DETERMINATION OF BEARING CAPACITY OF GRAVEL …igs/ldh/conf/2012/A.pdf · Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No A 102) ... ISMB 150/ISMC

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No A 102)

DETERMINATION OF BEARING CAPACITY OF GRAVEL BOULDER DEPOSIT FROM FOOTING LOAD TEST

Hari Dev, Scientist-C, CSMRS, [email protected] Rajbal Singh, Scientist-E, CSMRS, [email protected] G.V. Ramana, Scientist-B, CSMRS, [email protected] R.P. Yadav, Research Assistant, CSMRS, [email protected] ABSTRACT: Evaluation of bearing capacity of soil or gravel-boulder strata is a challenging task as it involves very cumbersome field arrangements of testing. The load displacement characteristics of gravel boulder deposits depends upon various factors, viz. type of soil, compaction/interlocking of soil molecules and more importantly size of boulders beneath the footing. Footing load test deals with the subsurface investigation in relation to design of foundations for single and multi-storeyed buildings, overhead water tanks, piers and abutments of bridges in boulder-gravel deposits. Load-settlement curve can also be utilized to determine the yield pressure and recommended allowable pressure for foundation design. The field tests are the only tool to get the in-situ characteristics of gravel boulder deposits. The laboratory tests conducted on the tiny samples do not represent the exact behaviour of such a complex matrix. The present paper discusses the results and interpretation of the in-situ footing load tests conducted at proposed colony at Sangaldan station yard of Udhampur-Srinagar-Baramulla Rail Link (USBRL) Project. INTRODUCTION The Jammu-Udhampur-Katra-Quazigund-Baramulla Railway line is the biggest project in the construction of a mountain railway since independence. From Jammu to Baramulla, length of the new rail line is 345 km. It passes through the young Himalayas, tectonic thrusts and faults. The proposed 950 m long Sangaldan Station yard is located between km 92.360 km and 93.310 km and is part of USBRL project. It involves cutting from 17 m to 55 m height. Geotechnical investigation revealed that rock is not available up to formation level of the station yard. This paper deals with evaluation of bearing capacity based on two footing load tests conducted at gravel boulder deposits on the slopes of sangaldan station yard of USBRL project in Jammu and Kashmir. GEOLOGICAL SETUP AT SANGALDAN Due to folding/shearing and presence of Murree thrust on the northern side of the yard area, rock units are closely fractured, shattered and broken in to small pieces (especially sandstone). With the passage of time, these rock units have undergone the process of weathering and erosion. Subsequently, heavy slide material have deposited on the

underlying rocks. In the course of time, slided rock material partly thrusted within soft and weathered clay stone/siltstone. Due to this, rock mass has become soft and at places appear like overburden. The soil matrix at the testing locations included the boulders even larger than 300 mm, generally mixed with fine (4.75 - 20 mm) to coarse (20 - 80 mm) gravels. BEARING CAPACITY OF SOIL-BOULDER DEPOSIT Bearing capacity is the ability of soil to safely carry the pressure placed on the soil from any engineered structure without undergoing a shear failure with accompanying large settlements. Considering the large size of aggregates involved large sized footing need to be used in the test. Scale effect of the size of plate or loading area has been depicted in Fig. 1. Steel plates are not of much use because of seating difficulties. Best results are obtained with cast in-situ concrete blocks or with pre-cast blocks, set with fresh mortar, so that there is perfect bond between the soil and the block. The size of the footing shall be such that it will span over several boulders so as to mobilize group action under load. To satisfy this criterion, the minimum size of the footing shall not be less than 10 times the average grain size to a minimum of 150 cm (IS 10042 - 1987). Safe bearing capacity may be determined by dividing the ultimate bearing capacity by a suitable factor of safety. Factor of safety is often determined to limit settlements to less than 25 mm and generally it ranges in between 2 to 4. The behaviour of soil-boulder deposit is governed by the size, quantity and distribution of the boulder in the filler.

Fig. 1: Illustration of scale effect in foundation design

Hari Dev, Rajbal Singh, G.V.Ramana, R.P.Yadav

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TESTING METHOD Bearing capacity of soil or gravel-boulder deposit can be determined either by conducting plate load test or cast in-situ footings. Because of the stress influence below the foundation, footing load test is generally preferred over the plate load test. In the present investigation, cast in-situ footing load tests were conducted in the proposed colony area on the slope at Sangaldan station yard. Soils with a large quantity of gravel-boulder deposit pose several problems in testing. Presence of large sized particles precludes the sampling by usual methods of soil sampling in true sense. Tests on disturbed samples are likely to yield unreliable results, as natural arrangements of the grains and matrix material are never achieved by re-compaction. Best results are obtained by properly chosen field tests. Therefore, the cast in-situ footing size was kept as 150x150x35 cm in accordance with the provisions of IS Code of practice for ‘Site Investigations for foundation in Gravel-Boulder Deposit’ (IS 10042: 1981). A sketch showing the cast in-situ RCC footing alongwith the position of dial gauges is shown in Fig. 2. Photographs of footings 1 and 2 are given in Figs. 3 and 4, respectively. Loading was provided through a kent-ledge arrangement of ISMB 150/ISMC 100 sections, MS plates, C.G.I. sheets and sand bags/boulders. Settlement at each increment of load was measured by four dial gauges and load versus settlement curves were plotted. The bearing capacity and the settlement of the foundation were determined with the help of load settlement curve for the actual footing.

a) Cross-Section

b) Plan

Fig. 2: Cast in-situ RCC footing and position of dial gauges

Fig. 3: Cast in-situ footing 1

Fig. 4: Cast in-situ footing 2 TESTING PROCEDURE The foot-loading testing block of 150x150x35 cm square in shape was used. The proposed foundation area was excavated upto around 2.0 m depth. The foot-loading test seated at the centre over a fine sand layer of 5 mm thick. A Kent-ledge was made by preparing a loading platform with ISMB 150/ISMC 100 and MS plates, sand bags and small rock boulders were placed over this platform to obtain a total load of around 35 tonnes. Load was applied on the footing by hydraulic jack of 200 ton capacity by reaction obtained from the Kent-ledge (Figs. 5). Load on the hydraulic jack was recorded by a calibrated load gauge placed between the jack and hydraulic pump. A seating load of around 1.6 tonnes (70 gms/cm2 as per IS 1888-1982) was first applied and released after some time. Loads were applied in the incremental intervals of 1/5th of the design load upto failure or until a settlement of 40 mm had occurred, which ever was earlier. At each loading increments, settlement was recorded at the time intervals of 1, 4, 10, 20, 40 and 60 minutes by means of 4 displacement dial gauges with 0.01 mm least count and 25 mm travel placed diametrically opposite direction on the footing. The settlements were recorded with reference to datum bar. The load increment was kept for not less than 1 hour or upto a time when the rate of settlement gets appreciably reduced to a value of 0.02 mm/min which ever is

150 cm

150 cm 60 cm

60 cm D3

D4

D1

D2

300 cm

300 cm

150 cm

60 cm

35 cm

50 cm

Ground Line

Cast In-situ RCC Footing

160 cm

Ground Line

Determination of Bearing Capacity of Gravel Boulder Deposit From Footing Load Test earlier, the next increment of load shall then be applied and the observations repeated. Load settlement curves for foot-loading tests were plotted from the obtained test data. TEST RESULTS The footings 1 and 2 were loaded to the extent of 30 tonnes of load in five increments of 6, 12, 18, 24 and 30 tons. The displacements of the foundation were recorded with the help of 4 digital dial gauges with an accuracy of 0.01 mm.

Fig. 5: Loading setup for footing load test Differential settlement must be limited to avoid cracking and other damage in structures. A typical allowable differential/span length ratio (δ/L) for steel and concrete frame structures is 1/500 where δ is the differential movement within a span length L (ASTM D 1194). Footing load test no. 1 (FLT-1) was conducted at 93941.761 m chainage and RL 1298.427 m. The load versus settlement plot for FLT-1 is given in Fig. 6. Load versus settlement curve for individual dial gauges was also drawn and is given in Fig. 7.

L OAD S E TTL EME NT C UR VE

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Fig. 6: Load versus settlement curve for FLT-1


‐26‐24‐22‐20‐18‐16‐14‐12‐10‐8‐6‐4‐20

0 3 6 9 12 15 18 21 24 27 30 33L oad (Tonnes)

Settle

men

t (mm)

Total S ettlement

S ettlement at D‐1




Fig. 7: Load versus settlement curve for individual dial gauges (FLT-2) It may be seen from Fig. 9 that the settlements recorded by dial gauge no. D4 was only 5.25 mm and was comparatively less than other dial gauges. The settlement from dial gauge D2 was the maximum i.e. 24.10 mm (Fig. 7). It may be attributed to the fact that a boulder was seen in one corner and dial gauge no. 4 was installed on that side only. The average settlement from all 4 dial gauges was 14.96 mm with a variation from 5.25 mm to 24.10 mm. Since, distinct failure has not occurred, the final load was taken as ultimate load based on the load settlement curve. Footing load test no 2 (FLT-2) was carried out at chainage 93254.726 m and RL 1306.837 m of proposed residential colony at Sangaldan station yard area. Load versus average settlement plot is shown in Fig. 8 and settlements recorded through individual dial gauges have been plotted against load in Fig. 9. There was not much variation of settlement recorded by the four individual dial gauges as compared with FLT-1. The settlements recorded by the four dial gauges range from 8.56 mm to 12.96 mm. Maximum settlement of 12.96 mm was recorded by dial gauge D1 and minimum of 8.56 mm by D4 (Fig. 10). The average settlement from all 4 dial gauges was 10.52 mm with a variation from 8.56 mm to 12.96 mm. Since, distinct failure has not occurred, the final load was taken as ultimate load based on the load settlement curve. In view of the above, ultimate bearing capacity of the strata corresponding to the maximum design load was calculated as follows: Ultimate load (design load) = 30 tons. Ultimate bearing capacity of formation at design load = 30/(1.5x1.5) = 13.33 T/m2

ISMB 150

Sandbags

MS Plates 500 x 500 x 2.5 cm ISMB 150

Upslope

Sandbags

Hari Dev, Rajbal Singh, G.V.Ramana, R.P.Yadav

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Fig. 9: Load versus settlement curve for FLT-2 The uneven displacements measured by the dial gauges at four corners indicated the differential settlement of the footing. The differential settlement of the block in one of the tests authenticated the presence of boulders underneath the footing. Presence of large boulders at the contact surface plays a major role in resistance to settlement. Since, distinct failure has not occurred, final load was taken as ultimate load based on the load settlement curve. Bearing capacity of boulder formation based on design load and load settlement curve may be taken as 13.33 t/m2. For better understanding or clear shear failure point of view, the test may be required to continue with higher load.


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S ettlement of D‐1




Fig. 10: Load versus settlement curve for individual dial gauges (FLT-2) CONCLUSIONS Based upon the test results of two footing load tests, the ultimate bearing capacity (at design load) of the geological formation at proposed residential colony Sangaldan station yard may be taken as 13.33 t/m2. However, since the failure had not occurred at the specified design load, the ultimate bearing capacity and safe bearing capacity of the formation may be higher which may be ascertained by applying higher load.

REFERENCES 1. ASTM D1194 (1989). Bearing Capacity of Soil for

Static Load and Spread Footings. Annual Book of ASTM Standards 04.08, 192±194.

2. IS 10042 (1982, 1987): Indian Standard code of Practice for ‘In-situ Footing load test on Gravel-Boulder Deposit’

3. IS 1888 (1982 – Reaffirmed in 2002): Indian Standard code of practice for method of load test on soils.

4. IS 8009 (1976) – Part I: Indian Standard Code of practice for calculation of settlement of foundation.

5. IS 6403:1981 Indian Standard Code of practice for determination of bearing capacity of shallow foundations

Proceedings of Indian Geotechnical Conference December Delhi (Paper No. A103)

STRESS-STRAIN RESPONSE OF FIBRE REINFORCED ADMIXED SOIL J.S.Dhanya, Post Graduate Student, Anna University, Chennai-600 025, India. E-mail: [email protected]. K. Ilamparuthi, Professor, Anna University, Chennai–600 025, India .E-mail: [email protected]. ABSTRACT: The present research work is proposed to study the stress-strain behavior of fiber reinforced admixed soil. Tests are conducted to quantify improvement in engineering properties of soil by stabilizing it using two major waste materials namely flyash and plastics. The fiber properties like fiber length and fibre content are kept as varying parameters with soil and artificially cemented soil and tested. The strength of the reinforced soil is compared with the unstabilized (virgin) soil to bring out the effectiveness of randomly oriented fiber inclusions. Inclusion of plastic fibres enhanced the soil strength irrespective of fibre content and length of fibre. Though addition of flyash brought out significant improvement in strength of admixed clay, it induced brittleness to the stabilized soil. However addition of flyash and fibres proved better combination by compensating the limitation of one material by the other. Among the combinations studied, 15% flyash and fibre content of 1.5% with fibre length of 20mm provided better performance. INTRODUCTION Fiber reinforcement is a technology to improve the strength of soils used as fill in geotechnical structures. Reinforcement interacts with the soil through friction and adhesion. This technique is highly suitable for stabilization of soil layers, repairing failed slopes, soil strengthening around the footings and earth retaining structures. Similarly, cementing materials or other chemicals are added to natural soil to improve one or more of its properties. The project is intended to study the stress-strain behavior of soil which is reinforced and admixed so as to improve its properties. To reinforce the soil the waste plastic bottle is cut into thin strips (referred as fibres hereafter) and mixed with soil along with flyash and the influence of flyash content, fibre content and fibre length are studied by conducting strength tests. Studies concerning flyash utilization for soil stabilization have been conducted in the past by many investigators like Maher et al (1993) and Consoli et al (2002). The results of direct shear tests performed by Sadek et al (2010) on sand specimens mixed with nylon fibers indicated the increase in shear strength with the addition of fibres. These results were supported by a number of researchers like Consoli et al (2002) and Al-Refeai (1991). The improvement in ductility characteristic of the soil with the addition of fibres was studied by Ranjan et al (1999). Choudhary et al (2010) indicated that fibre reinforced soil exhibit a higher CBR value than the unreinforced soil. Studies by Muntohar (2009) and Kumar et al (2007) on various types of fibre reinforced admixed soil indicted that upto some optimum fibre content and admixture content, both the tensile and compressive properties of the soil improved significantly. The present study investigated the effect of various factors such as flyash content, aspect ratio, fibre length and fibre contend on the stress-strain behavior of the soil through experiments. The test results thus obtained are interpreted and presented in this paper. EXPERIMENTAL WORK The soil used for this study is low plastic clay having specific gravity of 2.6, liquid limit 32% and plasticity index 20%.The

maximum dry density and optimum moisture content of the soil as determined from the standard proctor test were 18.7 kN/m3 and 13.7% respectively. For the present study waste plastic bottles are cut into fibres of length 10mm and 20mm having thickness of approximately 1.5mm. Flyash used as admixture in this study is composed of Silica (SiO2) 35.72%, Alumina (Al2O3) 21.33% and Calcium (CaO) 4.56%. A series of UCC and CBR tests are conducted on virgin soil and soil mixed with varying combinations of flyash content (5%,10% and 15%), fibre content (0.5%,1% and 1.5%) and fibre length (10mm and 20mm) at their corresponding MDD and OMCs. Effect of flyash and fibres on strength of the soil The addition of flyash reduced the liquid limit and plasticity index of the soil significantly. The liquid limit and plasticity index of the soil is reduced by 25% and 15%with addition of 15% flyash. Further standard proctor compaction tests conducted on the soil has shown marginal reduction in maximum dry density (MDD) and increase in OMC. This implies that more water is needed for compacting the soil with flyash mixture to achieve maximum density .

Fig 1 UCC test for soil stabilized with flyash

J.S.Dhanya & K.Ilamparuthi

UCC tests is carried out for virgin soil and soil stabilized with 5%, 10% and 15% flyash combinations at their corresponding OMC and dry density at a strain rate of 0.05mm/min. The stress-strain response are shown in Fig 1.The increase in the UCS with addition of flyash is attributed to change in fabric arrangement and mechanical binding of the particles of flyash and clay due to energy imparted to the soil-flyash mixture The axial stress of stabilized soil was mobilized at a lower axial strain that that of virgin soil. The results of UCC test conducted on soil reinforced with 0.5%, 10% and 15% fibre content using fibres of 10mm and 20mm lengths at their respective OMCs and MDDs revealed that there is a slight increase in axial strain corresponding to peak stress with increase in fibre content. It is further observed that the brittleness reduces and peak stress increases with inclusion of fibers as shown in Fig 2. The peak strength of the soil was increased by 77% with the addition of 1.5% fibres of 20mm length.This can be explained that the total contact area between fibers and soil particles increases while increasing the fiber content and consequently the friction between them increases. Further the slip or pullout of fibre is also less for the longer fibre.

Fig 2 UCC test for soil reinforced with 10mm fibre length

Fig 3 UCC test for soil reinforced with 20mm fibre length and 10% flyash content and varying fibre content

Fig 4 Influence of fibre content and flyash content on the UCC strength of the soil for 10mm fibres The curves obtained from UCC test results conducted on fibre reinforced admixed soil indicate that the mixing of flyash increases the peak compressive strength while the fibre addition induces ductile behavior to the soil. Thus the overall strength of the soil increases as in Fig 3.The addition of 1.5% of 10mm fibres along with the addition of 5%,10% and 15% flyash increased the peak strength of the virgin soil by 1.75, 1.85 and 1.97 times respectively as shown in Fig 4.It is also observed that the 20mm long fibres performed better with flyash stabilized soil than the 10mm long fibres Effect flyash and fibres on CBR characteristics of the soil The CBR test was conducted for virgin soil and soil stabilized with flyash under soaked condition at their respective MDD and OMCs. The load-penetration curves obtained from the CBR tests for soil mixed with 5%, 10% and 15% flyash are shown in Fig 5. The addition of flyash increased the soaked CBR load for a given penetration

Fig 5 Variation of soaked CBR value of virgin soil and soil stabilized with flyash

Stress-strain response of fibre reinforced admixed soil

It can be observed that the initial slope of the load-penetration curve is also improved due to the addition of flyash. This shows that the addition of flyash increases the stiffness of clayey soil. The soaked CBR value was increased by 82% due to the addition of 15% flyash. The increase in soaked CBR value was attributed to the increase in the content of finer particles which enhanced the mechanical binding in the stabilized mixture as well as from the chemical reaction due to the pozzolanic property of the flyash during the period of soaking.

Fig 6 Variation of soaked CBR value of virgin soil and soil reinforced with 20mm fibre length The CBR test was conducted for soil reinforced with 10mm and 20mm fibres for varying fibre content in soaked condition at the respective dry density and OMC of the soil-fibre mixtures and the penetration curves are presented in Fig 6. The soaked CBR value of virgin soil (i.e., 2.59%) increased approximately twice with the addition of 0.5% of 10mm and 20mm fibres as shown in Fig 7.The increase in the CBR is due the additional resistance by way of interlocking offered by the fibres present in the soil. It is further revealed that the increase in the CBR value is more pronounced for higher fibre content and fibre length .

Fig 7 Influence of fibre content on the soaked CBR value of virgin soil

It is observed from the CBR test conducted on fibre reinforced admixed soil that the the addition of 1.5% of 10mm fibres along with the addition of 5%, 10% and 15% flyash increased the strength of the virgin soil by 4.29, 4.86 and 5.51 times respectively. Similarly the addition of 1.5% of 20mm fibres in flyash admixed soil with content of 5%, 10% and 15% flyash increased the peak strength of the virgin soil by 4.79, 5.58 and 6.29 times as shown in Fig 8. It is also clear that the increase in the CBR value is more pronounced for higher fibre content and fibre length with the highest value obtained for 1.5% of 20mm fibres.

Fig 8 Influence of fibre content and flyash content on the soaked CBR value of the soil for 20mm fibres CONCLUSIONS The following conclusions were drawn from the UCC and CBR tests conducted on soil mixed with fiber and cementitious materials individually and also in combination: 1. The addition of flyash significantly improved the peak

uniaxial compressive strength of the soil. However the effect of brittleness increased with flyash addition. In case of fibre reinforced soil, there is a significant increase in peak strength with increase in fibre length and fibre content and slight increase in corresponding axial strain. In case of fibre reinforced admixed soil, the UCC strength is greater than that of individually fibre reinforced and flyash stabilized soils.

2. There is a significant increase in soaked CBR value for flyash stabilized soil due to mechanical binding in the stabilized mixture between clay and flyash particles as well as from the possible chemical reaction with clay particles due to the pozzolanic property of the flyash during the period of soaking. For fibre reinforced soil the increase in the CBR value is more pronounced for higher fibre content and fibre length with the highest value obtained for the fibre content of 1.5% with fibre length of 20mm.The cementitious behavior of the flyash and the frictional resistance offered by the fibres contribute higher CBR value to the fibre reinforced admixed soil

J.S.Dhanya & K.Ilamparuthi

The conclusions drawn here are valid for the clay of low plastic chosen for the study. This study needs to be extended for other soils particularly problematic soils to prove the effectiveness of using flyash and fibres in combination to enhance engineering response of the soil. This study is limited to laboratory; however field trial with monitoring over a year is needed to prove its effectiveness under varied climatic seasons. REFERENCES 1. AI-Refeai, T.O. “Behavior of granular soils reinforced

with discrete randomly oriented inclusions.” J. Geotextiles and Geomembranes,Vol.10(4) ,(1991), pp. 319-333

2. Choudhary A.K., Jha J.N. and Gill K.S. “A study on CBR behavior of waste plastic strip reinforced soil” Emirates Journal for Engineering Research, Vol. 15, No.1, 2010.

3. Consoli.N and Montardo.J, “Engineering Behavior of Sand Reinforced with Plastic Waste”, ASCE Journal of Geotechnical and Geoenvironmental Engineering, Vol. 128, No. 6, June 1, 2002.

4. Maher, M.H. and Gray, D.H. “Static response of sands reinforced with randomly distributed fibres.” Journal of Geotechnical Engineering, ASCE, Vol.116(7),(1990), pp.1661-1677

5. Kumar, R., Kanaujia, V.K. and Chandra, D “Engineering Behavior of Fibre-Reinforced Pond ash and Silty Sand”, Geosynthetics International, Vol.6 (6), (1999), pp.509-518

6. Muntohar A. S, “Influence of Plastic Waste Fibers on the Strength of Lime-Rice Husk Ash Stabilized Clay Soil”, Civil Engineering Dimension, Vol. 11, No. 1, March 2009, 32-40

7. Ranjan, G, Vasan, R.M. and Charan, H.D. “Probabilistic analysis of randomly distributed fibre-reinforced soil.”Journal of Geotechnical Engineering, ASCE, Vol.122 (6)(1996), pp.419-426.

8. Sadek.S, Najjar.S and Freiha.F “Shear strength of Fiber Reinforced Sands”, ASCE Journal of Geotechnical and Geoenvironmental Engineering, Vol. 136, No. 3, March 1, 2010

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No. A 104)

A STUDY ON THE INFLUENCE OF SOIL-MOISTURE MEASURING METHODOLOGIES ON SWCC

Malaya C., Assistant Professor, Assam Engineering College, [email protected] Sreedeep S., Associate Professor, Indian Institute of Technology Guwahati, [email protected] ABSTRACT: Measurement of the soil-moisture is of prime importance in many of the geotechnical and geoenvironmental applications. Several methodologies have been developed and used by the researchers for measuring soil moisture. While employing different methodologies for soil-moisture measurements, there is a possibility that various factors (viz., measurement methodology, type of the soil, range of soil moisture measurement, presence of salts or contaminants in the soil etc.) would influence the obtained results. Therefore, it becomes essential to investigate the uniqueness of moisture measurements obtained using some commonly adopted moisture measurement methodologies. The capacitance sensor is one of the most common methods of moisture measurement due to its low cost, simple measurement procedure, and robustness. The present study has used three capacitance sensors for moisture measurements to obtain the soil-water characteristic curve (SWCC) of a locally available fine-grained soil. The study indicates that the moisture measurements obtained using different methodologies may not be unique. As such, there is a need to calibrate the moisture measurement equipments for each type of soil to establish a new calibration equation, instead of using the manufacturer’s calibration so as to facilitate the moisture measurement of soil precisely. INTRODUCTION The amount of water present in the porous media such as soil and rock is commonly defined by using gravimetric water content (w) or volumetric water content (θ). θ is defined as the ratio of volume of water to the total volume of the porous media. Measurement of w needs soil samples to be extracted (destructive sampling) and requires a minimum duration of 24 h. for oven drying [1]. This has resulted in the emergence of θ as an important state variable, which can be measured instantaneously and non-destructively. Also, with the significant development in electronics, there has been a considerable increase in the development of θ measuring methodologies [2]. It is also felt that θ is a better representation of volumetric properties of the porous media [3]. In view of this, researchers have developed a lot of correlations, which are based on θ for determining various properties of the porous media such as its soil-water characteristics, volume compressibility, infiltration capacity, reactivity etc. Therefore, the measured θ value plays an important role in the modeling of several phenomenon related to agriculture, soil science, geotechnical and geoenvironmental engineering, waste management, watershed modeling etc. [4-7]. Several researchers have developed different methodologies such as time domain reflectometry (TDR), capacitance, impedance, theta, neutron probes for measuring θ [8-12]. Since, there are different methodologies for measuring volumetric water content measurement, there is a possibility that the inherent features of a methodology and measurement procedure would significantly influence the θ measurement and hence, the soil-water characteristics. To investigate this, some of the commonly adopted methodologies for measuring θ such as EC-5, EC-TE and SM200 were carefully evaluated

and compared by determining soil-water characteristic curve (SWCC) of a locally available fine-grained soil. The study brings out the fact that SWCCs established using different θ measuring methodologies may not be unique. Such an observation clearly highlights the need for careful evaluation of the SWCC before employing it for modeling unsaturated soil behaviour. The study indicates that it is very difficult to ensure the validity of a calibration equation for a wide range of soil type. There was no consensus in measured volumetric water content for the three probes when general calibration equation was adopted. The study recommends laboratory evaluation of all θ measuring probes before employing it for establishing SWCC. MATERIALS AND METHODS A locally available fine-grained soil designated as RS was used in this study. The soil is characterized for its specific gravity, grain size distribution, liquid limit and plastic limit by following the guidelines presented in the literature [13-15]. The details of the characterization are listed in Table 1. It can be noted that the soil is clayey soil of low compressibility. Table 1 Physical properties and classification of the soil, RS

Property Soil, RSSpecific gravity 2.62Particle size (mm)Sand (4.75-2 mm) 26Silt (0.075-0.002 mm) 67Clay (<0.002mm)) 7Liquid limit 46Plastic limit 27Soil classification (USSC) CL

Malaya C. & Sreedeep S.

Equitensiometer (Delta-T Devices, UK) was used to measure ψm and EC-5, EC-TE (Decagon Inc., USA) and SM200 (Delta-T Devices, UK) volumetric water content sensors were employed for measuring θ of soil. The equitensiometer consists of a precision soil moisture sensor, the ThetaProbe, whose measuring rods are embedded in a porous material, the equilibrium body. The porous material has a known, stable relationship between water content and matric suction. When the equitensiometer is inserted into the soil for ψm measurement, the matric suction within the equilibrium body rapidly equilibrates to that of surrounding soil. The water content of the equilibrium body is measured directly by the ThetaProbe and gives the output in milliVolt (mV) [16], and this is converted to the matric suction, ψm of the surrounding soil using the calibration curve supplied by the manufacturer (Delta-T Devices, UK). The EC-5 and EC-TE are low cost sensors manufactured by Decagon Inc., USA. These probes determine θ based on the dielectric constant or permittivity of the material in which they are inserted. EC-5 is two pronged where as EC-TE is a three pronged probe. The details of these probes are reported in the literature [2,17]. The sensor SM200 is a low cost frequency domain reflectometry probe manufactured by Delta-T Devices, UK. SM200 measures soil dielectric with a 100 MHz waveform. The sensor consists of a sealed plastic body attached to two stainless steel sensing rods. The probe is fitted with a waterproof connector, which is easily connected to extension cables. For measuring θ, the probe rods are inserted directly into the soil. The output from an SM200 is a simple analogue dc voltage. This output is converted into θ using the supplied general soil calibration equation. The details of this SM200 probe are reported in the literature [18]. The measurements using equitensiometer, EC-5, EC-TE and SM200 probes are performed employing the test set up as shown in Fig. 1(a). Figure 1(b) shows the arrangement of the probes in soil sample for suction and volumetric water content measurements.

(a)

(b)

Fig. 1 Diagrammatic representation of the laboratory test set

up

The test set up essentially consists of a perspex container with a base plate into which the soil is compacted. The ψm and θ measuring probes are connected to a computer through respective data loggers.

The air-dried soil is compacted at required compaction state in the perspex mold of 300 mm diameter and 160 mm height. The EC-5, EC-TE and SM200 probes are inserted directly into the soil, and the sharpened tip of the probes facilitates the insertion process. For inserting equitensiometer, dummy holes were made (whose size is smaller than the probe size) and then the probes inserted. It must be noted that the soil next to the probe has the strongest influence on the readings, and hence sufficient care has been taken to avoid air gap around the probes. The soil has been initially saturated so that ψm ≈ 0. The probes are inserted into the soil sample and the readings recorded over a period of time. The measured ψm and θ values are used for developing SWCC.

RESULTS AND DISCUSSIONS The variation of ψm and θ with time for continuous drying of RS sample is depicted in Fig. 2. It can be noted from the Fig. 2(b) that the initial θ measured by EC-5, EC-TE and SM200 vary significantly. However, the difference in volumetric water contents reduces for θ < 42 %. The suction and volumetric water content values obtained using equitensiometer (EQT), and EC-5, EC-TE and SM200 probes are plotted as depicted in Fig. 3. Each set of ψm and θ data points presented in Fig. 2 correspond to the same logging time. It can be clearly noted from the Fig. 3 that the SWCCs established using EQT and three volumetric water content probes deviate significantly. It is essential to investigate the implication of such differences on unsaturated soil behaviour modelling. The three SWCCs meet a suction value of 30 KPa. The difference between the SWCCs obtained using EC-5 and EC-TE is less beyond 30 KPa of suction. However, the SWCC obtained using SM200 sensor varies significantly beyond 30 KPa of suction. From the study, it is clear that different methodologies adopted for volumetric water content measurement do not yield a unique value. The study recommends further investigating the validity of a single calibration equation of EC-5, EC-TE and SM200 probes for all type of soils. The study also highlights

A Study on the Measuring Methodologies on Soil Moisture Measurements

the need to understand the implication of such differences in SWCC on unsaturated soil behavior modeling.

1 0 0 10 1 10 2 1 03 1 0 4 10 51 0 0

1 0 1

1 0 2

1 0 3

1 0 4

10 0 1 01 1 02 10 3 10 4 1 050 .2 0

0 .2 5

0 .3 0

0 .3 5

0 .4 0

0 .4 5

0 .5 0

0 .5 5

0 .6 0

(a ) (b )

E Q T

ψm

(KPa

)

T i m e (m i nu te s )

θ

S M 2 0 0 E C-5 E C-T E

Fig. 2 Variation of ψm and θ with time

100 101 102 103 1040.20

0.25

0.30

0.35

0.40

0.45

0.50

0.55

0.60

θ

ψm (KPa)

EQT-SM200 EQT-EC-5 EQT-EC-TE

Fig. 3 Drying SWCCs for the soil RS CONCLUSIONS This study deals with an investigation on the influence of soil-moisture measuring methodologies on the soil-water characteristic curve (SWCC) of a locally available soil. In this study, three low cost volumetric water content sensors, such as EC-5, EC-TE and SM200 sensors were used for continuous measurement of volumetric water content in a continuously drying soil sample from saturated state. For continuous measurement of suction, equitensiometer (EQT) was used. The study indicates that the SWCC established by using different moisture measuring methodologies employed

in this study may not yield a unique trend. Such an observation may be mainly attributed to the insufficient calibration equations of the moisture measuring methodologies supplied by the manufacturers. This asserts the need for a precise calibration equation and evaluating the generality of calibration equations for the better performance of volumetric water content measuring methodologies. REFERENCES 1. ASTM (2005), Standard test methods for laboratory

determination of water (moisture) content of soil and rock by mass, D2216-05, Annual Book of ASTM Standards, 04.08, ASTM International, West Conshohocken.

2. Kizito, F., Campbell, C.S., Campbell, G.S., Cobos, D.R., Teare, B.L., Carter, B., and Hopmans, J.W. (2008), Frequency, electrical conductivity and temperature analysis of a low-cost capacitance soil moisture sensor, Jl. of Hydrol., 352, 367-378.

3. Friedman, S.P., Robinson, D.A. (2002), Particle shape characterization using angle of repose measurements for predicting the effective permittivity and electrical conductivity of saturated granular media, Water Resour. Res., 38(11), 1236.

4. Eller, H. Denoth, A. (1996), A capacitive soil moisture sensor, Jl. of Hydrol., 185, 137-146.

5. Hillel, D. (1998), Environmental Soil Physics, Academic Press, Inc., San Diego, California.

6. Gardner, C.M.K., Dean, T.J. and Cooper, J.D.(1998), Soil water content measurement with a high frequency capacitance sensor, Jl. of Agric. Engng Res., 71, 395-403.

7. Thompson, R.B., Gallardo, M., Valdez, M.D. and Fernandez, M.D. (2007), Determination of lower limits for irrigation management using in situ assessments of apparent crop water uptake made with volumetric soil water content sensors, Jl. of Agricultural Water Management, 92, 13-28.

