DNV CN 30.4 Foundations.pdf

CLASSIFICATION NOTES NO. 30.4

DET NORSKE VERITAS

FOUNDATIONS

FEBRUARY 1992

Det Norske Veritas Classification AS V E R I T A S V E I E N 1, N - 1 3 2 2 H 0 V I K , N O R W A Y T E L : +47 67 57 9 9 00 F A X : +47 67 57 99 11

FOREWORD

Det norske Veritas is an independent Foundation with the objective of safeguarding life, property and the environ-ment at sea and ashore. Classification, certification and quality assurance of ships, offshore installations and industrial plants, as well as testing and certification of materials and components, are main activities.

Det norske Veritas possesses technological capability in a wide range of fields, backed by extensive research and development efforts. The organization is represented world-wide in more than 100 countries.

Classification Notes are publications which give practical information on classification of ships, offshore installa-tions and other objects. Examples of design solutions, calculation methods, specifications of test procedures, quality assurance and quality control systems as well as acceptable repair methods for some components are given as interpretations of the more general rule requirements. An updated list of Classification Notes available is given in the latest edition of the Introduction-booklets to the Rules for Classification of Steel Ships, Rules for Classification of Mobile Offshore Units and Rules for Classification of Fixed Offshore Installations)).

Det norske Veritas 1992 Computer Typesetting by Division Ship and Offshore, Det norske Veritas Classification A/S Printed in Norway by Det norske Veritas

02.92.2000

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CONTENTS

1. SOIL INVESTIGATIONS FOR FIXED OFFSHORE STRUCTURES 4

1.1 Introduction 4 1.2 Methods and techniques 4 1.3 Soil investigation for gravity type foundations . 7 1.4 Soil investigation for pile foundations 8 1.5 Soil investigation for jack-up platforms 8 1.6 Soil investigation for pipelines 9

2. AXIAL PILE RESISTANCE 10 2.1 Introduction 10 2.2 Resistance in cohesive soils 10 2.3 Resistance in cohesionless soils 13 2.4 Resistance in calcareous soils 15 2.5 Group effects 15 2.6 Effects of installation procedure 15 2.7 Effects of cyclic loading 15

3. LATERAL PILE RESISTANCE 16 3.1 Introduction 16 3.2 Piles in cohesive soils 16 3.3 Piles in cohesionless soils 18 3.4 Piles in calcareous soils 19 3.5 Modifications of py curves 19 3.6 Plastic analysis of piles 20

4. STABILITY OF GRAVITY BASE FOUNDATIONS 20

4.1 Introduction 20 4.2 Soil shear strength 21 4.3 Solution methods 26 4.4 Bearing capacity formulae 28

5. SETTLEMENT OF GRAVITY FOUNDATIONS 31

5.1 Introduction 31 5.2 Stress distribution theories 32 5.3 Settlement calculations 36 5.4 Time rate of consolidation 38

6. PENETRATION RESISTANCE OF SKIRTS 39 6.1 Introduction 39 6.2 Methods of calculation 40

7. SOIL-STRUCTURE INTERACTION 41 7.1 Introduction 41 7.2 Global dynamic foundation stiffness for GBS

type foundation 41 7.3 Soil reaction on structural foundation elements 43

8. FOUNDATION OF JACK-UP PLATFORMS 44 8.1 Introduction 44 8.2 Individual leg supported jack-up platforms . . . 45 8.3 Mat-supported jack-up platforms 49 8.4 Foundation restraints 50

9. REFERENCES 52

4 Classification Notes No. 30.4

1. Soil Investigations for Fixed Offshore Structures 1.1 Introduction 1.1.1 General 1.1.1.1 Guidelines for determination of soil investi-gation programme for gravity type foundations, piled foundations and foundations of pipelines are given in this Chapter. Brief descriptions of the various methods and techniques to be used in geophysical and geotechnical surveys are given.

1.1.2 Planning

1.1.2.1 The required amount of information with re-spect to soil properties normally changes during a field development. At an early stage the gathered data should be sufficiently detailed to demonstrate the feasibility of a given concept. Also, the information available at this stage facilitates the selection of the most favourable lo-cation for the structure within the development area. At a final stage the soil investigation should provide all nec-essary data for a detailed design of a specific structure at the specific location.

1.1.2.2 The soil investigation necessary for field devel-opment should normally be performed in progressive stages so that structural concepts can be developed with due regard to soil conditions. In order to optimize the extent of the soil investigation, planning should be done based on the results from previous findings. Factors as geological history, uniformity*of foundation deposits, size and type of structure etc. should be reflected in the extent of the site investigation.

1.1.2.3 The sequence of the soil investigation for a platform should be as follows:

Collection of available geological, geotechnical and foundation performance data for the area.

Carrying out of a geophysical survey at an early stage of the field development, comprising: - Bathymetry and seabed surveys - Sub-bottom profiling. This is to be supplemented with a feu seabed samples (e.g. gravity cores) and one or two soil borings.

When the type and location of platform have been determined, a detailed geotechnical investigation and topographical mapping and seabed survey of the ac-tual location should be carried out.

1.2 Methods and techniques

1.2.1 General 1.2.1.1 The soil investigations may be divided in:

Geological studies Geophysical surveys Geotechnical investigations.

Below, brief descriptions of the various survey methods and techniques are given. General sampling recommen-

dations and guidelines for planning of laboratory test programmes are also presented.

1.2.2 Geological studies

1.2.2.1 The geological study should be based on infor-mation about the geological history of the general area of field development. The purpose of such a study is to establish a basis for selection of methods and extent of the site investigation.

1.2.3 Geophysical surveys

1.2.3.1 The main purpose of the geophysical survey should be to extend the more localized information from borings and in situ testing to get an understanding of the seabed topography and the stratification within defined areas. As such, these surveys should give guidelines in se-lection of suitable platform sites within the exploration area.

1.2.3.2 Seabed topography and layering are investigated by means of seismic methods. Geophysical surveys are carried out by towed devices with specifed characteristics.

1.2.3.3 For determination of water depth and sea floor topography, high accuracy echosounders may be used together with vessel movement sensors (surface system). However, use of a towed fish with echosounder and pressure sensor will improve the accuracy significantly. By adding a side-scan sonar device to the towed system, any seabed obstruction or feature may be investigated in more detail by towing closer to the seabed. Manned or unmanned submersibles for visual/video surveys of the actual foundation area will complement the echosounder and sonar profiles. Echosounders with adequate high fre-quency response may detect gas seeps at the sea floor and particularly soft seabed deposits. Steel and iron objects may be detected with a marine proton magnetometer bottle, which measures the total magnetic field intensity along the tow line. Obstacles detected at the sea bottom shall be carefully mapped and identified.

1.2.3.4 The choice of an appropriate geophysical pro-filing system depends upon the required depth of pene-tration, the desired degree of resolution and the seismic response of the shallow formations. The resolution, i.e. the ability to identify the different sub-bottom layers, in-creases as the frequency of the transmitted and received signals increases. However, higher frequencies result in larger absorption losses in the ground and less pene-tration. The basic components of a seismic profiling sys-tem are a sound source, hydrophones and a recording unit. Typical operating characteristics for high energy systems are frequencies in the range 100400Hz capable of achieving penetrations down to about 300m depth with a resolution of some metres. A high resolution profiling system should contain a set of towed devices having dif-ferent frequency response. The necessary depth to which the investigation should extend depends on the geological formations and the type of structure.

1.2.3.5 Coarse grid surveys may give guidelines in se-lecting the optimum foundation site where detailed sur-veys must be done. By reducing the grid spacing, details of the geologic formations may be obtained for the most interesting area.

Classification Notes No. 30.4 5

1.2.4 Geotechnical surveys in general

1.2.4.1 The principal methods to be employed in a geo-technical investigation are:

Sampling for laboratory testing In-situ testing.

The geotechnical investigation at the actual platform site should secure all data necessary for the foundation de-sign. Options for modifications of the initial site investi-gation program in the course of the survey may be favourable. A qualified geotechnical engineer should therefore be onboard the survey vessel. The soil investi-gation should be tailored to the design methods used. To facilitate the interpretation of the test results, an overlap of information between the various methods em-ployed should be planned. The field and laboratory in-vestigations should establish the detailed soil stratigraphy across the site providing the following types of geotech-nical data for all important layers:

Data for classification and description of the soil Parameters required for a detailed and complete foun-

dation design.

1.2.5 Sampling without drilling

1.2.5.1 Grab samplers, gravity corers and bottom oper-ated corers may secure soil samples from the top soil layers. With present equipment these samples have usu-ally been found disturbed and consequently only useful for identification purposes.

1.2.5.2 Gravity corers consist essentially of a heavy torpedo-shaped body with a sampling tube (50100 mm diameter) attached in front. The basic method of opera-tion is to lower the corer on a wire until it is a few metres above the sea bottom. It is then released and allowed to fall to the bottom. In soft to firm clays the depth of pen-etration is 35m while no penetration may be experienced in dense or hard soils.

