock anchor strand break- ing, was reported by the jack operators. The instrumenta- tion at this point showed the beam to be creeping. A further attempt to continue the test resulted in the load increasing to 5170 kN. No further increase was possible due to more tendons breaking and the test was stopped for safety reasons while a safe destressing procedure could still be adopted.

7.4 RESULTS OF TEST 3

A load/displacement curve for the gauge under the load position is shown figure 17. It indicates that the beam's maximum load capacity was unlikely to have exceeded 6000kN. Strain profiles for this test are of little use because the top gauges could not be mounted directly to the beam. These were fixed to an added insitu concrete footpath which proved to be poorly connected to the structural concrete. Readings on the bottom gauges also proved unreliable due to the large number of concrete repairs to the beam. The limits of sound bonded concrete were difficult to determine.

7.5 TEST 4 (SOUTH MAIN BEAM)

Load was again applied in 200 kN increments and the logging procedures were the same as for Test 3. During the initial load increments there was little sign of any damage to the beam. Some of the existing cracks were

noted to be extending at a load of 600 kN and small flexural cracks were found at a load of 1000 kN. These cracks continued to extend over the next few load increments and the first shear cracks occurred at a load of 2000 kN (increment 11 ).

Over the next three loading increments the cracking increased markedly with the flexural and shear cracks lengthening considerably. The flexural crack directly under the load lengthened to over half the depth of the beam from increment 10 to increment 13, a load increase of 600 kN.

At a load of 3000 kN (increment 16 ) two more shear cracks were found on the short shear span side of the load position. The cracking pattern was now more extensive than had been the case with Test 3 at this level of load despite the beam appearing to be in better overall condltton.

The load was again increased which produced another shear crack at 3400 kN (increment 18) and evidence of the beam creeping at 3787 kN (increment 20).

The jacks reached the end of their stroke at a load of 3990 kN and were locked off and retracted. On inspection of the parapet a crack, similar to that which had occurred in Test 3. was found over the top of the pier. The jacks were restressed back to their lock off load and the test was continued.

14

Z

"10

O _J

6000

5000

4000

3000

2000

1000

0 L I i I 0 10 20 30 40

Displacement (mm)

Fig.17 Displacement at the load position: Test 3

50

The load continued to be increased in increments to 4970 kN (increment 27). During this increment the large central flexural crack that had formed at increment 13 opened considerably to approximately 10mm. At the road surface the crack over the pier had extended across the carriage- way towards the north side parapet.

Two further load increments brought the load to 5310 kN when as with Test 3 tendons started to break. The test was ended at increment 30 at a load of 5258 kN and the destressing procedure again adopted to release the estimated 4000 kN still locked in the tendons after the jacks had been retracted. On release of the rock anchor tendons the beam was seen to recover a considerable amount of displacement relative to the independent scaffold access tower adjacent to the parapet at mid span. A photograph of the beam after test is shown in figure 22.

7.6 R E S U L T S OF T E S T 4

As with Test 3 the shape of the load/displacement graph shown in figure 18 indicates that the beam's maximum load capacity was unlikely to have exceeded 6000 kN. The strain profiles (see figures 19-21) show high compressive strains in the beam haunches indicating that most of the section was in compression at these points. Beneath the load point a high tensile strain developed despite the gauge being close to a large flexural crack running from the bottom of the beam to the slab soffit (figure 23). The presence of such a crack would tend to result in a local reduction of strain in the adjacent concrete.

15

6000

5000--

4000

3000 2

2000

1000

0 I I I I 0 10 20 30 40

Displacement(mm)

Fig.18 Displacement at the load position: Test 4

50

-900 -800 -700 -600 o [ T

E~E 500 I

1000 E

1500 --

2000

O O 5252 kN

H 4721 kN

h__ -~ 3984 kN

" "-- 3198 kN

[] [ ] 2195 kN

998 kN

Strain (microstrain)

-500 -400 -300 -200 -100

I T T ] [ 100 200

Fig.19 Strain profi le through beam at pier 1 (gauges 3 and 6): Test 4

16

-500 Or---

5OO

T

Strain(microstrain)

1000

1 1500 2000

T 2500

E

¢-

E

m

200

400

600

800

O O 5252 kN

4721 kN

~ 3984 kN

3198 kN

O O 2195 kN

~ 998kN

1000

1200

Fig.20 Strain profile through beam at load point (gauges 2 and 5): Test 4

-900 0

Strain (microstrain)