8. Bell, J.P. and McCulloch, J.S.G. (1966), Soil moisture estimation by neutron scattering method in Britain, Jl. of Hydrol., 4, 254-263.

9. Topp, G.C., Davis, J.L. and Annan, A.P. (1980), Electromagnetic determination of soil water content: Measurements in coaxial transmission lines, Water Resour. Res., 16, 574–582.

10. Ledieu, J., Ridder, P.de., Clerck, P.de. and Dautrebande, S. (1986), A method of measuring soil moisture by time-domain reflectometry, Jl. of Hydrol., 88, 319–328.

11. Bell, J.P., Dean, T.J. and Hodnett, M.G. (1987), Soil moisture measurement by an improved capacitance technique, Part II. Field Techniques, evaluation and calibration, Jl. of Hydrol., 93, 79-90.

12. Evett, S.R. and Steiner, J.L. (1995), Precision of neutron scattering and capacitance type soil water content gauges from field calibration, Soil Sci. Soc. Am. Jl., 59, 961–968.

13. ASTM (2006), Standard test method for specific gravity of soil solids by water pycnometer, D854-06, 04.08,

Malaya C. & Sreedeep S.

Annual Book of ASTM Standards, ASTM International, West Conshohocken.

14. ASTM (2008), Standard practice for classification of soils for engineering purposes (Unified Soil Classification System), D2487-06ε1, Annual Book of ASTM Standards, 04.08, ASTM International, West Conshohocken.

15. ASTM (2005), Standard test methods for liquid limit, plastic limit, and plasticity index of soils, D4318-05, 04.08, Annual Book of ASTM Standards, ASTM International, West Conshohocken.

16. Soil Matric Potential Sensor (1999), User Manual EQ2-UM-1.3, Delta-T Devices, U.K.

17. Malaya, C. and Sreedeep, S. (2010), A study on the influence of measuring procedures on suction-water content relationship of a sandy soil, Jl. of Test. and Eval., ASTM, 38(6), Page count: 9.

18. Kodešová, R., Kodeš, V. and Mráz, A. (2011), Comparison of two sensors ECH2OEC-5 and SM200 for measuring soil water content, Soil & Water Res., 6(2), 102-110.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No. A105)

TRIAXIAL BEHAVIOUR OF QUARRIED AND ALLUVIAL ROCKFILL MATERIALS

N.P. Honkanadavar, Scientist ‘C’, CSMRS, New Delhi-110016, [email protected] Sandeep Dhanote, Scientist ‘B’, CSMRS, New Delhi-110016, [email protected] S.L. Gupta, Scientist ‘E’, CSMRS, New Delhi-110016, [email protected] Murari Ratnam, Director, CSRMS, New Delhi-110016, [email protected] ABSTRACT: In the present study, alluvial modeled rockfill material from Renuka dam project, Himachal Pradesh and quarried blasted modeled rockfill material from Salma dam project, Afghanistan were obtained and tested. The maximum particle size (dmax) of the prototype gradation rockfill material for Renuka dam and Salma dam is 1000 and 600 mm respectively. For testing, the size is scaled down to dmax of 80, 50 and 25 mm for both project materials using parallel gradation technique. Consolidated drained triaxial tests were conducted on both the modeled rockfill materials with the confining pressure (σ3) ranging from 0.4 to 1.2 MPa. All the tests were conducted for 87% relative density (RD). Stress-strain-volume change behaviour of both the modeled rockfill material is studied and presented. It is observed that the stress-strain behaviour is non-linear, inelastic and stress dependent for both the materials. From the volume change behaviour, it is observed that the material show dilation effect which decreases with increase in confining pressure (σ3) and dmax. Shear strength parameter, angle of internal friction (φ) is determined for all the dmax of both project materials using Mohr Coulomb failure criterion. Elastic parameters, modulus of elasticity (E) and Poisson’s ratio, ν were determined. Material constants (modulus number, k and modulus exponent, n΄) were also determined. From the study, it is observed that the φ increases with increase in dmax for alluvial and reverse trend is observed for quarried rockfill material. The E increases with increase in dmax and σ3 for alluvial rockfill material but E decreases with increase in dmax and it increases with increase in σ3 for quarried rockfill material. Modulus number, k increases and n΄ decreases with increase in dmax for alluvial and k decreases and n΄ increases with dmax for quarried rockfill materials. From the particle breakage analysis, it is observed that the breakage factor, Bg increases with increase in dmax and σ3 for both materials. However, the effect of σ3 and dmax is more on Bg for quarried rockfill material than alluvial rockfill material.

INTRODUCTION Rockfill materials are widely being used in the construction of rockfill dams because of their inherent flexibility, capacity to absorb large seismic energy and adoptability to various foundation conditions. The behaviour of the rockfill materials is of considerable importance for the analysis and safe design of rockfill dams. Rockfill materials consist of maximum particle size (dmax) up to 1200 mm. Rockfill material with such a large particle size is not feasible to test in the laboratory. Some kind of modelling technique is often used to reduce the size of particles so that the specimens prepared with smaller size particles can be tested. Among all existing modeling techniques, the parallel gradation technique (Lowe 1964) is most commonly used. The behaviour of rockfill materials is affected by particle size, shape, surface structure, mineral composition, individual particle strength, relative density etc. Therefore, it is very much essential to study the behaviour of rockfill material by simulating the exact field conditions so as to design the safe and economical rockfill dam. The behaviour of the alluvial rockfill material has been reported by number of researchers. Marsal. (1967), Marachi et al. (1969), Gupta (2000), Varadarajan et al. (2002a, 2003), Abbas (2003), Abbas et al. (2003), Honkanadavar (2010), CSMRS (2010) and CSMRS (2011) have performed laboratory tests on alluvial and quarried rockfill materials collected from different river valley projects from India and abroad. They concluded that stress-strain behaviour is non-linear, inelastic and stress level dependent

for both the materials. The volume change at failure increases with increase in confining pressure (σ3) and dmax for both the materials. This paper deals with the testing of quarried and alluvial rockfill materials obtained from Salma dam site, Afghanistan and Renuka dam site, Himachal Pradesh, India and study their behaviour under large size triaxial shear test. Tests were conducted on 25, 50 and 80 mm dmax with the varying σ3 ranging from 0.4 to 1.2 MPa. All the tests were conducted with 87% relative density. Consolidated drained triaxial shear tests were carried out for all the modeled dmax of both project materials and studied their stress-strain-volume change behaviour. The shear strength parameter, angle of internal friction (φ) is determined using Mohr Coulomb failure criterion and compared between alluvial and quarried rockfill materials. Using the initial portion of the stress-strain-volume change behaviour, modulus of elasticity, E and Poisson’s ratio, ν are determined using Hyperbolic relationship (Kondner 1963) for both the materials and compared. The material constants viz. modulus number, k and modulus exponent, n΄ are determined using Janbu’s relationship and are compared between both project materials. The effect of dmax and σ3 on particle breakage is also studied for both the materials and compared. EXPERIMENTAL INVESTIGATIONS AND DISCUSSION


Material Used To carry out this research work, quarried rockfill material from Salma dam site, Afghanistan and alluvial rockfill material from Renuka dam site, Himachal Pradesh, India have been considered. The rock type is Dolomite for Renuka and quartzite for Salma dam site. The dmax proposed in the construction of Renuka dam is 1000 mm and for Salma dam is 600 mm. Both material has been modelled into three dmax (25, 50 and 80 mm) using parallel gradation technique as shown in Figs. 1a and 1b for Salma and Renuka dam respectively to test in the triaxial specimen of size 381 mm diameter and 813 mm height.

Experimental Programme Consolidated Drained Triaxial Test Consolidated drained triaxial tests have been conducted on the modeled rockfill materials with σ3 varying from 0.4 to 1.2 MPa for 87% RD at Central Soil and Materials Research Station (CSMRS), New Delhi.

0

20

40

60

80

100

0.01 0.1 1 10 100

Perc

ent F

iner

dmax (mm)

Fig.1a: Prototype and modeled Grain Size Distribution Curves for Salma Dam Rockfill Material

0

20

40

60

80

100

0.01 0.1 1 10 100 1000dmax (mm)

Per

cent

Fin

er

Fig.1b: Prototype and modeled Grain Size Distribution Curves for Renuka Dam Rockfill Material

Stress-Strain-Volume Change Behaviour The stress-strain-volume change behaviour of both the modeled rockfill material for 87% RD has been studied (CSMRS 2010; 2011) and presented. From the stress-strain plots, it is observed that the behaviour is non-linear, inelastic and stress level dependent. The deviatoric stress and axial strain at failure increases with increase in dmax and σ3 for both materials. The volume change behavior shows compression during the initial part of shearing and dilation with further shearing which decreases with increase in dmax and σ3 for both materials. Typical stress-strain-volume change behaviour for 25 mm dmax of Renuka dam rockfill material is shown in Fig.2. Mean stress v/s deviator stress was plotted (Fig.3) and shear strength parameter, φ is determined from the best fit curve for all the dmax of both project materials and presented in the Table 1. From the table, it is observed that the φ−value increases with increase in dmax for alluvial rockfill material and reverse trend is observed for the quarried rockfill material. Determination of Elastic Parameters The value of modulus of elasticity, E is determined from the stress-strain-volume change behaviour using hyperbolic relationship (Kondner 1963) as

(σ1-σ3) = ε1/(a+bε1) (1)

where, (σ1-σ3) is deviatoric stress, ε1 is axial strain, a is material constant, inverse of modulus of elasticity, E and b is material constant, inverse of ultimate strength (σ1-σ3) ult. The value of ε1/(σ1-σ3) are calculated from the initial part of the stress-strain curves and are plotted against ε1. The intercept of the best fit line in the transformed plot is obtained as the value of the constant a. The reciprocal of the constant a gives the modulus of elasticity. The detail procedure is explained by Honkanadavar (2010). From the analysis, it is observed that E increases with increase in dmax and σ3 for Renuka dam alluvial material however, E decreases with increase in dmax and it increases with increase in σ3 for Salma dam quarried rockfill material.

0.0

0.5

1.0

1.52.0

2.5

3.0

3.5

0.0 2.0 4.0 6.0 8.0 10.0Axial Strain (%)

Dev

iato

r Stre

ss (

1--

3 ) M

Pa

(a) Stress-Strain Curves

Prototype 600 mm

Modeled 25 mm

Modeled 50 mm

Modeled 80 mm

Prototype 1000 mm

Modeled 80 mm

Modeled 25 mm

Modeled 50 mm σ3 = 1.2 MPa

σ3 = 0.4 MPa

σ3 = 0.8 MPa


-2.0

-1.8

-1.5

-1.3

-1.0

-0.8

-0.5

-0.3

0.0

0 2 4 6 8 10Axial Strain (%)

Vol

umet

ric S

train

(%)

(b) Axial strain-Volumetric strain Behaviour

Fig.2: Stress-Strain-Volume Change Behaviour for 25 mm dmax of Renuka Dam Rockfill Material

y = 1.5569xR2 = 0.9901

0

1

2

3

4

0 1 2 3

Mean Stress (σm = (σ1+2σ3)/3) MPa

Dev

iato

r Stre

ss ( σ

1-σ 3

) MP

a

Fig.3: Relation between Mean stress (σm) v/s deviator stress (σ1-σ3) space for the dmax of 25, Renuka dam The value of lateral strain is calculated from the initial part of axial strain and volumetric strain curve, and the value of Poisson’s ratio is determined as the ratio of lateral strain to axial strain. It is observed that the Poisson’s ratio decreases with increase in dmax for both the materials. The values of E and ν for different dmax and σ3 are presented in Table 2.

Table 1: Shear Strength Parameter of Rockfill Material

Name of Project

RD (%) dmax (mm) 25 50 80

Salma Dam, Afghanistan

87 43.35 42.39 40.84

Renuka Dam, India

87 37.13 38.37 39.76

Table 2: Values of E and ν of Rockfill Material

Name of Project

RD (%)

σ3 (MPa)

E (MPa) dmax (mm)

25 50 80 Salma Dam, Afghanistan

87 0.4 0.8 1.2

59.88 84.03 106.38

53.48 78.12 95.24

45.66 70.42 84.75

Poisson’s Ratio (ν) 0.32 0.31 0.30 Renuka Dam, India

87 0.4 0.8 1.2

65.238 86.46 129.36

72.37 98.45 136.88

80.965 103.578 145.227

Poisson’s Ratio (ν) 0.32 0.31 0.30 Determination of Material Constants The modulus of elasticity is expressed as a function of the confining pressure (Janbu 1963) as

E = kPa(σ3/Pa)n΄ (2)

where, k and n΄ are the modulus number and modulus exponent of the material respectively, σ3 is the confining pressure and Pa is the atmospheric pressure. Taking log on both sides, Eq. (2) can be written as log (E/Pa) = log k + n΄ log (σ3/Pa) (3)

Plots between E/Pa and σ3/Pa on log scale are made and k and n΄ are determined as antilog of intercept on y-axis and slope of a best fit line respectively. The detail procedure is explained by Honkanadavar (2010). The values of material constants for both the materials are presented in Table 3. Table 3: Values of k and n΄ of Rockfill Material Name of Project

RD (%) dmax (mm) Material Constants k n΄

Salma Dam, Afghanistan

87 25 50 80

264.18 236.50 218.32

0.694 0.713 0.729

Renuka Dam, India

87 25 50 80

374.37 385.21 410.68

0.466 0.469 0.462

It is observed that the value of k increases with increase in dmax for both materials whereas n΄ increases with increase in dmax for quarried and no trend is observed with dmax of alluvial rockfill material. Determination Breakage Factor The breakge of particles under different stress level is quantitatively expressed as breakage factor, Bg as proposed by Marshal (1965). Breakage factor is determined based on pre and post test grain size distribution curves for all the dmax of both project materials. The effect of σ3 and dmax on the breakage factor, Bg is studied by plotting dmax v/s Bg for different σ3 for both project materials (Fig. 4). From the analysis, it is observed that the breakage factor increases with increase in dmax and σ3 for both project materials. However, it is observed that the effect of dmax and σ3 on Bg is more on quarried rockfill material as compared to alluvial

σ3 = 1.2 MPa

σ3 = 0.8 MPa

σ3 = 0.4 MPa

6sinφ/3-sinφ = 1.5113 φ = 37.13o


rockfill material. This may be due to high interlocking present in the quarried rockfill material.

0

2

4

6

8

10

12

0 25 50 75 100

dmax (mm)

Brea

kage

Fac

tor,

B g (%

)

Fig..4: Variation of Breakage Factor with dmax and Confining Pressure. CONCLUSIONS The quarried and alluvial rockfill material was obtained from Renuka dam, Himachal Pradesh and Salma dam, Afghanistan respectively. Both the project material was modeled into dmax of 25, 50 and 80 mm and tested in the laboratory. From the stress-strain-volume change behaviour, it is observed that the behaviour of both materials is non-linear, inelastic and stress dependent. The stress at failure increases with increase σ3 for both material. Both material show dilation effect which decreases with increase in dmax and σ3. The shear strength parameter, φ increases with increase in dmax for alluvial rockfill material where as reverse trend is observed for quarried rockfill material. The value of E increases with increase in dmax and σ3 for alluvial rockfill material. E decreases with increase in dmax and it increases with increase in σ3 for quarried rockfill materials. Modulus number, k increases and n΄ decreases with increase in dmax for alluvial and k decreases and n΄ increases with dmax for quarried rockfill materials. From the particle breakage analysis, it is observed that the breakage factor, Bg increases with increase in dmax and σ3 for both materials. However, the effect of σ3 and dmax is more on Bg for quarried rockfill material than alluvial rockfill material. From the detail triaxial study, it is observed that both alluvial and quarried rockfill material behave differently. Therefore, it is very much essential to carry out detail triaxial laboratory investigations so as to design a safe and economical rockfill dam.

ACKNOWLEDGEMENT

The authors wish to thank all the staff of rockfill division, CSMRS for their help in conducting the tests. REFERENCES 1. Abbas, S.M. (2003), “Testing and Modeling the

Behaviour of Riverbed and Quarried Rockfill Materials”, Ph.D. Thesis, I.I.T. Delhi.

2. Abbas, S.M., Varadarajan, A. and Sharma, K.G. (2003),” Prediction of Shear Strength Parameter of

Prototype Rockfill Material”, IGC-2003, Vol-I, pp. 5-8, Roorkee.

3. CSMRS. (2010), “Large size triaxial shear test report on quarried rockfill material from Salma Dam Project, Afghanistan.

4. CSMRS. (2011), “Large size triaxial shear test report on alluvial rockfill material from Renuka Dam Project, Himachal Pradesh, India.

5. Gupta, A.K. (2000), “Constitutive Modeling of Rockfill Material”, Ph.D. Thesis, I.I.T. Delhi.

6. Honkanadavar, N.P. (2010), “Testing and Modelling the Behaviour of Modelled and Prototype Rockfill Materials”, Ph.D. Thesis, I.I.T.Delhi.

7. Kondner, R.L. (1963), “Hyperbolic Stress-Strain Response, Cohesive Soils”, J. of SMFE, ASCE, 89, SMI, pp. 115-143.

8. Lowe, J. (1964), “Shear Strength of Coarse Embankment Dam Materials”, Proc. 8th Int. Congress on Large Dams, Vol. 3, pp. 745-761.

9. Marachi, N.D., Chan, C.K., Seed, H.B. and Duncan, J.M. (1969), “Strength and Deformation Characteristics of Rockfill Materials”. Report No. TE. 69(5), Civil Engineering Department, University of California, Berkeley, USA.

10. Marsal, R.J. (1967), “Large Scale Testing of Rockfill Materials”, J. of Soil Mech. And Foundations Division, ASCE, 93(2), pp. 27-43.

11. Varadarajan, A., Sharma, K.G., Venkatachalam, K. and Abbas, S.M. (2002a), “Constitutive Modeling of Rockfill Materials from Tehri Dam, Uttaranchal”, Proc. IGC2002, Allahabad, India, Vol. 1, pp: 592-595.

12. Varadarajan, A., Sharma, K.G., Venkatachalam, K. and Gupta, A.K. (2003), “Testing and Modeling Two Rockfill Materials”, J. Geotech. And Geoenv. Eng., ASCE, 129 (3), pp. 206-218.

Alluvial for σ3=0.4 MPa Quarried for σ3=0.8 MPa For σ3=1.2 MPa

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No.A 106)

ROLE OF GROUND PENETRATING RADAR FOR SITE INVESTIGATIONS AND SITE CHARACTERIZATION – CASE STUDIES Rajesh Khanna, Scientist ‘B’, CSMRS, New Delhi, [email protected] S. L. Gupta, Scientist ‘E’, CSMRS, New Delhi, slgupta.nic.in

ABSTRACT: Ground Penetrating Radar (GPR) survey is the main geophysical non- invasive techniques for site investigations to detect geological, engineering and hydraulic features for river valley projects. GPR can also be used in congested urban areas for foundation investigations. GPR survey is quick and large area can be surveyed in short time. GPR survey was conducted at Aseana quarry site near Tehri Hydro Electric Project, Uttarakhand to find out the depth of over burden, to find the depth of bed rock for construction of anchor blocks for penstock at Ganwi H. E Project (H.P.), to locate the leakage point in the body of Matatila dam (Jhansi) around 35 cm dia. cast iron pipe (intake for water supply) and also at major intersections in thickly populated areas in Delhi to detect the underground utilities like water supply line, sewer line and live electrical cables present at the construction site to have advance knowledge so that it may not present hindrance during excavation and construction. GPR survey requires lot of expertise for analysing the data. The primary objective of this paper is to provide different applications of GPR for site investigations and site characterization including procedure and precautions to be taken into account for GPR survey. INTRODUCTION Investigation of sites for geotechnical purposes is usually carried out by drilling together with sampling and determination of rock properties. These results generally give a satisfactory amount of information and are important for planning and design purpose. The problem with this type of information is that it refers to a particular place, but geological conditions generally change to a considerable extent between two nearby bore holes. Among the various geophysical methods for subsurface investigations, GPR method is one of the convenient methods for solving geotechnical problems and large area can be investigated in minimum time. GPR survey gives continuous profile of the substrata and sometimes able to locate the seepage path to some extent. GROUND PENETRATING RADAR (GPR) GPR uses the principal of the reflection of electromagnetic waves to locate buried objects. The basic principal and theory of GPR are same as those used to detect the aircraft overhead but GPR uses a much broader band width and with transmitting and receiving antennas that are pointing downward towards the ground. Transmitting antenna transmits electromagnetic waves (25 MHz to 1GHz) into the ground. The wave spreads out and travels downwards until it hits an object that has different electrical properties from the surrounding ground. If the wave hits a buried object then part of the wave energy is reflected to the surface, while part of the energy continues to travels downwards. The waves that are reflected back to the surface is captured and recorded on a digital storage device for interpretation. In practice, GPR measurements are made by towing the antennas continuously over the ground. The antennas can be towed manually, or with a vehicle. A radar wave is

transmitted and received each time that the antenna has been moved a fixed distance across the ground surface. The information that is recorded while the receiver is turned on is called trace. A trace contains the reflections that have bounced back from the buried objects. It takes a specific amount of time for electromagnetic waves to make the round trip from the surface down to the reflector, and back to the surface. Travel time for the electromagnetic waves is measured in nanoseconds. Two ways travel time is greater for deep objects than for shallow objects. If the velocity of the wave in the subsurface is known, the time of arrival for the reflected wave recorded on each trace can be used to determine the depth of buried objects. It can locate any object that has electrical properties in contrast with the surrounding ground and is within the detection range of the radar waves. INSTUMENTATION GPR system used in this investigation is SIR – 10 system from GSSI, USA. It consists of (1) Control unit, (2) Display unit, (3) Antennae of centre frequency of 80, 300, 500 and 900 MHz. The antennas are monostatic (both transmitter and receiver in the same unit). The data can be displaced on colour monitor and can be recorded on 8 mm digital tape. After recording the data digitally it can be seen in the main frame system and can be transferred to a computer for further review and processing. ESTIMATION OF OVERBURDEN AT ASENA QUARRY OF TEHRI DAM PROJECT Tehri Dam is a multipurpose river valley project proposed to tap the vast potential of the river Bhagirathi, a tributary to river Ganges. The project envisages the construction of 260.5 m high earth and rockfill dam. The dam creates a reservoir area of 42 sq. km. and with gross and live storage capacity of 3540 million cubic metres respectively.

GPR survey was carried out in additional area of Asena quarry of Tehri Dam for the determination of thickness of overburden. The rip – rap material is to be collected from the quarry site to be placed at the dam. The approximate area of the ground with overburden is 22000 Sq. m (200 m × 110 m) between RL 950 and 1100 The rock formation at the querry site are thinly quartize and massive quartize with varying overburden and matrix material FIELD PROCEDURE The profile lines at different elevation of the quarry area were marked in consultation with the project authorities and GSI. The profile line lengths were restricted as per the site conditions. The antenna was pulled manually at a constant speed on profile lines with markings on predetermined distances of 5 m. GPR survey was carried out using 80 MHz antenna to investigate the geological variation up to a depth of 30 m. The location of these profile lines is presented in Fig.1 Five number of profile lines covering a total length of 130 m were surveyed. The signals were recorded on 8 mm exbyte tape. GPR records for the above mentioned profiles were interpreted using RADAN/RADPRINT software.

Fig. 1 Layout of GPR profile lines at Asena quarry of Tehri H. E. Project INTERPRETATION OF DATA Total of five profile lines numbered F 5, F 3, F 15, F 19 and F 20 of length 15 m, 15 m, 20 m, 15 m and 40 m respectively in the direction shown in the arrow were surveyed. From the interpretation of the data, as shown in Fig. 2, it is inferred that in case of profile F 15, the overburden consisted of boulders mixes with silty clay soil (matrix material) for a profile length of 0 to 3 m and is fractured rock for the remaining length up to a depth of 18 m. Similarly as shown in Fig. 3, in case of profile line numbered F 20, the overburden consists of rock fragments, the big boulders mixes with soil from 7 to 32 m along the profile length and is soil mixed with gravels for the remaining length up to a depth of 18 m. For a profile line F 5 as shown in fig. 4, it is inferred that from 0 to 5 m along the profile length the overburden is 12 m and from 5 m to 15 m the overburden is 18 m.

Fig. 2 GPR profile for F 15 at RL 1058 of Asena querry of Tehri H. E. Project

Fig. 3 GPR profile for F 20 at RL 1058 of Asena querry of Tehri H. E. Project

Fig. 4 GPR profile for F 5 at RL 1058 of Asena querry of Tehri H. E. Project LOCATION OF DEPTH OF BED ROCK AT GANWI H. E. PROJECT (H.P.) Ganwi H. E. Project is a runoff river scheme on river Ganwi, a tributary of river Sutlej, envisages generation of 22.5 MW

of power utilizing a discharge of 7.25 cumecs and a head of 385 m. The total length of penstock is 600 m with diameter 1.40 m which bifurcates in two branched of 1.20 m diameter each. The purpose of investigation was to ascertain the depth of overburden at anchor block locations of penstock for designing the safe foundation on competent rock and for this investigation GPR survey and seismic survey was carried out. FIELD PROCEDURE There were eight anchor blocks AB – 1 to AB – 8 along the penstock alignment. Due to steep slope, non-availability of approach, working space, due to the presence of loose overburden and difficult site conditions field work using GPR was carried out at anchor block location AB – 5 and AB – 7 using 300 MHz antenna. Longitudinal section along the penstock is shown in Fig. 5

Fig. 5 Longitudinal section along the Penstock alignment INTERPRETATION OF DATA At anchor block AB- 5, at an elevation of 1373.0 m and RD 258.8, at this location rock was exposed as 6.0 m deep bench was excavated. Two profile lines were run across the bench, strong reflection were seen at a depth of 5 m depth at one end of the anchor block and slopes down to7.7 m. These were observed as of bed rock. At AB -7, overburden comprised of silty clay mixed with boulders and vegetative cover and seems to be highly organic. GPR survey was carried out but nothing could be seen as the electromagnetic waves were absorbed by the conductive clay. GPR record of presence of bed rock is presented in Fig. 6. GPR INVESTIGATION AT MATATILA DAM PROJECT (U.P.) Matatila dam, built across the river Betwa is a multipurpose project. The height of the dam from the deepest point is 45.72 m and from the river bed is 33.53 m. The objective was to locate the leakage point in the body of the dam around 35 cm dia. Cast iron pipe(intake for water supply).

Fig. 6 GPR record showing bed rock profile at anchor block location AB -5 FIELD PROCEDURE OF DATA GPR survey was carried out using 80 MHz antenna for depth of 20 m. The antenna was moved manually at predetermined distances. Total of 42 profile lines were surveyed across and along the axis of the pipe from top of the dam. INTERPRETATION From the interpretation of the data with the help of RADPRINT software, it is inferred that a pipe is passing through the body of the dam as shown in the GPR print out in Fig. 8. However no anomaly was detected about the leakage through the body of the dam as well through the pipe. Strong reflections could also be seen nearby the pipe from the top to the bottom through the depth as shown in Fig.7 and Fig. 8

Fig. 7 GPR record showing location of intake pipe in the body of Matatila dam

Fig. 8 GPR record showing location of intake pipe and presence of reinforcement in the body of Matatila dam CONCLUSIONS These non- invasive techniques can be used for site investigation of major river valley projects and also in urban areas for site investigations. These non- invasive techniques are quick and large area can be surveyed in small time. As these are the indirect methods of investigations, the results of these investigations should be verified by direct or more than one indirect methods. The results of the GPR survey were matched with the log of drill holes along the GPR profile and found to be in agreement. ACKNOWLEDGEMENT Author is thankful to Director CSMRS for allowing to present the data of the project reports for this conference. REFERENCES 1. Paul, R. (1994), “SIR System – 10 (A) Training Notes”,

Geophysical Survey System Inc. North Salem, NH USA. 2. Woods, R.D. (1994), “Geophysical Characterisation of

Sites”, XII International Conference on Soil Mechanics and Foundation Engineering, New Delhi, India

3. H. Westerdahl., Kong, F.N. (1999), “Merits and Potential OF Geo Radar as a Subsurface Investigation Tool – Experience at NGI during the Past Ten Years”, Proceeding of the 5th International Symposium on Field Measurement in Geomechanics – FMGM 99, Singapore.

4. Danial., D.J. (1998), “Subsurface Penetrating Radar”, Institution of Electrical Engineers, London.

5. CSMRS., (2002), “ Report on Ground Penetrating Radar Survey at Ganwhi H. E. Project, jeori, H.P.” New Delhi, India.

6. CSMRS., (2003), “ Report on Ground Penetrating Radar Survey at Matatila Dam Project” Jhansi (M.P.).

7. CSMRS., (2004), “ Report on Ground Penetrating Radar Survey of Asena Quarry of Tehri H. E. Project, Tehri, Uttrakhand” New Delhi, India.


EXPERIMENTAL STUDY ON EFFECT OF INITIAL WATER CONTENT ON VOLUME CHANGE OF EXPANSIVE SOIL DURING SWELLING

S. P. Lajurkar, Asst. Professor, Priyadarshini Indira Gandhi College of Engineering, Nagpur, [email protected]. S. R. Khandeshwar, Asst. Professor, Yeshwantrao Chavan College of Engineering, Nagpur. M. S. Bhagat, Asst. Professor, Yeshwantrao Chavan College of Engineering, Nagpur. ABSTRACT: Expansive soils exhibits high volumetric changes due to seasonal water content change. This behavior of soil leads to many problems in construction activities and causes severe damages to structures fonded on such soil deposits. Prediction of vertical soil movements and ground heave in expansive soil deposit is an important requirement for taking appropriate measures in the design and construction of the structures at the site of expansive soil deposit. Considering this aspect “Limiting Unit Swell Potential” parameter is introduced for characterizing realistically the degree of swellability that can be exhibited by any soil mass pertaining to its formation and environmental change condition based on the water content change with respect to extreme physical states of the soil mass. While developing this parameter it is presumed that the rate of change in volume of any soil mass is same when it is allowed to swell from any initial water content to fully swollen saturated condition. In the present study, this presumption has been checked. INTRODUCTION Soil, in general are broadly categorized in two main groups viz. coarse grained soil and fine grained soil. The fine grained soils which comprise silt size and clay size particles constitute a very large proportion of natural ground deposit. Also the clay soils which are cohesive in nature occur over the large portion of the earth surface. Because of the presence of electrochemical activity in such clays, they exhibit distinctly different characteristics depending on the clay mineral structure of the particles. The clays in which the particles are of Montmorillonite mineral structure exhibit very undesirable engineering behaviour. Such clays in which there is predominance of Montmorillonite clay mineral particles are called as expansive or swelling clays [1]. This type of clay exhibit very complex and undesirable characteristics when used as engineering material. This soil is truly assumed the notoriety of being the most difficult and problematic soil for any civil engineering constructions. The technology associated with such deposit viz, the expansive soil technology has remained for long time an important domain of continued research and development. The studies undertaken under the present project falls in a category of continuation of expansive soil technology. Identification a Specific Aspect of Swelling Soil Behaviour Characterization of swelling soil is usually done by the following property parameter,

• Free swell Index • Swelling potential • Swelling pressure

In order to access the degree of expansivity of the given expansive soil deposit the swelling potential parameter is of great relevance. Unfortunately this parameter is not uniquely and precisely defined. However in the present literature the definition of swelling potential is the axial strain of a laterally

confined soil with its initial condition corresponding to Procter M.D.D. and O.M.C. produced by the soil on swelling under a very nominal surcharge pressure of 1 psi. It may be realized that the swelling potential of clay soil should act as a measure of the ability of swelling and degree to which such a soil might swell if its environment were changed in a definite way.

The offer said swelling potential definition does not characterize the swellability of expansive soil under its various formation and environmental condition. In order to rationalize this parameter, Golait and Khanzode [2], Golait and Wakhare [3] put forth a new concept of swelling potential of soil and proposed a parameter “Limiting Unit Swell Potential” for characterization realistically the degree of swellability that can be exhibited by any soil mass pertaining to its formation and environmental change condition. Important missing aspects in this present study try to accomplish its near perfection. Aims and Objectives of the Study The “Limiting Unit Swell potential” proposed by Golait and Khanzode [2], Golait and Wakhare [3] has been calculated based on the water content change with respect to extreme physical states of the soil mass. These states are fully shrunken dry soil mass with water content zero and the unrestrained fully swollen saturated condition with water content Wsat.

Based on this water content difference i.e. (Wf-0) the “Unit Swell Potential” is evaluated. As the “Limiting Unit Swell Potential” basically is the volumetric strain (volume increase) exhibited by the soil due to unit percentage change (increase) in water content. It is of interest to know whether this parameter for a given soil remains same when the soil swells from any initial water content state. The main aim of the present study is to thoroughly investigate this aspect.

Mrs. S. P. Lajurkar & Ms. M. S. Bhagat

LABORATORY INVESTIGATIONS As highlighted earlier, the primary purpose of the present work was to evaluate the swelling potential of given natural soil of different volume change characteristics at different initial water content. To achieve the above mentioned objective, a laboratory investigation programme was planned and the results have been analyzed. Planning of Testing Programme The testing programme was planned to reveal the following aspects

• How and at what rate the soil swell from different initial water content under unrestrained condition (i.e. under no surcharge pressure condition) till it becomes fully swollen mass with full saturation with water.

• How does the soil with different maximum swelling potential swell i.e. what is the rate of swelling with respect to change in water content for different clays.