1.2.5.3 Piston corers look like gravity corers but take longer samples. The piston remains near the top of the sediment by sliding up the sampling tube as the corer penetrates into the seabed. By this method samples ex-ceeding 40m in length have been taken. The tube diameter is 50-100 mm.

1.2.5.4 Vibratory sampling can provide soil samples up to 8m in length in soft to firm clays and loose sands while the length may be limited to 0.52m in hard clays. The sampling is carried out from a rig, lowered to the seabed and remotely controlled from the surface. The sampling tube has a diameter of 100270 mm.

1.2.6 Sampling from a drilled borehole

1.2.6.1 For sampling at greater depths drilling of a bo-rehole is recommended. The sampling device is then low-ered inside the drillstring to the bottom of the borehole at the depth where a sample is taken. The boring is made with a straight flush rotary drilling technique. Drilling mud may be needed to remove the cuttings and to stabi-lize the hole. The top of the drillstring is connected to a motion compensator in the crown of the derrick so that the drillstring is in constant tension. The maximum

available bit pressure is governed by the weight of the drill collars and the tension force required to avoid buckling of the drillstring above mudline.

1.2.6.2 The traditional sampling method is percussion sampling with a wireline tool consisting of a thin-walled tube and a sliding hammer. The percussive action of the falling weight produces clay samples which are signif-icantly disturbed.

1.2.6.3 The sample disturbance is reduced to some ex-tent by push sampling. Different techniques have been developed for this type of sampling:

a) The sample tube is latched into the drillstring and pushed into the soil by the weight of the drillstring (by reducing the tension load).

b) The sample tube is pressed into the soil by hydraulic jacks operating: either from the sea floor as part of a heavy jacking

unit providing the reaction force or within the drillstring near the bit with the re-

action force provided by friction between the bore-hole wall and inflatable packers inserted in the drillstring just above the bit.

Wherever possible push sampling should be preferred as compared to percussion sampling, especially in cohesive soils.

1.2.6.4 Rotary sampling tools are generally used for drilling and sampling in hard formations such as rock cemented sand, hard heavily-overconsolidated clays, and boulder type clays. A typical tool of this type is the ma-rine wireline double walled core barrel. Cores are taken by rotating outer barrel, while the non-rotating inner barrel is stationary around the core. After coring is com-pleted the inner tube and the core are recovered to the surface by use of the wireline assembly.

1.2.7 General sampling requirements

1.2.7.1 The sampling tools should be checked for proper operation and should be equipped with undamaged, properly machined sample retainers. Where sampling is carried out from the bottom or a borehole, care must be taken to achieve a clean borehole free from cuttings and debris at the time of sampling. If metallic tubes are used to secure and store undisturbed samples, only new tubes with proper cutting edge should be employed. The sampling operation should be conducted in such a way that damage to the sampler and disturbance of the soil samples are avoided.

1.2.8 In-situ testing

1.2.8.1 The cone penetrometer test (CPT) is the most commonly used in-situ testing method in offshore soil in-vestigations. The test is carried out either from an un-derwater rig without drilling of a borehole (e.g. Seacalf), from a seafloor-based jacking unit (e.g. Stingray) or down-the-hole without use of any seafloor unit (e.g. Wison). The test is carried out by pushing a 10cm2 cone at a pen-etration rate of 2cm/s into the soil. The cone tip simul-taneously measures the tip (cone) resistance and the


friction along a sleeve behind the tip. The results provide useful information, both quantitatively and qualitatively about soil strength and stress-strain characteristics. Piezo cone penetrometers, which incorporate also a pressure transducer at the tip to measure the pore pressure, are the most common today in offshore site investigations.

1.2.8.2 The Seacalf rig or similar equipment performs remotely controlled static cone penetrometer tests from the seabed using a hydraulic jacking system and a re-action force of 60260 kN provided by the ballasted frame of the rig. The depth of penetration typically ranges from about 20m in hard clays or dense sands to 3060m in soft normally consolidated clays. Continuous plots of cone resistance, sleeve friction and excess pore pressure as function of depth may be obtained.

1.2.8.3 The Stingray rig or similar equipment is an ocean floor, hydraulically powered drillstring reaction device typically weighing 230 kN. Sampling or in-situ testing can be performed on the seafloor or at any depth. It is designed to operate in water depths down to 900m. Cone testing is carried out in increments of max. 36m or until refusal. After each increment the cone is re-trieved by use of the wireline and the drillpipe is advanced to the depth penetrated by the cone. At this depth the cone penetrometer testing is resumed. This procedure is repeated until the complete depth of interest is tested. The Stingray is designed to accommodate cone penetrome-ter tests, vane tests, pressure-meter tests, load tests as well as soil sampling.

1.2.8.4 The Wison cone penetrometer system enables in-situ tests to be performed from the base of a borehole. The cone penetrometer is lowered inside the drillstring and latches to the drillcollar. The Wison is activated to push the penetrometer into the soil. After reaching maxi-mum depth (about 3m) or earlier if the total thrust ca-pacity is reached, the tool is depressurized. The drillstring is lifted to retract the test rod and the Wison unit subse-quently retrieved. Also the Wison CPT, with or without measurement of pore pressure, can be carried out in combination with wireline sampling or push sampling.

1.2.8.5 The Remote Vane is a wireline tool to be used for in-situ measurement of the soil's undrained shear strength. The instrument has two main sections the tool body and the motion compensating section. The lower portion of the tool body contains the test vane and a reaction vane, both of which are inserted into the soil. The operational sequence for performing a Remote Vane test begins by advancing the borehole to a depth approx-imately lm above the desired test depth and then sus-pending the drillpipe with the drill bit a couple of metres above the bottom of the hole. The tool is then lowered through the drillpipe until it rests on the bottom of the borehole, and the motion compensating unit is approxi-mately 80% retracted. While the tool rests on the bottom with its weight removed from the wireline, the pawls are activated to extend from the tool body. The drillpipe is then lowered until the open-center drill bit bears on the pawls, pushing the vane blade to the desired test depth. The drillpipe is again suspended off bottom and the test is performed. After the test is completed, the tool is re-trieved and the borehole is advanced to the next depth.

1.2.8.6 The pressuremeter is a dilatable cylindrical probe which is generally run into a borehole or sometimes driven into the soil. The test at a given depth consists of measuring the pressure-volume relationship during the dilation phase. Any pressuremeter test includes two suc-cessive operations, i.e. setting the probe in place and then expanding the cell together with data acquisition. This is a process capable of determining the static as well as the dynamic stress-strain characteristics of the soil.

1.2.8.7 Radioactive well logging is carried out by low-ering into a borehole a probe containing radioactive iso-topes. On its way down through the drillstring a recording is done on a paper trace which will give the wet density and/or the moisture content profile through the surveyed depth. The gamma radiation method uses a source of gamma rays inserted at a fixed distance from a Geiger type gamma ray detector. The induced gamma rays pass into the soil and the detector records the number of rays which undergo Compton scattering which is a measure of the wet density of the soil. This probe contains a source of high energy neutrons which pass into the surrounding soil and are reduced in energy especially by colliding with hydrogen nuclei. By providing a unit which detects low energy neutrons, a measure is obtained of the moisture concentration.

1.2.8.8 The measurement of in-situ shear wave velocity requires a system comprising a source generating shear waves, receivers, preferably 3-component. a recorder to measure travel times and a triggering system for trigger-ing the recorder. The source may be located either at the seabed with the receivers at different depths below the seabed or in one of two parallel boreholes with the re-ceivers in the other borehole. The former of these tech-niques is the most common for offshore applications. The receivers are located in a cone penetrometer unit which can be lowered by wireline and latched into the bottom of the drillstring. By this device the shear wave velocity can be measured as a function of depth. In a second step the small strain shear modulus (Gmax) of the soil can be calculated. The simultaneous measurement of the cone resistance makes this so-called seismic cone a very useful tool.

1.2.8.9 The dilatometer consists of a flat blade which is pushed into the undisturbed soil from the bottom of a borehole or from the seabed. For the offshore dilatome-ter, which is smaller than the onshore Marchetti device to fit inside the standard drillpipe used offshore, oil pres-sure is used to expand the membrane. Readings are taken of the membrane lift off pressure (at rest pressure, p0) and the 1-mm expansion pressure (pi). A filter located on the opposite side of the membrane centre allows pore pressure to be measured continuously. The dilatometer can be used to determine the lateral earth pressure in-situ and thus the earth pressure coefficient K(>. Tentative em-pirical relationships are also developed for derivation of other geotechnical design parameters.

1.2.8.10 As a guidance for assessment of a minimum setting depth of conductors, so-called hydraulic fracture testing is used. Equipment and procedures for this type of testing are still under development based on practical experience from various prototype testing in connection with offshore soil investigations.