-800 -700 -600 -500 -400 -300 -200 -100 0

I I I I I 100

E

r -

E

a3

500

1000

O O 5252 kN

4721 kN

L & 3984 kN

~ 3198 kN

[] O 2195kN

998 kN

1500

2000

Fig.21 Strain profile through beam at pier 2 (gauges 1 and 4): Test 4

17

Fig.22 South main beam after testing: Test 4

, > . " + ; D , ; ° ' ° ~ , ,

_ + , , > _

Fig.23 South main beam textural cracking: Test 4

18

8. ANALYSIS A N D A S S E S S M E N T

For each of the elements tested a range of different analyses can be made to assess the strength of the element. The following is a list of generic types of analysis which have been applied to Dornie Bridge.

Assessment to BD 21/84 and BD 44/90 Assessment calculations were prepared by W.A.Fairhurst & Partners for all parts of the bridge superstructure. These calculations use the HA loading given in BD 21/84 and Amendment 1.

Test predictions using BD 21/84 Using the strength values derived in the assess- ment without their partial factors predictions were made by W.A.Fairhurst & Partners of the capacities of the various elements under the test arrangement.

Best estimate limit analysis capacities Code methods need to be universal and they must be safe. Necessarily they miss many beneficial effects either because the effect may not always be present or because the effect cannot yet be quantified with sufficient confidence. The best estimate analysis ignores this caution and uses methods taken from the very wide range of pub- lished papers available. It also uses the material strengths as measured. The reference to "limit analysis" in the heading indicates that this current report covers only initial studies which looked at the conditions at failure.

Limit analysis predictions These use the "best estimate" analysis methods but the material properties are those which would be used for a BD 21/84 assessment. In this report they are not strictly predictions because they were made after the test. They are intended to represent predictions which could be made for a future, similar test.

The results of these analyses are given in Table 2 and their derivation is explained below.

8.1 TEST 1: CROSSBEAM

The BD 44/90 analysis predicts an ultimate load of 750 kN in combined bending and shear. In the test the beam failed in shear at a load of 2480 kN which is 3.3 times greater than the code prediction.

The "best estimate" calculation below gives an ultimate load that exceeds the code prediction for two principal reasons: firstly because end restraint forces have been allowed for and secondly because there was some additional strength in the reinforcement.

A strut and tie model has been used for the analysis (figure 24) as advocated by Schlaich, Schafer & Jennewein (1987) and Schlaich & Schafer (1991). This is able to model the restraining forces in the end regions, Schlaich's D regions. The analysis is performed in pencil on the drawing board. A pattern of struts and ties is guessed. Limit values for certain gravity loads and reinforcement tie forces are calculated and then equilib- rium is checked using a Maxwell-Bow diagram (Coates, Coutie and Kong 1990). Both geometric and equilibrium requirements are satisfied graphically. An iterative procedure is followed with slight changes being made to angles and positions in order to maximise the load capacity. In responding to the graphical constraints the assessor derives a direct feeling for the real constraints. A model which is close to the limit, if not on the limit, can be drawn quite quickly and the failure load is predicted.

To maximise the capacity of the model the struts should be kept as narrow as possible so their width is chosen to give a compressive stress of, say, the cylinder strength of the concrete. These widths must be calculated and then the drawing is adjusted to accommodate them.

In this case the diagram has been drawn for the known failure load with the half load drawn as 1240 kN. The bottom reinforcement of the beam is assumed to be just short of its yield because there were no large flexural cracks during the test. The length da is shown to be 1250 kN.

TABLE 2

Predictions from analyses

Test 1 Test 2 Test 3 Test4

BD 21/84 prediction (kN)

Best estimate capacity (kN)

Limit analysis prediction (kN)

Observed Failure load (kN)

750 1250 1130 1130

2460 2572 2940* 2940*

1620 1955

2480 2850 >5174 >5258

* Ignores 3 dimensional actions.