With the above mentioned views, five different soils having swellability reducing from high to low were selected. Also as the swell-shrink property is exhibited by only fine fractions in soil material, only fraction of soil passing through 75µ sieve was taken for investigations. Six Gang Consolidometer was used for the investigations.

Tests Conducted The following test series were conducted on various soil samples: Consistency Property Tests In this, various conventional consistency properties like liquid limit, plastic limit and shrinkage limit were determined. Free Swell Index Tests Series In this, the free swell index values were obtained by various available methods and approaches viz.

1. Free Swell Index as per A. Sridharan et al. method [4] 2. Free Swell Index as per Golait and Kishore method [5] 3. Free Swell Ratio by S. Prakash et al. method [6]

Volume Change (Swelling) Test Series Under this series, each sample is tested at different initial water content at zero surcharge pressure intensities to evaluate the volume change exhibited by the soil. The main body of the work for the intended purpose of the study pertains to the third test series mentioned above and based on the results of this test series, the analysis is done for investigating the swell characteristics of soils of different degree of swellability at different initial water content.

Material Used in Laboratory Investigations Black Cotton Soil (BCS) The soil used for the investigations was collected from a black cotton soil deposit in Nagpur area. Bentonite Commercial bentonite available in the local market was used for mixing with native black cotton soil to enhance swelling characteristics. Gray Clay Gray clay used in preparation of soil samples of lower expansivity.

Soil Mixes for Laboratory Investigations In order to study the effect of initial water content on the overall swelling behaviour of field soil deposit, the soil mixes of different swellability characteristics, consisted of soil fraction with soil particles less than 75 micron size were prepared. In actual cases of natural soil the finer active fraction (F) may also possesses the swellability characteristics of different magnitude depends on the type and amount of clay minerals, fineness of clay particle etc. In order to incorporate these varying swellability characteristics of finer fraction the artificial prepared material was used. It consists of mixture of <75µ local black cotton soil, commercial bentonite and gray clay. Considering the varying degree of expansivity of finer fraction (F) various mixes were prepared by adding different percentage of bentonite and gray clay as shown in Table 1. In this way finer fraction (F) material with 30% bentonite and 70% of 75µ black cotton soil fraction is considered as fine active fraction with largest swellability, where as finer fraction (F) containing 100%, 75 µ gray clay is considered have least swellable material.

Table 1 Composition of sample mixes

Sr. No. Designation Soil Mixes 1 A BCS+30% BENTONITE 2 B BCS+15% BENTONITE 3 C 100% BCS 4 D BCS+30%GRAY CLAY 5 E 100% GRAY CLAY

Sample Preparation The soil sample black cotton soil and gray clay was collected from a site. Then it was kept for air drying for 24 hours. After 24 hours it was soaked with water to form slurry and the slurry was sieved through 75 µ. Then the slurry passing through 75 µ was kept steady for 1 to 2 days in the tub. The water from the tub was removed by siphoning. The wet soil was oven dried at 105°C. Then it was pulverized with help of mallet. The pulverized soil was then sieved through 1 mm.

Experimental Study on Effect of Initial Water Content on Volume Change during Swelling

sieve. The soil sample thus obtained was stored for further use.

Commercial bentonite available in local market was used for preparing finer fraction (F). Individual samples were prepared by considering weights of different material as indicated in Table 1. On all the above mentioned five samples of soils swelling volume change behaviour test were performed.

ANALYSIS AND INTERPRETATION OF RESULTS The results are analyzed and interpreted with respect to the following:

1. Index property parameters of the soil investigated 2. Swelling characteristics of the soils Index property Parameters The test results for five different soils designated as A, B, C, D, and E are presented in Table 2. Table 2 Index property parameters

MIX A B C D E

Liquid Limit (wL) 73.50 67.50 57.50 47.80 42.13

Plastic Limit (wP) 36.99 32.56 29.08 27.02 23.18

Plasticity Index(IP) 36.51 34.94 28.42 20.78 18.95

Shrinkage Limit, (ws) 12.90 13.75 15.82 19.98 20.79

Shrinkage Index, (Is) 60.60 50.88 41.68 27.82 21.34

FSI(AS) (cc/gm) 1985 2.89 2.12 1.67 1.49 1.41

FSR (Prakash et. al) 2009 1.955 1.55 1.318 1.11 0.98

FSI(GK) (%) 1990 278.99 251.12 232.12 224.36 207.90

Specific Gravity 2.620 2.650 2.680 2.660 2.690 It is clearly seen from above values of various property parameter that the soil A is of highest swellability among these soils and soil E is of lowest swellability. The continuous decrease in swellability is exhibited in the following order A B C D E. As per this sequence of soil type the values of WL, Ip, FSI and FSR, show gradual decrease. Parameters Considered in Characterizing Volume Change The main environmental cause of volume changes exhibited by fine grained soil (with appreciable percentage of Montmorillonite type of clay particles) is the change in water content. On increasing the water content, fine grained soil swells and the decrease in water content causes shrinkage. The volume of soil at any stage of water content can better be represented by specific volume of soil mass having unit mass or weight of soil particle (1gm) at any stage of water content

and thus the specific volume has unit like cubic cm per gram (cm3/gm). This is the most convenient parameter for characterizing the volume change nature for different water content states.

The other parameter specifying more realistically the volume change characteristics is the ‘Limiting Unit Swell Potential’. This concept of ‘Limiting Unit Swell Potential’ was proposed by Golait and Khanzode [2], Golait and Wakhare [3]. It essentially signifies the volumetric strain exhibited by soil due to unit percentage increase in water content. Thus these two parameter specific volume (v) and ‘Limiting Unit Swell Potential’ (Psu)0 are extensively used in analyzing and interpreting the test results from the investigations.

From the investigation the following values for all soils are calculated 1. Initial water content (wi) 2. Initial dry unit weight ( γdi) 3. Water content on full swelling (wf) 4. Dry unit weight on full swelling (γdf) 5. Initial specific volume (vi) 6. Specific volume on full swelling (vs) 7. Specific volume on full shrinkage (vd) 8. Water content change on full swelling (wf-wi) 9. Volumetric strain on full swelling (ΔHf/H)

Specific Volume on Swelling and Shrinking Considering the vd and vf values, it is seen for all the samples that they are almost the same irrespective of initial water content from which the soil swells on wetting or shrinks fully on drying. Hence, the average values of the specific volume for five soils are given in Table 3. Table 3 Average Values of the specific volume

Soil vd (cm3/gm)

vf (cm3/gm)

Specific volume change from shrunken to

swollen state (cm3/gm) A 0.5015 0.770 0.269

B 0.5090 0.725 0.216

C 0.516 0.704 0.188

D 0.522 0.694 0.172

E 0.5490 0.688 0.138 It shows that the specific volume of fully shrunken dry soil mass (vd) increases as the swellability of soil decreases. However, specific volume of fully swollen soil (vf) decreases as the swellability of soil decreases. In other words the highly swelling soil will have small dry specific volume and large swollen specific volume. The magnitude of volume change that the soil can exhibit from fully shrunken dry state to fully swollen condition can be characterized by specific volume change from vd to vf. These values of (vf - vd) for soil A, B, C, D and E are 0.269, .216, 0.188, 0.172 and 0.138 respectively as shown in Table 3. This indicates that soils arranged in descending

Mrs. S. P. Lajurkar & Ms. M. S. Bhagat

order of swellability are A>B>C>D>E. This is agreement with Index property parameters as given in Table1. Volume Change Characteristics on Swelling.

Limiting Unit Swell Potential’ (Psu)0 Tests were conducted on five soil samples of different degree of swellability at various initial water contents and they were allowed to swell fully for further water content change. The ‘Limiting Unit Swell Potential’ values (Psu)0 were worked out from the equation as under.

These values for all the soils with respect to their initial water content have been shown in Table 4.

It may be noted that for a given soil, irrespective of swelling freely from different initial water content states, the (Psu)0 value is almost the same. It shows that swelling takes place at the same rate with change in water content irrespective of any initial water content state. The values of (Psu)0 are 1.727, 1.051, 0.962, 0.918 and 0.883 for soil A, B, C, D and E respectively. By its definition (Psu)0 characterize the swelling potential and the degree of swellability of soil on increasing water content. Higher the (Psu)0 higher is the degree of swellability. These values confirm the following descending order of swellabilty of soils A>B>C>D>E.

Table 4 Values of Limiting Unit Swell Potential obtained for five different soil samples at different initial water content

Soil wi (Psu)0 Avg (Psu)0

A

35.870 1.766 30.476 1.740 1.727 19.380 1.728 12.771 1.672

B

23.793 1.024 21.326 1.003 1.015 16.825 1.157 11.842 1.020

C

26.250 0.965 24.121 *3.844 0.962 18.902 *4.005 10.703 0.958

D

30.172 0.919 22.059 0.918 0.918 13.043 0.921 12.724 0.913

E

24.286 0.882 20.00 0.885 0.883 17.47 0.883

15.217 0.88 * indicate inconsistent values therefore neglected

Correlation between ‘Limiting Unit Swell Potential’ (Psu)0 and Specific Volume Change (vf -vd) It is seen from the Fig. 1 that though as expected the specific volume change increase with increasing ‘Limiting Unit Swell

Potential’, the relationship between them is not in direct proportion i.e. no straight line relationship exists between these two parameters signifying the volume change behavior of soils.

CONCLUSIONS The laboratory investigation program conducted under the present study involves five soil samples of different degree of swellability. All these samples were tested for studying volume change behaviour during swelling from different initial water content and a noteworthy conclusion arrived from the result obtained from this investigation;

1. The values of ‘Limiting Unit Swell Potential’ for the

given soil sample at any initial water content are more or less similar i.e. whatever be the initial water content the rate of change in volume (increase) due to the unit percentage change (increase) in water content are similar i.e. the presumption which has been made while developing ‘Limiting Unit Swell Potential’ (Psu)0 is proved to be correct.

REFERENCES 1. Katti, R. K. (1979), Search for solutions to problems in

black cotton soils. First Indian Geotechn. Soc. Annu. Lect., Indian Geotech. Journal, 9 (1), 1–88.

2. Golait, Y. S. and Khanzode, R.M. (1995), Prediction of ground Heave of swelling clay deposit under natural environmental conditions’, Procededing. International Symp. on Compression and Consolidation of Clayey Soils, HIROSHIMA, Japan, Vol. 1, 769-775.

3. Golait, Y.S. and Wakhare, A.S. (1998), Unit swell potential concept for expansive soils and its simple evaluation’, Eighth Australia New Zealand Conference on Geomechanics Hobart, Vol. 1, no. 2, 171-178

4. Sridharan, A., Rao, S.M. and Murthy, N.S. (1988), Liquid limit of kaolinitic soils, Geotechnical Journal, Vol. 2, 191-198.

5. Prakash, K., Sridharan, A., Prasnna, H. S. and Manjunatha, K. (2009), ‘Identification of soil clay minerology by free swell ratio method’, Indian Geotechnical Conference, Vol.1, 27-30.

6. Golait, Y.S., Kishore, M.P. (1990), A new approach for free swell index and evaluation of swelling potential for construction sites’. Proceedings of Indian Geotechnical Conference, IGC-90, Bombay, 485-489.

ΔHf (Psu)0 = ------------- x 100% H (wf-wi) Fig. 1 Relation between ‘Limiting Unit Swell Potential’

(Psu)0 and Specific Volume Change (vf -vd)


TESTING AND MODELING THE BEHAVIOUR OF ALLUVIAL ROCKFILL MATERIAL

N.P. Honkanadavar, Scientist ‘C’, CSMRS, New Delhi-110016, E-mail: [email protected] K. G. Sharma, Professor, Indian Institute of Technology Delhi, New Delhi-110016, E-mail: [email protected] ABSTRACT: Rockfill materials are being used widely in the construction of rockfill dam to store the river water. These materials are used because of their inherent flexibility, capacity to absorb large seismic energy and adoptability to various foundation conditions. Rockfill material, obtained from Noa Dehing dam site, Arunachal Pradesh is tested. The maximum particle size of the material used in the dam is 600 mm. For testing, the size is scaled down to smaller sizes of 4.75, 10, 19, 25, 50 and 80 mm maximum particle size (dmax) using parallel gradation technique. Drained triaxial tests are carried out with a specimen size of 381 mm diameter and 813 mm height with varying confining pressure (σ3) from 0.2 to 0.8 MPa. All the dmax are tested for 87% and 75% relative density (RD). The index properties of the rockfill materials viz. unconfined compressive strength (UCS) and uncompacted void content (UVC) are determined. Stress-strain-volume change behaviour of the modelled rockfill material is studied and presented. Using the laboratory test results, Material parameters viz. elastic, ultimate, phase change, hardening and non-associative parameters are determined. Procedures were developed to predict the above parameters using the basic index properties of the rockfill material. Using the predicted material parameters, Stress-strain-volume change behaviour for all the dmax is back predicted using hierarchical single surface (HISS) model and compared with the observed stress-strain-volume change behaviour. From the comparison, it is observed that the predicted stress-strain-volume change behaviour of modeled alluvial rockfill material match closely with the observed results for all the dmax tested with 87% and 75% relative density. Using the developed procedures, material parameters for the prototype rockfill material were determined and predicted the stress-strain-volume change behaviour. It is observed that the predicted behaviour of prototype rockfill material follows similar trend as that of modeled rockfill material. Therefore, HISS model appears to be suited to characterize the behaviour of alluvial rockfill materials.

INTRODUCTION Rockfill materials are extensively being used all over the world in the construction of rockfill dams for harnessing the water resources. The behaviour of the rockfill materials is of considerable importance for the analysis and safe design of rockfill dams. Rockfill materials consist of maximum particle size (dmax) up to 1200 mm. Rockfill material with such a large particle size is not feasible to test in the laboratory. Some kind of modelling technique is often used to reduce the size of particles so that the specimens prepared with smaller size particles can be tested. Among all existing modeling techniques, the parallel gradation technique (Lowe 1964) is most commonly used. The behaviour of the alluvial rockfill material has been reported by number of researchers. Marsal (1967), Marachi et al. (1969), Gupta (2000), Abbas (2003), Abbas et al. (2003) and Honkanadavar (2010) have performed laboratory tests on alluvial rockfill materials collected from different river valley projects from India and abroad. They concluded that stress-strain behaviour is non-linear, inelastic and stress level dependent. The volume change at failure increases with increase in confining pressure (σ3) and dmax for the alluvial rockfill materials. Stress-strain-volume change behaviour of alluvial rockfill material has been characterized by many researchers using HISS models (Varadarajan et al. 2002a, 2003; Gupta 2000; Abbas 2003; Honkanadavar 2010). From the laboratory test

results, they determined the material parameters and back predicted the stress-strain-volume change behaviour using HISS model and compared with the observed behaviour. From the predicted and observed results they found that both observed and predicted results match closely. This paper deals with the testing of the rockfill materials obtained from Noa Dehing dam site, Arunachal Pradesh and study its stress-strain-volume change behaviour tested with 87% and 75% relative density. Tests were also conducted to determine the index properties viz. UCS, UVC and RD. Procedures were developed to predict the material parameters using the basic index properties of the rockfill material where UCS represents the strength of the rock from which rockfill materials are derived and it is independent of dmax. UVC includes the effect of gradation, shape, size and surface texture of the rockfill materials and it is dependent on dmax. RD represents the relative compactness of the rockfill materials. Procedures were developed to predict the material parameters. Using the developed procedures, material parameters were predicted and stress-strain-volume change behaviour for all the dmax is back predicted using the HISS model and compared with the observed stress-strain-volume change behaviour of modeled rockfill material tested with 87% and 75% RD. Developed procedures were also used to predict the material parameters for the prototype rockfill material and back predicted its stress-strain-volume change behaviour.


EXPERIMENTAL INVESTIGATIONS AND DISCUSSION Material Used To carry out this research work, rockfill material from Noa Dehing dam site, Arunachal Pradesh has been considered. The rock type is Dolomite, fine grained textured inequigranular rock and blue to dark gray in colour. The dmax proposed in the construction of the Noa Dehing dam is 600 mm. The material has been modelled to six dmax (4.75, 10, 19, 25, 50 and 80 mm) using parallel gradation technique as shown in Fig. 1 to test in the large size triaxial specimen of size 381 mm diameter and 813 mm height. Experimental Programme Determination of Index Properties It is known that the behaviour of granular materials is dependent on RD, σ3, individual particle strength, dmax, shape, surface texture and mineralogy. The individual rockfill particle strength can be represented by UCS of the rock from which rockfill material is derived. Three cylindrical NX (54 mm diameter) size rock core specimens were tested as per IS: 1943-1979 and average value of UCS is obtained as 85.32 MPa (Honkanadavar 2010). Shape, size, surface texture and gradation of the material are represented by a basic characteristic known as UVC for coarse material (ASTM C1252-98, Alhrich 1996). The apparatus has been fabricated to determine UVC of rockfill material (Honkanadavar 2010). The UVC apparatus is designed to test the modelled rockfill materials of dmax = 4.75, 10 and 19 mm. To determine the UVC for dmax of 25, 50, 80 and prototype (600 mm) rockfill material, following procedure has been adopted.

Fig. 1: Prototype and modeled Grain Size Distribution Curves

Three modeled rockfill materials of dmax = 4.75, 10 and 19 mm were obtained using parallel gradation technique and

they were tested to determine the UVC. The dmax v/s UVC has been plotted on semi-log graph and then the UVC for 25, 50, 80 and 600 mm dmax is determined using a best fit linear extrapolation as UVC = -0.04ln (dmax) + 0.548 (1) The determined index properties are given in Table 1.

Drained Triaxial Test Consolidated drained triaxial tests have been conducted on the modelled rockfill materials with confining pressure varying from 0.2 to 0.8 MPa for 87% and 75% RD at Central Soil and Materials Research Station (CSMRS), New Delhi. Stress-strain-volume change behaviour for all the dmax is studied (Honkanadavar 2010). From the stress-strain plots, it is observed that the behaviour is non-linear, inelastic and stress level dependent. For the same dmax and σ3, the deviatoric stress at failure increases with increase in RD (results not reported). The volume change behavior shows compression during the initial part of shearing and dilation with further shearing which decreases with increase in dmax and σ3. For the same dmax and σ3, the volumetric strain at failure decreases with increase in RD (results not reported). Typical stress-strain-volume change behaviour of 4.75 mm modelled rockfill material tested with 87% RD has been presented in Fig. 2. Mean stress v/s deviatoric stress was plotted and shear strength parameter, φ is determined for all the dmax tested with 87% and 75% RD and presented in Table 1. Using the standard procedures, material parameters viz. elastic, ultimate, phase change, hardening and non-associative parameters are determined from the laboratory test results (Honkanadavar 2010).

Table 1: Index Properties of Rockfill Material

Properties RD (%)

dmax (mm)

4.75 10 19 25 50 80

UVC 0.48 0.45 0.43 0.42 0.39 0.37 UCS (MPa) 85.32

φ (Degree) 87 36.3 38.6 39.8 40.8 42.6 43.9 75 34.5 36.8 38.2 39.4 41.5 42.8

PREDICTION OF MATERIAL PARAMETERS USING INDEX PROPERTIES Following procedures have been developed to predict the elastic parameters (E and ν) and the non-dimensional parameter B of the modeled rockfill materials using index property.

21 T

a3T

ir )P/()UVC(DE/E σ= (2)

43 Ta3

Tir )P/()UVC(D/ σ′=νν (3)


(a) Stress-Strain Behaviour

(b) Volume Change Behaviour

Fig. 2: Stress-Strain-Volume Change Behaviour of Riverbed Rockfill Material of 4.75 mm for Noa Dehing Dam Tested with 87% RD. where, E and ν are the modulus of elasticity and Poisson’s ratio of rockfill and Eir and νir are the modulus of elasticity and Poisson’s ratio of intact rock from which rockfill materials are derived. D and D΄ are the coefficients and T1, T2, T3 and T4 are the exponents. Uniaxial compression tests were conducted on the NX (54 mm diameter) size rock cores and Eir and νir were determined from the axial strain v/s axial stress and lateral strain v/s axial stress plots at 50% failure stress respectively (Honkanadavar 2010). It is known that the factors viz. relative density, UCS, gradation, dmax, particle shape and surface texture affect the shear strength parameter of the granular materials. Therefore, to incorporate these factors, a non-dimensional parameter B has been related to the index properties viz. UCS, UVC and RD. The relationship of B with index properties is proposed as

321 ppp )RD()UVC()P(CB =′ (4)

where, P is the normalized UCS (Ratio of UCS of the material to the maximum UCS of the material among all the rockfill materials i.e. UCS/UCSmax). Using the developed computer programme, the coefficients and exponents in Eqs. 2-4 are determined using a least squares fitting technique. For determining the coefficients and exponents, total nine project materials viz. Noa Dehing dam and Lower Jehlum Project (Honkanadavar 2010) and from Ranjit Sagar dam, Tehri dam (old Dobatta and new Dobatta quarries), Western Yamuna Canal (Bridge site and Silt Ejector site), Kol dam and Shah Nehar projects (Abbas 2003) have been considered. Substituting the values of coefficients and exponents in Eqs. 2-4 becomes

52.0a3

991.04ir )P/()UVC(10x839.1E/E σ= −− (5)

031.0

a3615.0

ir )P/()UVC(04.2/ −σ=νν (6)

351.0164.0218.0max )RD()UVC()UCS/UCS(995.0B =′

(7) Substituting the values of Eir, UVC, νir and σ3, E and ν of rockfill material can be determined for any dmax using Eqs. 5-6. Substituting the normalized UCS (UCSmax= 168.68 MPa), UVC and RD in Eq. 7, B′ value can be determined for any dmax. Using Eqs 5-6, Elastic parameters were predicted for all the dmax tested with 87% and 75% RD and compared with the elastic parameters determined from the laboratory tests. It is observed that both the results match closely (Honkanadavar 2010). Therefore, these procedures have been adopted to determine the elastic parameters of modeled and prototype (600 mm) rockfill material. Other material parameters of modeled rockfill material viz. ultimate, phase change, hardening and non-associative parameters were determined by the laboratory tests. Plots were drawn between these parameters v/s B′ and parameters were predicted for different dmax using best fit curve (Honkanadavar 2010).

PREDICTION OF STRESS-STRAIN-VOLUME CHANGE BEHAVIOUR USING HISS MODEL

Using the developed procedures, material parameters were determined. Using the predicted material parameters, stress-strain-volume change behaviour of modeled and prototype alluvial rockfill material of Noa Dehing dam has been back predicted for all the dmax tested with 87% and 75% RD. The predicted behaviour of modeled rockfill material matches closely with the observed behaviour of modeled rockfill material for both RD. The typical observed and predicted behaviour of modeled and predicted prototype rockfill material of different dmax for Noa Dehing Dam project tested with 87% RD (σ3=0.8 MPa) is shown in Fig. 3.


(a) Stress-Strain Behaviour

(b) Volume Change Behaviour

Fig. 3: Observed and Predicted Stress-Strain-Volume Change Behaviour of different dmax for Noa Dehing material tested with 87% Relative Density (σ3 = 0.8 MPa).

CONCLUSIONS The alluvial rockfill material from Noa Dehing dam site, Arunachal Pradesh has been considered in the present research work. The material has been modelled into dmax of 4.75, 10, 19, 25, 50 and 80 mm and tested in the laboratory under drained triaxial test conditions for different confining pressures varying from 0.2 to 0.8 MPa and RD of 87% and 75%. The index properties viz. UCS and UVC have been determined. Procedures were developed to predict the elastic and non-dimensional parameters using the index properties viz. UCS, UVC and RD of the alluvial rockfill material. The predicted and determined (from laboratory tests) elastic parameters were compared and observed that both values match closely. Other material parameters were related with the non-dimensional parameter B and predicted the values for modeled and prototype material using best fit extrapolation technique.

Using the predicted material parameters, the stress-strain-volume change behaviour of modeled and prototype alluvial rockfill material is predicted using HISS model based on elasto-plasticity. The predicted stress-strain-volume change behaviour of modeled rockfill material is compared with the observed behaviour. From the comparison, it is observed that both the results match closely. It is also observed that the predicted behaviour of prototype rockfill material follows similar trend as that of modeled rockfill material. Therefore, this model appears to be suited to characterize the behaviour of alluvial rockfill materials.

ACKNOWLEDGEMENT The authors wish to thank Shri Murari Ratnam, Director, CSMRS for allowing to carry out the laboratory tests. Thanks are also to the staff of rockfill division, CSMRS for their help in conducting the laboratory tests. REFERENCES 1. Abbas, S.M. (2003), Testing and Modeling the

Behaviour of Riverbed and Quarried Rockfill Materials, Ph.D. Thesis, I.I.T. Delhi.

2. Abbas, S.M., Varadarajan, A. and Sharma, K.G. (2003), Prediction of Shear Strength Parameter of Prototype Rockfill Material, IGC-2003, Vol-I, pp. 5-8, Roorkee.

3. Ahlrich, R.C. (1996), Influence of Aggregate Prpoerties on Performance of Heavy-Duty Hot Mix Asphalt Pavements, Transportation Research Record 1547, Transportation Research Board, National Research Council, Washington D.C.

4. ASTM C1252 (1998), Standard Test Method for Uncompacted Void Content ASTM Standard.

5. Gupta A.K. (2000), Constitutive Modelling of Rockfill Materials Ph.D. Thesis, I.I.T. Delhi.

6. Honkanadavar, N.P. (2010), Testing and Modelling the Behaviour of Modelled and Prototype Rockfill Materials, Ph.D. Thesis, I.I.T.Delhi (Submitted).

7. Lowe, J. (1964), “Shear Strength of Coarse Embankment Dam Materials”, Proc. 8th Int. Congress on Large Dams, Vol. 3, pp. 745-761.

8. Marachi, N.D., Chan, C.K., Seed, H.B. and Duncan, J.M. (1969), Strength and Deformation Characteristics of Rockfill Materials. Report No. TE. 69(5), Civil Engineering Department, University of California, Berkeley, USA.

9. Marsal, R.J. (1967), Large Scale Testing of Rockfill Materials, J. of Soil Mech. And Foundations Division, ASCE, 93(2), pp. 27-43.

10. Varadarajan, A., Sharma, K.G., Venkatachalam, K. and Abbas, S.M. (2002a), Constitutive Modeling of Rockfill Materials from Tehri Dam, Uttaranchal, Proc. IGC-2002, Allahabad, India, Vol. 1, pp: 592-595.

11. Varadarajan, A., Sharma, K.G., Venkatachalam, K. and Gupta, A.K. (2003), Testing and Modeling Two Rockfill Materials”, J. Geotech. And Geoenv. Eng., ASCE, 129 (3), pp. 206-218.

, , , , , Observed Predicted

Combined legend for Fig 3 (a) and 3 (b)

4.75 mm 10 mm 19 mm 25 mm 50 mm 80 mm 600 mm Predicted

, , , , , Observed Predicted

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No. A111)

INFLUENCE OF BENTONITE PROPORTIONS ON SWELLING CHARACTERISTICS OF CLAY MIXES

J. M. Kate, Professor, Datta Meghe Institute of Engineering, Technology & Research, Wardha, India, [email protected] Sunil Kumar, Former M. Tech student, Indian Institute of Technology Delhi, New Delhi, India M. P. Bhorkar, Asst. Prof., G. H. Raisoni College of Engineering, Nagpur, India, [email protected] ABSTRACT: Extremely fine grained fat clay commercially known as ‘Bentonite’ has extensive utilities in construction industries, Geotechnical explorations and ground improvement techniques. The present experimental study has been undertaken to understand swelling and swelling pressure behavior of high grade bentonite and kaoline clay mixed in different proportions by weight, in the context of suitability of such mixes for various utilities. The studies clearly indicate considerable influence of bentonite on swelling and swelling pressure behavior and related properties of clay mixes. The findings of the present study alongwith a comparison between clay mixes with low grade bentonite and the recommendations given may prove to be useful in construction industry where bentonite is used as a single material or mixed with other clays. INTRODUCTION The properties of bentonite clays such as very high volume change (swelling and shrinking) & swelling pressure, high cohesion, extremely low permeability, thixotrophy, etc. make it extremely useful material in construction industry for specific civil engineering construction, subsurface exploration and ground improvement practices. It is extensively used during the construction of diaphragm walls, used in slurry form as bore hole stabilizer , fine clay grout, drilling mud, etc. Processed bentonite clay is invariably adopted as a sealing / lining material for industrial waste repositories and landfills to prevent percolation of liquid pollutants in to the surroundings. It may be used as a single material or in combination with other clays mostly on economy considerations. Processed kaoline clay is chosen as another clay for preparing mixes with bentonite as it possesses negligible swelling and relevant properties. In view of numerous utilities and different specification requirements for ech type of utility, very recently Kate et.al. [1] conducted experimental study to understand the volume change & swelling pressure behavior of clay mixes of low grade bentonite with kaoline clay. The present study is identical to theirs, with the difference of quality of bentonite used. Herein, high grade bentonite is used instead of low grade bentonite to understand the influence of quality. The findings have been discussed in the light of industrial applications of this study. The recommendations pertaining to suitability of clay mixes for any particular job utility have been given. LITERATURE REVIEW Overview The swelling phenomenon in clays is attributed mainly to the presence of montmorillonite clay mineral. Bentonite clay contains sodium montmorillonite as its predominant

mineral constituent. The governing property of bentonite to satisfy specification requirements is different in different utility jobs. However, all such specification criteria are directly related with degree of expansivity & swelling potential, and all these can conveniently be assessed/ predicted through swelling/ volume change behaviour of expansive clays [1]. Kate et. al. [1] recommended clay mix with at least 70% bentonite (low grade) for its use in construction of diaphragm walls, sealing/ lining for industrial waste respositories, etc. Whereas, they reccomended clay mixes between 50% to 30% bentonite for the work where used in slurry form e.g. fine clay grouts, borehole stabilizer, drilling mud, etc. Engineering Behaviour of Expansive Clays Experimental investigation carried out by Kate [2] on two extreme (low and high) grades of Indian bentonites to assess their suitabilities for utilities in civil engineering construction works. The relevant properties determined were mineralogy, chemical compositions, swelling, swelling pressure, viscosity, gel strength and coefficient of permeability. He reported that, even without any chemical additive both these extreme grade bentonites satisfy minimum specification requirements for their uses in almost all engineering utility works. Based on synthesis of data available in literature, Sridharan and Prakash [3] summarized broad guidelines to predict degree of expansivity of clays from routine tests. Accordingly, from the magnitudes of liquid limit, plasticity index, colloidal clay contents, shrinkage limit, free swell index, etc. the degree of expansivity (low, medium, high or very high) can be predicted. The synthesis and analysis of data from literature on swelling, swelling pressure and other relevant engineering characteristics of expansive clays were carried out by Sunil Kumar [4]. He demonstrated very good correlations amongst free swell

J. M. Kate, Sunil Kumar and M. P. Bhorkar

index, maximum percentage swell and corresponding swelling pressure of such clays. EXPERIMENTAL PROGRAMME The clays chosen, clay mixes prepared, various tests conducted, the procedure adopted, etc. in the present laboratory study are briefed herein. Clays and Clay Mixes Commercially available processed high grade bentonite is chosen for this investigation. The results of tests namely x-ray diffraction, differential thermal analysis, chemical analysis and scanning electron microscopy adopted on it’s samples inferred presence of sodium montmorillonite as predominant clay mineral [2]. The bentonite samples exhibited base exchange capacity and specific surface area of 112 meq/ 100 g and 153 m2/g respectively. The maximum dry unit weight and optimum moisture content values corresponding to Standard Proctor tests were 12.3 kN/m3 and 43 % respectively. Kaoline clay is another clay used in this study to prepare mixes with this high grade bentonite. The relevant test results adopted pronounced the presence of mostly kaolinite clay mineral [2]. The base exchange capacity and specific surface area of samples were 18 meq/ 100 g and 14m2/ g respectively. Standard Proctor test provided the values of 16.9 kN/m3 and 20 % respectively for maximum dry unit weight and optimum moisture content. Representative clay mixes were prepared by thoroughly mixing absolutely dry (zero moisture content) bentonite and kaoline clay in desired proportions by weight. The proportions of bentonite and kaoline clay in these mixes were 100:0, 80:20, 70:30, 50:50, 30:70 and 0:100, which for brevity are referred in the text as clay mix A, B, C, D, E and F respectively. Tests Performed Both the clays (bentonite, kaoline clay) and clay mixes have been tested for their relevant engineering properties following the procedure as per Indian standard (IS) codes of practice for various laboratory tests. These tests include hydrometer analysis and Atterberg’s limits such as liquid limit, plastic limit, plasticity index and shrinkage limit, etc. Free Swell Index (FSI) tests on clays and clay mixes were conducted adopting the procedure suggested in IS: 2720, part 40 [5]. The maximum percentage swell and swelling pressure were determined as per the procedure given in International standard TC-6 [6]. Accordingly, the maximum percentage swell is the ratio of heave (under applied stress of 5kPa) to the initial thickness of clay sample on it’s full saturation in Oedometer. The percent swell has been obtained under each applied stress followed by determination of swelling pressure (which corresponds to magnitude of applied stress at zero percent swell) from the plot of percent swell versus applied stress (on log 10 scale). Swell and swelling pressure tests have been

conducted on all these clay samples remoulded by static compaction method at initial dry unit weight of 12.3kN/m3 and zero percent initial moisture content. The initial moisture content of zero percent has been selected as it always provides maximum possible value of swelling pressure corresponding to chosen initial compaction dry unit weight. RESULTS AND DISCUSSION Engineering Properties The engineering properties of clays and their mixes are produced in Table 1. Table 1 Engineering properties & free swell index of Clay

Mixes

Property Clay Mix

A B C D E F

Proportions (%) Bentonite 100 80 70 50 30 0 Kaoline Clay 0 20 30 50 70 100

Size Distribution(%) Silt (0.075-0.002mm) 12 19 24 30 38 46

Clay (<0.002 mm) 88 81 76 70 62 54

Colloidal Clay (<0.001 mm) 55 49 42 38 35 24

Atterberg Limits(%) Liquid Limit 610 390 240 135 105 52 Plastic Limit 150 120 55 45 40 29 Plasticity Index 460 270 185 90 65 23 Shrinkage Limit 5.2 6.5 8.2 9.1 9.8 17.1

Classification CH CH CH CH CH CH

FSI (%) 780 610 480 320 290 32 The Table 1 clearly shows distinct trend of changes in all the properties brought due to decreasing percentage of bentonite. For example, the shrinkage limit continuously increases from 5.2% for 100% bentonite to 17.1% for 0% bentonite (100% Kaoline clay). Such trends of either continuous decrease or increase in magnitudes of property with decrease in bentonite percentage are distinctly noticeable. The percentages of colloidal clay (<0.001 mm) in these mixes >27% indicate that, they possess very high degree of expansivity [3]. Similarly, very high values of liquid limit (>60), plasticity index (>32 %) and extremely low values of shrinkage limit (<10) substantiate that these clay mixes characterizes high degree of expansion [3]. All the clays

Influence of Bentonite Proportions on Swelling Characteristics of Clay Mixes

and clay mixes are classified (‘A’ line classification) as CH i.e. clays with high plasticity. The values of free swell index (FSI) >200 exhibited by clay mixes A, B, C, D & E demonstrate very high degree of their expansivity [3]. The kaoline clay (sample F) shows atmost no expansivity. Maximum Swell and Swelling Pressure The plots of swell versus applied stress for these clay mixes are illustrated in Fig. 1. The swell continuously decreases with increasing applied stress as represented by variational curves in Fig. 1. It is interesting to note that beyond certain value which corresponds to swelling pressure, the swell enters into negative zone indicating shrinking (compression) of sample. The magnitudes of swelling pressure at no volume change condition have been deduced from these curves.