1.2.9 Laboratory investigations 1.2.9.1 The recovered soil samples should be described both onboard and later in the onshore laboratory. The description should comprise estimates of:

Grain size distribution Colour and smell Consistency Laminations Carbonatic reaction Other relevant information.

1.2.9.2 The samples should be properly cleaned, marked, sealed and stored. Storage, handling and trans-portation of the samples should be as gentle as possible to avoid or limit disturbance.

1.2.9.3 The onboard laboratory testing should normally comprise the following determinations:

Water content Unit weight Undrained shear strength of cohesive samples by me-

ans of pocket penetrometer, torvane, miniature vane, fall cone and UU triaxial test

Carbonatic reaction Grain size distribution of selected cohesionless samples Liquid and plastic limits of selected cohesive samples.

Recently even more advanced laboratory tests, e.g. oe-dometer tests, direct simple shear tests etc. have been performed with success onboard the vessel. An experienced geotechnical engineer or an engineering geologist should be present on board during sampling and laboratory testing.

1.2.9.4 The onshore laboratory testing should be carried out on representative samples which shall as closely as possible be subjected to the same stress conditions as ex-perienced offshore. It is essential that initial stress condi-tions, overconsolidation ratio and stresses induced by the structure and environment are realistically reflected. A combined static/cyclic test programme should allow de-termination of stiffness, damping and strength of the soil under the range of load conditions to be covered by the design. The random nature of wave and earthquake loading re-quires that special attention should be paid to the load simulation technique used in the testing. The chosen procedure should reflect the effect of the stress level and load duration on the development of pore pressure and shear strain. The types of tests which should normally be considered in the planning of a programme are the following:

Classification and index tests such as: - Unit weight of sample - Unit weight of solid particle - Water content - Liquid and plastic limits - Grain size distribution - Salinity - Carbonate content.

Permeability tests. Consolidation tests. Static tests for determination of shear strength pa-

rameters: - Triaxial tests (UU, ClUa, CAUa, CAUp, oedo-

triax, Kq) - Direct simple shear (CCV).

Test for determination of remoulded shear strength (type UU, CIU).

Cyclic tests for determination of strength and stiffness parameters: - Triaxial tests (CIUc, CAUc) - Direct simple shear (CCVc) - Resonant column (ClUrc) - Shear wave velocity measurement.

1.3 Soil investigation for gravity type foundations

1.3.1 General

1.3.1.1 The soil investigation for a gravity type founda-tion should give basis for a complete foundation design comprising evaluations of:

Stability Settlements Penetration and retrieval resistance of skirts Local contact stresses Dynamic response of foundation soil.


1.3.2.1 The minimum depth of sub-bottom profiling should correspond to the depth to rock or the width of the largest base dimension. The required accuracy for sea bed topography measurements is normally in the order of 0.10.2m for relative elevations. This is outside the capacity of echosounders operated from the sea surface subject to wave and wind effects. Alternative methods include submarines or remotely controlled underwater vehicles equipped with differential pressure transducers or echosounders. Any obstructions e.g. large boulders discovered during seismic and topographic surveys within the foundation area should be accurately located. A fairly close grid of seismic profiles (50100m spacing) over the actual area for correlation with other investigation results will reduce the number of borings to a minimum.

1.3.3 Geotechnical surveys

1.3.3.1 As basis for all foundation analyses an extensive investigation of the shallow soil deposits should be per-formed. The minimum depth should be deeper than any possible critical shear surface. Further, all soil layers in-fluenced by the structure from a settlement point of view should be thoroughly investigated.

1.3.3.2 The extent of shallow borings with sampling should be determined based on type and site of structure as well as on general knowledge about the soil conditions in the area of platform installation. Emphasis should be given to the upper layers and potentially weaker layers


further down. Sampling intervals should not exceed 1.01.5m. A number of seabed samples (gravity cores or equivalent) evenly distributed over the area should also be taken for evaluation of scour potential.

1.3.3.3 In addition to the borings, shallow CPTs dis-tributed across the installation area should be carried out. The number of CPTs depends on size and type of struc-ture and soil conditions. If the soil conditions are very irregular across the foundation site, the number of CPTs will have to be increased. The shallow CPTs should give continuous graphs from mudline to the maximum depth of interest.

1.3.3.4 For settlement evaluations and determination of dynamic response of the foundation soil, investigation of the soil to a greater depth is necessary. The depth should not be less than that corresponding to the largest base dimension of the structure. The investigation should consist of one boring with continuous CPT and at least one boring with sampling close to the CPT hole. The sampling interval is to be determined from the CPT re-sults but should not exceed 3m.

1.3.3.5 If, during the course of the soil investigation, a weaker layer is encountered, along which a critical failure surface can be expected, special emphasis should be put on investigation of this layer.

1.3.3.6 Special tests such as plate loading tests, pressu-remeter tests and shear wave velocity measurements should be added where relevant.

1.4 Soil investigation for pile foundations

1.4.1 General

1.4.1.1 The soil investigation for a pile foundation should give basis for a complete foundation design com-prising evaluations of:

On-bottom stability of unpiled structure Lateral pile capacity Axial pile capacity Pile drivability predictions.

1.4.1.2 The extent of the soil investigation is dependent on type and size of the structure and the consistency and degree of uniformity of the foundation soil.


1.4.2.1 In 1.2.3 the demands to a geophysical survey are described. As for gravity foundations it is essential to carry out a coarse grid geophysical survey at an early stage of the field development. The minimum depth of seismic profiles should be the anticipated depth of pile penetration plus a zone of influence of about ten pile di-ameters.

1.4.2.2 A topographic survey of the selected area should be carried out. Remotely controlled underwater vehicles with video camera and differential pressure transducers will provide data of sufficient accuracy. Any obstructions, within the foundation area, discovered during the seabed surveys shall be accurately mapped.


1.4.3.1 For on-bottom stability and lateral pile analyses shallow cone penetration tests should be carried out from mudline to 2030m depth. In addition, shallow borings with sampling should be performed for better determi-nation of characteristics of the individual layers identified by the cone penetration tests. The sampling interval should not exceed 1.01.5m.

1.4.3.2 A number of seabed samples (gravity cores or equivalent) evenly distributed over the area should be taken for evaluation of scour potential.

1.4.3.3 For axial pile capacity analysis at least one down-the-hole CPT boring giving a continuous CPT profile and one nearby boring with sampling should be carried out. The minimum depth should be the antic-ipated penetration of the pile plus a zone of influence. The zone of influence should be sufficient for evaluation of the risk of punch through failure. The sampling inter-val should be determined from the CPT results but should not exceed 3m.

1.4.3.4 If no potential end bearing layers or other dense layers which may create driving problems are found, the above scope of sampling and in-situ testing are sufficient.

1.4.3.5 In case potential end bearing layers or other dense layers are found, additional cone penetration test-ing and sampling should be carried out in order to deter-mine the thickness and lateral extension of such layers within the foundation area. Use of rotary sampling tools may be recommended in very hard or dense formations.

1.5 Soil investigation for jack-up platforms

1.5.1 General

1.5.1.1 For general site assessment and evaluation of the foundation behaviour of a jack-up rig, adequate geotech-nical and geophysical information should be available, including information about:

Seafloor topography and sea bottom features. Soil stratification and classification. Characteristics for soil in various strata.

1.5.1.2 The most important soil parameters are the soil shear strength parameters, i.e. undrained shear strength or the effective stress parameters cohesion (alternatively attraction) and angle of internal friction. As found rele-vant in each case, it may also be required to determine other characteristics such as grain size distribution, rela-tive density, unit weight and in-situ small strain shear stiffness Gmax-

1.5.1.3 The soil investigations may be based on a com-bination of the following types of information:

General geological knowledge about the area Geophysical investigations (bathymetry, sidescan so-

nar, shallow seismic) Sampling and laboratory testing In situ testing, e.g. cone penetration test.



1.5.2.1 Geophysical investigations required for a site assessment includes bathymetric survey, seabed surveys with side scan sonar or high resolution multibeam echo-sounder and shallow seismic surveys. The various surveys are normally performed in parallel from one survey vessel using multipurpose tow equipment. Shallow seismic with digital recording will have to be performed separately, however. At the selected location, the line spacing should be suffi-ciently small to detect all features of interest, such as seabed irregularities or debrises, variations in subsoil strata including erosion channels etc. Interlining within the area of most interest may be decided based on initial survey of a wider area with coarser spacing. Depending on the general irregularities detected by the first survey, the line spacing for the detailed central survey can be decided. Typical spacings may be 100 x 250m for a coarse grid ana 25 x 50m for a finer grid.

1.5.2.2 The purpose of the seabed survey is to detect seabed irregularities or debrises, as reefs, iceberg plough-marks, pockmarks, wrecks or other debrises. For de-tection of pipelines, cables or other metallic debris at or slightly below the seabed, magnetometer surveys may be required.