19

\

lr x ~

b

Maxwell-Bow Diagram of forces / / [ g - ~ H

f,h

l C

Half load

= 1240 kN

Loading pad

+ I d

Fig.24 Strut and tie model for Test 1

The greatest uncertainty in the diagram is the magnitude of the restraint forces V and H mobilised at X. The diagram was constructed with these as unknowns which were derived by drawing. The parapet is stiff enough to collect vertical dead load on its own line. V has a value of 960 kN which is 19% more than the deadweight arriving at that column. The yield strength of just one column bar, the 32mm bar at the apex of the column section, is enough to cover this deficit although the drawings show it with very little anchorage length above the top of the column.

The six horizontal reinforcing bars which anchor at X have a combined yield strength of 730 kN and an ultimate strength of 1350 kN. The value of H is 2050 kN. The difference must be collected by lateral spanning action in the deck. The edge beam is heavily reinforced so this is probably not a limitation.

On the diagram a chain-dotted line has been drawn from the edge of the loading plate at an angle of 22 degrees to the horizontal. This is the shallowest angle which is

allowed when using the variable strut inclination method in the draft Eurocode EC2 (Commission of the European Communities, 1989). Beams represented by strut-and-tie models with strut angles shallower than this may fail prematurely in shear. This is the phenomenon which causes the "valley of diagonal failure" (Kani (1966), Russo, Zingone and Puleri (1991)). The term "valley" is used to indicate the region on a plot of test results where the strengths fall short of the flexural capacity. Even though Kani's work was related only to beams without shear links it indicates that, in general, a separate prediction is required for shear failure.

A similar situation is considered by Nielsen (1984). A concentrated load is carried on a beam with a shallow strut model which he shows in his figure 5.19. The angle shown is much less than 22 degrees. However he uses it with an empirical concrete strength which depends on the shear span ratio. He does not give the formula but it presumably recognises the premature failure that can occur with shallow struts.

20

There are a plethora of theories for predicting shear strengths in reinforced concrete beams. The 'Variable strut inclination method" from EC2 was chosen because it was known to give good predictions, it was simple and it links easily with the strut-and-tie model. Shear is assumed to be carried by a lattice truss of links and diagonal compression struts in the concrete. Hence all the shear is carried through the links and the method cannot be applied to members without links.

For the shear failure check it was assumed that the force ef is shared between all the shear reinforcement which is anchored both sides of the chain-dotted line. There are eight sets of links which are circled in figure 24 and one diagonal bar. Together these have an estimated yield strength of 1050 kN which is very close to the value on the diagram of 1060 kN. Hence the best estimate load capacity is 2460 kN. Note that the EC2 method has been followed by using the yield stress in the assessment despite the fact that many of the bars were ruptured in the test and must have reached a significantly higher stress. The method has been calibrated using yield stresses and is hence only valid when used with yield stresses.

For the limit analysis prediction the analysis was re- worked using a reinforcement yield strength of 230 MPa which is that suggested in BD 21/84 (DTp 1984).'The resulting diagram is indicated in chain dots on figure 24. It is governed by yield in the links and gives a predicted failure load of 1620 kN, 65% of the test value.

8.2 TEST 2: DECK SLAB

The BD 44/90 prediction of 1250 kN is based on a yield line analysis. This assessment was made after the test because W.A.Fairhurst & Partners had not been given the reinforcement details for the slab. In the test the slab failed in punching at a load of 2850 kN which is 2.3 times greater than the prediction.

For the "best estimate" analysis the yield line capacity was considerably enhanced by including the effect of compressive membrane action (CMA) which is due to the restraint forces on the slab around its perimeter. CMA also enhances the punching shear capacity but it is difficult to quantify this. Regan & Braestrup (1985) devoted a 240 page document to punching shear. Their only reference for the enhancement is Hewitt & Batchelor (1975) who give a method which can be used when the restraint forces are known. When the restraint forces are unknown they offer some empirical restraint factors for point loads. The interpretation of these for the finite line load used at Dornie would be a problem. Regan & Braestrup (1985) state "A substantial proportion of the test results reported in the literature as punching shear failures exhibit ultimate loads which do not differ signifi- cantly from the flexural capacity". The mechanisms of the two failures are closely linked. For the present study a CMA enhanced yield line analysis has been used.