Fig. 1 Applied stress versus Maximum percentage swell The swelling pressures exhibited by clay mix A, B, C, D and E are 780, 610, 480, 320 and 290 kPa respectively and the corresponding maximum swells are 32.5, 29.7, 25.4, 20.9 and 17.1 percents. These values of swelling pressures as well as maximum swell (%) fall under very high expansivity category for all the clay mixes A, B, C, D & E. The swelling pressure, maximum swell and free swell index as a function of bentonite percentage demonstrates that all these increase with increase in bentonite as illustrated by curves in Fig. 2. The relationship between swelling pressure (Psw) and bentonite percentage obtained through regression analysis of data is expressed by Eq. 1.

20.02( ) 2.4( ) 17SWP B B= + + (1) Wherein, B denotes the bentonite percentage.

Fig. 2 Variation of swelling characteristics with bentonite FSI and Swelling Pressure A synthesis of available data on FSI and corresponding swelling pressure for various expansive clays extracted from most recent literature has been carried out in the present study. Various points on the plot of FSI versus swelling pressure illustrated in Fig. 3 correspond to such data from several investigators [4].

Fig. 3 Experimental and review synthesized data The curve in this figure has been derived as a best fit representing the entire data, and its regression analysis provides the correlation expressed by Eq. 2.

0.7254.46( )SWP FSI= (2) The experimental results of the above expansive properties obtained in this study for high grade bentonite alongwith data on low grade bentonite from Kate et. al. [1] is also illustrated in Fig. 3. The experimental values of swelling pressure of these clay mixes correlates with FSI as given in Eq. 3.

0.814.1( )SWP FSI= (3)

J. M. Kate, Sunil Kumar and M. P. Bhorkar

It is seen in Fig. 3 that both the curves exhibit nearly the same trend of power law relationships. It is interesting to note that these curves show significant deviations at lower values of swelling pressures, which tends to reduce with increasing values, and beyond swelling pressure of 250 kPa both these curves overlap. This reflects on the distinct behaviour of bentonite having predomonantly sodium montmorillonite as its mineral constituent and its mixes as compared to other expansive clays. The synthesized data from the literature corresponds mostly to clays having combination of montmorillonite and illite as their mineral constituents. CONCLUSIONS The findings of the present experimental study lead to the following conclusions. (i) The prediction through direct criteria namely FSI,

percent swell and swelling pressure and also from semi-direct criteria such as colloidal clay contents and Atterberg’s limits, all these clay mixes fall in the category of very high expansivity.

(ii) The synthesized data from literature and present experimental data demonstrate a power law relationship between FSI and swelling pressure covering a very wide range of their magnitudes.

(iii) This study clearly demonstrates that the quality of bentonite is very important criterion in clay mixes, as the earlier study for low grade bentonite [1] has shown these mixes (>70 % bentonite) only of high expansivity whereas, the ither mixes (<70% bentonite) moderate expansivity.

RECOMMENDATIONS The quality of bentonite plays very important role in its use in clay mixes with kaoline clay. The high quality bentonite having its proportion only 30 % with kaoline clay is recommended for its use in construction of diaphragm wall, sealing/ lining for industrial waste repositors, etc. This bentonite even with much smaller proportion may be used for the work in slurry form e.g. fine clay grouts, bore hole stabilizer, drilling mud, etc.

REFERENCES 1. Kate, J.M., Sunil Kumar and Bhorkar, M.P. (2012),

Volume change characteristics of clay mixes with different Bentonite percentages, 2nd Int. Conf. On Geotechnique, construction materials and Environment (GEOMAT-2012) Nov. 14-16, Kuala Lumpur (accepted).

2. Kate, J M, (1989), Mineralogical and Engineering Behaviour of Indian Bentonites, 2nd World Congress on Non-metallic Minerals, Beijing, China, Oct., pp. 393-398.

3. Sridharan A and Praksh K, (2000), Classification Procedures for expansive Soils, J. of Institution of Civil Engineers, Geotechnical Engineering, Vol. 143, pp. 235-239.

4. Sunil Kumar (2007), An Experimental Study on the Behaviour of Expansive Clays treated with different Fly Ashes, Unpublished M. Tech. Thesis CE Deptt., IIT Delhi, May, pp.1-158.

5. Indian Standard IS: 2720, Part 40 (1997), Determination of Free Swelling Index of Soils, Bureau of Indian Standards, New Delhi, pp. 1-17.

6. Technical Committee TC- 6, (1993), Evaluation of Swelling Pressure and corresponding Heave of Expansive Soils by constructing Swell percentage versus applied total stress diagram, Report of Technical Committee (ISSMGE) on expansive soils, CBIP, New Delhi, pp. 1-26.


PERFORMANCE OF NANOSIZE PARTICLES IN SOIL GROUTING Stalin V.K, Professor, Department of Civil Engineering, Anna University, Chennai – 600 025, E-mail: [email protected]. Balakumaran G, Post Graduate Student, Anna University, Chennai – 600 025, E-mail: [email protected]. ABSTRACT: An attempt is made in this investigation to incorporate nanosize particles in soil grouting technique. Permeation experiments were conducted in column test for Nano silica and Bentonite solution for varying particle size (coarse, medium and fine sand), varying density (loose and dense state) and varying viscosity of grout solution (5%, 10% and 15%). Time taken for steady value of grout solution were measured and correlated with density of sand, particle size and viscosity of the solution. It is found that at any given density and viscosity of grout solution, the higher volume of soil is successfully grouted using Nano solution unlike Bentonite. The coefficient of permeability of nano size particles grouted sample is 3 to 4 orders lower than that of ungrouted specimen unlike bentonite solution. INTRODUCTION Nanoscience is the study of phenomena and manipulation of materials at atomic, molecular and macromolecular scales, where properties differ significantly from those at a larger scale. According to the number of nanoscale dimensions, soil nanoparticles in general have three different forms: nanoplatelets (e.g., some platy clay mineral particles with one dimension at nanoscale), nanowires or nanotubes (i.e., fibrous particles with two dimensions at nanoscale), and nanodots (i.e., nanoscale at all three dimensions). The size of the nano particle plays a significant role in behavior of soil exhibiting different properties. A typical example is the significantly different properties exhibited by a cohesive soil consisting of dominantly clay particles and by a coarse-grained sandy soil. This change in behavior is caused by two main reasons: much increased surface area and high reactivity. In the present investigation, attempt was made to enhance the practitioner’s knowledge in the rheological properties and flow characteristics of nanosize particles for improving the application of nanoparticles based permeation grouting in the stabilization of soil. In the last decades, researches have been carried out to study the significance of nanosize particles in soil behaviour and also in rocks and concrete. Zhang et al. (2003) [1] discussed the effects of Fe-oxide cementation on compressibility and rate of consolidation of a weathered old alluvium. Marzieh Kadivar et al. (2011) [2] discussed the application of nanotechnology in geotechnical engineering, in which the concept of nanotechnology as well as the new concept of nanosol is explained. Brian Green (2006) [3] developed a high density cementitious rock matching grout using nanoparticles. Morsky et al. (2010) [4] studied the effect of Nano clay on the mechanical properties and microstructure of Portland cement mortar. The nano-clay used in this research was nano-kaolin. Despite the significant research about the influence of nanosize particles on physical and mechanical properties of mortars or concretes, there have not been significant researches about their efficacy when utilized in grouting purposes.

MATERIALS Sand Three different sand samples (sample 1, sample 2, and sample 3) were used in this investigation. The particle size distribution curve is shown in figure 1. The samples are classified as ‘SP’ category and are used as the material for studying the effectiveness of Nanosize particles in soil stabilization. The relative density of sand is varied as 30%, (Loose state) and 70%, (dense state).

Fig. 1 Particle size distribution curves of sand samples used Nano particles and Bentonite Grouting Solution

. Fig. 2 SEM images of Nano (Left) and Bentonite (Right) Sample In the present study, Nanoparticles and Bentonite were selected as two different grout solutions. The SEM images of the original sample of Bentonite and the Nano Silica solution

Stalin V.K and Balakumaran G

are shown in the figure 2. (magnified 15,000 times). It generally shows that the Nano solution samples have greater portion of nano size particles in which the 60-90 nm particles are clearly evident when compared to Bentonite. From the SEM results, it is clearly evident that particle sizes are very much finer compared to Bentonite. EXPERIMENTAL SETUP The test apparatus consists of an inlet tank, transparent acrylic mould, control valve and flexible connection hoses. The sample mould as shown in figure 3 was constructed of 4-inch (10-cm) Inner diameter and height of 200mm transparent acrylic pipe with a wall thickness of 8 mm. End caps with openings for inflow and outflow of grout solution were used on the top and bottom end of the mould. The overall grout flow was supplied from the supply tank, connected to the top end cap of the acrylic mould. Furthermore, the test apparatus has control valve to regulate the flow rate and pipes to ensure connections between equipment.

Fig. 3 View of Column Test Conducted on Soil Sample Permeating Nano and Bentonite Solution

To understand the effectiveness of the Nanosize particles in stabilization of soil, permeation tests were conducted in the model test setup for varying viscosity of nano solution. The test was also carried out for varying particle size and density of sand sample. In order to compare the results of the Nano solution, experiments were also conducted on the Bentonite solution. The details of the test programme are shown in table 1. The Nano and Bentonite solution with varying viscosity (5%, 10%, 15%) was sent through the pipe and the cumulative volume of grout solution were noted at regular interval till the known quantity of solution spreads into the sample. RESULTS AND DISCUSSSION

In order to evaluate the performance of Silica Nanosize particles as grout solution, permeation tests were conducted for Nano solutions in the model experimental setup for varying density of sand (loose and dense state),varying particle size (coarse, medium and fine sand) and varying viscosity of grout solution (5%, 10%,15% by weight). Effect of State of Soil (Relative Density) on Grouted Volume of Nano and Bentonite Solution For better understanding of flow rate of nanosize particles and Bentonite grout solution and grouted volume at a given time, column tests were conducted in a model tank for varying viscosity of solution, density of sand and particle size. The density of the soil to be grouted (otherwise void ratio) plays a crucial role to determine the grout pressure to be selected and also the viscosity of grout solution. In this investigation, the loose and dense state of the sand was taken as 30% and 70% respectively.Further, it is seen from figure 4, the peak value of grouted volume reached at 150 sec itself for 5% Nano solution in loose state, 225 sec in dense state and where as the peak value is seen only at 210 sec and 375 sec respectively for loose and dense state when Bentonite solution was used. Also, the grouted volumes for loose and dense are 637 ml and 738 ml respectively for 5% nano solution and the same is 623ml and 728 ml for Bentonite solution. Thus, the increase in density of soil sample holds good for Nano solution as reflected in higher grouted volume compared to Bentonite solution

Fig. 4 Effect of density on Grouted Volume and Permeation Time for 5 % Nano and Bentonite Solution in Coarse Sand Effect of Particle Size on Grouted Volume for Nano and Bentonite Solution To understand the effect of particle size on the permeation efficiency of Nano particles in comparison with the Bentonite solution, flow test was conducted for coarse sand (4.75mm - 2.36mm), medium sand (2.36 mm – 425micron), fine sand (75 micron – 425micron).Coarse sand and fine sand can have same void ratio, but the size of void need not be the same. Coarse sand may have bigger voids of fewer numbers and on the other side fine sand will have smaller voids larger number. In this case, the size of void is a matter for changing the permeation characteristics of a given solution. Among the two grout solutions, the Nano solution provides higher

Performance of Nanosize Particles in Soil Grouting

grouted value at lesser time in coarse sand and fine sand in loose state or dense state. For example, the peak value of grouted volume is 685ml for 150 sec (figure 5) for coarse sand in loose state, whereas fine sand in loose state yields grouted volume of 700ml only at 390 sec and medium sand lies in between (i.e 260 sec for 650 ml).

Fig. 5 Effect of Particle Size on Grouted Volume and Permeation Time for 5 % Nano Solution in Loose State In the case of Bentonite solution, the peak value of grouted volume reached at 210 sec (figure 6) for coarse sand corresponding to 600ml of grout, 225 sec corresponding to 400 ml of grouted volume for medium sand and 225 sec for 350 ml of grouted volume. This means, that there is an abrupt reduction in permeation of Bentonite solution with reference to particle size and whereas, Nano particles gives higher grout volume as wells less time interval.

Fig. 6 Effect of Particle Size on Grouted Volume and Permeation Time for 5% Bentonite Solution in Loose State Effect of Viscosity of Grout Solution on Grouted Volume Apart from the particle size and density of soil to be grouted, it is equally important to understand the viscosity of the grout solution which normally depends on the in-situ conditions, with reference to water table, stratification of soil etc. To understand the effect of viscosity of grout solution, three different viscosities were selected as 5% (by weight of solution), 10 % and 15%.For example,50 g of dry Nano particles are mixed with 1000ml of water for 5% viscosity,100g and 150 g for 10% and 15% viscosities

respectively. Figure 7 presents the peak time of grouted volume for 5 %, 10%, and 15% Nano solution is 150 sec, 240 sec and 285 sec respectively for Coarse sand in loose state. The grouted volume corresponding to peak time was 654ml, 803ml and 849 ml for 5%, 10% and 15% Nano solution respectively.

Fig. 7 Effect of Viscosity of Nano Solution on Grouted Volume and Permeation Time for Coarse Sand in loose State In the case of Bentonite solution, the grouted volume corresponding to peak time was 623 ml at 210sec, 446 ml at 195 sec and 468 ml at 225 sec for 5%, 10% and 15% viscosities respectively (figure 8). It was found that the progression of Nano solution through the soil columns occurs most rapidly when the viscosity is about 5% by weight of dilution. By the time the viscosity increases to about 10% and 15% respectively, the viscosity of the Nano solution has decreased the initial flow rate and the travel time required has increased dramatically.

Fig. 8 Effect of viscosity of Bentonite Solution on Grouted Volume and Permeation Time for Coarse Sand in Loose State Hydraulic Conductivity of Grouted Soil Permeability of soils is one of the most sensitive properties which need to be determined for important geotechnical engineering applications such as lanfill design, filter design, dewatering etc. Permeability coefficient is influenced by many factors such as type of soil, soil strata, hydraulic head, degree of saturation, pore fluid characteristics etc,. Of them, the influence of pore fluid characteristics is very important

Stalin V.K and Balakumaran G

espescially for grouting methods of soil improvement, employed for either arresting the seepage or formation of gel as stable layer to give raise to higher strength and lower comprssibility, which depends cementitious components present in grout solution. The constant head permeability tests were conducted to determine the permeability co-efficient of grouted and ungrouted samples to understand the effect of grouting.The permeability co-efficient values (k) were calculated by using the following equation k = QL/Aht (1) Where Q = Discharge in cm3, L = Length of soil sample in cm, A = Cross sectional area in cm2,

h = pressure head in cm, t = time in sec Table 1 Co- efficient of Permeability Values Before and After Grouting for Varying Soil Type and Viscosity of Solution

Type of Sample

Relative Density

Co-efficient of Permeability (cm/sec)

Before Grouting

After Grouting

5% 10% 15%

Coarse Sand

Loose (30%)

9.78 x10-1

5.78 x10-3

3.45 x10-4

6.23 x10-5

Dense (70%)

7.84 x10-2

8.67 x10-4

2.65 x10-5

7.86 x10-6

Medium Sand

Loose (30%)

8.56 x10-2

3.80 x10-4

4.55 x10-5

1.24 x10-6

Dense (70%)

3.69 x10-3

4.21 x10-4

9.35 x10-5

6.73 x10-6

Fine Sand

Loose (30%)

4.34 x10-3

8.74 x10-4

5.88 x10-5

9.23 x10-5

Dense (70%)

9.34 x10-3

2.78 x10-5

8.22 x10-5

7.45 x10-6

The ‘k’ value of ungrouted coarse sand, medium and fine sand are 7.82 x 10-2 cm/sec, 3.62 x 10-3 cm/sec and 9.34 x 10-3 cm/sec respectively in dense state and the same increased to 9.78 x 10-1cm/sec, 8.56 x 10-2 cm/sec and 4.34 x 10-

3cm/sec repectively in loose state. On the otherside, the ‘k’ value of nano grouted soil (5%) is 5.78 x 10-3 cm/sec, 3.84 x 10-4 cm/sec and 8.74 x 10-4 cm/sec where as 1 to 2 fold lower than the same case of ungrouted soil sample. For 10% and 15% nano solution, the order of permeabilty rduction (table 1) is 3 to 4 fold comparing grouted sample to ungrouted soil sample. CONCLUSIONS Based on the analysis of grouted volume and co-efficient of permeability for different time interval, the following general conclusions may be drawn,

1. As the density of soil increases, the time of permeation increases and grout volume decreases both for Nano and Bentonite solution but, however, upon using Nano solution, the grouted volume was always high and permeation time low compared to Bentonite solution.

2. As the particle size increases (coarse sand, medium and fine sand), the permeation time decreases and grouted volume increases for Nano solution much higher than Bentonite solution.

3. As the viscosity of the Nano or Bentonite solution

increases, the permeation time increases and grouted volume decreases. But, however, Nano particle solution, the changes are significant compared to Bentonite solution.

4. The co-efficient of permeability of sand after permeation

of nano particles showed a 3 to 4 order of magnitude decrease in hydraulic conductivity from the range between 10-2 and 10-3 cm/s to the range between 10-5 and 10-6 cm/s, compared to ungrouted sand.

Thus it could be summarized that, at any given density, particle size and viscosity of solution, the performance of Nano size solution is always superior compared to Bentonite solution as reflected by its high grouted volume at lesser permeation time. It is hence recommended that the Nano size particles may be used for effective grouting of soil improvement to reduce the permeation time of grout and also enhances the volume of grout. . REFERENCES 1. Zhang, G., Germaine, J. T., and Whittle, A. J (2003)

“Effects of Fe-oxide Cementation on the deformation characteristics of a weathered old alluvium in San Juan, Puerto Rico.” Soils and Foundations, Vol. 43, No. 4, 119-130.

2. Marziegh kadivar, Kazem barkhordari, Mehdi kadivar

(2011) “Nanotechnology in Geotechnical Engineering” Advanced Materials Research Vols. 261-263 (2011) pp 524-528.

3. Brian Green (2006) “Development of a high density

cementitious Rock- matching Grout using Nano particles” Nanotechnology of concrete: Recent developments and Future Perspectives, Vol. 2, 72-74.

4. Morsky M.S, S.H.Alsayed and M.Aqel. (2007) “Effect

of Nano clay on mechanical properties and microstructure of ordinary Portland cement mortar” International journal of civil & Environmental Engineering Vol. 10.

Performance of Nanosize Particles in Soil Grouting

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No A114)

EFFECT OF COMPACTIVE ENERGY ON COMPACTION CHARACTERISTICS OF SOILS Soumya Rosalint Jolly, M. Tech Student, College of Engineering Trivandrum, [email protected] P. Vinod, Associate Professor, College of Engineering Trivandrum, [email protected]

ABSTRACT: The compaction test has been widely used and accepted for characterizing soil compatibility for field compaction control. This paper presents a time saving method to predict the compaction characteristics of soils based on reduced compactive energy. The compactive effort is an important parameter that depends on weight of the hammer, height of hammer fall, number of layers, number of blows per layer and volume of the compacted soil. In the present study compactive energy is reduced by means of reduction in the number of blows and reduction in the number of layers. The results of the study indicate that compaction characteristics bears a unique relationship with compactive enegy for different soils. This makes it possible to predict the maximum dry density and optimum moisture content of soil through the performance of compaction test with much reduced compactive effort. INTRODUCTION Modification of soils at a site to improve their engineering properties is essential in many applications, such as highways, railway subgrades and airfield pavements. Mechanical compaction is used as a practical means of achieving the desired strength, compressibility and hydraulic conductivity characteristics of soils. Proper compaction of materials ensures the durability and stability of earthen constructions. Compaction of soil is a simple process of densification by mechanical manipulation. The moisture content-dry density relationship for the soil obtained from laboratory compaction tests forms the basis for the assessment of the suitability of a soil for compaction specifications. It takes considerable time and effort to determine the compaction characteristics of soils by the standard methods. Hence many attempts have been made in the past to correlate compaction characteristics of soils with easily determinable soil parameters, such as plasticity index and grain-size distribution characteristics. Correlations between the plastic properties of inorganic finegrained soils of alluvial nature and the Proctor’s dry density at various placement moisture content in terms of liquid limit and plasticity index to liquid limit ratio (PI/LL) have been developed [1]. It has been established that by defining a soil by its liquid limit and percentage fraction coarser than 425μm, the path of compaction for a specific compactive effort can be predicted through a simple density-water content-liquid limit relationship [2]. Experimental study has been conducted to determine the compaction characteristics of fine grained soils through a simple index test, the plastic limit [3]. A mini compaction apparatus primarily for use in fine grained soils, which requires only about 1/10th volume of soil needed for the standard and modified Proctor test has been used to determine the compaction characteristics [4]. An empirical model has been developed to estimate dry density of pond ash, using multiple regression analysis, in terms of compaction energy, moisture content, and specific gravity

[5]. A time - and cost-effective method has been introduced to predict standard Proctor compaction characteristics of non-gravel and gravelly soils, using a small compaction apparatus [6]. In the present investigation an attempt has been made to study the effect of compactive energy on compaction characteristics of soils. The compactive effort is a measure of the mechanical energy delivered to the soil mass. A commercially available clay (kaolinite), a lateritic soil and a naturally occuring clay, namely, kuttanad clay are used for the present study. Compactive energy is reduced by means of reduction in the number of blows and reduction in the number of layers. The present study attempts to express the compaction characteristics of soils in terms of energy ratio (the ratio of mechanical energy imparted to the soil per unit volume to the standard mechanical energy imparted to the soil per unit volume for standard heavy compaction) by emperical equations for different soils, through a detailed experimental investigation. This can make it possible to predict maximum dry density and optimum moisture content of soil through the performance of compaction test with much reduced compactive effort. MATERIALS USED Compaction tests were conducted using a commercially available clay, kaolinite obtained from English Indian Clays Limited, Trivandrum, India, a locally available lateritic soil collected from the premises of College of Engineering, Trivandrum, India and a naturally occuring clay collected from Kuttanad region in Kerala, India. Hydrometer analysis, liquid limit, plastic limit, specific gravity and compaction characteristics were determined as per BS standards [7]-[10]. The basic and index properties of the soils determined from laboratory experiments are summarized in Table 1.

Soumya Rosalint Jolly & P. Vinod

Table 1 Properties of kaolinite, lateritic soil and kuttanad clay

Properties Value

Kaolinite Lateritic soil

Kuttanad clay

Specific gravity 2.51 2.48 2.72 Clay size content (%) 72.0 52.0 24.0 Silt size content (%) 28.0 31.5 51.8 Sand content (%) - 16.5 24.2 Liquid Limit (%) 62.0 53.6 92.0 Plastic Limit (%) 33.0 36.5 38.0 Plasticity Index (%) 29.0 17.1 54.0 OMC (%) 24.9 18.9 23.4 γdmax (g/cc) 1.480 1.718 1.494

EXPERIMENTAL PROGRAMME Conventional water content-dry density relationship for kaolinite, lateritic soil and kuttanad clay samples were found out using heavy compaction test. The test consists of compacting the soil in five layers in a standard mould of 1000cm3, each layer being given 27 blows of 4.5kg rammer with a free drop of 450mm. Then dry densities of the compacted soil were plotted against corresponding moisture contents. A smooth curve was drawn through the resulting points and the co-ordinates of the peak point on this curve (optimum moisture content and maximum dry density) were determined. The compaction tests were repeated for reduced compactive energy by a similar procedure. The reduction in compactve energy was imparted by reducing the number of blows and reducing the number of layers from five to three. The number of blows per layer required for each energy ratio is shown in Table 2. Table 2 Compaction tests under reduced compactive energy

Energy ratio (%)

No. of blows per layer When no. of blows

alone reduced When no. of layers reduced from 5 to 3

15 4 6 30 8 13 45 11 19 60 15 25 75 19 31 90 23 38 100 25 42

RESULTS AND DISCUSSION In this section the experimental results on effect of compactive energy on the compaction characteristics of kaolinite, lateritic soil and kuttanad clay are presented The discussion on the results obtained from the experiments has been made in this section with reference to relevant figures and tables. Discussion on development of empirical model by regression technique has also been made in this section.

Compaction Characteristics of Soils Under Reduced Compactive Energy Heavy Compaction tests were conducted for all the soils for both conventional and reduced compactive energies. The compaction characteristics of these soils under reduced compactive energies by reduction in number of blows is shown in Tables 3-5 and that by reduction in number of layers is shown in Tables 6-8. Table 3 Compaction characteristics of kaolinite under reduced compactive energy by reduction in no. of blows

Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.334 26.6 0.90 1.07 30 1.453 24.9 0.98 1.00 45 1.623 24.2 1.10 0.97 60 1.820 23.7 1.23 0.95 75 1.881 23.2 1.27 0.93 90 1.910 22.7 1.29 0.91100 1.914 22.2 1.29 0.89

Table 4 Compaction characteristics of lateritic soil under reduced compactive energy by reduction in no. of blows

Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.562 23.1 0.91 1.22 30 1.635 22.5 0.95 1.19 45 1.662 20.8 0.97 1.10 60 1.699 19.7 0.99 1.04 75 1.712 19.6 1.00 1.0490 1.740 19.4 1.01 1.03 100 1.749 19.2 1.02 1.02

Table 5 Compaction characteristics of kuttanad clay under reduced compactive energy by reduction in no. of blows

Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.383 28.1 0.93 1.20 30 1.426 26.7 0.95 1.14 45 1.478 25.5 0.99 1.09 60 1.506 24.8 1.01 1.06 75 1.522 24.3 1.02 1.04 90 1.537 234 1.03 1.00100 1.566 23.2 1.05 0.99

Effect of compactive energy on compaction characteristics of soils

Table 6 Compaction characteristics of kaolinite under reduced compactive energy by reduction in no. of layers

Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.301 36.9 0.88 1.48 30 1.362 34.9 0.92 1.40 45 1.658 30.9 1.12 1.24 60 1.791 26.4 1.21 1.06 75 1.850 25.0 1.25 1.00 90 1.889 21.7 1.28 0.87 100 1.900 21.2 1.28 0.85

Table 7 Compaction characteristics of lateritic soil under reduced compactive energy by reduction in no. of layers Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.538 23.1 0.90 1.22 30 1.616 22.8 0.94 1.21 45 1.637 21.3 0.95 1.1360 1.675 18.4 0.97 0.97 75 1.708 18.3 0.99 0.97 90 1.715 18.0 1.00 0.95 100 1.715 16.9 1.00 0.89 Table 8 Compaction characteristics of kuttanad clay under reduced compactive energy by reduction in no. of layers Energy Ratio (%)

(γdmax-R) (g/cc)

wR (%)

(γdmax-R) / γdmax

(wR) / w

15 1.315 0.274 0.88 1.1730 1.394 0.264 0.93 1.13 45 1.437 0.253 0.96 1.08 60 1.463 0.241 0.98 1.03 75 1.465 0.234 0.98 1.0090 1.509 0.229 1.01 0.98 100 1.531 0.229 1.03 0.98

γdmax = maximum dry density corresponding to Standard Heavy Compaction

w = optimum moisture content corresponding to Standard Heavy Compaction γdmax-R = maximum dry density corresponding to compaction with reduced compactive energy wR = optimum moisture content corresponding to compaction with reduced compactive energy It can be seen that as the compactive energy increases, the maximum dry density increases For a given moisture content, increase in amount of compaction energy results in closer

packing of soil particles and hence there is an increase in dry density. The decrease in OMC with increase in compaction energy may be due to the reduction in friction between the particles of the soil, enhancing the tendency of the particles to come closer as a result of which air voids decreases. Empirical Models The experimental results indicate that compaction characteristics bears a unique relationship with compactive energy. Hence an attempt has been made to develop empirical models to estimate the compaction characteristics of soils taking into consideration the effects of compactive energy. Multiple regression analysis is performed to develop the empirical model. The possibility of different models such as logarithmic, exponential, and polynomial were investigated from which polynomial model was identified to be the best fit based on its coefficient of determination (R2). The R2 value represents the proportion of variability in the data explained or accounted for by the regression model. The equations of the trend lines along with the respective values of R2 for different soils are presented in Figs. 1-4.

Fig. 1 (γdmax-R) / γdmax vs energy ratio curve for soils under reduced compactive energy by reduction in no. of blows

Fig. 2 (wR) / w vs energy ratio curve for soils under reduced compactive energy by reduction in no. of blows

Soumya Rosalint Jolly & P. Vinod

Fig. 3 (γdmax-R) / γdmax vs energy ratio curve for soils under reduced compactive energy by reduction in no. of layers

Fig. 4 (wR) / w vs energy ratio curve for soils under reduced compactive energy by reduction in no. of layers

The empirical models developed for the determination of compaction characteristics of the soils is a three parameter polynomial equation. It can be observed that the coefficients of determination, R2 value are greater than 0.8 for all the correlations. CONCLUSIONS This paper discusses the influence of compactive effort on the compaction characteristics of three typical soils such as a commercially available clay (kaolinite), a locally available lateritic soil and a naturally occuring kuttanad clay . Based on the results, empirical models have been developed to estimate maximum dry density and optimum moisture contents of different soils as a function of energy ratio. The empirical equation obtained is a three parameter polynomial model for compaction characteristics of soils. Based on the values of coefficient of determination, it seems that the relationships represent the experimental results well. These simple empirical models can be of use to practicing engineers for preliminary estimate of compaction characteristics of soils. REFERENCES 1. Sudhenu Saha and Chattopadhyay, B. C., (1988),

Prediction of Compaction Characteristics from Index Properties, Indian Geotechnical Journal, 18(4).

2. Pandian, N. S., Nagaraj, T. S. and Manoj, M., (1997), Re-examination of compaction characteristics of fine-grained soils, Geotechnique, 47(2) pp. 363-366.

3. Gurtug, Y. and Sridharan, A., (2002), Prediction of Compaction Characteristics of Fine-grained Soils, Geotechnique, 52(10), pp. 761-763

4. Sridharan, A. and Sivapullaiah, P. V., (2005), Mini Compaction Test Apparatus for Fine-Grained Soils, Geotechnical Testing Journal, 28(3).

5. Ashis Kumar Bera, Ambarish Ghosh and Amalendu Ghosh, (2007), Compaction Characteristics of Pond Ash, Journal of Materials in Civil Engineering, ASCE, 19(4), pp. 349-357.

6. Avirut Chinkulkijniwat, Ekachai Man-Koksung, Anuchit Uchaipichat and Suksun Horpibulsuk, (2010), Compaction Characteristics of Non-Gravel and Gravelly Soils Using a Small Compaction Apparatus, Journal of ASTM International, 7(7), pp. 1-15.

7. BS 1377 (Part 2)-1990, Hydrometer Test. 8. BS 1377 (Part 2)-1990, The Atterberg Limits: Liquid

limit with Cassagrande Cup and Plastic limit. 9. BS 1377 (Part 2)-1990, Specific Gravity Test. 10. BS 1377 (Part 4)-1990, Proctor Test.