1.5.2.3 The purpose of shallow seismic surveys is to de-termine the soil stratigraphy to a depth of interest as in-dicated in 1.5.3.4 and to detect any possible presence of shallow gas concentrations. The determination of soil stratigraphy requires correlation with soil boring data within the surveyed area. Even when a boring is per-formed at the location, a shallow seismic survey should be available to be able to show whether irregularities exist within the foundation area that give other foundation conditions than that determined by the boring, e.g. detect possible erosion channels or general variations in layer thicknesses of importance.

Analog records may be used for determination of soil stratigraphy whereas registration of shallow gas will re-quire digital recording. The equipment characteristics (energy and frequency) should be chosen to fit the ex-pected soil conditions and the correspondingly required depth for determination of soil stratification.


1.5.3.1 The required extent of the geotechnical surveys is dependent on the variability of the soil conditions in the area, and on possible problems foreseen for the subject jack-up platform at the given location. In the planning for site specific soil investigations, any existing information should be made available, such as general geological knowledge about the area, results from possible previous geophysical investigations, borings and/or in situ testing.

1.5.3.2 As a minimum at each platform location, one should normally provide either one borehole with sampl-ing and laboratory testing, or carry out in-situ testing. Such testing can be omitted provided that:

1) Existing knowledge about the general geology of the area (history of deposits) together with geophysical

surveying can justify extrapolation from documented soil profiles nearby the platform location.

2) It can be documented that, based on a possible range of soil conditions derived from existing soil data, the platform can be safely operated during installation (preload) as well as during normal operations and pull-out phases, see Chapter 8.

1.5.3.3 If the area, within which the platform is to be located shows irregular soil conditions, it may be neces-sary to perform more than one boring/ in-situ testing in order to verify the variations within the foundation area. For such events the uncertainties in positioning should be considered. Special concern should be given to the possibility of bur-ied erosion channels with soft infill material.

1.5.3.4 The design shear strength profile should be es-tablished to a depth below which the soil conditions have no influence on the foundation behaviour. For platforms supported by individual leg foundations (spud cans) the required depth of the documented soil profile will nor-mally be one to two spud-can diameters below the antic-ipated penetration of the spud can. For mat supported foundations, usually only the upper few metres are of in-terest, except at locations with very soft clays where a deepseated failure may be relevant to study, see 8.3.1.4.

1.5.3.5 In areas with high potential of scour, grain size distribution tests should be performed on samples from the upper 23m in order to improve evaluations of scour potential.

1.6 Soil investigation for pipelines 1.6.1 General

1.6.1.1 The site investigation for a pipeline typically consists of a shallow seismic profiling survey of the wide lay barge anchoring corridor, a detailed bathymetric sur-vey of the 100150m wide construction corridor and fi-nally a geotechnical investigation comprising cone penetration tests (CPT), push sampling, vibro coring, gravity coring etc. To define the various soil deposits along a proposed pipeline route, the emphasis is put on the shallow seismic profiling results. In-situ testing and sampling should subsequently be performed for determi-nation of the soil properties in these deposits.


1.6.2.1 Total water depth is needed to determine ex-ternal water pressure on the pipe and wave effects on the bottom sediments. The trenching, laying and burying methods will also be dependent on water depth. The seabed topography will influence the support conditions of the pipe, the formation of free spans and the stability of the seabed itself. Consequently, surveys with precise echosounders and sidescan sonar are usually required. The accuracy of such measurements will directly influence the degree of conservatism in the design of the pipeline itself.

1.6.2.2 Especially in areas of highly variable seabed to-pography, the limitations of the echosounder may neces-sitate more accurate mapping methods. Profiling with


small submarines may improve the accuracy compared with that of surface vessels. Seismic profiling is necessary to define the extent and variations of the various soil deposits along the pipeline route. The equipment used should give good resolution for the shallow layers down to about 10m depth for definition of erodable materials, applicability of trenching methods and stability of the pipeline itself. Deeper penetration should be recommended for identification of strata out-cropping at other locations along the route.

1.6.3 Geotechnical surveys 1.6.3.1 A sufficient number of samples should be se-cured from each major surface deposit to identify the soil or rock. Several types of shallow sampling techniques are now available for this purpose, see 1.2.5. In addition CPTs and/or vane shear tests should be performed.

1.6.3.2 A laboratory should be available onboard for the necessary soil classification and index testing, see 1.2.9.

1.6.3.3 In special cases the seabed conditions should be documented by use of TV or photos. 1.6.3.4 To complement the above surveys, measure-ments of seawater temperature and currents should be taken.

2. Axial Pile Resistance 2.1 Introduction 2.1.1 General

2.1.1.1 Different methods for axial pile capacity calcu-lations are given in this Chapter. 2.1.1.2 Axial pile resistance is composed of two parts:

The accumulated skin resistance The end resistance.

2.1.1.3 Piles carrying their loads mainly through mobi-lized end bearing resistance are called end bearing piles, while the term friction piles is used for piles carrying their loads mainly through mobilized shaft friction. 2.1.1.4 The pile resistance may be assessed using total or effective stress analysis depending on which analysis best represents the actual conditions.

2.1.1.5 Irrespectively of the method applied for calcu-lation of the skin resistance, the effects of factors such as procedure of pile installation (driven or drilled piles), type of drilling mud and grout, length and geometry of pile (cylindrical or with increased base diameter), etc. have to be considered. 2.1.1.6 The axial pile resistance may be determined ac-cording to one or preferably a combination of the fol-lowing methods:

Load testing of piles Static pile formulae Dynamic pile formulae (driven piles only)

Semi-empirical methods based on in-situ tests.

2.1.1.7 Dynamic pile formulae, herein understood as those based on the wave propagation theory, are not ac-cepted as the only method for determination of pile re-sistance. 2.1.1.8 The axial pile resistance should be calculated in accordance with one or preferably by different methods (see 2.2.1.1).

2.1.1.9 The methods to be applied should be developed based on tests resembling the present situation with re-spect to soil conditions, determination of soil parameters, pile size, loading etc.

2.1.1.10 Where grout is relied upon to transfer loads from one pile element to another or from the pile ele-ments to the foundation soil, the surfaces are to be free from rust scale etc. which can reduce the capacity for load transfer. Furthermore the grout itself is to have stress-strain characteristics permitting the transfer of such loads.

2.2 Resistance in cohesive soils 2.2.1 General

2.2.1.1 The design of offshore piles in cohesive soils is based largely on the experience with onshore piles. The methods developed are empirical and subject to the limi-tations and uncertainties in the database, see /1,2,3,4/.

2.2.1.2 It is generally recognized that the pile pene-trations and axial loads encountered offshore are often greater than those covered by the database. There is also a need to extend the database by conducting field pile tests in soil types more relevant to offshore conditions.

2.2.1.3 During the last decade, considerable research has been put into trying to understand the changes which occur in the soil due to installation of a pile by driving, during reconsolidation of the soil mass after installation, and finally during application of a combination of static and cyclic loads, typical for offshore piles, see e.g. /5,6,7/.

2.2.1.4 The interaction between a driven pile and the surrounding soil during axial loading depends basically on the factors mentioned in 2.2.1.3. The effect of cyclic loading on the shaft friction depends e.g. on:

The mobilization of soil shear stresses due to the static pile load

The additional shear stresses in the soil caused by cy-clic loading

The loading history The number of cycles at the various load levels The loading rate compared to the rate in static pile

load testing.

2.2.1.5 For long flexible offshore piles, failure between pile and soil may occur close to mudline even before the soil near the pile tip is mobilized at all. This means that considerable slip between the upper part of the pile and the surrounding soil may occur before the lower part of the pile has reached failure. In a strain softening soil the measured static capacity of a pile will thus be less than the predicted capacity assuming an ideal (rigid) pile, which


mobilizes the peak skin friction simultaneously down the whole pile shaft, see /3/. This so-called length effect is important also with respect to the effect of cyclic loading.

2.2.1.6 The degradation of the skin friction due to cyclic loading becomes significant once relative slip occurs be-tween the pile and the soil, increasing in magnitude and importance with increasing degree of overconsolidation of the soil and particularly when two-way cyclic shear stresses (reversed slip) are imposed on the slip surface.

2.2.1.7 The loading rate during wave loading is about two orders of magnitude greater than during conventional static pile load testing. This relative increase in loading rate may partly compensate for the effect of cyclic de-gradation on the shaft capacity. When cyclic resistance is determined based on cyclic tests, the rate effect is ac-counted for through the use of a realistic cyclic period in the test.

2.2.1.8 No rational analytical design method exists, which can capture the effects of all factors of significance for the prediction of the axial resistance of piles in clay. This has led to the introduction of design philosophies based on extensive use of in-situ testing, including field pile tests, combined with the necessary supporting labo-ratory testing, to assist in the development of site specific pile design parameters. The extrapolation from small scale test results to prototype pile and load conditions may, however, require special considerations, which should be documented in detail in each case.

2.2.1.9 In the following, some methods for prediction of the static axial resistance of driven piles in clay are shortly described. Due to the uncertainties in the predic-tion methods, the pile capacity should normally be pre-dicted based on more than one method. The effect of cyclic loading should be assessed based on the actual loading conditions with due consideration of the soil and pile properties.