The method which allows for CMA enhancement is taken from Rankin, Niblock, Skates & Long (1991). This covers distributed loads on a restrained slab but the failure mechanism and internal work equations can be applied to the line load used at Dornie. Their expression (2) is

modified because the diagonal yield line will not be at 45 degrees and the reinforcement is different in the two directions. If the length of the line load is LL then expres- sion (2) becomes:

Internal work done = 4x ~x x(My + My')

+4x Lx x(Mx +Mx') Ly-LL

where:

Lx is the short span of the slab Ly is the long span of the slab LL is the length of the line load

Mx, Mx', My and My' are the sagging and hogging moment capacities about the x and y axes.

Substituting in the values for Dornie with the reinforce- ment yield strengths and concrete strengths as in section 3 and separating the internal work done for bending and arching action gives Ib=1890 and la=1300. Summing these two gives a failure load of 3190 kN. This assumes the yield line pattern extends the full width between main beams with Ly=4.775m. Reworking the calculation with different Ly values gives a lowest failure load of 2572 kN when Ly=2.8m. This is taken as the best estimate and is 10% less than the failure load in the test. The separate components are Ib=1344 and la=1228.

The analysis was reworked using the mean ultimate strengths of the reinforcing bars in place of the yield strengths. This gave a lowest failure load of 3328 kN which is 17% over the test load.

For the limit analysis prediction the reinforcement yield stress and concrete cube strength were taken as the BD 21/84 (DTp 1989) values of 230 MPa and 15 MPa respectively. These give a prediction of 1955 kN, 31% less than the test load.

8.3 TESTS 3 AND 4: MAIN BEAMS

These beams were not taken to failure but their load/ displacement graphs suggest that they were close to failure. A limit analysis has been made for Test 4 using a strut-and-tie model and a Maxwell-Bow diagram (figure 25). This analysis considers the behaviour of one beam on its own. Lengths fg and ad have been drawn to represent the net yield strengths of the beam reinforce- ment after deductions of about 10% were made for the forces under dead load. For convenience the dead load was then omitted from the analysis. The applied load ia is 2940 kN which is only 55% of the maximum load in the test. If the diagram were drawn for a larger load then the reaction hi at an adjacent pier would require too large a horizontal component.

The additional capacity in the real bridge is derived from the interaction between the two main beams. Evidence of this interaction is clear in some of the strain plots, for example figure 21. The expected strains due to the hogging moment over the pier have superimposed on them the strains due to transverse compatibility moments.

21

~ o 0 ~ c o ~

. b w

E Q~

O • E

. D d

L~

o ~ LL

0 ~

t ' -

2 2

9. DISCUSSION

The tests show that there were large reserves of strength in the structure which appear to have been derived mostly from the longitudinal end restraints on members and partly from the excess strength in the reinforcement.

It is difficult to quantify the effect of end restraint and in the interests of simplicity design code methods are usually based on the assumption that members are unrestrained longitudinally. Hence researchers, who are usually providing data for the code writers, traditionally take great care to achieve frictionless end supports in their test rigs. The imperatives of bridge assessment are different from those for design and there will be many cases where the "longitudinal end restraint reserve" can be invoked with great economic benefit. This reserve is only available when the assessor has sufficient under- standing of the behaviour of real structures at failure and this understanding is derived from experience, usually the experience of others reported in documents like this.

The strut-and-tie method has been used for the analysis of Tests1 and 4. It allows an assessor to represent his or her intuitive understanding of the restraints available in non-standard situations. The method is particularly well suited when there is a dominant concentrated load, as in the tests. It would need to be adapted for the distributed loads used in assessment.

Test 2 shows the importance of "Compressive membrane action" (CMA). This is the name given to the specific case of longitudinal end restraint in slabs loaded out of plane. Despite the publication of a large number of papers on this topic over several decades Rankin, Niblock, Skates & Long (1991) report in their introduction: "At present, no rationally based design method which recognises the strength enhancing effects of CMA has gained accept- ance among practising engineers". This may be largely due to the difficulty of quantifying restraint. Test 2 has shown that the failure load exceeds the prediction based on rigid end restraint for the panel. This prediction is nearly twice that for the case usually considered by codes with the lateral restraint released but the rotational restraint maintained.

It is important to recognise that CMA and other longitudi- nal end restraint effects in beams can only be invoked in cracked reinforced concrete members which are not over-reinforced. The benefits arise because, on cracking, the neutral axis migrates towards the compression face and the plane of the original neutral axis position extends. It is resistance to this extension which generates the beneficial forces.