ONE DIMENSIONAL COMPRESSION BEHAVIOUR OF ROCKFILL MATERIALS

N. P. Honkanadavar, Scientist ‘C’, CSMRS, New Delhi-110016, E-mail: [email protected] Sanjeev Bajaj, Scientist ‘B’, Central Soil and Materials Research Station, New Delhi-110016, E-mail: [email protected] S. L. Gupta, Scientist ‘E’, Central Soil and Materials Research Station (CSMRS), New Delhi-110016, E-mail: [email protected] Murari Ratnam, Director, Central Soil and Materials Research Station, New Delhi-110016, E-mail: [email protected] ABSTRACT: In the present study, alluvial modeled rockfill material from Renuka dam project, Himachal Pradesh, India and quarried blasted modeled rockfill material from Salma dam project, Afghanistan were obtained and tested. The maximum particle size (dmax) of the prototype gradation rockfill material for Renuka dam and Salma dam is 1000 and 600 mm respectively. For testing, the size is scaled down to dmax of 80, 50 and 25 mm for both project materials using parallel gradation technique. One dimensional compression (Oedometer) tests are carried out with a specimen size of 1000 mm diameter and 600 mm height. All the tests were conducted for the vertical stress (σ1) ranging from 0.637 to 3.82 MPa and 87% relative density (RD). Stress-strain behaviour of both the modeled rockfill material is studied and presented. It is observed that the stress-strain behaviour is nonlinear and inelastic for both the materials. The deformation modulus, M and elastic modulus, E was determined for all the dmax of both alluvial and quarried rockfill material. It is observed that M and E increases with increase in dmax for alluvial and quarried rockfill material. However, for the same vertical stress, the value of M is higher than the value of E for both the materials. The coefficient of earth pressure increases with decrease in dmax for both materials. The void ratio decreases with increase in vertical stress but it increases with decrease in dmax. The coefficient of permeability, K increases with decrease in dmax. From the particle breakage analysis, it is observed that the breakage factor, Bg increases with increase in dmax and σ1 for both materials. However, the effect of σ1 and dmax is more on Bg for quarried rockfill material than alluvial rockfill material. INTRODUCTION Rockfill consists of gravels, cobbles and boulders obtained either from the natural riverbed or by blasting the parent rock quarry. Alluvial rockfill material is rounded to subrounded in shape and quarried rockfill material is angular to subangular in shape. These materials are used in the construction of rockfill dams because of their inherent flexibility, capacity to absorb large seismic energy and adoptability to various foundation conditions (Abbas 2003; Abbas et al. 2003a). The behaviour of the rockfill materials is of considerable importance for the analysis and safe design of rockfill dams. Prototype rockfill material consists of maximum particle size (dmax) up to 1200 mm. Material with such a large particle size is not feasible to test in the laboratory. Therefore, some kind of modeling technique is often used to reduce the size of particles so that the specimens prepared with smaller size particles can be tested. Among all existing modeling techniques, the parallel gradation technique (Lowe 1964) is most commonly used. The behaviour of rockfill material is affected by particle size, shape, surface structure, mineral composition, individual particle strength, relative density, gradation (Gupta 2000). Therefore, it is very much essential to study the behaviour of rockfill material by simulating the exact field conditions so as to design the safe and economical rockfill dam.

The behaviour of alluvial and quarried rockfill material has been reported by number of researchers. Venkatachalam (1993), CSMRS 2010), Honkanadavar (2011) and CSMRS (2011) have performed one dimensional compression (Oedometer) tests on alluvial and quarried rockfill materials collected from different river valley projects from India and abroad. They concluded that stress-strain behaviour is non-linear and inelastic for both the materials. The behaviour of these materials is also dependent on dmax and shape of the particles (Venkatachalam 1993). This paper deals with the testing of quarried and alluvial rockfill materials obtained from Salma dam site, Afghanistan and Renuka dam site, Himachal Pradesh, India and study their behaviour under one dimensional compression test. Tests are carried out on 25, 50 and 80 mm dmax with the varying vertical stress ranging from 0.367 to 3.82 MPa. All the tests were conducted for 87% relative density. One dimensional compression (Oedometer) tests were carried out for all the modeled dmax of both project materials and studied their stress-strain behaviour. Using the stress-strain behaviour, deformation modulus, M, elastic modulus, E and Poisson’s ratio, ν are determined for both the materials and compared. The effect of dmax and shape of particles on coefficient of earth pressure, K0, voids ratio, e, coefficient of permeability, K and breakage factor, Bg were also studied on both alluvial and quarried rockfill materials.


EXPERIMENTAL INVESTIGATIONS AND DISCUSSION Material Used Quarried rockfill material from Salma dam site, Afghanistan and alluvial rockfill material from Renuka dam site, Himachal Pradesh, India have been considered. The rock type is Dolomite for Renuka and quartzite for Salma dam site. The dmax proposed in the construction of Renuka dam is 1000 mm and for Salma dam is 600 mm. Both materials have been modeled into three dmax (25, 50 and 80 mm) using parallel gradation technique as shown in Figs. 1a and 1b for Salma and Renuka dam respectively to test in the one dimensional compression test with specimen size of 1000 mm diameter and 600 mm height.

Fig. 1a: Prototype and modeled Grain Size Distribution Curves for Salma Dam Rockfill Material

Fig. 1b: Prototype and modeled Grain Size Distribution Curves for Renuka Dam Rockfill Material

Experimental Programme One Dimensional Compression (Oedometer) Test One dimensional compression tests were carried out on the modeled rockfill materials with σ1 varying from 0.637 to 3.82 MPa for 87% RD at Central Soil and Materials Research Station (CSMRS), New Delhi. In this test, during loading and unloading stages, vertical displacement is allowed where as the lateral displacement is restrained the specimen was tested under saturation condition. During loading the specimen was subjected to different stress levels ranging from 0.637 to 3.82 MPa. At each stress level, M, E, Void ratio, e and coefficient of earth pressure were determined for all the dmax of both materials. Stress-Strain Behaviour The stress-strain behaviour of both the modeled rockfill material tested with 87% RD has been studied (CSMRS 2010; 2011). From the stress-strain plots, it is observed that the behaviour is non-linear and inelastic for both modeled rockfill material. The axial strain at failure increases with increase in dmax for both materials. Typical stress-strain behaviour of quarried (Salma dam) rockfill material is shown in Fig.2.

Fig. 2: Relation between Axial Strain and vertical Stres (σ1) for Salma Dam Rockfill Material

Determination of Material Parameters The relation between applied vertical stress (σ1) and the observed vertical strain, ε1 is given as

a

a

11 Pa

1m1

⎟⎟⎠

⎞⎜⎜⎝

⎛ σ⎟⎠⎞

⎜⎝⎛

⎟⎠⎞

⎜⎝⎛=ε (1)

where, m is modulus number, a is the modulus of exponent, Pa is the atmospheric pressure and σ1 is the vertical stress. Plotting vertical stress v/s observed vertical strain for different loading stages of a dam, modulus number, m and modulus of exponent, a is determined (CSMRS 2010, 2011) for both materials and compared. From the comparison, it is observed that m and a increases with increase in dmax for both alluvial and quarried rockfill materials.


The deformation modulus, M, elastic modulus, E and Poisson’s ratio, ν are determined for both alluvial and quarried rockfill material using the following relationships (CSMRS 2010, 2011).

( )a1

a

1a P

PmM−

⎟⎟⎠

⎞⎜⎜⎝

⎛ σ= (2)

( )( )

( ) M1

211Eν−

ν−ν+= (3)

( )0

0

K1K+

=ν (4)

where, K0 is the coefficient of earth pressure (σ3/σ1). σ3 is measured with the help of sensors attached to the specimen cell. It is observed that the modulus of elasticity and elastic modulus increases with increase in vertical stress for both materials. It is also observed that for the same vertical stress, the Modulus of elasticity and elastic modulus increases with increase in dmax for both materials. Typical values of Modulus of elasticity and elastic modulus for 25 mm dmax of quarried rockfill material (Salma dam) are presented in the Table 1. The coefficient of earth pressure, K0 decreases with increase in dmax for both materials. Table 1: Deformation and Elastic Modulus for 25 mm dmax

of Quarried (Salma Dam) rockfill Material.

Vertical Stress (s1)

MPa

Deformation Modulus, (M) MPa

Elastic Modulus (E)

MPa

0.637 66.016 46.134

1.273 80.155 56.014

1.910 89.810 62.761

2.546 97.344 68.027

3.183 103.632 72.420

3.820 109.069 76.220 Voids ratio, e was determined for different vertical stress and dmax of both materials and studied its variation (CSMRS 2010, 2011). It is observed that the value of voids ratio is more for alluvial rockfill material than quarried rockfill material. This may be due to the presence of angular particles and higher interlocking in the quarried rockfill material. Voids ratio decreases with increase in σ1 and it increases with decrease in dmax. Typical variation of voids ratio with vertical stress and dmax for quarried (Salma dam) rockfill material is given in Fig. 3.

Permeability study was carried out to determine the coefficient of permeability, K at different stress level using constant head permeability test for both alluvial and quarried rockfill material to study the pervious nature of rockfill material (CSMRS 2010, 2011). K varies from 1.05e10-2 to 7.47e10-3 cm/sec for both materials. It is also observed that the K decreases with increase in σ1 and dmax for both project materials. Typical variation of K with voids ratio and dmax of quarried (Salma Dam) rockfill material is shown in Fig. 4.

Fig. 3: Variation between Vertical Stress, Voids Ratio and dmax for Quarried (Salma Dam) rockfill Material

Fig. 4: Variation between Coefficients of Permeability, Voids Ratio and dmax Breakage of particles study was carried out based on pre and post-tests grain size distribution curves for quantifying the breakage factor, Bg. The breakage is quantitatively expressed as breakage factor, Bg as proposed by Marshal (1965). The detail procedure to calculate the breakage factor, based on pre and post test grain size distribution is given by Honkanadavar (2010a). The variation of breakage factor v/s maximum particle size is plotted for all the dmax of both project materials (CSMRS 2010, 2011). From the analysis it is observed that the Bg increases with increases in dmax for both materials. The Bg varies from 6.5% to 11.0 % for dmax from 25 to 80 mm of both materials. Particle breakage is more for quarried (angular to subangular ahape) rockfill material as compared to alluvial (rounded to subrounded shape) rockfill material. This may be due to the presence of higher interlocking in the quarried rockfill material. Typical


variation of Bg with dmax for quarried (Salma Dam) rockfill material is shown in Fig. 5.

Fig. 5: Variation of Bg with dmax for quarried (Salma Dam) rockfill material CONCLUSIONS In the present study, quarried modeled rockfill material from Salma dam site, Afghanistan and Alluvial modeled rockfill material from Renuka dam site, Himachal Pradesh, India have been considered and tested. Field investigations were carried out and derived an average prototype gradation curve. Based on the average prototype gradation curve, modeled gradation curves were derived using parallel gradation technique with dmax of 25, 50 and 80 mm. These modeled materials were collected and tested in the laboratory under one dimensional compression (Oedometer) test for different vertical stress ranging from 0.637 to 3.82 MPa. All the tests were carried out for 87% relative density. Laboratory tests are carried out for saturated specimen of size 1000 mm diameter and 600 mm height under one dimensional compression test and studied the stress-strain behaviour under different stress level on different dmax. Following conclusions were drawn from the present study: • The deformation and elastic modulus increases with

increase in vertical stress and dmax for both alluvial and quarried rockfill materials.

• The coefficient of earth pressure decreases with increase in dmax.

• Voids ratio decreases with increase in σ1 and it decreases with decrease in dmax.

• The coefficient of permeability, K decreases with increase in σ1 and dmax for both alluvial and quarried rockfill materials.

• Breakage factor, Bg for quarried rockfill material is observed more than alluvial rockfill material. This may be due to the presence of angular particles and higher interlocking in the quarried rockfill material.

• Both alluvial and quarried rockfill material are pervious in nature.

• Size and shape of particles influence the behaviour of rockfill materials. Therefore, it is very much essential to carry out detail one dimensional compression test so as to

evolve the correct design parameters and use in the design for safe and economical rockfill dam.

ACKNOWLEDGEMENT The authors wish to thank all the staff of rockfill division, CSMRS for their help in conducting the tests. REFERENCES 1. Abbas, S. M. (2003), Testing and Modeling the

Behaviour of Riverbed and Quarried Rockfill Materials, Ph.D. Thesis, I.I.T. Delhi.

2. Abbas, S. M., Varadarajan, A. and Sharma, K. G. (2003), Prediction of Shear Strength Parameter of Prototype Rockfill Material, IGC-2003, Vol-I, pp. 5-8, Roorkee.

3. Lowe, J. (1964), Shear Strength of Coarse Embankment Dam Materials, Proc. 8th Int. Congress on Large Dams, Vol. 3, pp. 745-761.

4. Gupta A. K. (2000), Constitutive Modeling of Rockfill Materials, Ph.D. Thesis, I.I.T. Delhi.

5. Venkatachalam, K. (1993), Prediction of Mechanical Behaviour of Rockfill Materials, Ph.D. Thesis, I.I.T. Delhi.

6. CSMRS (2010), One Dimensional (Oedometer) Compression test report on quarried rockfill material from Salma Dam Project, Afghanistan.

7. Honkanadavar, N. P., Gupta, S. L. and Sanjeev Bajaj (2010), Deformability charactistics of quarried rockfill material, International Journal of Earth Sciences and Engineering, pp. 128-131.

8. CSMRS (2011), One Dimensional (Oedometer) Compression test report on alluvial rockfill material from Renuka Dam Project, Himachal Pradesh, India.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No.A-119)

PERMEABILITY AND CONSOLIDATION CHARACTERISTICS OF RECONSTITUTED KAOLIN CLAY

B. A. Mir, Assoc. Professor, Dept. of Civil Engg., NIT Srinagar, Kashmir, E-mail: [email protected], [email protected] Ashish Juneja, Assoc. Professor, Dept. of Civil Engg., I. I. T. Bombay - 400076, Mumbai, [email protected]

ABSTRACT: Kaolin clay has been widely used both in fundamental studies of soil behavior and in physical model tests. In this paper, permeability and consolidation behavior of reconstituted consolidated kaolin clay specimen wrapped with radial filter cage as side drain and installed with central sand column is examined in a series of conventional oedometer and triaxial consolidation tests. By using side drains, consolidation time reduced to less than 24 hours compared to more than 6-7 days required to consolidate the specimen without the side drains. However, by using sand drains, a consolidation time of about 100minutes compared to 24hours was required to consolidate the specimen with side drains. The test results show that permeability and compressibility characteristics of soft kaolin clay can be improved by choosing a suitable method of characterization. INTRODUCTION Soft clay deposits usually have a low bearing capacity and undergo excessive settlement over a long period of time. One of the important problems geotechnical engineers are often to deal with is the accurate and reliable measurement of permeability and compressibility characteristics of soft soils. The coefficients of permeability and consolidation are an essential soil parameters required for seepage, settlement, stability calculations and for predicting the rate of settlement of soft soils. Their importance is further increased in case of environmental problems, such as waste disposal, seepage of leachate and detrimental effects on the surrounding ground due to contamination. Therefore, it is necessary to improve the permeability and consolidation characteristics of soft soils before commencing construction activities in order to prevent ground contamination, excessive differential settlement and to reduce consolidation time. The permeability and compressibility of remoulded clays can be used as a frame of reference for the behaviour of a natural, undisturbed sample. Similar work to examine the mechanical properties of reconstituted clay has been carried out recently and in the past [1, 2, 3]. The properties of such type of soil are referred to as intrinsic, as these parameters were felt to be unique and inherent for a given soil type [4] and the behaviour of a reconstituted sample depends on the effort used in the reconstitution process. This paper examines the permeability and consolidation characteristics of reconstituted Kaolin clay. The results of the coefficients of permeability and consolidation in the vertical and horizontal directions are determined from the laboratory oedometer and triaxial consolidation tests on reconstituted kaolin clay. Conventional trixaial tests have been performed on 100mm diameter and 200mm long clay samples wrapped with filter paper as side drain and installed with sand column in centre to simulate the successive consolidation behaviour at different confining pressures varying from 50kPa to 575kPa. Side drains of filter king paper were used to

accelerate the rate of radial drainage of specimens. Kawakami [5] compared different filter paper cage positions in his tests and observed that by leaving a 5mm gap between the bottom porous stone and filter paper cage gave better results than if no gap was present. A similar procedure was adopted in the present study. By using side drains, consolidation time reduced to less than 24 hours compared to more than 6-7 days required to consolidate the specimen without the side drains. However, by using sand drains, a consolidation time of about 100minutes compared to 24hours was required to consolidate the specimen with side drains. The results show that the dissipation of excess pore water pressure occurs faster in the radial direction due to the greater coefficient of soil permeability in the horizontal direction and the reduced drainage path. Thus, the main function of filter paper and sand drains application is to accelerate soil consolidation by shortening the drainage path and activating radial drainage, thereby reducing post-construction settlement. Thus, the test results show that the permeability and compressibility characteristics can be improved by choosing a suitable method of characterization of soft kaolin clay. SAMPLES PREPARATIONS AND EXPERIMENTAL PROGRAM The experimental program consisted of 2 standard oedometer tests, 10 triaxial consolidation tests on homogeneous clay and 10 triaxial consolidation tests on composite clay (e.g. clay specimen installed with sand column in the centre). To erase the structure and load history of the clay, the natural soil was remolded at water content of 1.5 times the liquid limit into slurry form. De-aired clay slurry was consolidated in a 250mm diameter and 450mm long stainless steel cylindrical mould on the laboratory floor, first under its own self-weight and later under surcharge of 194 - to 352kN/m2 applied in stages on top of the clay surface using a custom designed pneumatic load frame. Upon completion of the 1-D consolidation, the block of clay was extruded and trimmed

B. A. Mir, A. Juneja

into three 100mm diameter specimens. Up to 3 triaxial specimens could together be prepared using this mould (Fig 1). For outward radial consolidation, a filter paper of which long strips are cut out in parallel is wound round the specimen. Its upper end is tucked over the upper porous disc by Lp/2.5 (5mm), while the lower portion of a depth of 5mm of the specimen is kept bare of the filter paper (Fig. 2) to avoid a “short circuit” between the back pressure and pore pressure systems [6]. For sand column, the specimens were held in split cylindrical moulds and a smooth PVC casing slowly pushed along its length to form a cylindrical hole at the centre. The hole was backfilled with fine sand (d50 = 0.3mm) compacted in layers using a pneumatic compactor (Fig. 2). Diameter of the sand column varied between 25- and 80mm in the specimens. This corresponds to an area replacement ratio, as [7] that ranges between 6.25- and 64%. After preparing samples, isotropic consolidated triaxial tests were performed on 100mm diameter and 200mm long specimens wrapped with filter paper as side drain and installed central sand column. The samples were first saturated and then consolidated isotropically [8]. Similarly, Oedometer (1-D consol) tests [9] were conducted on the specimens prepared from slurry consolidated samples. Table 1 shows the properties of the clay used in this study, and Table 2 shows details of the specimens prepared for triaxial testing. Fig. 1 Preparation of reconstituted specimen from clay slurry sample Fig. 2 Lay-out for radial cage for homogeneous clay sample and composite clay sample

Table 1 Properties of kaolin clay

Property Value Clay size (%) <0.002mm 75 Specific gravity 2.64 Liquid limit (%) 49 Plastic limit (%) 26 Shrinkage limit (%) 16 pH value 7.7 Maximum dry unit weight (kN/m3) Optimum moisture content (%)

14.75 27.5

SiO2 (%) 55.76 Al2 O3 (%) 38.63 Fe2O3 (%) 2.15

Table 2a Experimental program for homogeneous samples

Test No

1-D σv

(kN/m2)

Mean effective stress, p′ (kN/m2) Preconsoli-

dation pressure,

p′o (kN/m2)

OCRp'o / p'At the end

of 1-D loading

At the end of

consolidationP-1 194 141 500 500 1 P-2 211 153 300 300 1 P-3 277 160 50 160 3.2 P-4 277 159 30 159 5.3 P-5 277 165 100 165 1.7 P-6 352 267 50 267 5.34 P-7 352 253 100 253 2.53 P-8 211 153 100 153 1.53 P-9 211 153 300 300 1 P-10 211 149 50 149 3

Table 2b Experimental program for composite samples

*: σ′v =Vertical stress at end of 1D loading, #: ds

= Equivalent diameter of sand column, $: p′o =Preconsolidation pressure=

( )3

21 0kv +σ **: p′ = Mean effective stress at end of consolidation, In this table, σv is the 1-D vertical stress used for consolidating slurry in cylindrical mould on the laboratory floor.

Mean effective stress towards the end of 1-D loading was estimated using Ko (0.56) obtained from undrained shear test results.

OCR is defined as the ratio of preconsolidation pressure to the mean effective stress at the end of consolidation.

Test No

σ′v* (kPa) ds

# (mm) p′o $ (kPa) p′** (kPa) OCR

S1 404 25 285 100 2.85 S1 404 25 285 150 1.9 S1 404 25 300 300 1 S2 211 30 450 450 1 S2 211 30 200 200 1 S2 211 30 149 50 3 S3 211 40 375 375 1 S3 211 40 575 575 1 S3 211 40 149 75 2 S4 211 80 149 150 1

200

mm

314mm

Cut-off openings

Drainage strips

Porous stone Overlap

5mm

L0

Radial cage lay-out Porous stone

Rad

ial c

age

Sand column

Soil

sam

ple

Composite sample lay-out

5mm cut-off

Permeability and consolidation characteristics of reconstituted kaolin clay

RESULTS AND DISCUSSIONS

Isotropic triaxial consolidation of reconstituted specimens with outward radial drainage Triaxial consolidation tests were performed on 200mm long and 100mm diameter cylindrical specimens prepared from remoulded and reconsolidated commercially available kaolin clay. Change in volume of the cylindrical specimen during consolidation using side drains was measured using automatic volume change measuring apparatus. By using side drains, consolidation time reduced to less than 24 hours compared to more than 6-7 days required to consolidate the specimen without the side drains. The coefficient of consolidation and coefficient of permeability was determined at consolidation stage of triaxial tests using specific drainage conditions. Figure 3 shows the average degree of consolidation, Uavg plotted against time during isotropic consolidation.

Fig. 3 Variation of average degree of consolidation with time using filter paper as side drain The permeability versus effective stress relationship for specimen in triaxial cell with radial cage is shown in Fig. 4. Also superimposed in the figure are the findings from oedometer test results. Fig. 4 Effective stress vs coefficient of permeability and coefficient of consolidation with side drain

The Isotropic and 1-D consolidation lines for unloading and reloading are shown in Figure 5.

Fig. 5 Isotropic and 1-D consolidation lines Mean effective stress during loading in oedometer tests was calculated using the equation [10]:

)'sin.(p 'V

' φσ 6701−= (1)

where φ/ is the effective angle of friction obtained after post consolidated undrained tests. Likewise, p' on the unloading-reloading line (URL) was calculated using the equation [11]:

( )( )( ) ( ){ }⎥⎦

⎤⎢⎣

⎡−−

−+=

OCR'sin..exp'sin..

''p v lnφφ

σ1850930

1670330 (2)

The compression curve obtained from isotropic consolidation is a straight line, which can be represented by the equation:

/plnvv λλ −= (3)

where λν (=2.19) is the intercept of NCL with p′ = 1kN/m2

Using the test results, the critical state parameters were deduced as: λ = 0.10, κ = 0.01, where- λ is the slope of normal consolidation line, κ is the slope of swelling line. As can be seen from Fig.5, the slope of 1-D compression line and NCL are in good agreement, thereby lending further confidence in the selected value of λ. Isotropic triaxial consolidation tests on clay specimens installed with sand column (inward radial drainage) The procedure to back saturate the composite specimens was similar to that adopted for homogeneous specimens. The specimens were normally consolidated to p′ between 50 and 575kN/m2 in different tests. Table 2b also shows the overconsolidation ratio (OCR) of the specimens after isotropic consolidation. Consolidation due to horizontal drainage was computed using Barron’s [12] theory. Barron [12] arrived at the solutions for the excess pore pressure at any radial distance from the drain and at any time during

Mean effective stress, p' (kN/m2)

1.741.761.781.801.821.841.861.881.901.921.94

1.96

Spec

ific

volu

me,

v (=

1+e

)

1-D consol.-I1-D consol.-IIIsotropic consol.-P8Isotropic consol.P9

1 5 10 50 100 500 1000

10 100 1000

Mean effective stress, p' (kN/m2)

1x10-12

1x10-11

1x10-10

1x10-9

Coef

f. of

per

mea

bilit

y, k

(m/s)

k:Triax

Ch:Triax0

2

4

6

Coef

f. of

con

sol.

(m2 /

year

)

k:OedCv:Oed

0 100 200 300 400 500

Time (minutes)

1

0.8

0.6

0.4

0.2

0

Uav

g

TestStd. Clay Test: 300kPa P2:300kPa P6:500kPa P8:100kPa

}Clay Testswith radialcage

B. A. Mir, A. Juneja

consolidation. It is seen that by using sand drains, a consolidation time of about 100minutes compared to 24hours was required to consolidate the specimen with side drains (Fig.6). The results show that the dissipation of excess pore water pressure occurs faster in the radial direction due to the greater coefficient of soil permeability in the horizontal direction and the reduced drainage path. Fig. 6 Variation of average degree of consolidation with time using sand column The coefficient of horizontal permeability (kh) is computed from the following expression

wvhh mck γ××= (4)

where Ch is the coeff. of hor. Consolidation, mv is coeff. of volume change compressibility, and γw is the unit weight of water. The variation of kh and Ch with eff. Stress is shown in Fig.7. It is seen that permeability and consolidation characteristics have been considerably improved by using sand drain. Fig. 7 Effective stress vs coefficient of permeability with sand column

CONCLUSIONS The permeability and consolidation behaviour of kaolin clay was studied using 20 triaxial consolidation tests on slurry consolidated specimens with different drainage conditions. It is seen that specimens with sand drain (inward radial consolidation) is more effective to improve permeability and consolidation characteristics. The results show that the dissipation of excess pore water pressure occurs faster in the radial direction due to the greater coefficient of soil permeability in the horizontal direction and the reduced drainage path. REFERENCES 1. Rampello, S. and Silvestri, F. (1993), The stress-strain

behavior of natural and reconstituted samples of two overconsolidated clays, Proceedings of the Int. Symposium on Geotechnical Engineering of Hard Soils-Soft Rocks, Athens, A.A. Balkema, Rotterdam, Vol. 1, pp. 769-78.

2. Coop, M.R., Atkinson, J.H. and Taylor, R.N. (1995), Strength and stiffness of structured and unstructured soils,” Proceedings of the llth European Conference on Soil Mechanics and Foundation Engineering, Danish Geotechnical Society, Bulletin 11, Copenhagen, Vol. 1, pp. 55-62.

3. Siddique, A. and Safiullah, A.M.M. (1995), Permeability characteristics of reconstituted Dhaka clay, Jl. of civil engineering division, the Institution of engineers, Bangladesh, vol. CE23, No.1, pp. 103-115.

4. Burland, J. B.(1990), On the compressibility and shear strength of natural clays, Geotechnique 40, No. 3, pp. 329 – 378.

5. Kawakami, H. (1964). The measurement of horizontal coefficient of consolidation by a triaxial testing apparatus. Soils and Foundations 4(2), pp. 45-54.

6. Head, K. H. (1986), Manual of soil laboratory testing, Vol 3: Effective stress tests. London, pp. 743-1238.

7. Aboshi, H., Ichimoto, E., Enoki M. and Harada, K. (1979), Composer: method to improve characteristics of soft clays by inclusions of large diameter sand column. Proceedings of the International Conference on Soil Reinforcement: Reinforced Earth and other Technique, Paris, Vol. 1, 211-216 (1979).

8. ASTM D4767 – 04, Standard test method for consolidated undrained triaxial compression test for cohesive soils, Annual Book of ASTM Standards, Vol. 04.08, ASTM International, West Conshohocken, PA.

9. ASTM Standard D 2435-04, Standard test methods for 1-D consolidation properties of soils using incremental loading, Annual Book of ASTM Standards, Vol. 04.08, ASTM International, West Conshohocken, PA.

10. Wroth, C. P. (1984), The interpretation of in-situ soil tests, Geotechnique 34(4), pp. 449-489.

11. Mayne, P. W. and Kulhawy, F. H. (1982), Ko- OCR relationship, Journal of Geotechnical Engineering Division, ASCE (108), 851-872.

12. Barron, R. A. (1948). Consolidation of fine-grained soils by drain wells. Transactions ASCE, 113, 718-754.

0 20 40 60 80

Time (minutes)

1

0.8

0.6

0.4

0.2

0

Uav

g

TestStd. Clay Test: 300kPa S3: 300kPa S15:150kPa S18:75kPa

}Clay Testswith SandColumn

10 100 1000

p' (kN/m2)

1.0

k *

10- 9

(m/s)

Clay specimensinstalled withsand column

k v/s p'Ch v/s p'

20

24

28

32

36

Coef

f. of

con

sol.,

Ch

(m2 /

year

)

4.0

2.0

3.0

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No. A 120)

EXPERIMENTAL STUDY ON THE INFLUENCE OF MOULDING WATER CONTENT ON THE PERMEABILITY BEHAVIOUR OF COMPACTED SOIL

B. A. Mir, Assoc. Professor, Dept. of Civil Engg., NIT Srinagar, Kashmir, [email protected], [email protected] Insha, F. khan, B. Tech. Final Year Student, Dept. of Civil Engg., NIT Srinagar, Kashmir; [email protected] W. Ali, B. Tech. Final Year Student, Dept. of Civil Engg., NIT Srinagar, Kashmir; [email protected] I. A. Bhat, B. Tech. Final Year Student, Dept. of Civil Engg., NIT Srinagar, Kashmir; [email protected]

ABSTRACT: This paper presents the results of an experimental study carried out on clayey samples to investigate the influence of moulding water content on the permeability behaviour of the compacted clay. The permeability tests were conducted on clay samples compacted at maximum dry density and with varying water content (±OMC) using the falling head permeameter. The test results show a high influence of moulding water content on the permeability values due to highly flocculated soil structure on dry side of optimum compared to wet side of optimum. The permeability tests were also conducted with varying dry density (±MDD) and at OMC, which resulted in pronounced decrease in permeability upto MDD. Beyond MDD, the variation is less sensitive. The discussion leads to the possible conclusion that to achieve minimum possible coefficient of permeability, there is a need for soil water content at placement to be at or slightly above OMC and compacted at or beyond MDD. INTRODUCTION Permeability is an important engineering property of soil and is essentially used as a quality control parameter for for seepage, settlement and stability calculations [1]. Permeability of fine-grained soils is also one of the key factors in predicting contaminant migration under the ground. Contaminant flow quantities and patterns can only be estimated accurately with the reliable values of permeability. As fine-grained soils are used effectively and extensively as impervious boundaries in geotechnical projects, predicting accurate permeability can evaluate and minimize the future hazardous potential. Although, it would be ideal to determine the permeability in the field due to several required conditions such as getting a representative sample, relatively large scale of a test, and non-homogeneity and anisotropy of natural soil sample, laboratory tests of permeability, especially in compacted soils are widely used because they can be done within less time and with less expense. There have been many studies that indicate the correlation of values between field tests and laboratory tests [2, 3, 4]. Furthermore, with laboratory tests, the study on the effect of changes in design parameters and the interaction between the soil and various contaminants can be also performed more conveniently [3, 5]. This paper presents the results of an experimental study carried out on clayey samples to investigate the influence of moulding water content and dry density on the permeability behaviour of the compacted clay. The permeability tests were conducted on clay samples compacted at maximum dry density and with water content as optimum moisture content (OMC), at the dry side of optimum (-2.5%OMC, -5%OMC) and at the wet side of optimum (+2.5%OMC, +5%OMC) using the falling head permeameter. The test results show a high influence of moulding water content on the permeability values. It has been seen that for the same void ratio, a specimen compacted

dry of optimum and saturated results in a higher permeability (3.24*10-9m/s), whereas the specimen wet of optimum has a lower permeability (5.31*10-10m/s). This is because the soil structure compacted on the dry side of optimum is randomly oriented (flocculated) than in the wet-side of optimum, which has a more parallel particle orientation (dispersed), mainly characterised by long strings with face-to-face contacts between the particles. The results show that the change in structure with water content appears to be drastic around the moisture state of OMC where the permeability falls from 10-7 to 10-9m/sec. Beyond OMC, it is seen that there is a small change in permeability values. Mitchell et al. [1] also reported a high influence of moulding water content on the permeability values and showed that the lowest permeability occurs at water contents slightly (2-4%) in the wet side of optimum water content. The permeability tests were also conducted with varying dry unit weight (±MDD) and at constant water content (OMC). It is seen that there is a pronounced decrease in permeability upto MDD and beyond MDD, the permeability is almost unaffected by the changes in dry unit weight. Therefore, the main objective of this study is to investigate the effect of variation of moulding water content (both dry side & wet side of optimum) and void ratio on the permeability behaviour of compacted fine-grained soil. MATERIALS USED For the present work, disturbed and undisturbed soil samples were collected from a site at three locations in Lal Bazar where ground was already excavated to about a depth of 12ft (4mts.) for the purpose of laying underground drains. All the tests were carried out as per the relevant Indian Standards [6-11]. The physical properties of materials used are listed in Table 1. Particle size distribution analysis revealed that the soil samples contained more than 70% silt+clay content. The particle size distribution curves are shown in Fig. 1. It is seen

B. A. Mir, Insha F. Khan, Willayat Ali, I. A. Bhat

that test specimens consists of well graded clayey silt with small fractions of fine sand. The values of consistency limits are given in Table 1. The values of liquid limit and plastic limit are useful in the classification of soils. Based on test results, the soil is classified as clayey silt of medium plasticity (MI). The compaction characteristics i.e. optimum moisture content (OMC) and the maximum dry density (MDD) of the soil samples were determined using the Proctor standard compaction test. Figure 2 shows the Standard Proctor’s compaction curves for the test materials. Table 1 Physical properties of soil samples

Property S-1 S-2 S-3 Gravel size (%) Sand size (%) Silt+clay (%) Clay size (%) Coeff. of uniformity, Cu Coeff. of curvature, Cc

0 2

73 25 29 1.5

2 6 75 17 25 2.8

37 74 16 20 1.5

Suitability factor, Sn >50 >50 >50Specific gravity 2.67 2.63 2.65Liquid Limit (%) Plastic Limit (%) Plasticity Index (%) PlasticityIndex, PIA (%) Shrinkage Limit (%)

36.5 24.9 11.6 12.0 13.2

35.5 25.4 10.1 11.30 14.8

35.125.1 10.0 11.0 12.4

Classification MI MI MIFree swell ratio (%) 1.1 0.9 0.9Activity, Ac 0.42 0.66 0.67MDD (kN/m3) 17.4 17.60 18.0OMC (%) 19.5 19.00 18.0In-situ coeff. of permeability (m/s) 1.88*10-9 2.41*10-9 3.32*10-9

Coeff. of permeability from sieve analysis (m/s)

1.22*10-9 2.16*10-9 4.86*10-9

Fig. 1 Particle size distribution curves for under ground drainage soil at Lal Bazaar

Fig. 2 Standard Proctor’s compaction curves for under ground drainage soil at Lal Bazaar EXPERIMENTAL PROGRAM Permeability tests were conducted in rigid-wall compaction mould permeameters using the falling head method. A falling head permeameter is more popular for fine-grained soils with low permeability. This method can measure much smaller volumes of flow. Within a time period, t, a decrease of water level in a volumetric tube with the cross-sectional area, a, from the height h1 to h2 is measured. Tests were run at a constant temperature of 20±1°C. Specimens were placed in a cell and saturated from the bottom until water ponded at the top. Then, the coefficient of permeability, k (m/s) is determined by using the Darcy’s equation:

( ) 2

1

21 hhLn

ttAaLk−

= (1)

where, a = cross-sectional area of the standpipe (m2); L = length of the soil specimen (m); A = cross-sectional area of the specimen (m2); t1 = initial test time (min); t2 = final test time (min); h1 = initial hydraulic head (m) at time t1; and h2 = final hydraulic head (m) at time t2. All the permeability tests were conducted in two stages. The detailed experimental program is given in table 2.