2.2.2 Resistance of piles in compression

2.2.2.1 The pile resistance, R, is composed of two parts, one part being the accumulated skin resistance, Rs, and the other part the end resistance, Rp:

soils, i.e. that large diameter piles develop a smaller unit end resistance than do small diameter piles in the same soil. The displacement required to mobilize the unit end resistance will be an order of magnitude greater than that required to mobilize the skin resistance, which should be considered in the pile capacity predictions, especially where the pile end resistance represents a substantial part of the total axial pile resistance. 2.2.2.3 For piles in mainly cohesive soils, the average unit skin friction, fs, may be calculated according to:

Total stress methods, e.g. the a-method, see /8/. Effective stress methods, e.g. the jff-method, see /9/. Combined total/effective stress methods, e.g. the

A-method, see /10,11/.

Existing alternate methods, which are based on sound engineering principles and are consistent with industry experience, may be used in practice. An upper limit of 200 kPa is recommended for the unit skin friction on the basis of previous North Sea experi-ence. Justification of higher values will require special documentation.

2.2.2.4 According to the a-method, in its simplest form, the average unit skin friction in layer i is given by:

fsi = a c u

where

a = a multiplier which is correlated with cu and is equal to or less than 1.0, decreasing with increasing cu and depth of pile penetration.

cu = undrained shear strength based on UU triaxial tests.

2.2.2.5 Based on /8/ the a-factor may be calculated from the equations:

a=0.5^" 0 ' 5 l.O

still with the constraint than a


cu = average undrained shear strength from UU triaxial tests over the embedded pile length, L.

p'0 = average effective overburden pressure over the embedded pile length.

Reference is made to Fig. 2.1 for determination of ap and F.

Fig. 2.1 Criteria for capacity prediction, see /4/.

2.2.2.7 Based on a complete reevaluation of the pile test database, a revised a-method, called NCL1 (New Clay Method 1), is proposed in / l / , where the average unit skin friction over the embedded pile length is given on the form:

where

a = adhesion factor. cu = average undrained shear strength over the embed-

ded pile length. FC = soil strength correction factor. F l = pile penetration correction factor.

The soil strength correction factor, FC , is the ratio of the shear strength determined using consolidated-undrained triaxial compression tests on samples of high quality to the shear strength measured using some other techniques. The unit end resistance is according to the NCL1-method:

qP = 9C U F C

where

cu = undrained shear strength at the level of the pile tip.

2.2.2.8 According to /l/ , values for Fc should be ob-tained from local experience. In the absence of such data, a value of 1.1 for unconfined compression tests on sam-ples of high quality, 1.8 for unconfined compression tests on samples taken with typical driven samplers, and 0.7 for in-situ vane shear tests are suggested in /I /. Final design should not be based on unconfined compression tests or on samples taken with driven samplers.

2.2.2.9 Table 2.1 shows the relationship between a and cu-Fc.

Table 2.1 Relationship between a and cuFc. Cu'Fc (kPa)

0 29 58 240 oo

a 1.0 1 . 0 0.5 0.3 0.3

2.2.2.10 The correction factor for pile penetration FL varies as shown in Table 2.2.

Table 2.2 Relationship between L and Fi.. L(m) 0 30 53 oo FL 1.0 1.0 1.8 1.8

2.2.2.11 In the j?-method, as proposed in /9/, the unit skin friction, fsi, is related to the effective stress parame-ters K and 5 as follows:

where

K = average coefficient of earth pressure on pile shaft. tan S = average coefficient of friction between soil and

pile shaft. p'0 = effective overburden pressure.

For piles in normally consolidated clays inducing no ap-preciable change in lateral ground stress conditions, it may be assumed that:

K = 1 sin '

If it is further assumed that failure takes place in the re-moulded soil close to the shaft surface, the remoulded, drained angle of shearing resistance may be used for q>' along with


2.2.2.12 Other methods where the unit skin friction is considered a function of the effective overburden pressure is proposed in /12,13/.

2.2.2.13 In the A-method, see /10/, the total shaft re-sistance, Rs, is calculated from the expression:

Rs=^(^m + 2cm)A s

or

f s + 2cm)

where

fs = average unit skin friction along pile shaft. am = mean effective overburden pressure between the

mudline and the pile tip. cm = mean undrained shear strength along the pile

shaft. As = pile shaft area. A = dimensionless coefficient (see Fig. 2.2).

2.2.2.14 The pile length effect on the average unit skin friction predicted by the a-, /?- and A-methods is investi-gated in / l l / by correlating the predicted shaft capacity with relevant pile load test data. Ref. / I I / describes a modified A-method, which accounts for the pile length effect.

2.2.3 Resistance of piles in tension

2.2.3.1 For piles in cohesive soils, the pile-soil friction may be equal in tension and in compression.

2.2.3.2 No resistance from the soil below pile tip should be accounted for.

2.2.3.3 For piles with an increased base diameter the resistance is to be based on the shaft resistance or on the resistance of the base. The two contributions are, how-ever, not to be assumed to act simultaneously at the lower part of the pile.

2.3 Resistance in cohesionless soils

Fig. 2.2 Frictional coefficient A as function of depth.

2.3.1 General

2.3.1.1 Prediction of the axial capacity of driven off-shore piles in cohesionless soils (silicious sands and silts) often requires extrapolations beyond the boundaries of the database with respect to pile size, pile penetration, pile load and soil conditions. This is especially the case with piles designed for many North Sea locations, where the soils may be dense to very dense and often overconsol-idated.

2.3.1.2 Due to the uncertainties in the database, the pile design parameters should be conservatively assessed. Limiting values are normally defined for the unit skin friction and the unit end resistance. These limiting values are a function of the soil conditions, the quality and ex-tent of the soil investigation, the method of pile installa-tion, etc.

2.3.2 Resistance of piles in compression

2.3.2.1 For piles in cohesionless soils the unit skin fric-tion, fs, may be taken as:

fs = Kp'0 tan S < fj

where

K = coefficient of lateral earth pressure. = 0.8 for open-ended piles. = 1.0 for closed-ended piles.

p'0 = effective overburden pressure at the point in ques-tion.


Table 2.3 D< driven piles in 2A (1987) /8/;

>sign parameters for axial resistance of cohesionless silicious soil (source: API RP

Density Soil de-scription

5 (de-grees)

fi (kPa)

Nq ( - )

qi (MPa

Very loose Loose Medium

Silt Sand-silt 2> Silt

15 48 8 1.9

Loose Medium Dense

Sand Sand-silt 2> Silt

20 67 12 2.9

Medium Dense

Sand Sand-silt 2>

25 81 20 4.8

Dense Very dense

Sand Sand-silt 2>

30 96 40 9.6

Dense Very dense

Gravel Sand

35 115 50 12.0

1) The parameters listed in this table are intended as guidelines only. Where detailed information such as in-situ cone penetrometer tests, strength tests on high quality soil samples, model tests or pile driving perform-ance is available, other values may be justified.

2) Sand-silt includes those soils with significant fractions of both sand and silt. Strength values generally increase with increasing sand fractions and decrease with increas-ing silt fractions.

2.3.2.2 The limiting unit skin friction should normally not be taken greater than 120 kPa. In case of carbonate, granular soils, see 2.4.

2.3.2.3 The unit end resistance of plugged piles in cohesionless soils, qp, may be taken as:


shown to be justified, the coefficient of lateral earth pressure, K in 2.3.2.1, may be taken equal to 0.5 for piles in tension.

2.3.3.2 For piles with an increased base diameter, the resistance is to be based on the shaft resistance or on the resistance of the base. The two contributions are, how-ever, not to be assumed to act simultaneously at the lower part of the pile.

2.4 Resistance in calcareous soils

2.4.1 Driven piles

2.4.1.1 The axial capacity of driven piles in calcareous soils is calculated according to the same principles as adopted for piles in sands, except that the limiting unit skin friction and end resistance values are typically smal-ler. For guidance reference is made to /18/.

2.4.1.2 Factors of importance for assessment of limiting unit end and skin friction values are among others, the degree of cementation, grain crushability, relative density, compressive strength and carbonate content.

2.4.2 Drilled and grouted piles

2.4.2.1 Skin friction of drilled and grouted piles in cal-careous sands is usually higher than the friction mobilized by driven piles in the same formations. For guidance ref-erence is made to /18/.

2.4.2.2 Pile shaft resistance is limited by the shear strength of the pile/grout interface, the soil/grout inter-face or the soil itself

2.4.2.3 For cemented calcareous soil the ultimate shaft shear is often related to the unconfined compressive strength of the cemented soil.

2.4.2.4 Relationship between ultimate shaft shear and rock unconfined compressive strength to be used in ca-pacity calculations should be developed based on general experience from the location or pile load tests.