When modelling real behaviour at failure the question arises as to which reinforcement stress should be used, the yield stress or the rupture stress? In general, if the reinforcement ruptures, then clearly its rupture stress has been reached and this is the stress that should be used in the analysis. However, there are several exceptions which together justify the usual practice of using the yield stress in assessments.

In flexural members the large strains that precede rupture will impose correspondingly large strains on the concrete which is taken well into the region of strain softening and failure. Most of the extra stress available cannot be converted into extra moment.

Another exception is demonstrated in the analysis of Test 1. The shear strength used was derived using a codified method which required that the shear reinforcement was assumed to be acting at its yield stress. A variant of the EC2 method can be envisaged which could have pre- dicted the same failure load by using the rupture stress together with a steeper limiting truss angle. The evidence of the test is that such a variant might be a better model because the links did rupture and the angle of the shear failure was significantly steeper than the 22 degrees implied in EC2.

A third exception relates to punching failure and hence Test 2. As stated in 8.2 the mechanisms for punching failure in flexure and shear are closely linked. At the onset of flexural yield the strain patterns are altered and the punching shear failure is triggered. Hence it is yield that governs the capacity.

10. CONCLUSIONS

The conclusions from the testing and analysis are as follows:

1. The span tested was capable of supporting loads far in excess of the full assessment loading. Any failure that occurred in service would have been of a local nature and would have been highly unlikely to cause danger to the public. However the decision to replace the structure was based on the high cost of concrete repairs and the condition of the half joints and bascule span.

2. Despite their appearance the main structural ele- ments proved to have considerable reserves of strength beyond those usually considered in assessment. This can be attributed to end restraints on members, structural interaction between elements and reserves of reinforce- ment strength.

3. Realistic rules which quantify these effects could significantly increase the assessed strengths of bridges of this type, many of which remain in service. Bridges can benefit even without such rules because the results of these tests add to the totality of vicarious experience on which every assessment engineer depends when making judgements.

11. FUTURE WORK

Analyses are planned using a computer model to predict the response of the structure under increasing load. To simulate failure it would be necessary to use a non-linear

23

analysis. The load/displacement and load/strain graphs could then be compared with the predictions.

For the main beam tests 3 and 4 a more detailed analysis would enable the three dimensional effects, described in section 8.3, to be further explored.

12. ACKNOWLEDGEMENTS

The work described in this report was carried out in the Bridges and Ground Engineering Resource Centre of TRL. The Authors would like to thank the TRL site team for their professionalism in carrying out the tests, to programme, in the limited time available, and also the Scottish Office Roads Department for making the bridge available for testing.

REGAN, P E and BRAESTRUP, M W (1985). Punching shear in reinforced concrete, a state of art report. Bulletin d'lnformation No 168, Comite Euro-lnternational du Beton, Lausanne.

RUSSO, G, ZINGONE, G and PULERI, G (1991). Flexure-shear interaction model for longitudinally rein- forced beams. ACl Structural Journal, (January-Febru- ary).

SCHLAICH, J, SCHAFER, K and JENNEWEIN, M (1987). Towards a consistent design of structural con- crete. PCl Joumal (May-June).

SCHLAICH, J and SCHAFER, K (1991). Design and detailing of structural concrete using strut-and-tie models. The Structural Engineer, Vol 69, No 6 (March).

13. REFERENCES BRITISH STANDARDS INSTITUTION (1984). BS 5400: 1984, Steel, Concrete and Composite Bridges; Part 4 Code of Practice for the design of concrete bridges. London: BSI.

BRITISH STANDARDS INSTITUTION (1990). BS EN 10 002-1: 1990, Tensile testing of metallic materials; Part 1 Method of test at ambient temperature. London: BSI.

COATES, R C, COUTIE, M G and KONG, F K (1980). Structural analysis. Walton-on-Thames, UK. Nelson.

COMMISSION OF THE EUROPEAN COMMUNITIES (1989). Design of concrete structures. Eurocode EC2: Part 1 (revised final draft).

DEPARTMENT OF TRANSPORT (DTp) (1984). The assessment of highway bridges and structures. Depart- mental Standard BD21/84 (1984), including Amendment 1 (1989).

DEPARTMENT OF TRANSPORT (DTp) (1990). The assessment of concrete highway bridges and structures. Departmental Standard BD 44/90.