(a) (b) Table 2 Laboratory experimental program

Stage-I: At MDD and varying water content dry & wet side

of OMC

Stage-II: At OMC and varying dry unit

weight (±MDD)

w (%) γd (kN/m3)

w (%) γd (kN/m3)

-5% OMC =14.0MDD= 17.36 kN/m3

OMC =

19%

0.9* MDD-2.5% OMC =16.5 0.95* MDDOMC (19%) =19.0 At MDD +2.5% OMC =21.5 1.05* MDD+5% OMC =24.0 1.10* MDD

0.0001 0.001 0.01 0.1 1 10 100

Particle size (mm)

0

20

40

60

80

100

Perc

enta

ge fi

ner (

%)

S-1S-2S-3

Size S-1 S-2 S-3 D10 = 0.00045 0.0006 0.0009 D15 = 0.0009 0.0018 0.0021 D20 = 0.0015 0.0028 0.0034 D30 = 0.0030 0.0050 0.0050 D50 = 0.0085 0.0120 0.0120 D60 = 0.0130 0.0150 0.0180 D85 = 0.0240 0.0400 0.0500 Cu = 29 25 20 Cc = 1.5 2.8 1.5 Sn = > 50 >50 >50 Sn = Suitability factor for back-fill material (>50: unsuitable)

Clay size Silt size Sand size Gravel size

0.002 0.075 0.425 4.75(mm)

Line of optimums

0 5 10 15 20 25 30

Water content (%)

15

15.5

16

16.5

17

17.5

18

18.5

19

Dry

den

sity

(kN

/m3 )

S-1S-2S-3ZAV: G=2.65

Experimental study on the influence of moulding water content on the permeability behaviour of compacted soil

RESULTS AND DISCUSSIONS Permeability characteristics Permeability is one of the factors that determine shear strength and is a measure of water or air movement through the soil. Water content and permeability are interrelated and have a significant influence on the suitability of a soil material for use as a fill, subgrade, or foundation material. Water content (w) is one of the most important factors affecting the properties and behaviour of fine-grained material. In the present study, the permeability tests were conducted on in-situ samples collected from the project site. The coefficient of permeability was also computed from particle size (D10) using the following eqn:

210D*Ck = (2)

where, C = 4-8 (1/mm-s) for very fine well graded or with appreciable fines less than 75micron sieve; D10 = particle size (mm) such that 10% of the soil is finer than this size. The test results are summarized in table 1. The permeability tests were also conducted on compacted samples as per experimental program (Table 2) using falling head permeability method and the test results are described as below. Effect of moulding water content on the permeability behaviour of compacted fine-grained soil In order to investigate the effect of moulding water content on permeability of compacted soil, samples of clayey-silt were prepared using static compaction over a range of water contents at constant density (MDD) as shown in Fig. 3. Fig.3 Variation of coeff. of permeability with varying water content (dry of optimum, at OMC and wet of optimum) at constant maximum dry unit weight

As can be seen from Fig. 3, there is a slight increase in permeability with an increase in moulding water content on dry side of optimum. This is understandable due to the fact that initially the range of water content on dry side of optimum is considerably less with highly flocculated soil structure thereby increasing the permeability. Similar results have also been reported by Mitchell et al. [1, 12, 13]. But as the moulding water content is increased, the coefficient of permeability decreases and the specimen prepared wet of the optimums had a permeability nearly one order of magnitude less than the permeability of samples prepared dry of optimum. This can be argued by the fact that on wet side of optimum, the soil has the tendency of increased dispersion with increasing water content. Due to higher degree of dispersion, the soil gets compacted easily and permeability is decreased. Hence, this indicates that compaction dry of optimum renders a more open structure with much larger voids than in the wet-side compacted sample with a more parallel particle orientation. However, samples compacted in the laboratory are not likely in the same state of compaction with soils compacted in the field. Therefore, the variables that influence the structure of laboratory-compacted samples should be carefully controlled (14). Effect of varying dry unit weight on the permeability behaviour of compacted fine-grained soil In order to investigate the effect of varying dry unit weight on permeability of compacted soil, samples were compacted at different densities but at constant moulding water content (OMC) as shown in Fig. 4.

Fig. 4 Variation of coeff. of permeability with dry unit weight and void ratio at optimum moisture content It is seen that there is the pronounced decrease in permeability with an increase in dry density at a given water content (OMC) upto MDD. Beyond MDD, the variation is

14 16 18 20

Dry unit weight, γd (kN/m3)

1E-010

1E-009

1E-008

1E-007

Coef

f. of

per

mea

bilit

y, k

(m/s)

γd v/s k

MDD

e v/s k

0.3 0.4 0.5 0.6 0.7

Void ratio (e)

12 14 16 18 20 22 24 26

Water content (%)

1E-010

1E-009

1E-008

Coef

f. of

per

mea

bilit

y, k

(m/s)

Water content v/scoeff. of permeability

OMC

B. A. Mir, Insha F. Khan, Willayat Ali, I. A. Bhat

less sensitive. Similar trend is observed for variation of coefficient of permeability with void ratio at optimum water content as shown in Fig. 4. It can be seen that permeability decreases with decreasing void ratio and e-log k relationships are non-linear. Similar results have also been reported by other researchers [15]. The permeability is also known to vary with e2, e2/(1+e) and e3/(1+e) as shown in Fig. 5. Thus, for a given soil, the greater the void ratio, the higher is the value of the coefficient of permeability. Fig. 5 Variation of coefficient of permeability with e2, e2/1+e and e3/1+e CONCLUSIONS Based on the experimental findings of this research the following conclusions can be drawn: 1. The soil sample is dominantly clayey silt of medium

plasticity. 2. The values of coefficient of permeability for compacted

clay soil studied herein are sufficiently low as to make seepage problems non-critical in the drainage design.

3. It is observed that the permeability is dependent on the moulding water content; the density and the void ratio; and the compactive effort.

4. Initially, there is a slight increase in permeability with an increase in moulding water content on dry side of optimum. But as the moulding water content is increased, the coefficient of permeability decreases and the specimen prepared wet of the optimums had a permeability nearly one order of magnitude less than the permeability of samples prepared dry of optimum.

5. Compaction dry of optimum renders a more open structure with much larger voids than in the wet-side compacted sample with a more parallel particle orientation.

6. There is the pronounced decrease in permeability with an increase in dry unit weight at a given water content (OMC) upto MDD. Beyond MDD, the variation of permeability with varying unit weight is less sensitive.

FUTURE SCOPE The effect of varying compactive effort on the permeability behaviour of compacted clays can be studied in detail. This will give the exclusive effect of compactive effort on the soil permeability and hence on the soil fabric. Since the soil sample moulded at the water content corresponding to maximum permeability is having highest degree of flocculation, the effect on permeability will be a measure of degree of flocculation or degree of dispersion of the soil fabric corresponding to changing compactive effort.

REFERENCES 1. Mitchell, J.K., Hooper, D.R., and Campanella,

R.G.(1965), Permeability of compacted clay, Jl. of Soil Mech. Fd., Divn., ASCE, Vol. 91, No. SM4, pp. 41-65.

2. Benson, C.H., Zhai, H. and Wang, X. (1994), Estimating Hydraulic Conductivity of Compacted Clay Liners, Jl. of Geotech. Engg, Vol. 120, No.2, pp. 366-387.

3. Daniel, D.E., Anderson, D.C. and Boynton, S.S. (1985), Fixed-wall versus flexible-wall permeameters, Hydraulic Barriers in Soil and Rock, ASTM STP 874, ASTM, Philadelphia, pp. 107-126.

4. Day, S.R. and Daniel, D.E. (1985), Hydraulic conductivity of two prototype clay liners, Jl. of Geotech. Engg., Vol. Ill, No.8, pp. 957-970.

5. Houston, S.L. and Randeni, J.S. (1992), Effect of clod size on hydraulic conductivity of compacted clay, Geotech. Testing Jl., Vol. 15, No.2, pp. 123-128.

6. IS: 2720-part 1 (1980), Indian Standard Code for preparation of soil samples, Bureau of Indian Standards, New Delhi.

7. IS: 2720-part 4 (1985), Method of test for soils: Determination of grain size distribution, Bureau of Indian standards, New Delhi.

8. IS: 2720-part 5 (1985), Method of test for soils: Determination of Atterberg limits, Bureau of Indian standards, New Delhi.

9. IS: 1498 (1970). Method of test for soils: Classification and identification of soils for general engineering purposes, Bureau of Indian standards, New Delhi.

10. IS: 2720-part 7 (1980), Method of test for soils: Determination of compaction characteristics, Bureau of Indian standards, New Delhi.

11. IS: 2720-part 17 (1986), Method of test for soils: Determination of permeability, Bureau of Indian standards, New Delhi.

12. Lambe, T. W. (1958), The engineering behaviour of compacted clay, Jl. of Soil Mech. and Fd. Divn., Vol. 84, No. SM 2, pp. 1655-1 1655-35.

13. Dunn, R.J. and Mitchell, J.K. (1985), Fluid conductivity testing of fine-grained soils, Jl. of Geotech. Engg., Vol. 110, No. 11, November, pp. 6481665.

14. Benson, C.H., and Daniel, D.E. (1990), Influence of clods on hydraulic conductivity of compacted clay, Jl. of Geotech. Engg., Vol. 116, No.8, pp. 1231-1248.

15. Tavenas, F., Jean, P., Leblond, P. and Lerouell, S. (1983), The permeability of natural soft clays, Can. Geotech. J., Vol. 20(4), pp. 645-660.

0 0.2 0.4 0.6

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1E-010

1E-009

1E-008

1E-007

Coef

f. of

per

mea

bilit

y, k

(m/s)

e2 v/s ke2/1+e v/s ke3/1+e v/s k


LABORATORY STUDY ON PERMANENT DEFORMATION BEHAVIOUR OF SOIL ADMIXED WITH BAGASSE ASH UNDER REPEATED TRIAXIAL LOADING

Aditya Kumar Anupam Research Scholar, Civil Engg. Dept., IIT Roorkee Email: [email protected] Praveen Kumar Professor of Civil Engineering Department, IIT Roorkee and Email: [email protected] G.D. Ransinchung R.N. Assistant Professor of Civil Engineering Department, IIT Roorkee Email: [email protected] ABSTRACT: This work presents the laboratory study conducted on bagasse ash as soil stabilizer. Using bagasse ash, the deformation behaviors of subgrade soil-mixtures under repeated triaxial loading was conducted. Resilient strain, permanent strain and resilient modulus were ascertained at different moisture contents and confining pressures for a constant deviator stress level. Admixing of bagasse ash reduces the maximum dry density but increase the water demands. The variations of elastic deformation and resilient modulus were significantly influenced by bagasse ash contents. Better performances were observed for bagasse ash admixed soil-mixtures irrespective of number of load applications. The results obtained from the laboratory confirm the potential use of subgrade soil admixed with bagasse ash with respect to the permanent deformation. KEY WORDS: Permanent deformation, elastic deformation, resilient modulus and repeated triaxial loading. INTRODUCTION Flexible pavements commonly consist of upper asphalt layers over base and/or sub-base which are together compacted over a suitable soil subgrade. As it is well known, soil subgrade provide the most important structural element in road pavements because in most cases, when the soil subgrade does not have sufficient bearing capacity, the rutting phenomenon takes place mainly in the granular base and subbase layers causing progressive fatigue cracking of bituminous layers. In the case of clayey soil, the interfacial strength is low resulting in an early failure of the interface before the full strength of bagasse ash (stabilizer) can be mobilized. Thus, the strength of bagasse ash may be largely underutilized due to the failure of the interface. The interaction between the soil and bagasse ash is even more important in the case of the admixed soil structures subjected to cyclic loads such as in pavement subgrade. The strength of soil under cyclic loading depends on a number of factors such as the number of cycles, frequency of loading and the magnitude of the cyclic stress [1]. In particular, the global response of subgrade results in resilient and permanent deformations when subjected to repeated loading. Resilient deformations are related to the stiffness characteristics of the material that should be sufficiently high in order to avoid the fatigue cracking of top layer. On the other hand, the gradual accumulation of permanent deformations, although they are very small during each loading cycle, could lead to deteriorate the structure due to excessive rutting. Therefore, sufficient bearing capacity pavement subgrade should experience accumulation of permanent deformations that during its service life will eventually cease resulting in a stable and basically resilient response in order to avoid premature failure. A review of these issues [2, 3] described how observations from repeated load triaxial testing had demonstrated the existence of a threshold deviator stress level, below which the magnitude of accumulated plastic shear strain is quite small. The same phenomenon is apparent in the theoretical concept

of shakedown, derived from structural engineering and first applied to pavements [4]. In this context, the main aim of this research study is focused on a better understanding of the permanent deformation behaviour under repeated loading of soil subgrade admixed with bagasse ash materials. The addition of fly ash and bagasse ash for different soil types increases CBR linearly. However, in case of soil-BA mixture the rate of increment is nearly constant after 25 to 30 % of ash content [5]. This research analysis is based on the results of a laboratory investigation carried out by means of repeated load triaxial tests. EXPERIMENTAL PROGRAM Materials Soil Clay of medium compressibility (A-7-6) soil is used for this study. The index properties such as liquid limit, plastic limit, plasticity index and other important soil properties as per AASHTO and United States soil classification systems are presented in Table 1. Table1 Physical Properties of Soil

Properties Values Optimum moisture content (%) 17 Dry density (gm/cc) 1.85 Specific gravity 2.674 Liquid limit (%) 46 Plastic limit (%) 21 Plasticity index 25 Unified soil classification CL AASHTO soil classification A-7-6

Bagasse ash (BA) Bagasse ash is an agricultural by-product of sugar manufacturing. When juice is extracted from the cane sugar,

Aditya Kumar Anupam, Praveen Kumar & G.D. Ransinchung R.N. the solid waste material is known as bagasse. After the extraction of all economical sugar from sugarcane, about 40-45% fibrous residue is obtained, which is reused in the same industry as fuel in boilers for heat generation leaving behind 8 -10 % ash as waste, known as sugarcane bagasse ash (SCBA). This ash is considered as waste and disposed in an environmentally debilitating manner. Mill-fired BA was collected from Laxmi Sugar Mill, Roorkee, India. The uncontrolled boiler fired ash was black in colour due to 8.4% carbon content. Based on the Thermo Gravimetric Analysis (TGA), the carbon weight loss of BA is higher at 550°C. Beyond this temperature, the loss of weight of carbon is nearly constant. This burning process could bring down the carbon content to 3.4%. Hence, for the present study, BA heated at temperature of 550°C was used as a stabilizing agent for further laboratory investigations. The physical and chemical properties of bagasses ash are given in Table 2. Table 2 Physical and Chemical Properties of bagasse ash Physical Properties Chemical Properties

Property Value Constituents % by weight

Type Class F Ignition loss 5.7 Specific gravity 1.85 SiO2 66.12 Liquid limit 48 Al2O3 12.46 Plastic limit Non-

plastic Fe2O3 4.12

Optimum moisture content (%)

28 CaO 6.84

Maximum dry density (g/cm3)

1.4 MgO 3.18

Specific surface (cm2/g)

9430 Na2O3 1.3

Lime reactivity (kg/cm2)

42 K2O 2.05

Loss on ignition (%) 5.7 SO3 0.94 Compaction The major parameters analyzed in this study are effect of bagasse ash content on the deformation behavior of soil of different properties. Soil admixed with bagasse ash of different percentage viz. 5, 10, 15, 20 and 25% were added to determine the engineering behavioral aspects. Test was performed at optimum moisture content conditions and also extended to the pure soil specimen as reference tests to make comparison between the strength behaviors of bagasse ash admixed soil and to that of pure soil. Each admixed soil sample was compacted at Maximum Dry density (MDD) and Optimum Moisture content (OMC). In order to obtain these parameters, heavy compaction test was employed for the mentioned mixture proportions as per IS: 2720 (Part 8). The results for OMC and MDD for soil admixed with bagasse ash are as shown in Table 3. Test procedure A repeated load triaxial apparatus, consisting essentially of a main cyclic load device and a removable pressure chamber, allowed a constant confining pressure and a cyclic axial load to be applied on the specimen through the test. During the

test, two LVDTs measured the axial deformations at the top of the specimens. Table 3 OMC and MDD soil admixed with bagasse ash

Percentage Bagasse Ash OMC (%)

MDD (gm/cm3)

0 17 1.89 5 19 1.78

10 20 1.74 15 21 1.68 20 22 1.62 25 23.5 1.57

Load and deformation data were then stored by a proper data acquisition system. All tests were performed in undrained conditions. Cyclic triaxial strength tests were conducted under undrained conditions to simulate essentially undrained field conditions during moving traffic. The cyclic loading generally causes an increase in the pore-water pressure in the specimen, resulting in a decrease in the effective stress and an increase in the cyclic axial deformation of the specimen. Repeated triaxial were conducted on sample of sized 51mm diameter and 102mm height in conventional triaxial cell. The repeated compressive deviator stresses were applied at two different confining pressures. The frequency of load application in all tested were kept 70 cycle per minute; this was fixed based of traffic density [6]. The loads were applied upto 10,000 cycles and behavior of various parameters such as resilient strain, permanent strain and resilient strain was observed at different cycles. In the present study, the cyclic triaxial tests were conducted as per ASTM D5311 on soil and soil admixed with bagasses ash of different percentage viz. 5, 10, 15, 20 and 25%. Stress-strength ratio in this study is called deviator stress levels (DSL) and can be defined as the ratio of the σd of repeated load triaxial test (RLTT) to the soil strength obtained from undrained triaxial (σs). Three levels of σ3 (50, 100 and 150 kPa) at 0.50 DSL were applied. TEST RESULTS AND DISCUTION TGA study From TGA study, it has learnt that maximum carbon weight loss of BA was maximum at 5500C temperature and could bring down the carbon content upto 3.4 percent. Beyond this temperature, the weight loss was very negligible. MDD and OMC Fig.3 shows that optimum moisture content increases with the increase of BA content. This increase was approximately linear with the ash content. In contrast, admixing of BA lead to reduction in dry density (Table 3). This reduction was more significant for higher percentage BA admixing. This reduction maintained almost linear with the ash content (Table 3). This phenomenon is likely attribute to lighter unit weight of BA than untreated soil. Resilient Strain Resilient strain is affected by confining pressure applied. At the same deviator stress, resilient strain decrease with the

Laboratory study on permanent deformation behaviour of soil admixed with bagasse ash under repeated triaxial loading

increase in confining pressure (Figs. 1-3). It was also observed that the resilient strain increases with no. of load cycle irrespective of confining pressure applied. Marked improvements in deformation was distinctly noticeable for BA admixed soil samples in comparison to the performance of untreated soil sample. This improvement is most probably due to formations of cementitious material within the mixture system resulting in better densification of admixed soil sample.

Fig. 1 Resilient Strain at 50kPa confining pressure and 0.50 DSL

Fig. 2 Resilient Strain at 100kPa confining pressure and 0.50 DSL

Fig. 3 Resilient Strain at 150kPa confining pressure and 0.50 DSL Permanent Strain Figs.4-6 shows the variation of permanent strain with applied deviator stress and number of load cycle at three different confining pressures. It was observed that the permanent strain increases with the increase in no. of load cycle applications, because each load application contributes a small increment

to the accumulation of strain. Inclusion of BA improves the permanent strain considerably (Figs. 4-6). This improvement was more significant for soil samples containing higher percentage content of BA. This phenomenon helps in predicting the performance of pavement deformations and fatigue characteristics. Based on the present investigation, BA admixed soil samples show lesser permanent strain and this signifies that admixing BA would not only improve the deformation resistance but also would increase the fatigue cracking resistance.

Fig. 4 Permanent Strain at 50kPa confining pressure and 0.50 DSL

Fig. 5 Permanent Strain at 100kPa confining pressure and 0.50 DSL

Fig. 6 Permanent Strain at 150kPa confining pressure and 0.50 DSL Resilient Modulus Resilient response of subgrades can be quantified by resilient modulus (Er) that is a stress-strain relationship like modulus of elasticity. However, Er is determined from a repeated load triaxial compression test and is based on only the recoverable

Aditya Kumar Anupam, Praveen Kumar & G.D. Ransinchung R.N. portion of the strain [7, 8]. It is expressed as the ratio of axial repeated deviator stress (σd) to the recoverable axial strain (σr) and considered as an indication of load-carrying capacity of the subgrades. Figs. 7-9 show the trend of resilient modulus with number of load cycles and different percentage of bagasse ash. The resilient modulus decreases with the increase of number of cycles and confining pressures. When the percentage of BA increases, the resilient modulus also increases. The increase of resilient modulus was noted as a function of BA content. Higher resilient modulus was observed for higher percentage content of BA.

Fig. 7 Resilient Modulus at 50kPa confining pressure and 0.50 DSL

Fig. 8 Resilient Modulus at 100kPa confining pressure and 0.50 DSL

Fig. 9 Resilient Modulus at 150kPa confining pressure and 0.50 DSL CONCLUSION 1. Based on TGA study, ideal temperature to remove

maximum possible carbon content is 5500C. Beyond this temperature, the loss of carbon content was negligible.

2. Admixing of bagasse ash lead to reduction in maximum dry density of soil-mixtures but increase the optimum moisture content. The increased in moisture content was attributed to higher surface area possessed by bagasse ash and reduction in maximum dry density is probably due to cation exchange resulting in particle floculation and aggregation.

3. The tests fulfilled with the Clay of medium compressibility soils used in this work proved that the total permanent strain was highly influenced by the stress state and the bagasse ash content of the test specimens.

4. Admixing of bagasse ash showed significant improvements in resilient strains and permanent strains in comparison to untreated soil. This decrease in strain is attributed to formation of cementitious material within the mixture system resulting in better densification of admixed soil sample. From the present study, it is preliminary concluded that admixing of bagasse ash would significantly improve the pavement life.

5. Resilient modulus is an excellent measurement under pavement materials service condition in the pavement structure. When the percentage of bagasse ash increases, the resilient modulus also increases.

6. From this study, the bagasse ash may be effectively utilized in soil to get improvement in rut depth at maximum number of load cycles and thus improvement in the bearing capacity. Baggase ash addition in soil can also be effectively used as the subgrade materials for the roads, back filling, and improvement of soil bearing capacity of maximum no of load repetition.

REFERENCES 1. Seed, H.B., Chan, C.K., 1966. “Clay strength under

earthquake loading conditions.” Journal of Soil Mechanics and Foundry Engineering Division, ASCE 92, 53–78.

2. Brown, S.F. (1996) “36th Rankine Lecture: Soil Mechanics in Pavement Engineering Geotechnique” 46(3): 383–426.

3. Brown, S.F. (2004) “Design considerations for pavement and rail track foundations.” Proc. Inst. Seminar on Geotechnics in Pavement and Railway Design and Construction, Athens: 61–72.

4. Sharp, R.W. & Booker, J.R. (1984) “Shakedown of pavements under moving surface loads.” Journal of Transportation Engineering, ASCE 110(1): 1–13.

5. Anupam, A. K., Kumar P. and G. D. Ransinchung. R. N. (2012) “A Comparative Study Of Sugar Cane Bagasse Ash & Fly Ash For Use In Pavement Construction” International Conference on Highway Engineering April 18-20, Bangkok, Thailand.

6. Kumar, P. and Singh, S. P. (2008). “Fiber-reinforced fly ash subbases in rural roads.” Journal of Transportation Engineering ASCE, Vol. 134 (4), 171-180.

7. Yoder EJ, Witczak MW (1975). “Principles of pavement design” A Wiley-Interscience Publication, Newyork, USA.

8. Elliot RP, Thornton SI (1988). “Resilient modulus and AASHTO pavement desing.” Trans. Res. Record.1192: 1-7.


GEOTECHNICAL CHARACTERIZATION FOR NEW SPILLWAY AT KABRAI DAM BY GEOPHYSICAL SURVEY

Birendra Pratap, Scientist ‘B’ CSMRS, New Delhi, e-mail: [email protected] Alex Varughese, Scientist ‘C’ (I/c) CSMRS, New Delhi, e-mail: alexvarughese@ gmail.com ABSTRACT: Engineering geophysical methods are widely accepted for deciphering subsurface stratigraphy including for inferring the quality of rock. This information is required for deciding and designing the foundations for the civil structures economically. This project envisages the construction of new spillway at pre occupied by old spillway for Kabrai dam, arjun sahayak project, district, Mahoba, Uttar Pradesh. The geophysical survey comprises seismic refraction and electrical resistivity were conducted in consultation with executive engineer, Maudaha dam construction division-I, Mahoba, U.P. with project engineers. The aim of the survey was to provide and evaluate the thickness and quality of the overburden and characteristics of the bedrock profile, along the surveyed seismic lines. The survey results are presented in P-wave velocity profiles with reduced level (RL). These profiles may assist to decipher the subsurface stratigraphy, bedrock quality and weak zone in bed rock with depth in the area of interest. The geoelectrical sounding (GS) delineate the unconsolidated formations overlying on the basement rock. It was found that the P- wave velocity of the first layer is 700/1000- 1400 m/sec, which is the general P- wave velocity of overburden and highly weathered rock. The P- wave velocity of the second layer is 4900 -5300 m/sec. From geophysical interpretation it can be confirmed that the first layer (top surface layer) consists of overburden and highly weathered rock. Second layer consists of bed rock of very good quality. The depth of the first layer varies from 5 m to 12 m. It was also observed that the P- wave velocity of the second layer is 2500-4100 m/sec, which is the P- wave velocity of weak zone in bed rock. This paper presents the results of seismic refraction survey and geoelectrical sounding, discussion thereof and conclusions/recommendations arrived at. The geophysical methods employed at the project site proved to be very useful for geotechnical site characterization for construction of new spillway quickly and economically. Key words: Seismic refraction, geoelectrical sounding, non-destructive test 1. INTRODUCTION There is an increasing requirement for geophysical surveys carried out during geotechnical investigations to provide direct information about bed rock profile and rock quality or other geotechnical parameters for deciding and designing the foundations of civil structures of any project. Since spillway in a project is an important structure, the foundation should be geotechnical sound to sustain the force of water current and high stress of water pressure of the reservoir. The arjun sahayak project envisages to diverting 73.6 cumec water from Lachura dam to arjun reservoir through 38.6 km. long feeder canal and 62.32 cumec water from arjun reservoir to Kabrai dam through 31.3 km long feeder canal. The Kabrai dam was built in the year 1995 across Magaria and Kulahari rivers between Rachiya and Dharaun hills. The existing earthen dam is 15.24 m high and 2.24 km long. In order to accommodate the additional water in Kabrai dam, it is proposed to raise the height of the existing Kabrai dam by 12.0 m. At present, the Kabrai dam has a spillway with a flankscape on its left side. Raising the earthen section of the dam by 12.0 m will require construction of a new spillway with an increased height at the pre occupied by the old spillway. In order to assess the depth to fresh bed rock and subsurface

rock mass condition in the area of interest. The geophysical seismic refraction survey in conjunction with electrical resistivity survey conducted at the proposed site for construction of new spillway. In total, 595 m consisting of 5 spreads of seismic refraction survey, four across spillway site and one parallel to the old spillway (CSMRS Report, 2010) were conducted. For the seismic refraction survey work, blasting and hammers were used as the energy source. One geoelectrical sounding conducted along the seismic profile No. 110002. 2. SUBSURFACE GEOLOGY The project area falls in Bundelkhand province of Indian Peninsula. Bundelkhand Province is a triangular segment composed of rocks of Bundelkhand Granitoid Complex (BGC) comprising undifferentiated granite-granodiorite, quartzo-feldspathic, gneisses and enclaves of meta-sedimentary and meta-volcanic rocks, quartz reefs and dykes of dolerite. The existing geochronological data indicate that the rock were cratonised mainly in Late Archaen period. Besides, rock exposure of Bundelkhand Granitoid Complex, the area comprises sediments of banda older alluvium and newer allivium. The banda older alluvium lying over the Bundelkhand Granitoid Complex (BGC) consists of red to deep brown sand with gravel lenses, silt and clay with kankar. The newer alluvium lies over banda alluvium and is

Birendra Pratap & Alex Varughese

further divided into channel alluvium and terrace alluvium. The channel alluvium and terrace alluvium comprises of unconsolidated quartz, red sand, silt and clay. The rock is hard, massive and moderately jointed. The rock exposed at the upstream and downstream portions of the proposed new spillway site is medium to coarse grained, pinkish grey granite belonging to Bundelkhand Granitoid Complex (GSI Report, 2009). The rock in general is fresh to slightly weathered but at places particularly at the base of Dharaun hill is highly weathered. Besides the rock exposures in the area, particularly D/S of the existing dam axis is covered with the alluvium comprising silty and clay of banda alluvium. Since, the old spillway structure already exists at the proposed site for new spillway hence major part of the area is covered with the anthropogenic structure and material like fill material, boulders, concrete and pitching. Rock at the site is sparsely exposed. 3. GEOPHYSICAL SURVEY Geophysical survey of this project were conducted in consultation with executive engineer, Maudaha dam construction division-I, Mahoba, U.P. with project engineers. The main objective this survey was to determine the depth to bed rock profile, quality of the bed rock and delineate the unconsolidated formations overlying on the basement rock. The field survey was conducted by seismic refraction and electrical resistivity methods. 3.1 Seismic Refraction Method In seismic refraction survey, elastic waves are generated artificially by explosive source or by mechanical hammer.

The source should ideally provide a pulse of duration of not more than a few milliseconds with high amplitude. In an elastic homogeneous ground the source inducts instantaneous deformation, which causes deformation in the vicinity of the source point. By virtue of elastic behavior of the rock, the elastic deformation propagates in all direction in spherical wave fronts. On critical refraction the refracted wave travels along the interface and sends out secondary waves to the surface. These returning waves are recorded in the seismic refraction survey. A typical seismic profile consists of 24 numbers of vertical geophones uniformly spaced at a distance of 5 m. Explosive source/mechanical hammer is used for creation of seismic waves. Five shots are used for recording the seismic waves depending upon the topography and layout of the profile. The quantity of the explosive charge used range from 0.5 kg to 5 kg depending upon the site conditions. The shots are fired with zero delay detonators. The seismic equipment is actuated automatically by make and breaks circuit or by a trigger geophone planted a distance of less than 0.5 m from the shot point. The seismic waves were trace recorded in the floppy disc and a hard copy print was taken. The seismic refraction survey were conducted using the ABEM TERROLAC- 24 channel seismograph consisting two 12 - conductor cables, amplifiers, geophones, seismic cables, seismic energy source, blaster and recording device. A total five seismic lines, four across old spillway and one parallel to the old spillway were conducted along three seismic profiles. One geoelectrical sounding conducted along the seismic profile No. 110002 is shown in Figure 1.

Fig. 1 Location map of Geophysical survey at Kabrai dam, District, Mahoba, U.P.

Geotechnical Characterization for New Spillway at Kabrai Dam by Geophysical Survey

The seismic line was surveyed with 24 geophone spreads. The seismic surveys were performed with 5 m geophone spacing. For each spread of 24 geophones, 5 shots executed. Two parallel seismic profiles of length 240 m each with 5 m geophone interval were conducted across the old spillway. The first seismic profile consists of two seismic lines 110001 & 110002 and second geoseismic profile consists of seismic lines 220001 & 220002. Third seismic profile of length 115 m with same interval consists seismic line 330001 conducted along the road site, nearly parallel to dam axis. The locations of each seismic profile lines are shown in Figure 1. The geoseismic section of profile lines is shown in Figure 2- 4.