2.4.2.5 The contribution of the pile tip to the total pile capacity is dependent on a clean bottom hole.

2.5 Group effects

2.5.1 General

2.5.1.1 The group resistance of piles depends on factors such as pile spacing, type and strength of soils, sequence of soil layers, pile installation method, etc. The know-ledge of the behaviour of full-scale pile groups relative to the behaviour of individual piles in the same group is limited and conservative assumptions are therefore re-commended for the calculation of pile group resistance.

2.5.1.2 In estimating pile group resistance from a cal-culated single pile resistance, special considerations are required in each case in order to account for:

Method of pile installation Weak deposit underlying a bearing layer of limited

thickness Negative skin friction along pile shaft.

2.5.1.3 In addition to the possible limitation of the group resistance, closely spaced piles will also influence the displacements of the individual piles which is of im-portance to consider for the interaction between the structure and the pile foundation. This can in principle be done by calculating the displacement of the soil sur-rounding one pile due to the loading from the other piles in the pile group. These displacements may be calculated based on elastic halfspace solutions for constant or steadily increasing shear modulus, ref. e.g. /24/. The un-certainties related to the selection of appropriate equiv-alent soil moduli should be considered, and the choice should be related to the general stress level in the soil volume within and outside the pile group.

2.5.2 Pile groups

2.5.2.1 For a given geometry and number of piles in a group, a transition zone of pile spacing exists within which the failure mechanism gradually changes from pier failure at small spacings to individual pile failure at larger spacings.

2.5.2.2 In case of a pier failure, the axial resistance of the pile group consists of skin friction along the outer perimeter of the group plus end bearing of the pier.

2.5.2.3 Solid piers enclosing all soil within a pile group envelope (minimum pier circumference) as well as hollow piers (minimum pier area) should be considered when relevant. Limitations in tip resistance for the pier due to limitations in allowable displacements should be considered.

2.5.2.4 In the above calculations the unit skin friction fsi can be taken equal to the undrained shear strength cu in clay and tan


2.7.1.2 The effects of cyclic loading are most significant for piles in cohesive soils, in cemented calcareous soils and in finegrained cohesionless soils (silt), whereas these effects are much less in medium- to coarsegrained cohe-sionless soils. The remoulding of the soil due to pile in-stallation and the subsequent time dependent reconsolidation of the soil are important factors in the evaluation of the effects of cyclic loading in finegrained soils.

2.7.2 Evaluation of the cyclic effects

2.7.2.1 The most important factors to be considered in modelling of cyclic axial loading of piles are:

Type of cyclic loading (one-way vs. two-way, load-controlled vs. displacement-controlled) and number of cycles (at various stress levels)

Soil properties and variation of soil strength and stiff-ness with depth

Pile flexibility and pile length Static stress distribution along the pile before cyclic

loading Compatibility in terms of both cyclic and average dis-

placements and stresses.

See also 2.2.1.42.2.1.6.

2.7.2.2 For cohesive soils comprehensive research has been performed with respect to the analysis of piles sub-jected to combined static and cyclic loadings. Reference is made to /6,19/ for guidance on how to assess the effects of cyclic loading. Due to the uncertainties involved in modelling and analyzing the effects of cyclic loading the design methods proposed in the literature are normally based on a theoretical framework, which has been cali-brated against the results from small to large scale pile tests in various types of soil.

2.7.2.3 For calcareous soils the effects of cyclic loading on the capacity of both driven and drilled and grouted piles may be significant and should be evaluated from case to case for local conditions. For guidance, reference is made to /18/.

3. Lateral Pile Resistance 3.1 Introduction

3.1.1 General

3.1.1.1 This note deals with the analysis of laterally loaded piles for offshore platforms. For such piles having diameters typically 1.02.5m the most severe loading conditions arise from cyclic wave loads. Special problems related to other types of piled structures subjected to lat-eral loads will be mentioned only briefly.

3.1.1.2 The most common method for analysis of lat-erally loaded piles is based upon the use of so-called py curves. According to this method lateral load-deflection (p-y) curves are specified for simulation of the mobiliza-tion of resistance from the surrounding soil when the pile deflects. The pile is then divided into elements and the

lateral resistance of each node level will be characterized by a py curve, see Fig. 3.1.

3.1.1.3 In the construction of the py curves consider-ation should be given to the type of soil, the type of loading, the remoulding due to pile installation, the effect of scour, the effect of mud slide forces etc.

3.1.1.4 In the absence of more definite criteria for con-structing py curves semi-empirical methods supported by the results of a few well documented full-scale tests are currently in use, see e.g. /8/. The methods given in the following for calculating the lateral pile resistance in clay and sand agree largely with the recommendations given in /8/. The basis for these recommendations is found in /20,21,22/.

3.1.1.5 The designer should satisfy himself that the pile foundation is safe under the influence of both static and cyclic lateral loads, with due consideration of the actual safety requirements.

3.1.1.6 The scope of the soil investigation should be sufficiently extensive to reveal important variations in the soil properties in both the lateral and the vertical di-rections down tc a depth of at least 10 pile diameters. For evaluation of the depth of scour special knowledge about the properties of the upper soil layers is required.

3.2 Piles in cohesive soils 3.2.1 Lateral resistance in soft clay

3.2.1.1 For static lateral loads the ultimate lateral re-sistance per unit length pu for piles with diameter D in soft clay has been found to vary between 8cu D and 12cu D except at shallow depths, where failure occurs in a different mode due to minimum overburden pressure. Cyclic loads cause deterioration of lateral resistance be-low that for static loads.

3.2.1.2 In the absence of more definitive criteria the following is recommended: pu increases from 3ca D to 9cu D as X increases from 0 to XR according to:

Pu = (3cu + / X ) D + J cu X (a)


and

pu = 9cu D for X > XR (b)

where

pu = ultimate resistance per unit length (kN/m). cu = undrained shear strength for undisturbed clay soil

samples (kPa). D = pile diameter (m). y' = effective unit weight of soil (kN/m3). J = dimensionless empirical coefficient with values in

the range 0.250.50. The upper limit holds for soft normally consolidated cohesive soils.

X = depth below soil surface (m). XR = depth below soil surface to bottom of reduced re-

sistance zone in m. For a condition of constant strength with depth Eqs. (a) and (b) are solved si-multaneously to give:

Fig. 3.2 Criteria for predicting py curves. (a) short-time static loading, (b) equilibrium under initial cyclic loading and (c) reloading after cycling.

6 D

Where the strength varies with depth, Eqs. (a) and (b) may be solved by plotting the two equations, i.e. pu vs. depth. The point of first intersection of the two equations is taken to be XR. These empirical relationships may not apply where strength variations are erratic. In general, minimum values of XR should be about 2.5 pile diameters.

3.2.2 Load-deflection (py) curves for soft clay

3.2.2.1 The py curves for piles in soft clay are gener-ally non-linear. For the short-term static load case they may be generated from Table 3.1 (see Fig. 3.2 a).

Table 3.1

P/Pu y/yc 0 0

0.5 1.0 0.72 3.0 1.00 8.0 1.00 oo

In Table 3.1 the following applies:

p = actual lateral resistance (kN/m). y = actual deflection (mm). yc = 2.5 ec D (mm). c = strain which occurs at one-half the maximum

stress in laboratory undrained compression tests of undisturbed soil samples.

The form of the pre-plastic portion of the static resistance curve (up to point e in Fig. 3.2 a) can be approximated with the parabola:

1/3

= 0.5 ( ^ Pu \ yc /

3.2.2.2 For the case where equilibrium has been reached under cyclic loading, the py curves may be generated from Table 3.2 (see Fig. 3.2 b).

Table 3.2 x > x R x < x R

P/Pu y/yc P/Pu y/yc 0 0 0 0

0.5 1.0 0.5 1.0 0.72 3.0 0.72 3.0 0.72 oo 0.72 X/XR 15.0

0.72 X/XR oo

For this case the parabolic shape according to Eq. (3.2.2.1) is followed up to point d (at y/yc = 3) in Fig. 3.2 b from which point the shape is a straight line up to point f (at y/yc = 15). Depending on the ratio X/XR, the value of p/pu may vary between 0 and 0.72 as shown in the figure. For deflections y > 15yc the shape is a straight horizontal line.

3.2.2.3 For reloading after cyclic loading the py curve in Fig. 3.2 b may be modified to account for a possible gap between the soil and the pile due to previous (more intensive) cyclic loading (see Fig. 3.2 c).


3.2.3 Lateral resistance in stiff clay

3.2.3.1 For static lateral loads the ultimate unit lateral resistance pu of stiff clay (cu > 100 kPa) would vary be-tween 8cu and 12cu as for soft clay.

3.2.3.2 Due to rapid deterioration under cyclic loadings the ultimate resistance will be reduced to something con-siderably less and should be so considered in cyclic design.

3.2.4 Load-deflection (py) curves for stiff clay

3.2.4.1 While stiff clays also have non-linear stress-strain relationships, they are generally more brittle than soft clays. In developing stress-strain curves and subse-quent py curves for cyclic loads, good judgement should reflect the rapid deterioration of lateral resistance at large deflections for stiff clays.