HEWITT, B E, and BATCHELOR, B deV (1975). Punch- ing shear strength of restrained slabs. Journal of the Structural Division, American Society of Civil Engineers, ST9 (September).

KANI, G N J (1966). Basic facts concerning shear failure. ACI Journal, Proceedings V. 63, No.6 (June).

NIELSEN, M P (1984). Limit analysis and concrete plasticity. Prentice-Hall, New Jersey.

RANKIN, G I B, NIBLOCK, R A, SKATES, A S and LONG, A E (1991). Compressive membrane action enhancement in uniformly loaded, laterally restrained slabs. The Structural Engineer, Vol 69, No 16 (August).

24

14. APPENDIX A: REINFORCEMENT DETAILS

<

0 J~

E

o "

g 7 -

c 0

. c - O

0

0

U_

rn

C~l ~ I OJ II II II

0 m

II II II - - ' n n l ~

II II II II

~ ' ~

I I I I I I I I a D a o

<

t ~

=

r n

C . m m

0

II

C3

b

0

I '7 111 I--

...1

n -

n 7

U_ 0

T

Z I i i ._.1

~ism

0-o J

I ....

POo~ I-o

• c

¢ l

. , o g

~ 9

~ Q

~m _0

I n7

Z o I - - O I I I

00

I

O Z O I-- O W O3

o ~

7 ~

, ~ G,I

E Z o -~ f- o O W om

E

, ,O

C

X " O

C

25

oD

n~

-g

0

co

° I =

. .~ II ~ c~

~o

z 0 I-- (..) W if? ,._1

z

z 0 _ J

°

f

W a

-r o.~

. . ~ . _ ~

I I

n,-

7o

0 n "

~J

JO

=o0

I b

"o

rn

W -~ r7

7 -

_o 0

0

" 0

7O

a~ O.

7O

..0

n7

II1

¢,.

0 o

I

UJ - -

W , X U

W " 0

z

~c ,'~

a

, i x

"0 c

Q .

2 6

d

1/4" links all round

4" c/c

b c d (~

/ J

3" 4" 5" 6" 7" 8" 8" 10" c/c

d /

J

3 bars a D = 13/8" L = 23' 0" 2 bars b D = 11/4" L = 27' 0" 1 b a r c D=11/4" L = 2 7 ' 0 "

3 x 2 bars d D = s/8" L = 18 '3 "

3/8"links (2 sets)

Appendix A. Cross beam reinforcement details - Dornie Bridge

Printed in the United Kingdom for HMSO DdK63500 5/93 C5 G542 10170 27

MORE ZNFORMATI©N FROM TRL T R L has published the fo l lowing other reports on this area of research:

RR198 Tests on a half-scale prestressed-beam and slab bridge deck, K H Withey, Price Code B

RR310 Load test on a jack arch bridge with cast iron girders, A F Daly and S J Raggett,

Price Code B

If you would like copies, photocopy and fill in the slip below. There is a 20% discount if you take all the reports listed above. Prices include postage and are correct at the time of publication. Please see the enclosed letter for current price code values and handling charge. Enquiries to TRL Library

Services, 0344 770783 or 770784.

To: Publicat ion Sales, TRL Library, PO Box 304, Wokingham, Berkshire, RG11 6YU.

Please send m e the fol lowing T R L reports (state report Nos and quantity)

N a m e . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

Address .............................................................

Pos tcode ............................................................

Te lephone ..........................................................

Credi t card address (if different f rom above) ...

PAYMENT:

• I enclose a cheque for £ ..............................

payable to Transport Research Laboratory

• Please debit my Deposit Account

n o . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

• Please debit my Credit Card by £ ................

• Credit card no .............................................

Expiry date ....................................................

Signature .......................................................

TRL's researchers and Laboratory facilities are available at competitive rates.

Our 300 scientists and engineers include many world-class experts on highways design and maintenance, transport structures, traffic systems, vehicle safety, road safety and the

environment . . . . . . . . . . . . . . . . . . . ~ - - -

TRL facilities include a 2.3 mile test track, large structures test halls for static and fatigue testing, a dynamic pavement test facility, advanced computer systems and a large specialist library with on-line access to worldwide information.

If you are planning a project where we may be able to help, contact TRL at Crowthorne, Berkshire RG11 6AU, telephone 0344 770004, fax 0344 770356.