The analysis of seismic data has been carried out on the basis of seismic velocities recorded at each site and the depths derived adopting the recorded velocities using Plus–Minus method (Sjogren, 1984). Different seismic velocities seen from the field records have been interpreted by considering the variation in seismic velocities within the different types of Sub-surface material (Varughese et al, 2011) and taking into account the local geology. The interpretation procedure consisted of picking up of the first arrival of P-wave from the field records. The Time distance graph was drawn from all the shot points. The interpretation of P- wave velocity and depth of subsurface layers was performed by using plus minus method. The depth of the refractor was determined at each geophone location. Results of seismic refractions survey are shown in Table 1. The interpreted results are presented in the form of geoseismic section are given in Figures 2-4. 3.2 Electrical Resistivity Method Geophysical resistivity surveys are usually designed to measure the electrical resistivity of sub-surface materials by making measurements at the earth surface. An electrical field is imposed in the ground by a pair of electrodes at varying spacing expanding symmetrically from a central point, while measuring the surface expression of the resulting potential field with additional pair of electrodes at the appropriate spacing. For an array of current electrodes A and B, and potential electrodes M and N, the apparent resistivity ‘ρa

’ is expressed by the equation:

Table 1 The interpreted results of Seismic Refraction survey

S. No. Profile No. Depth of Ist Layer (m) Depth of 2nd Layer (m) Range of Seismic Velocity (m/sec) Ist Layer 2nd Layer/ bed rock

1 110001 &

110002

5.0 to 12.0 5.0-12.0 to - - - - - 700-1400 4900-5300

2 220001 &

220002

7.0 to 9.0 7.0-9.9 to - -- - - - 1000-14000 4900-5300

3 330001 5.0 to 8.0 5.0-8.0 to - -- - - - 500-1250 2500-4100

Birendra Pratap & Alex Varughese

Where R = resistance {R = ΔV/I}, I is the current introduced in the earth and ΔV is the potential difference between the potential electrodes. One Geoelectrical Sounding (GS) with Schlumberger electrode configuration was carried out. The Terrameter SAS-300 Manufactured by M/s ABEM, Sweden was used for electrical resistivity survey. The sounding was conducted with the maximum current electrode separation (AB) of 90 due to the spread constraint. Current electrode separation were expanded in steps of 1.0 m up to AB/2 = 12 m and then in steps of 5 m up to AB/2 = 45 m onwards, with appropriate MN separation. The values of apparent resistivity (ρa: product of resistance and geometric factor) in ohm-m were plotted against the related half-current electrode separation on double logarithmic scale paper of moduli 62.5 mm. The curve was carefully smoothened for the interpretation. Preliminary quantitative interpretation of VES curve was attempted by semi empirical ‘Auxiliary Point’ method with the help of two-layer master curves and auxiliary point charts of Orellana–Mooney. The interpreted result gives the resistivity of different layers and the depth of various interfaces underneath. The data was also processed and interpreted on IPI2Win to verify the manually interpreted result. Any deviation of the computed curve from the related field curve was modified keeping in view of the local geology to arrive at a realistic model (CWPRS, 2005). The interpreted true resistivity of the field GS curve indicates three sub-stratum geoelectrical layers. The resistivity

sounding curve obtained in the area is H or HA types. The resistivity of the 1st layer is 125 Ohm-m with thickness of 1.22 m, represent top surface layer/overburden. Resistivity of the 2nd layer is 448 Ohm-m and thickness is 11.8 m indicating weathered rock/ hard semi-consolidated formation. The 3rd layer is having resistivity of 9722 Ohm-m representing hard rock formation/bed rock. The true resistivity values, thickness and depth range are given in table 2. 4.0 CONCLUSIONS The geophysical Seismic Refraction and Electrical Resistivity survey conducted at Kabrai dam, helped in deciding and designing the foundation of new spillway at pre occupied by old spillway. The seismic survey results are presented in P-wave velocity profiles with reduced level (RL). These profiles revealed two subsurface layers; the first layer (top surface layer) consists of overburden and highly weathered rock and second layer consists of bed rock of very good quality. The geoelectrical sounding field curve interpretation delineates the three sub-stratum geoelectrical layers. The results of geophysical survey correlated with drilling results for accurate understanding. It is demonstrated that geophysical survey can be effectively used for geotechnical site characterization quickly and economically. ACKNOWLEDGEMENTS

The authors are really grateful to Director and Joint Director/Scientist ‘E’, Central Soil and Materials Research Station (CSMRS), New Delhi for his kind cooperation and permission to publication of this paper. The authors are also grateful to S/Sh. N.N. Singh and S. K. Kaushik for their hard work in completing field investigation.

Table 2 The interpreted results of Geoelectrical Sounding data

GS No True Resitivity (Ohm-m)

Thick-ness (m)

Depth Range ( m bgl )

Inferred Geology/sub-surface layers

1 ρ1 = 125 1.22 0.00-1.22 Top surface layer ρ2 = 448 11.8 1.22-11.8 Hard semi-consolidated formation ρ3 = 9722 ----- 11.8- --- Hard formation/bed rock

REFERENCES [1] CSMRS (2010), Report on Seismic Refraction Survey

for New Spillway at Kabrai dam, Arjun Shayak Project, District Mahoba Utter Pradesh.

[2] GSI (2009), Report on Geotechnical Investigations for the new spillway for Kabrai dam Arjun Shayak Project, District Mahoba, Utter Pradesh.

[3] Sjo’gren, Bengt (1984), “Shallow Refraction Seismic” Chapman& Hall Atlas Copco, ABEM AB (1983) Terraloc Seismic System, Bromma Sweden.

[4] Varughese, Alex, Pratap, Birendra and Gupta, S.L. (2011), Subsurface Investigation by Seismic Refraction Survey on Bouldary Bed- A Case Study, Proc. of Third Indian Rock Conference, INDOROCK-201, 13-15 Oct. 2011, Roorkee, 145-149.

[5] CWPRS (2005), A Short Course on Geophysical Investigations for Engineering Projects.


EXPANDED POLYSTYRENE (EPS) GEOFOAM: A MATERIAL BEHAVIOR A. H. Padade, Research Scholar, Indian Institute of Technology Bombay, email: [email protected] J. N. Mandal, Professor, Indian Institute of Technology Bombay, email: [email protected] ABSTRACT: In this present paper an attempt has been made to understand the behaviour of the EPS geofoam subjected to different types of loading conditions. The mechanical properties for different densities such as 0.15, 0.20, 0.22 and 0.30 kN/m3 of EPS geofoam are presented in this paper. The compressive strength, tensile strength, flexural strength and shear strength are evaluated using ASTM test specifications. The detail analysis of these strength parameters has been carried out. The relationships of the strength parameters with varying density of EPS geofoam have been established. These relationships are very useful for determination of the various strength parameters EPS geofoam.

INTRODUCTION Expanded polystyrene (EPS) geofoam is a super lightweight material. It is used as a compressible inclusion behind the retaining walls to reduce the lateral earth pressure and as a lightweight fill material over soft soils to reduce the overburden pressure. EPS geofoam is available in the form of blocks. The advantage of EPS geofoam due to its very low density, it can be handled, transported and placed at construction site very easily and no skilled labour is required. Also this EPS geofoam material is environment friendly which does not pollute the soil as well as ground water. The of EPS geofoam material under short term and long term compression loading conditions was discussed and suggested that the linear elastic limit of this material lies between 1 – 2% of strain [1]. Feasibility characteristic of EPS geofoam was investigated to determine the mechanical properties of EPS geofoam [2]. Centrifuge modelling of retaining wall was carried out to reduce the lateral earth pressure with geofoam as compressible inclusion behind the retaining wall [3]. Stability analysis of embankment made up of EPS geofoam blocks on soft soils was performed without and with reinforcement at separation level [4]. Centrifuge modelling tests was performed to improve the bearing capacity of soft soil using EPS geofoam [5]. Slope stability analysis of embankment constructed using EPS geofoam was carried out using Artificial Neural Network [6]. Drained triaxial test was carried out to study the of EPS geofoam under shear to develop elastoplastic model [7]. Deformability and tensile strength of EPS geofoam was investigated under short term loading condition. A mathematical model was obtained by using the thickness of EPS geofoam from which the parameters such as modulus of elasticity, relative ultimate strain and ultimate strength of the material can be obtained [8]. Tensile of EPS geofoam was carried out with densities 12, 20 and 25 kg/m3 and observed that the material was characterized by brittle failure and tensile strength increases with increase in the density [9]. Micro and macro mechanical properties of EPS geofoam carried out by conducting uniaxial and triaxial tests [10].

INVESTIGATION In the present study, the mechanical properties such as compressive strength, tensile strength, flexural strength and shear strength of EPS geofoam material with densities 0.15, 0.20, 0.22 and 0.30 kN/m3 was determine by using American standards of Testing Material (ASTM) specifications. The relationship between compressive strength at different strain level, tensile strength, flexural strength and shear strength with different densities of EPS geofoam have been established and presented in this paper. The uniaxial compressive strength (UCS) test was also performed on the EPS geofoam and the results are presented in this paper. RESULTS AND DISCUSSIONS Compressive strength The compressive strength test was performed on the EPS geofoam with different density as per ASTM D1621-2010 [11]. The compressive strength obtained at 5% and 10% strain levels and the yield strength was also obtained. The test results are shown in the Table 1. Table 1 Compressive strength of different density of EPS geofoam at various strain levels

Density of EPS geofoam (kN/m3)

Compressive strength (kPa)

@ 5% @ 10% @ yield

0.15 57.03 62.3 53.26 0.20 84.08 91.9 79.84 0.22 100.32 110.27 94.85 0.30 132.33 146.97 124.03

Figure 1 shows the relation between compressive strength values to various densities of EPS geofoam. The correlation of the parameters is established and Eqs. 1-3 are developed for compressive strength at various strain levels. It was observed that the compressive strength of the

A. H. Padade & J. N. Mandal

Fig. 1 Relationship between compressive strength at different strain level and density of EPS geofoam EPS geofoam varies linearly with the density of EPS geofoam.

85.192.564%5@ −= gc γσ (1)

48.158.500%10@ −= gc γσ (2)

4.0411.469@ −= gyieldc γσ (3)

Where, σc = Compressive strength (kPa); γg = Density of EPS geofoam (kN/m3)

Initial tangent modulus The initial tangent modulus was also calculated from the compressive strength test for different sizes of the test specimens with same density. It was observed that with all the possible sizes of the specimen the initial tangent modulus value shows very little variation given in Table 2. Table 2 Initial tangent modulus with different sizes of the specimen of different denity of EPS geofoam Size of the specimen (mm)

Density of EPS geofoam (kN/m3) 0.15 0.20 0.22 0.30

Initial tangent modulus (kPa)

50×50×50 2271 3924 5256 7121 100×100×100 2603 4277 5796 7566 150×150×150 2134 4178 5557 7909 57.5 ф × 50 ht 2416 3863 5755 7384 115 ф×100 ht 2702 4115 5321 7427 170 ф×150 ht 2702 4063 5361 7877

The relationship of initial tangent modulus with EPS geofoam is shown in Fig. 2 and the Eq. 4 gives the linear variation of the modulus with increase in the density of EPS geofoam material.

Fig. 2 Relationship between initial tangent modulus of different sizes of specimen and density of EPS geofoam

537234228 −= giE γ (4)

Where, Ei = Initial tangent modulus (kPa); γg = Density of EPS geofoam (kN/m3)

Tensile, flexural and shear strength Tensile, flexural and shear strength tests were carried out on the EPS geofoam with different density as per ASTM D1623-2009 [12], ASTM C203-2005 [13] and ASTM C273-2007 [14] respectively. The test results are given in Table 3. Table 3 Tensile, flexural and shear strength of different density of EPS geofoam

Density of EPS geofoam (kN/m3)

Tensile strength

(kPa)

Flexural strength

(kPa)

Shear strength

(kPa) 0.15 154.89 149.9 83.65 0.20 216.4 211.3 98 0.22 244.54 240.6 121.570.30 407.78 277 139.27

The relationship of various strength with density of EPS geofoam is shown in Figs. 3-5 respectively.

Fig. 3 Relationship between tensile strength and density of EPS geofoam

Expanded polystyrene (EPS) geofoam: a material behavior

Fig. 4 Relationship between flexural strength and density of EPS geofoam

Fig. 5 Relationship between shear strength and density of EPS geofoam A linear relationship was found in tensile and shear strength whereas the non linear relationship was found in flexural strength. The equations for tensile, flexure and shear strength are given in Eqs. 5-7.

9.1161714 −= gt γσ (5)

Where, σt = Tensile strength (kPa); γg = Density of EPS geofoam (kN/m3)

19930254880 2 −+−= ggf γγσ (6)

Where, σf = Flexural strength (kPa); γg = Density of EPS geofoam (kN/m3)

5.923.377 −= gS γ

(7)

Where, S = Shear strength (kPa); γg = Density of EPS geofoam (kN/m3) Uniaxial compressive strength (UCS) test The uniaxial compressive strength test was performed on the EPS geofoam of 75 mm diameter and 150 mm height with different density. The Mohr circles of various densities of EPS geofoam material has been plotted as shown in Fig. 6.

Fig. 6 Mohr circles from UCS test for different density of EPS geofoam The cohesion value of EPS geofoam material obtained from UCS test is given in Table 4. Table 4 Cohesion of different density of EPS geofoam from UCS test

Density (kN/m3) Cohesion (kPa)0.15 29.43 0.20 42.94 0.22 44.15 0.30 66.09

The cohesion value was then plotted against density as shown in Fig. 7. The cohesion can be calculated by using Eq. 8.

Fig. 7 Relationship between cohesion and density of EPS geofoam

.97269.241 −= gC γ (8)

Where, C = Cohesion (kPa); γg = Density of EPS geofoam (kN/m3) CONCLUSIONS A comprehensive study has been conducted to know the behavior of EPS geofoam under various possible loading conditions. Form the study the following conclusions are made.

A. H. Padade & J. N. Mandal

1. Compressive, tensile and shear strength of EPS geofoam increases linearly with the increase in the density of material. Whereas a non linear relationship was observed in case of flexural strength.

2. The initial tangent modulus values calculated from compressive strength test on different sizes of the EPS geofoam are observed to be very nearer. The initial tangent modulus increased linearly with increase in the density of material.

3. The equations developed in the present study for various strength parameters can be used for determining the approximate strengths of EPS geofoam for a given density.

4. A linear relationship was observed between cohesion and density of EPS geofoam through unconfined compressive strength test.

REFERENCES 1. Horvath, J.S. (1994), Expanded polystyrene (EPS)

geofoam: An introduction to material behaviour, Geotextiles and Geomembranes, 13, 263-280.

2. Mandal, J. N. (1998), A report on feasibility study of geofoam geosynthetics, Packshield Industries, Mumbai, IIT Bombay, pages 46.

3. Mandal, J. N. and Nimbalkar, S. S. (1999), Reduction of lateral earth pressure by compressible inclusion by centrifuge modelling, Proce. of IGC-1999, Dec 1999, Calcutta, 63-67.

4. Mandal, J.N. and Nimbalkar, S.S. (2000), Centrifuge tests for bearing capacity of EPS on soft soils, Foundation Problems and Case Studies, IGS Indore Chapter, Oct- 2000, 60-74.

5. Nimbalkar, S. S. (2000), Some studies on geofoam geosynthetics, M. Tech Thesis, IIT Bombay, pages 232.

6. Mandal. J. N. And Nimbalkar. S. S. (2004), Slope stability analysis by artificial neural network, International Conference on Geosynthetics and Geoenvironmental Engineering, IIT Bombay, India, 468-471.

7. Wang, H and Leo, C. J. (2006), A simple elastoplastic hardening constitutive model for EPS geofoam, Geotextiles and Geomembranes, 24, 299-310

8. Gnip, I.Y., Vejelis, S., Kersulis, V. and Vaitkus, S. (2007), Deformability and tensile strength of expanded polystyrene under short term loading, Polymer Testing, 26, 886-895.

9. Kang, Y., Li, X. and Tan, J. (2008), Uniaxial tensile of EPS, J. Cent. South University of Technology, 202-205.

10. Ossa, A. and Romo, M.P. (2010), Dynamic characterization of EPS geofoam, Geotextiles and Geomembranes, 29, 40-50.

11. ASTM D1621 (2010), Standard test method for compressive properties of rigid cellular plastics, ASTM, Philadelphia, PA, U.S.A.

12. ASTM D1623 (2009), Standard test method for tensile and tensile adhesion properties of rigid cellular plastics, ASTM, Philadelphia, PA, U.S.A.

13. ASTM C203 (2005), Standard test methods for breaking load and flexural properties of block-type thermal insulation, ASTM, Philadelphia, PA, U.S.A.

14. ASTM C273 (2007), Standard test method for shear properties of sandwich core materials, ASTM, Philadelphia, PA, U.S.A.


A STUDY ON THE STRENGTH AND COMPRESSIBILITY BEHAVIOUR OF ORGANIC

CLAYS FROM KUTTANAD, KERALA S. Chandrakaran, Professor, NIT Calicut, [email protected] C. J. Archana, Graduate Student, NIT Calicut, [email protected] ABSTRACT: This paper presents the study on the strength and compressibility behaviour of soil samples collected from Thakazhy, Thayankary, Kandamkary, and Champakulam villages belonging to Kuttanad taluk in Kerala. The influence of organic matter on the strength and compressibility behaviour is also presented. The compression indices and the liquid limit of soils increased with increase in organic content. Also, the influences of various pore fluids like water, NaCl, CaCl2 and FeCl3. 6H2O solutions of different concentrations on the index and strength properties are presented. Regression equations for compression index based on simple parameters obtained in the laboratory are developed. INTRODUCTION Kuttanad clays are characterised by their high compressibility and low shear strength behaviour. The soil in this region is generally silty clay of high compressibility with moderate to high organic content. Since the region lies below mean sea level the mineral constituents in the pore fluid also affects the soil behaviour. In this paper, the study on the strength and compressibility behaviour of Kuttanad clays is presented. Soil samples collected from Thakazhy, Thayankary, Kandamkary, and Champakulam villages belonging to Kuttanad taluk in Kerala are used for the study. The influence of organic content on the index properties, strength and compressibility behaviour is presented. Also, the influences of different pore fluids like water, sodium chloride(NaCl), calcium chloride(CaCl2) and ferric chloride(FeCl3. 6H2O) solutions on the index and strength properties are studied. Empirical regression equations for the prediction of compression index of these soils based on the simple parameters that can be easily obtained in the laboratory are presented. EXPERIMENTAL PROGRAMME Materials Materials used for the study include clayey soil obtained from four regions in Kuttanad taluk viz., Thakazhy, Thayankary, Kandamkary, and Champakulam. Eight undisturbed samples, two from each region were collected in iron sampling tubes of 100mm diameter. The samples collected were properly sealed on both sides to avoid loss of matter and moisture. Also sufficient quantities of loose samples were collected from the sites. In order to vary the organic content 5%, 10% and 20% starch solution were used. In order to estimate the organic content in the prepared soil specimens, 30 % hydrogen peroxide solution was employed. 0.25N, 0.5N, 0.75N and 1N solutions of NaCl, CaCl2 and FeCl3.6H2O and water were used as the different pore fluids.

Methods For all the eight undisturbed samples physical andindex properties tests were carried out initially. Then undisturbed soil samples were prepared for consolidation tests and oedometer tests were conducted. Also three soil samples from each of the four locations were prepared on previously air dried soils, with varying organic content by the addition of 5%, 10% and 20% starch solution and allowing for a curing period of one day. Organic content and Atterberg limits of these samples were determined as per the standards. Also remoulded samples were prepared and oedometer tests were conducted for these modified soil samples. Four soil samples, each from Thayankary, Kandamkary and Champakulam were prepared by mixing with 0.25N, 0.5N, 0.75N and 1N solutions of NaCl, CaCl2 and FeCl3.6H2O and allowing for a curing period of one day. Another three soil samples, from each of these three locations were prepared by mixing with 5%, 10% and 20% starch solution and allowing for a curing period of one day. All these samples were prepared on oven dried soils and used for Atterberg limits tests and unconfined compressive strength tests. RESULTS AND DISCUSSIONS Undisturbed Soil Samples Physical and Index Properties of the Undisturbed Soil Samples The physical and index properties tests of the undisturbed soil samples were conducted. The results of natural water content (wn), bulk density (γ), dry density (γs), organic content, specific gravity and Atterberg limits tests are summarized in Table 1 and 2. The soil classification is also done as per USCS based on the plasticity index. All the samples except that from Thakazhy fall under the classification of clay of low compressibility. This may be due to the higher percentage of sand found in these soils.

S. Chandrakaran, C. J.Archana

Table 1 Physical properties of undisturbed soil samples Location wn % γ % γd % OC % GThakazhy 27.79 19.50 15.20 2.65 2.50Thakazhy 23.07 21.10 17.10 1.25 2.53Thayankary 38.53 18.50 13.30 5.52 2.61Thayankary 39.14 18.70 13.40 3.55 2.60Kandamkary 38.53 19.00 13.70 2.95 2.74Kandamkary 29.05 19.70 15.20 3.38 2.73Champakulam 49.64 17.90 11.90 6.65 2.62Champakulam 46.30 18.30 12.50 7.66 2.62

Table 2 Index properties of undisturbed soil samples

Location wL % wp % ws % Ip % USCS classification

Thakazhy 37 23 17 14 CLThakazhy 39 24 16 15 CLThayankary 61 28 10 33 CHThayankary 73 29 8 44 CHKandamkary 59 28 23 31 CHKandamkary 57 27 21 30 CHChampakulam 73 28 26 45 CHChampakulam 72 28 24 44 CH

Consolidation Tests Standard oedometer tests were conducted for 8 undisturbed samples keeping the pressure increment ratio as 2 and the samples were loaded to a maximum pressure of 1600 KPa and unloading was done to one-fourth of the applied pressure in sequence. The compression indices, swell/expansion indices of the specimens were obtained and are tabulated in Table 3. Table 3 Results of consolidation tests on undisturbed soil samples

Location Initial VoidRatio, eo

Compression Index, Cc

Swell Index,Cs

Thakazhy 0.57 0.094 0.010Thakazhy 0.51 0.091 0.010Thayankary 1.57 0.525 0.049Thayankary 1.64 0.485 0.046Kandamkary 1.14 0.201 0.023Kandamkary 1.13 0.291 0.039Champakulam 2.11 0.597 0.043Champakulam 1.87 0.621 0.044

Disturbed Soil Samples Disturbed samples were collected from Thakazhy, Thayankary, Kandamkary and Champakulam. Both air dried and oven dried samples were used for the tests. The air dried samples were subjected to study on variation of Atterberg limits and compressibility behaviour by varying the organic content. Oven dried samples were used for the study of variation of Atterberg limits and unconfined compressive strength by employing different chemical solutions as pore fluids and also the effect of organic content over the Atterberg limits and UCC strength is studied.

Tests on Air Dried Soil Samples Disturbed samples collected from Thakazhy, Thayankary, Kandamkary and Champakulam were used for the study on the compressibility behaviour with varying organic content utilising 5, 10 and 20 % starch solution to the original air dried soil sample. The results of index properties tests and consolidation tests as shown in Table 4 reveal that the liquid limit increases with percentage increase in the organic content and for some soils it reaches a peak value and then drops [1, 2]. The compression index and the initial void ratios of these soils were found to increase as the organic content in the soil improved. Table 4 Index properties of air dried soil samples amended with organic content

Location %Starch

% OC

wL %

wp

% ws %

Ip % Cc

Thakazhy 5 3.68 45 24 23 21 0.21310 6.26 58 23 22 35 0.23620 11.74 52 31 29 21 0.271

Thayankary 5 7.47 134 54 34 81 0.45410 11.56 136 55 36 81 0.52120 19.39 128 56 39 72 0.700

Kandamkary 5 6.23 106 56 52 50 0.43610 10.74 106 54 52 52 0.44020 16.84 107 51 50 56 0.530

Champakulam 5 10.19 99 43 32 56 0.37610 13.10 107 42 31 65 0.39920 20.51 114 39 31 75 0.466

Tests on Oven Dried Soil Samples The Atterberg limits and unconfined compressive (UCC) strength values of modified soils from Thayankary, Kandamkary and Champakulam subjected to different concentrations of pore fluids like such as water, NaCl, CaCl2 and FeCl3. 6H2O solutions of varying concentrations were determined. Also the Atterberg limits and UCC strength for these soil samples mixed with different percentages of starch solution were determined. The values obtained are given in Table 5. Table 5 Index properties of oven dried soil samples amended with organic content and with chemicals as pore fluids

Site Chemical

% OC wL %

wp

% ws % UCC Name Concen

tration

Thay

anka

ry

Water - 4.23 59 32 30 61.94Starch 5% 5.38 64 42 33 91.07

10% 8.13 65 41 35 52.9420% 13.72 66 38 34 124.13

NaCl 0.25N - 59 37 34 70.030.5N - 60 35 33 67.70

0.75N - 60 36 33 63.011N - 58 38 33 57.32

CaCl2 0.25N - 60 34 32 75.340.5N - 58 31 30 73.59

0.75N - 57 31 31 71.881N 57 33 29 70.15

A study on the strength and compressibility behavior of organic clays from Kuttanad, Kerala

Site Chemical

% OC wL %

wp

%ws % UCC Name Concen

tration

FeCl3 0.25N - 64 44 32 79.99 0.5N - 66 36 33 95.80 0.75N - 62 33 31 99.65 1N - 55 34 32 95.83

Cha

mpa

kula

m

Water - 4.03 48 23 21 121.79Starch 5% 4.55 49 23 21 119.34

10% 8.05 54 24 21 167.64 20% 13.27 55 26 23 158.29

NaCl 0.25N - 45 23 21 159.28 0.5N - 51 21 20 157.58 0.75N - 52 21 20 150.14 1N - 50 22 20 139.36

CaCl2 0.25N - 53 29 22 144.56 0.5N - 43 24 22 138.18 0.75N - 40 22 21 135.49 1N - 41 23 20 135.16

FeCl3 0.25N - 47 24 22 78.55 0.5N - 53 24 22 159.80 0.75N - 53 25 22 151.88 1N - 48 28 23 85.51

Kan

dam

kary

Water - 3.77 69 37 36 110.09Starch 5% 4.22 60 46 43 58.26

10% 6.59 64 44 41 60.61 20% 10.34 71 41 40 45.46

NaCl 0.25N - 45 39 38 82.64 0.5N - 52 38 35 84.44 0.75N - 53 38 35 84.41 1N - 51 40 37 83.82

CaCl2 0.25N - 54 39 37 63.97 0.5N - 53 41 39 70.42 0.75N - 49 42 40 70.08 1N - 43 43 41 65.70

FeCl3 0.25N - 47 42 39 45.38 0.5N - 54 36 35 64.03 0.75N - 51 34 33 62.59 1N - 43 34 33 47.76

The liquid limits of all the samples show a decreasing trend towards higher chemical concentration pore fluids used. At higher concentrations of the pore fluids the diffused double layer is suppressed causing decrease in liquid limit. The replacement of water with the above mentioned salt solutions as the pore fluids results in an increase in the ion concentration which causes reduction in the double layer thickness and this, in turn, causes a reduction in the antiparticles repulsion and an increase in the attraction, resulting in the increase in cohesion. The UCC strength is found to be lower towards higher concentration of electrolytes [3, 4, 5]. This may be due to the disturbances of structural bonds in the lattice assembly, possibly due to the high suppression of repulsive forces. The UCC strength of soils from Kandamkary and Champakulam were found to show a decreasing trend at

higher % starch solution added to them. But For soil from Thayankary, there was an initial drop in the UCC strength but at higher percentages of starch solution the strength was found to be improved.

SIMPLE REGRESSION EQUATIONS FOR COMPRESSION INDEX OF SOILS FROM KUTTANAD Regression analysis gives a best fit curve for the relation between the dependant and the independent variables [6, 7, 8, 9, 10]. Regression analyses were performed on results of compression index as dependant variable and other properties like liquid limit (wL), initial void ratio (eo), natural water content (wn), plasticity index (Ip) and dry density of soils (γd) as the independent variable. The test results for undisturbed samples given in Table 1- 3 were used for the purpose. The regression equation obtained in each of the combination along with the obtained correlation coefficient (r) and coefficient of determination (r2) is given in Table 6. Table 6 Recommended Regression Equations Parameter Varied

Relation r r2

Liquid Limit, wL

Cc = 0.0139 wL – 0.454 0.906 0.822

Initial Void Ratio,eo

Cc = 0.3654 eo – 0.118 0.968 0.937

Natural Water Content, wn

Cc = 0.0209 wn – 0.4 0.875 0.766

Plasticity Index, Ip

Cc = 0.0163 Ip – 0.158 0.916 0.840

Dry Density, γd Cc = 1.95 – 1.13 γd 0.868 0.753γw/ γd Cc = 2.332 (γw/ γd) – 1.319 0.882 0.778 The most appropriate equation for the compression indices of undisturbed samples from Thakazhy, Thayankary, Kandamkary and Champakulam is given in Eq. 1

(1)

which has a r2 value = 0.937, where Cc is the compression index and eo is the initial void ratio. The observed and predicted range of values for compression index for the above equation is shown in Table 7. Table 7 Comparison of Compression index obtained in the study with the predicted values for compression equations obtained from the regression equation Cc = 0.3654 eo – 0.118

Sl. No. Observed Compression Index

Predicted Compression Index

1 0.094 0.0892 0.091 0.0673 0.525 0.4564 0.485 0.4815 0.201 0.2996 0.291 0.2957 0.597 0.6538 0.621 0.565

S. Chandrakaran, C. J.Archana

CONCLUSIONS The natural water content of undisturbed soil samples collected from Thakazhy, Thayankary, Kandamkary, and Champakulam vary from 23 % to 50% with organic content varying from 1.25% to about 7.66%. The soil samples from all the places except Thakazhy fall in the USCS classification of clay with high compressibility, and that from Thakazhy being clay with low compressibility due to the higher percentage sand content in the soil collected from Thakazhy. Consistency limits tests conducted on modified soils from Thakazhy, Thayankary, Kandamkary and Champakulam, whose organic content has been varied by the addition of starch solutions, indicate that there is an increase in the liquid limit and plasticity index, while plastic limit and shrinkage limit decrease with increase in % organic content. Consolidation tests on these modified soils indicate that with increasing % organic content the compression index was found to increase. When different concentrations of NaCl, CaCl2 and FeCl3.6H2O solutions were used as the pore fluid in oven dried soil samples, the liquid limit of all the samples show a decreasing trend as the pore fluid concentration increased. At higher concentrations the diffused double layer is suppressed causing decrease in liquid limit. The liquid limits of oven dried samples amended with organic content show an increasing trend whereas the plastic limits and shrinkage limits were found to be decreased. The UCC strength of the oven dried samples mixed with NaCl, CaCl2 and FeCl3.6H2O solutions increased marginally and then decreased with the increase in their respective concentrations. Regression analysis of results for compression index and other basic soil parameters for undisturbed soil samples from the four locations gave way to a new empirical relationship between compression index and initial void ratio which suited the results in the best possible way.