3.3 Piles in cohesionless soils

3.3.1 Lateral resistance in sand

3.3.1.1 The ultimate lateral resistance per unit length pu of sand has been found to vary from a value at shallow depths determined by Eq. (a) following to a value at greater depth determined by Eq. (b) following. At a given depth the equation giving the smallest value of pu should be used as the ultimate resistance:

pus = (C,H + C 2 D ) 7 ' X

pud = C3 Dy'X

(a)

(b)

where

Pus = shallow ultimate resistance (kN/m).

Pud = deep ultimate resistance (kN/m).

/ = effective unit weight of soil (kN/m3). X = depth (m).

4>' = angle of internal friction of sand (degrees).

Ci, C2, C3 = coefficients determined from Fig. 3.3 as a function of 0.9 for static loading

ultimate resistance at depth H (kN/'m). initial modulus of subgrade reaction (MN/m3), in-creasing linearly with depth, determined as a function of the angle of internal friction 4>' from Fig. 3.4. actual lateral deflection (mm), depth below soil surface (m).

3.3.2 Load-deflection (py) curves for sand

3.3.2.1 The lateral soil resistance-deflection (py) re-lationships for piles in sand are also non-linear and in absence of more definitive information they may be ap-proximated at any specific depth X by the following ex-pression:


3,5 Modifications of py curves 3.5.1 Pile group effects

3.5.1.1 The influence of one pile on the behaviour of another in a group of piles should be considered when the centre to centre distance between the piles (pile spacing) is 8 pile diameters or less.

3.5.1.2 The analysis may be run as a single pile analysis as outlined herein provided that the py curves are cor-rected for shadow effects on the p-values and dis-placements effects on the y-values as a result of the group action. Modification of the y-values to account for the group ef-fect, may be done by superimposing the interaction effects calculated according to the theory of elasticity.

3.5.1.3 For further details on pile group analysis refer-ence is made to relevant literature, e.g. /23,24/.

3.5.2 Scour

3.5.2.1 Scour will lead to complete loss of lateral resist-ance down to the depth of scour and should be considered so in the construction of the py curves for the soil layer susceptible to scour, see Fig. 3.5.

3.4 Piles in calcareous soils

3.4.1 General

3.4.1.1 The materials in question have grain sizes rang-ing from clay up to gravel and appear with different de-gree of cementation. The materials are often porous and are usually quite variable with depth with dry densities in the range 1319 kN/m3. Cemented material may have unconfined compressive strengths ranging from 0.55 MPa. These materials may be characterized as weak rocks with brittle failure at less than 1 % strain. For low strain conditions their behaviour is controlled by the inter-par-ticle cementation and by the deformability of the rock mass (intact rock resistance). At larger strains, the inter-particle bonding breaks down and it may be expected that the stress-strain behaviour will principally be controlled by the frictional properties of the material (residual re-sistance).

3.4.1.2 Sound engineering judgement and a basic un-derstanding of material behaviour under static and cyclic loading are necessary attributes for design of piles in cal-careous soils. For guidance reference is made to relevant literature, see e.g. /18/.

3.5.2.2 Scour will also reduce the effective stress, p'0, further down which should be considered by using a mo-dified mudline level in the construction of py curves. This has been demonstrated in Fig. 3.5. In sand this will reduce the value both of the k-parameter and the ultimate lateral resistance, pu, defining the py curve for a certain pile element. The reduction of effective stress in relation to total scour depth should be decided in each case based on expected shape of scoured surface.

3.5.3 Reloading

3.5.3.1 The modification due to reloading is based on the assumption that the design extreme lateral load gen-erates a space between the pile and the surrounding soil. For subsequent loading (fatigue limit state and servicea-bility limit state) the effect on the pile response of this space should be considered by introducing an initial de-


flection yv for subsequent loads, see Fig. 3.6. Conserva-tively the slope of the unloading branch in the extreme load cycle is taken equal to the initial slope, k, of the loading branch. This gives for reloading the load-defor-mation curve shown in Fig. 3.6. Where the space is ex-pected to be closed by time, it may be acceptable not to include such a gap.

Fig. 3.6 py curves for extreme load (a) and subsequent loads (b).

3.5.4 Long-term loading

3.5.4.1 Structures as piled anchors can be subjected to long term static loads. Tests in both clay and sand has shown that long term loads can give deformation 23 times greater than for short term static conditions. The increase will be greater with higher stress level.

3.6.2 Approach

3.6.2.1 The ultimate situation is illustrated in Fig. 3.7. Plastic hinges are assumed to develop in the upper fixed end and at some depth below mudline. It is assumed that the pile is sufficiently long that the lower end is prevented from rotating. The fully mobilized earth pressure may be calculated according to 3.2 and 3.3 for the clay and sand, respectively.

3.6.2.2 The equilibrium of the pile can be described by:

F = P + S (a)

My l = P xj + S x2 + My2 (b)

P = fi (x2) (c)

x, = f2 (x2) (d)

where (see also Fig. 3.7):

F = design load. P = resultant of soil resistance. S = shear force in plastic hinge no.2. Myi = yield moment in plastic hinge no.l. My2 = yield moment in plastic hinge no.2. xi = distance from hinge no.l to resultant P. x2 = distance between the two hinges.

3.6 Plastic analysis of piles

3.6.1 General

3.6.1.1 In a plastic analysis of piles, plastic hinges are assumed to develop in the pile (Fig. 3.7) along with a fully mobilized earth pressure between the two hinges. This analysis may be used for the ultimate limit state and the progressive collapse limit state under certain conditions.

The analysis should be supplemented with a load-deflec-tion analysis according to 3.5.3.

J- i-

Fig. 3.7 Plastic hinges in pile and fully mobilized earth pres-sure.

3.6.2.3 When evaluating the yield moments, Myi and My2, the axial force N (Fig. 3.7) must be taken into ac-count. A reduction in the axial force due to skin friction (NiN2 in Fig. 3.7) can be found in 3.2.

3.6.2.4 When the yield moments Myi and My2 and the magnitude and distribution of the soil reaction are known, Eqs (a) through (d) may be used to solve the un-knowns (P, S, xi and x2). F must be given the lowest value for which the equations have real solutions.

4. Stability of Gravity Base Foundations 4.1 Introduction 4.1.1 General

4.1.1.1 Requirements to foundation stability are often the most decisive factor for determination of foundation area, foundation embedment (i.e. skirt penetration depth) and submerged weight for a structure with gravity type foundation. It is therefore essential in an optimal design process to give high emphasis to foundation stability cal-culations. This clause gives recommendations on how to perform stability analysis for various soil and loading conditions.

4.1.1.2 The foundation stability is most commonly solved by limiting equilibrium methods, ensuring equilib-rium between driving and resisting forces as explained further in 4.3. Using limiting equilibrium methods several failure surface(s) will have to be analysed in order to find the most critical failure.


Alternatively finite element methods may be used being able to seek the critical failure surface as part of the analysis. 4.1.1.3 General bearing capacity formulae for idealized conditions will normally be too rough for foundation stability calculations of an offshore gravity base platform. For comparison with more elaborate analyses such a simple calculation could, however, be valuable in an early stage of design. For gravity foundations with relatively small areas, as e.g. mudmat foundation for temporary support of jackets or foundations for small subsea structures, bearing capacity formulae may be acceptable. For this purpose some bearing capacity formulae are given in Section 4.

4.1.1.4 In accordance with the general definition of an ULS limit state condition, foundation failure should be defined as a situation when the deformations of the soil become so large that damage may occur to the structure or vital equipment as conductors/casings or risers. This should be considered when shear strength parameters are selected for stability analyses. For total stress analyses undrained shear strength should thus be defined at an acceptable strain level: average (permanent) strains, cy-clic strains or a combination of these. For effective stress analyses, failure is defined by reaching a shear stress characterized by:

T = c'tgP

7m where

long as the design event does not take place during in-stallation. In clayey soils consolidation takes longer time dependent on permeability, compressibility and thickness of clay layers. The degree of consolidation to be accounted for at the time of the design event should be documented by calculations or analyses as appropriate for the actual case.

4.2.1.3 For a homogeneous layer the degree of consol-idation can be calculated from the time function:

T = t-cv

where

& = normal stress on the shear surface. = angle of friction at failure.

ym = material coefficient.

4.1.1.5 The effect of cyclic loading on the shear strength should be accounted for as applicable. This will normally be required for traditional gravity base platforms where wave loading is governing for the foundation stability. For total stress analysis a cyclic strength should be de-fined as described in 4.2. For effective stress analysis the pore pressure generation due to cyclic loading, should be accounted for.

4.2 Soil shear strength 4.2.1 General 4.2.1.1 Choice of shear strength should be based on re-levant high quality laboratory and/or in-situ tests. One should consider (as applicable):

In-situ stress level Effective static stresses at time of the design event Type of loading, e.g. static loading or cyclic load his-

tory Effect of several loading scenarios, e.g. effects of

storms preceding the design storm.

4.2.1.2 Consolidation of pore pressures generated due to submerged weight of the structure can be allowed for. In sands full consolidation can normally be assumed as

t = time from application of load. cv = coefficient of consolidation, h = half thickness of layer in case of two-way free

drainage.

= thickness of layer in case of one-way drainage.

Degree of consolidation can be read from Fig. 5.10.

4.2.2 Total stress analyses 4.2.2.1 The undrained shear strength should be deter-mined from laboratory tests being able to simulate the actual stress conditions. Different type of tests will be re-quired to simulate different stress conditions as illustrated on Fig. 4.1 (from /25A/). Thus triaxial compression and extension tests as well as direct simple shear tests will normally be required, to define active shear strength, passive shear strength and direct shear strength. Selection of shear strength for a given portion of a possible shear surface may be done as described in 4.3.6.

t:me

Triaxial tests

Simple shear tests

Fig. 4.1 Example of loading of soil elements along a potential failure surface in the foundation beneath an offshore gravity platform. Simplified.


4.2.2.2 Static undrained shear strength may be used for cases where the governing load has a mainly static char-acter. Rate effects may be included for loads of short duration as e.g. peak collision forces. The rate effects should then be properly documented by tests. 4.2.2.3 The effect of cyclic (wave) loading on the shear strength should always be considered. Cyclic loading may cause a build-up of pore pressures leading to a reduction of shear strength. The general effects are described by se-veral authors, see /26,27/. For soil subjected to cyclic loading one may define two undrained shear strengths:

1) Static shear strength reduced for the effect of cyclic loading.

2) Cyclic shear strength, being defined as the sum of static and cyclic stress that causes failure for a given number of cycles.

In cases where cyclic effects are important, the cyclic shear strength is recommended. As basis for stability analysis using cyclic shear strength, diagrams should be developed showing combinations of average shear stress, ia, and cyclic shear stress, Tcy, that lead to failure for various number of cycles as illustrated on Fig. 4.2 (from /28/). The cyclic shear strength if,cy = (ra -I- Tcy)r may be deter-mined from the diagrams in Fig. 4.2 and presented in di-agrams as illustrated in Fig. 4.3 (from /28/). The failure mode, i.e. the combination of average and cyclic shear strains at failure, ya and ycy are indicated along the curves. It should be emphasized that these figures are example diagrams, not necessarily applicable for a specific design. An approach for cyclic stability analysis utilizing such cyclic strength diagrams is described in 4.3.7.

x

V f f ' v c 0.00 0.05 0.10 0.15 0 20

Mum ii Ulii in

0.0 0.2 0.4 0.6 o 8

T a / S * 5

a) b) Fig. 4.2 Example of combinations of average shear stress ia and cyclic shear stress rcy that cause failure for various number of cycles for a) triaxial tests and b) direct simple shear tests. Example for normally consolidated Drammen clay, see /28/.


a)

b) .0 0.2 0.4 0.6 0.8 1.

x / s DSS

Fig. 4.3 Example of cyclic strength as a function of average shear stress for various number of cycles to failure for a) triaxial tests and b) direct simple shear tests. Derived from Fig. 4.2, see /28/.

4.2.2.4 Total stress analysis using cyclic shear strength is very suitable for analysis of clayey soils, which nor-mally can be considered undfained for all load effects during a design storm. If considering the effect of the ULS design storm only, the event should be considered to take place the first stormy season after installation of the structure. The beneficial effect of dissipation of pore pressures caused by platform submerged weight as de-scribed in 4.2.1.2 and 4.2.1.3 can be taken into account. For big structures on very thick deposits of clay, it might be required to investigate the effect of storms occurring prior to the design storm, e.g. considering the design storm to come at the end of the design life. The deteri-orating effect of several storms prior to the design storm may be counteracted by the beneficial effect of further consolidation of pore pressures from platform submerged weight. It is allowed to account for this effect.

4.2.2.5 The load duration of a single wave may be so short that even a sand will be basically undrained during the time of the design wave. Very dense sand, as often encountered offshore, may ex-perience large negative excess pore pressures (dilative be-haviour) and corresponding high undrained shear strengths when sheared to failure under undrained load-ing conditions. One should be careful, however, to rely upon the high undrained shear strength which depends upon large negative pore pressures. The possibility for such high negative pore pressures to dissipate should be considered, and so should the strain level required to

reach the corresponding stress conditions. Finally, one should not rely upon a high static shear strength with-out considering the effect of cyclic loading as discussed in 4.3.6.3.

4.2.2.6 In approximate analyses the following formulae which depend on limited dilation will normally be ac-cepted for determination of undrained shear strength of very dense sand. Active triaxial strength SUA characterized by increase of vertical stress:

SuA = sin '

1 sin 7 h

Active triaxial strength characterized by decrease of hor-izontal stress:

_ sin 4>' ^uA - i , - i, v 1 4- sin '

1 sin '

Passive triaxial strength characterized by decrease of ver-tical stress:

SuP = " sin '

1 + sin 4V


Direct shear strength SuD for horizontal parts of shear surface:

SUD = ff'vtg^' where

tr'v f = angle of friction at failure. 7m = (required) material coefficient.

= 1.2 according to the Rules /25/.

4.2.3.2 When working with effective stresses all con-tributions to pore pressures must be included:

Initial in-situ pore pressures, which may be different from hydrostatic pressures if the soil is not fully con-solidated for its own weight.

Pore pressure due to installation, including effect of added weight of structure and effect of skirt pene-tration, accounting the dissipation taking place after time of installation until the design event.

Pore pressure due to cyclic loading, including build-up and dissipations as relevant.

Pore pressure due to transient loading.

4.2.3.3 In-situ pore pressures should result from the soil investigations. Deviation from hydrostatic pressures are only to be expected for thick clay deposits with high de-positional rate, e.g. delta deposits. 4.2.3.4 Pore pressures due to weight of structure is de-termined based on initial pore pressures calculated as for transient loading described in 4.2.3.6 and based on stan-dard consolidation analyses.

4.2.3.5 Pore pressure build-up due to cyclic loading may be derived from pore pressure contour diagrams derived from undrained cyclic test. An example of pore pressure contour diagrams is shown on Fig. 4.4. Each curve shows number of cycles required at various stress levels to reach a given pore pressure. Curves need to be established for the specific soil. For low to moderate stress levels and pore pressure response the model given in /29/ may be used:

Ru =A crd d N d u where

Ru = cumulative pore pressure resistance against re-peated loading.

A


Fig. 4.4 Example of pore pressure contour diagram.

4.2.3.6 Pore pressures due to transient undrained load-ing can be defined from the following formula:

A u =A


real safety ievel depending on whether the soil behaves dilatant or contractant when approaching ultimate failure (Fig. 4.5). This should be considered and discussed when selecting the design soil parameters for an effective stress analysis.

4.2.3.8 Use of effective stress formulations in stability analysis for undrained conditions has been presented in /30/, where general bearing capacity diagrams are given also accounting for the effect of pore pressures. General formulations for the limit state stress conditions, that can be used for general limiting equilibrium solutions are given in /31/. These formulations account for the stress path defined by the dilatancy parameter D and the effect of the intermediate principal stress


tration etc.). Some typical failure modes to be considered are:

Sliding along base of skirt tip Sliding along soft layer below skirt tip Sliding at base with local failure around skirt tips Conventional deepseated bearing failure Deepseated failures governed by moment equilibrium

with centre of rotation above or below the foundation base.

The latter modes become increasingly important for plat-forms with high ratio of overturning moment to horizon-tal force. The above modes are illustrated on Fig. 4.7.

SLIDING AT SKIRT TIPS

ttump.,

SLIDING ALONG SOFT LAYER

LOCAL FAILURE ALONG SKIRTS

M I SAGA.

DCEP-SEATEO BEARING

CAPACITY FAILURE

MOMENT EQUILIBRIUM CENTRE LOCATED ANYWHERE

MOMENT EQUILIBRIUM CENTRE BELOW FOUNDATION B A S E

Fig. 4.7 Example of possible failure modes.

4.3.5 Side resistance

4.3.5.1 Limiting equilibrium solutions are based on plain strain-stress condition providing resistance per unit length of the foundation. This resistance is multiplied with the actual length of the foundation, and the 3-D ef-fect is included by calculating a resistance on the two vertical sides of the failure surface. For total stress anal-ysis of structures on soil with relatively constant shear strength, the side shear resistance at failure can be taken as 0.4 times the undrained direct simple shear strength. For soil with increasing undrained shear strength with depth one should use a lower value, however. For the soil resistance below the base of the foundation, i.e. below skirt tip, it may in this case be relevant to use zero side resistance (compare with Table 4.2 showing decreasing shape factors for high rate of shear strength increase with depth). Any chosen side shear resistance should be well docume

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