Scope for Further Work The soil available in Kuttanad exhibits unusual behaviour. Many researchers have carried out studies to understand the behaviour of these soils. The organic nature of the soil, and the presents of many salts and minerals in its composition gives it a dynamic behaviour when subjected to loads. In the present study on the compressibility and strength behaviour of these organic clays are done. For undisturbed samples only compressibility studies are made. This may be extended to the study of strength characteristics also. The behaviour may be re-examined for both air dried and oven dried samples. The use of different pore fluids and variations in their respective concentrations can also be employed in the study of compressibility behaviour. Pore fluids other than chlorides may be employed. In the present study, curing effect is not considered. This may be extended so that the effect of curing

on the compressibility and strength behaviours can be understood. REFERENCES

1. Krizek R. J., Max W. G., and Paul L. H. (April 1975): “Organic Content and Engineering Behaviour of Typical Maintenance Dredging”, Fourth South East Asian Conference of Soil Engineering, Kuala Lumpur, Malaysia, pp. 6 -15

2. Puppala A. J., Pokola S. P. ,Intharasombat N., Williammee R. (November 2007): “Effects of Organic Matter on Physical, Strength and Volume Change Properties of Compost Amended Expansive Clay”, Journal of Geotechnical and Geoenvironmental engineering, Vol 133(11), pp. 1449-1461

3. Mohamed Zahry Othman and Faridah Shafii (1998): “The Effects of Electrolytes on the Liquid Limit of Clay”, Journal of Civil Engineering, Vol.11, No.1, pp.7-19

4. Seracettin ARASAN and Temel YETIMOGLU (2008): “Effect of Inorganic Salt Solutions on the Consistency Limits of Two Clays”, Turkish Journal of Engineering and Environmental Sciences, Vol.32, pp.107-115

5. Yahia E.A. Mohamedzein and Mohammed H. Aboud (2006): “Compressibility and Shear Strength of a Residual Soil”, Geotechnical and Geological Engineering Journal, Vol. 24, pp. 1385-1401

6. Amr S. Azzous, Raymond J. Krizek and Ross B. Corotis (June 1976): “Regression Analysis of Soil Compressibility”, Soils and Foundations, Japanese Society for Soil Mechanics and Foundation Engineering, Vol. 16 (2), pp.19-29

7. Gil Lim Yoon and Byung Tak Kim (November 2006): “Regression Analysis of Compression Index for Kwangyang Marine Clay”, KSCE Journal of Civil Engineering, Vol. 10 (6), pp 415-418

8. Sengupta D. P. and Juneja P. L. (1965): “Prediction of Compression Index from Soil Properties”, Indian Journal of Technology, Vol. 3(10), pp. 328-331

9. Vinod P. and Bindhu J. (July 2010): “Compression Index of Highly Plastic Clays - An Empirical Correlation”, Indian Geotechnical Journal Vol. 40 (3), pp.174- 180

10. Wroth C. P. and Wood D. M. (May 1978): “The Correlation of Index Properties With Some Basic Engineering Properties of Soils”, Canadian Geotechnical Journal, Vol. 15 (2), pp.137-144


IMPLICATION OF SAMPLE QUALITY IN SETTLEMENT ANALYSIS OF FIELD CASES S. A. Degago, PhD, Norwegian Public Roads Administration, [email protected] G. Grimstad, Associate Professor, Oslo and Akershus University College of Applied Sciences, [email protected] ABSTRACT: Settlement analysis of field cases is normally studied based on parameters interpreted from laboratory samples influenced by varying degrees of sample disturbance. Such disturbance is more pronounced in natural soft clays and could significantly affect the inherent engineering properties of the soil. Hence, it is vital to understand the role of sample quality in relation to soil characterization for long-term settlement analyses. In this work, this has been numerically illustrated using a simple creep model based on various parameter sets derived from laboratory samples depicting effect of sample disturbance and by assessing the implication of these parameter sets on field prediction of a test fill with long-term measurements. INTRODUCTION Natural clays, especially soft marine clays, are characterized by significant creep deformation, destructuration and anisotropy among other features [1,2]. The numerical analyses of such natural clays are normally based on results obtained from laboratory tests. Hence, it is crucial that the laboratory samples have the desired level of quality to give acceptable prediction of field performances. In addition, when reducing a real problem into a numerical idealization that can readily be analysed, it is also vital that soil parameters are interpreted with special emphasis on the numerical model and its underlying assumptions as well as the nature of the problem to be dealt with such as the expected stress range and associated time considerations. In Scandinavia, one typically finds clays in normally consolidated state, where the only factor giving an apparent over consolidation ratio (OCR) is due to aging [1]. Determination of OCR is significantly affected by sample disturbance and its correct determination crucially lies on the sample quality [3,4]. Different quality of the sample will also give different compressibility with respect to stress and time actions. Sample disturbance is hardly avoidable in any soil sample extracted from in-situ. However, the degree of disturbance varies and ideally one aims to acquire a sample with highest quality. Still, it is common to encounter situations where field settlement analysis has to be based on samples of low quality. In such instances, it is important to study the implications of using parameters derived from samples of low quality. Hence, in this work such aspects are discussed from numerical analyses perspective. Laboratory samples depicting effect of sample disturbance are used as a basis for various analyses. In addition, an application example is also presented based on a well-documented test fill from Sweden. SOME MODELLING ASPECTS OF NATURAL CLAYS Constitutive models for clays that are used today are often based on the modified Cam-Clay model (MCCM) [5]. The MCCM was originally developed to model simple elasto-plastic behaviour of soils under triaxial stress and strain conditions. Various modifications and extensions are applied

to MCCM to account various features of clays such as anisotropy, destructuration and creep. Typical modifications of the MCCM include, (1) rotating the yield/reference surface to account for anisotropy [6]; (2) accounting for unstable structure by associating the yield/reference surface with a destructuration formulation by Gens and Nova [7]; and (3) modelling creep and rate dependency by controlling the size of the reference surface using concepts developed by, e.g. Šuklje [8], Perzyna [9] or Janbu [10]. Common for the extensions to the MCCM is the increased complexity that these features bring with them as well as the need for extra soil parameters which may require special laboratory tests. Hence, it is important to distinguish the benefits of using advanced models as compared to simple models. It is also vital to identify cases in which proper use of a simple model could give a better understanding of the problem than using a more advanced model. In this way one can focus on certain selected aspects of soils and grasp the overall picture of the resulting soil responses. Accordingly, such approach has been used in this study. IDEALIZATION This work focuses on settlement analyses of field cases with respect to sample quality. The subject of the study is essentially on clays with significant potential to undergo creep deformations. In addition, such clays usually exhibit anisotropic behaviour with unstable structure. For this illustration, an elasto-viscoplastic soil model available in the FE code Plaxis is selected. This model is referred to as the soft soil creep (SSC) model [11]. SSC is an extension of MCCM based on the isotache concept [8]. In addition the model is isotropic and does not take into account effect of destructuration. However, for the test fill considered in this work, an isotropic formulation is considered sufficient and the effect of destructuration is investigated based on assessing effect of various combination of a certain parameters. By keeping anisotropy and destructuration out of the picture, this work focuses on the creep parameters and their implications as in the isotache formulation adopted in the SSC model.

S. A. Degago & G. Grimstad

KEY PARAMETRS GOVERNING CREEP To properly calculate the development of settlement and pore pressure histories, a creep model must be used. Creep stain rate as a function of time, as defined in the isotache concept [8], is expressed in Eq. 1.

t*με =& (1)

where ε& is the strain rate, t is time and μ* is the modified creep parameter. The isotache concept furthermore uniquely relates effective stress state (peq), reference pre-consolidation stress (po

eq) and volumetric creep strain rate (εv

vp). This relationship, as used in the SSC model, is given in Eq. 2.

***

0

* μκλ

με

−

⎟⎟⎠

⎞⎜⎜⎝

⎛⋅=

eq

eqvpv p

pt

& (2)

where λ* and κ* are the modified compression indexes for virgin compression and recompression line respectively. τ is a reference time corresponding to the specified OCR or (po

eq/peq). Typical value of τ is one day as standard incremental oedometer tests are ran with one day increments. As can be seen from Eq. 2, in addition to μ* the ratio (λ*– κ*)/μ* and OCR are very important for the calculated strain rate. Rearranging Eq. 2, and combining it with the isotropic hardening rule for the reference surface, one could find an expression for the age of the clay, tage corresponding to a given OCR, as

***

μκλ

τ−

⋅= OCRtage (3)

For a certain (λ* – κ*)/μ*, Eq. 3 can be used to estimate either the age of a clay implied when an OCR corresponding to τ is known; or, an OCR corresponding to a certain reference time τ and age of the clay, tage. For example, Waterman and Broere [12] suggests values of (λ* – κ*)/μ* to be in the range 5 (“for soils with considerable amount of creep”) to 25 (“for soils with little creep”). This implies that a standard one day increment oedometer test conducted on a sample taken from a 500 year old clay would give OCR = 1.6 or 11.3 for (λ*– κ*)/μ* = 25 or 5 respectively. Accordingly, soils with considerable amount of creep are expected to show OCR as high as 11 which seem unlikely. A more practical range for (λ*– κ*)/μ* would be 15–50. Further Waterman and Broere [12] suggest that for (λ*– κ*)/μ* > 25 that creep could be ignored. This might be correct in some cases where the applied stress increase gives a stress situation well above the initial pre-consolidation stress. However, in many cases the stress increase is rather moderate and the new situation is around the pre-consolidation stress, making the contribution from creep more significant, regardless of the ratio. For OCR = 1.0 the creep rate becomes independent of the ratio and the actual value for μ* is more important, see Eq. 2.

In situations where the main interest is the final settlement after a certain period of time, an elsto-plastic model can also be used by selecting a single isotache that in average meets the final expected combination of strain and stress states. An isotache selected in this way would typically yield a lower OCR and lower λ*. Equation 4 shows the corrected over consolidation ratio, OCRcorr, which should be used in an elasto-plastic analysis when the true undisturbed OCR is known. By coincidence such approach, Eq. 4, changes the parameters in the same way as is typical to sample disturbance. This is one of the main reasons for explaining why some researchers [13] have been successful in predicting long-term settlements by disregarding creep despite the soil showing significant effect of creep [4, 14].

OCRt

OCRcorr ⋅⎟⎠⎞

⎜⎝⎛=

− ***κλ

μ

τ (4)

THE VÄSBY TEST FILL The Väsby test fill was designed and constructed by the Swedish Geotechnical Institute (SGI) in 1947.The test fill has been monitored with extensive instrumentations and there exists a detailed documentation of all measurements [15]. The Väsby test fill consists of a 30 m x 30 m square and 2.5 m high gravel fill constructed within 25 days. The applied total stress due to the fill was 40.6 kPa. The Väsby test fill site consists of soft sediments of glacial and post-glacial origin. The soft soil layer under the fill is 14 m thick. The ground water table, which is located at an average depth of 1 m from the surface, has a hydrostatic pressure distribution. Settlement history was measured below the centre of the fill, for 57 years after construction of the fill. In order to investigate the effect of sample disturbance, Leroueil and Kabbaj [4] conducted incremental oedometer tests on the Väsby clay. They took samples with 200 mm Laval sampler and compared it with results from the 50 mm Swedish sampler (conducted in 1967). Even though the samples were extracted from a similar depth of ca. 4.2 m, they show significant effect of sample disturbance where a distinct feature of sample disturbance was that it results in lower OCR and λ*. In Fig. 1 the two different laboratory curves are compared. The OCR for the 50 mm is estimated to be 1.1 and for the 200 mm to be 1.6.

Fig. 1 Laboratory incremental odeometer results on samples taken at 4.2 m (after Leroueil & Kabbaj [4])

Implication of sample quality in settlement analysis of field cases

NUMERICAL ILLUSTRATIONS Some of the creep aspects highlighted earlier are numerically illustrated based on six data sets, Table 1 & 2. Data set 1, 2 and 3 are characterized by low OCR as interpreted from the tube sampler and set 4 and 5 are characterized by high OCR as interpreted from the block sampler. In set 6, an even higher OCR is considered based on age considerations. A constant average permeability of 4.0e-5 m/day is used for all sets. Simulation of laboratory tests In Fig. 2 results from laboratory oedometer test simulations are presented. The maximum possible stress increase that can be obtained in the field is about 40 kPa, which means that the results above 70 kPa are of little interest. Furthermore due to buoyancy effect and stress distribution this value will be reduced to less than 50 kPa. This must be kept in mind while evaluating laboratory simulation results (Fig. 2) with respect to laboratory measurements (Fig. 1). Some researchers have reported constant μ*/λ* ratios for wide range of soft clays [17]. However, for models that assume a linear relationship between strain and log stress (constant λ*), like SSC model, this will imply a constant μ*. In reality both λ* and μ* are stress dependent (due to destructuration effect) and the ratio must be evaluated over the applied stress increment. Hence a constant value is not expected, unless the stress increments are small. Therefore adjustments to the ratio can be justified. For set 1, λ* was interpreted from the 50 mm test data and μ* = 0.06 λ* was used from literature [13]. Set 2 and set 3 are, respectively, μ* and λ* variations of set 1. Accordingly, set 1 fits the 50 mm test data best while set 2 and set 3 over predicting the deformations slightly and significantly respectively. The reason for this is that set 2 has a higher creep potential (higher μ*) while set 3 is significantly softer (higher λ*). However, (λ*– κ*)/μ* is higher for set 3 (with a lower μ*), implying that creep has relatively less importance. For set 4, λ* was interpreted from the 200 mm test data and μ* = 0.06 λ* was adopted. Set 5 and set 6 are variations of set 4. The λ* value for set 5 is chosen such that for an OCR of 1.6, a clay age of tage ≈ 500 years is obtained (Eq. 3). In set 6 the OCR is increased such that the age of the clay is tage = 500 years. From Fig. 2, set 6 under predicts the deformations, while both set 4 and 5 over predict deformation with set 5 giving the highest over prediction of the 200 mm test data.

Fig. 2 Odeometer simulation using data sets in Tabel 1 & 2.

Table 1 Data sets with low OCR (OCR = 1.10) Analyses sets Set 1 Set 2 Set 3Modified swelling index κ* 0.030 0.030 0.030Modified compression index λ* 0.191 0.191 0.357Modified creep index μ* 0.011 0.021 0.011Over-consolidation ratio OCR 1.10 1.10 1.10(λ* – κ*)/μ* 14.6 7.7 29.7Initial creep rate ε& (yr-1) 0.995 3.691 0.236“Age of clay” tage (yr) 0.011 0.006 0.047

Table 2 Data sets with high OCR (OCR = 1.60 & 2.18)

Analyses sets Set 4 Set 5 Set 6Modified swelling index κ* 0.030 0.030 0.030Modified compression index λ* 0.357 0.571 0.357Modified creep index μ* 0.021 0.021 0.021Over-consolidation ratio OCR 1.60 1.60 2.18(λ* – κ*)/μ* 15.6 25.8 15.6Initial creep rate ε& (yr-1) 0.005 4e-5 4e-5“Age of clay” tage (yr) 4.13 497 500

Simulation of the Väsby fill The material sets, Table 1 & 2, are adopted in the simulations of the Väsby fill. An axisymmetric model of very fine mesh was used. In addition, an updated-mesh and updated-water pressure procedures were adopted to account for the effect of large deformations and buoyancy respectively. Figure 3 gives settlement history below the centre of the fill along with the corresponding measurements of surface settlements. Interestingly parameter set 1 and 3 only slightly over predicts the measured settlements despite their low OCR values. Set 2 significantly over predicts the settlements due to the implied highest initial creep rate of all the other data sets. Set 3 gave the best field prediction even though it gave significantly softer response in the oedometer simulation (Fig. 2). This is because the initial creep rate for set 3 is lower than for set 1 and 2. The ratio (λ* – κ*)/μ* is highest for set 3 which means that creep is less important in the far field deformations and for situations where the pre-consolidation stress is exceeded. Set 4 also gave excellent prediction while 5 and 6 under predicts the surface settlements. The field under prediction by set 6 is small compared to the significantly stiffer response observed in the corresponding oedometer simulations. It is important to evaluate far field settlements as it provides a benchmark for controlling analyses results and to evaluate how much of the total settlement under the fill is related to the actual loading. In reality the far field settlement is close to zero. Prediction of such far field settlements can realistically be achieved if age considerations are taken into account in the selection of the parameters. The far field surface settlements, 50 m from the centreline, as implied by the analyses sets are given in Fig. 4. Set 1, 2, 3 and 4 gave significant far field settlements with set 2 giving the highest settlement. Set 5 and 6 gave the lowest and more realistic far field settlements due to the reasonable age considerations implied by the data sets. Overall, based on Fig. 3 & 4, parameter set 5 gave the best surface settlements predictions.

S. A. Degago & G. Grimstad

Fig. 3 Measured and simulated surface settlement histories below the center of the fill

Fig. 4 Simulated far field surface settlement histories FINAL REMARKS AND CONCLUSIONS Good estimation of settlement of embankments relies first of all on good sample quality. The effect of sample quality, by selecting different data sets, for the SSC model, was studied based on laboratory and field measurements. The calculations shows that good back calculations of the surface settlement at the centre of the embankment can be made with various sets of parameters. However, the far field settlement can only be realistic when the age of the clay is taken into consideration. A correct combination of the (λ* – κ*)/μ* ratio and OCR is important for overall sound creep settlement analyses. This means that for realistic ratios (λ* – κ*)/μ*, valid for the experienced stress interval, satisfactory far field settlement predictions can be justified at the same time as the settlement below a fill is satisfactory predicted. It is vital to understand the role of parameters using simple models before resorting to advanced models. An example of such advanced model that includes creep, anisotropy and destructuration is the n-SAC model [18]. Such a model puts even more demand on the sample quality for calibration of input parameters. In fact, with proper use, a simple model like the SSC model can give successful predictions and give the user control on certain key input parameters. However, to further improve predictions additional aspects of natural clays such as anisotropy and destructuration should be accounted. A project is initiated in Norway, led by the Norwegian Geotechnical Institute (NGI), to enhance models for settlement analyses. The project is financed by the Research Council of Norway and the industry. The Norwegian University of Science and Technology (NTNU) and Oslo and Akershus University College of Applied Sciences (HiOA) will among others work together on this task.

ACKNOWLEDGEMENTS The work described in this paper is partially supported by the Research Council of Norway through the International Centre for Geohazards (ICG). Their support is gratefully acknowledged. This is ICG contribution No. 400. REFERENCES 1. Bjerrum, L. (1967), Engineering geology of Norwegian

normally-consolidated marine clays as related to settlements of buildings. Géotechnique, 17(2): 81-118.

2. Burland, J. B. (1990), On the compressibility and shear strength of natural clays. Géotechnique, 40(3): 329-378.

3. DeGroot, D. J., Poirier, S. E. & Landon, M. M. (2005), Sample disturbance - Soft clays. Studia Geotechnica et Mechanica, XXVII (3-4): 107-120.

4. Leroueil, S. & Kabbaj, M. (1987), Discussion of 'Settlement analysis of embankments on soft clays' by Mesri & Choi. ASCE, 113(9): 1067-1070.

5. Roscoe, K. H. & Burland, J. B. (1968), On the generalized stress–strain behaviour of wet clay. Engineering Plasticity, Cambridge, 535-609.

6. Dafalias, Y. F. 1986, An anisotropic critical state soil plasticity model, Mech. Res. Commun. 13(6): 341–347.

7. Gens, A. & Nova, R. 1993. Conceptual bases for a constitutive model for bonded soils and weak rocks, Geotech. Hard Soils–Soft Rocks, Balkema, Rotterdam.

8. Šuklje, L. (1957), The analysis of the consolidation process by the Isotaches method. Proc. 4th Int. Conf. Soil Mech. Found. Engng, London, 1: 200-206.

9. Perzyna, P. (1963), Constitutive equations for work-hardening and rate sensitive plastic materials. Proc. Vibration Problems, 4(3): 281-290.

10. Janbu, N. (1969), The resistance concept applied to deformations of soils. Proc. 7th Int. Conf. Soil Mech. Found. Engng, Mexico City, 1: 191-196.

11. Stolle, D. F. E., Vermeer, P. A. & Bonnier, P. G. (1999), Consolidation model for a creeping clay. Canadian Geotechnical Journal, 36(4): 754-759.

12. Waterman, D. & Broere, W. (2005), Practical application of the soft soil creep–Part III. Plaxis Benchmarking, p. 22. http://kb.plaxis.nl/publications

13. Mesri, G. & Choi, Y. K. (1985a), Settlement analysis of embankments on soft clays. ASCE, 111(4): 441-464.

14. Degago, S. A., Nordal, S., Grimstad, G. & Jostad, H. P. (2011), Analyses of Väsby test fill according to creep hypothesis A and B. 13th Int. Conf. of IACMAG, Melbourne, 1: 307-312.

15. Larsson, R. & Mattsson, H. (2003).Settlements and shear increase below embankments. SGI, Report 63, 88p.

16. Mesri, G. & Godlewski, P. M. (1977), Time and stress-compressibility interrelationship. ASCE, 103(GT5): 417-430.

17. Grimstad, G. & Degago, S. A. (2010). A non-associated creep model for structured anisotropic clay (n-SAC). 7th European Conf. NUMGE, Trondheim, Norway, 3-8.


A LABORATORY STUDY ON THE EFFECT OF STRAIN-RATE ON SHEAR STRENGTH PARAMETERS OF COHESIVE SOIL

KINGSHUK DAN, Research Scholar, Jadavpur University, [email protected] R. B. SAHU, Professor, Jadavpur University, [email protected] ABSTRACT: The rate of deformation of structure is a function of rate of loading or unloading of soil. Generally at working load rate of deformation of soil is higher initially which reduces with time. So, the shear strength corresponding to the peak deviator stress as obtained from triaxial tests conducted at a typical strain rate is very often misleading. In the present study a series of undrained triaxial tests were carried out on artificially compacted locally available clay at five different strain rates to study the effect of strain rate on strength parameters. With the increase of strain rate peak deviator stress were found to increase which are similar to the results reported by various investigators. The friction angle were found to reduce with the increase in strain rate while the trend was reverse for cohesion. INTRODUCTION Stability analysis using limit equilibrium method is commonly used to calculate permissible load on soil. Working load is estimated by dividing maximum load by a suitable factor of safety. Effect of time on the strength and deformation characteristics of soft clays have been the object of numerous investigations since late fifties and early sixties. Most of these studies were by means of strength tests determining shear strength of saturated clays. Some studies were made on effect of strain rate on the shear strength of clay [2, 9] which was well supported by some reported data [1, 3]. Further experimental and theoretical studies disclosed that at high rates of testing the pore pressure distribution is highly non uniform and the error involved in the estimation of the effective stresses on the basis of single measurement of pore pressure, either at the centre or at the end of the sample, is a major one and quite unacceptable even in the routine testing. Though theoretical [5] and experimental [4] development enabled investigators to determine the permissible strain rate still it depends on the deformation characteristics of soil. Later it was observed that at different magnitudes of deviator stress, if deformation would occur without change in structure, effective stress or temperature then strain rate would become the only variable which is shown in Equation 1 [7].

( ) 131 ln•

+=− εσσ ba (1)

Where 1

•

ε = strain rate and a and b = constants.

In late seventies relationship between deviator stress and rate of axial strain was investigated using strain controlled triaxial tests to study the effect of confining pressure on the strength characteristics with the variation of strain rate of soil [8]. Results from some laboratory tests suggested that at low strain rates, the gradient of the shear stress-log (strain rate) relationship would decrease markedly [6]. This threshold strain rate had been reported to be about 0.2% per hour for Haney clay and about 0.05% per hour for plastic Drammen clay. With this background a series of undrained triaxial tests were carried out on locally available clayey soil compacted at two different moisture contents to determine effect of strain rate on shear strength parameters. TEST PROGRAMME Routine tests were carried out to determine liquid limit, plastic limit and grain size distribution of soil used in the present investigation. Liquid limit = 50.5%, Plastic limit = 24.5% Grain size distribution: Sand = 0%, Silt = 58% and Clay = 42%. The optimum moisture content and maximum dry density of soil were 18.4% and 1.675 gm/cc. Two sets of triaxial tests were carried out to determine shear strength parameters of soil at different strain rate, cell pressure and moisture content as shown in Table 1. Table 1 Test Programme

Moisture content

Strain rate (mm/min)

Cell pressure (kPa)

Nos. of tests

18.4% 21.4%

1.5 0.405 0.081

0.0356 0.0162

150 200 250

30

Kingshuk Dan, R.B.Sahu

RESULTS AND DISCUSSIONS Stress - Strain Curve Two typical stress-strain curves at strain rate 1.5 mm/ min for moisture content 18.4% and 21.4% are presented in Figure 1 and Figure 2 respectively. Peak deviator stress is maximum for cell pressure 250 kPa and minimum for cell pressure 150 kPa in all the cases.

Moisture Content=18.4%, Strain rate=1.5 mm/ min

0

100

200

300

400

500

600

700

0 2 4 6 8 10 12

Axial strain (%)

Dev

iato

r Stres

s

(kPa

) 150 kPa

200 kPa

250 kPa

Fig. 1 Stress-strain curve from triaxial test for moisture content 18.4%

Moisture content = 21.4%, Strain rate =1.5 mm/ min

050

100150200250300350400

0 2 4 6 8 10 12

Axial strain (%)

Dev

iato

r stres

s (k

Pa)

150 kPa

200 kPa

250 kPa

Fig. 2 Stress-strain curve from triaxial test for moisture content 21.4% Peak Deviator Stress The variation of peak deviator stress (σ1–3)max, as obtained from stress-strain curves, with strain rates at moisture content 18.4% and 21.4% are presented in Figure 3 and Figure 4. From Figure 3 it may be seen that peak deviator stress for all the three cell pressure increases with the increase in strain rates for moisture content 18.4%. The straight lines for different cell pressure indicate rate of increase of (σ1–σ3)max is same for different cell pressures for the range of strain rate in the present investigation. From the Figure 4 it is seen that for moisture content 21.4% peak deviator stress increases upto a strain rate 0.0356 mm/min, then it decreases upto 0.081 mm/ min and beyond which it again increases. The variation of (σ1–σ3)max with strain rate can be represented by a set of straight lines to show that the rate of increase of (σ1–σ3)max is linear and nearly parallel for different cell pressure.

Moisture content = 18.4%

0

100

200

300

400

500

600

700

0.01 0.1 1 10

Strain rate (mm/ min)

Peak

dev

iato

r stres

s

(kPa

) 150 kPa

200 kPa

250 kPa

Fig. 3 Relationship between peak deviator stress and strain rate for moisture content 18.4%

Moisture content= 21.4%

050

100150200250300350400

0.01 0.1 1 10

Strain rate (mm/ min)

Peak

dev

iato

r stres

s

(kPa

)

150 kPa

200 kPa

250 kPa

Fig. 4 Relationship between peak deviator stress and strain rate for moisture content 21.4% This increase in peak deviator stress is due to the fact that excess porewater pressure developed in the specimen during shearing does not dissipate at higher strain rate leading to a non uniform distribution of excess pore water pressure in the soil sample which ultimately increases the peak deviator stress. Comparing slope of the straight lines in Figure 3 and Figure 4 it is further observed that the rate of increase has been reduced with the increase in moisture content of the soil. This is due to the reason that increases in moisture content decreases degree of saturation reducing peak deviator stress of the soil specimen. Shear Strength Parameters Using the deviator stresses for different strains Mohr circle diagrams are plotted from which shear parameters can be easily obtained. The shear parameters for different strain and strain rate are presented in Table 2 and Table 3. In these two tables cohesion ‘c’ is in kPa and friction angle ‘φ’ is in degree. From the Table 2(a) and Table 2(b) it is quite clear that ‘c’ and ‘φ’ increases with the increase in strain for different strain rates. For moisture content 18.4% at strain rate 1.5, 0.405, 0.081 mm/ min peak cohesion was mobilized at 2-3% while peak angle of shearing resistance was developed at 5-8% strain. For strain rate 0.0356 mm/ min both peak

A Laboratory Study on the effect of Strain Rate on Shear Strength of Cohesive Soil

cohesion and peak angle of shearing resistance were developed at 5% strain. For strain rate 0.0162 mm/ min peak cohesion is lesser and peak angle of shearing resistance is higher in comparison with other strain rates. Table 3(a) and Table 3(b) shows that at moisture content 21.4% for strain rates 0.405, 0.081, 0.0356 and 0.0162 mm/ min both peak cohesion and peak angle of shearing resistance were developed at nearly 5% strain. For strain rate 1.5 mm/ min peak cohesion was mobilized at 6-7% strain. Table 2(a) Shear strength parameters at moisture content 18.4%

Strain rate (mm/min) 1.5 0.405 0.081

Strain (%) c φ c φ c φ

1 80 13 65 11 35 13

2 115

12.5 90 12

.5 65 12.5

3 110 15 90 15 95 11.3

5 110 18 85 18

.5 95 16

7.5 125 18 95 20 95 18

10 135 18 95 20

.5 95 18.5

Effect of Degree of Saturation The variation of peak deviator stresses with degree of saturation for cell pressures 150, 200, 250 kPa and strain rates 1.5, 0.405 and 0.0162 mm/ min are shown in Figure 5. From these figures it is seen that for a certain increase in degree of saturation the reduction in peak deviator stress is higher at higher strain rate which reduces with the decrease in strain rate. Table 2(b) Shear strength parameters at moisture content 18.4%

Strain rate (mm/min) 0.0356 0.0162

Strain (%) c φ c φ 1 30 9 25 14 2 45 11.5 30 203 55 13 10 245 80 14.5 10 25.5 7.5 80 17 10 27 10 80 17.5 10 27

Table 3(a) Shear strength parameters at moisture content 21.4%


1.5 0.405 0.081

Strain (%) c φ c φ c φ

1 20 9 30 7 30 6 2 35 10 40 8 40 7 3 65 9 50 8 50 8 5 85 8 50 10.2 50 10 7.5 90 11 56 13.2 65 12 10 90 11 63 13.5 65 12

Table 3(b) Shear strength parameters at moisture content 21.4%


0.0356

0.0162

Strain (%) c φ c φ 1 45 4 30 5 2 60 5 50 5 3 55 7 55 6 5 65 7.5 60 8 7.5 80 8.1 60 11.8 10 70 12 60 12.8

Strain rate = 1.5 mm/ min

0100200300400500600700

60 70 80 90 100Degree of saturation (%)

Peak

dev

iato

r st

ress

(kPa

)

150 kPa

200 kPa

250 kPa

Kingshuk Dan, R.B.Sahu

Fig. 5 Change in peak deviator stress due to increase in Degree of saturation CONCLUSION From the present study following conclusions may be drawn: • The peak deviator stresses were found to increase with

the increase in strain rate. The increasing patterns for different cell pressures are nearly similar and these rates of increase were observed to decrease with the increase in moisture content.

• For soil specimen at OMC cohesion is mobilized at lesser strain while friction angle takes higher strain to mobilize. At higher moisture content ‘c’ and ‘φ’ were observed to mobilize simultaneously.

• At very low strain rate friction angle was found to be higher and cohesion to be lower with respect to r strain rates. So, it is appropriate to take ‘c’ and ‘φ’ corresponding to the actual strain rate as well as strain that may develop in the soil during and after construction.

REFERENCES 1. Bjerrum, L., Simmons, N.E., and Tarbela, I., (1958),

“The Effect of Time on the Shear Strength of a Soft Marine Clay”, Proceedings of the Brussels Conference on Earth Pressure Problems, Brussels, Belgium, Vol. I, pp. 148-158.

2. Casagrande, A., and Wilson, S. D.,(1951), “Effect of Rate of Loading on the Strength of Clays and Shales at Constant Water Content”, Geotechnique, London, England, Vol. II, 1951, pp. 251-263.

3. Crawford, C.B., (1959), “The Influence of Rate of Strain on Effective Stress in a Sensitive Clay”, American Society for Testing and Materials Special Publication No. 254, pp. 36-61.

4. Gibbs, H., J., et. al., (1960), “Shear Strength of Saturated Clays”, ASCE Research Conference on Shear Strength of Cohesive Soils, Boulder, Colo.

5. Gibson, R., E., and Henkel, D., J., (1954) “Influence of Duration of Tests at Constant Rate of Strain on Measured Drained Strength”, Geotechnique, London, England, Vol. 4, 1954, pp.6-15.

6. Graham, J., Crooks, J., H., A., and Bell, A. L., (1983), “Time Effects on the Stress-Strain Behaviour of Natural Soft Clays”, Geotechnique.

7. Mitchell, J., K., (1964), “Shearing Resistance of Soils as a Rate Process”, ASCE Research Conference on Shear Strength of Cohesive Soils, Boulder, Colo.

8. Roy, N., and Sarathi, P., (1976) “Strain Rate Behaviour of Compacted Silt” Proc. ASCE.

9. Taylor, D., W., (1943), “Ninth Progress Report on Shear Strength to U.S. Engineers”, Massachusetts Institute of Technology, Cambridge, Mass., 1943.


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Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No. A 134.)

EMPIRICAL CORRELATION & STABILIZATION OF EXPANSIVE SOIL USING FURNACE ASH FOR SURAT REGION

K. S. Berawala, Assistant Professor, [email protected]

ABSTRACT: Expansive clay soils, those that change significantly in volume with change in water content are the cause of distortions to structures that cost taxpayers several billion dollars annually in the India. This paper is based on some of the key advances developed over the past 60 years in improving our understanding of the nature and methods of modifying and stabilizing expansive clay soils. Hence to improve the strength of expansive soil of Surat region, furnace ash used as the additive which increase the stability of soil and decrease the swelling of soil. As furnace ash is high in silica, calcium, and other minerals is provides the necessary homogenous mass for performing the required test. Different tests are carried out with varying percentage of furnace ash to check the effect on swelling pressure and on basic properties. New correlations of swelling pressure have been carried out for the furnace ash stabilized expansive soil of Surat region by statically analysis using linear regression analysis method. This statistical analysis is carried out in order to obtain the most suitable relationships. New correlations are proposed for prediction of swelling pressure using liquid limit, plasticity index, shrinkage index, free swell index and percentage of furnace ash in different combinations for expansive soil of Surat city in the Gujarat state of India. If any research organization needs to examine, they can use this result for a quick solution for this region, unless there is extensive change in geological formation of the strata. INTRODUCTION Expansive soils, also called swelling soils, are those whose volume change takes place while it comes in contact with water i.e. expand during the rainy season due to intake of wear and shrink during summer season. An expansive soil covers nearly 20% of the landmass in India and includes almost the entire Deccan plateau, Western Madhya Pradesh, parts of Gujarat, Andhra Pradesh, Uttar Pradesh, Karnataka, and Maharashtra. The swelling soils are commonly known by the name of black cotton soils. A major concern in geotechnical engineering is identification of expansive soils and estimation of their swelling magnitudes when subjected to changes in environment. Surat city which is also called diamond city situated in state of Gujarat, India also contain black cotton soil. Generally main properties of expansive soil are basic index properties and swelling properties like free swell index and swelling pressure which directly affect the bearing capacity and strength of foundation lying on such a soil. Typical behavior of swell and shrink of expansive soil generate many problem like cracking in foundation. RANGE OF SOIL PROPERTIES [4K. S. Berawala and C. H. Solanki, 3Holtz W. G. and Gibbs H. J.] By From the previous work of identification of soil for Surat region of Gujarat state, India, the following properties of soil are collected. These properties are then compared with already identified range of properties of expansive soil given by Katti for country India. The average range of soil properties and comparison is shown in table below:

Table 1: Range of Soil Properties & Comparison

Most of the ranges of properties given by Katti are matched with performed properties for Surat city. So on that basis soil of Surat city is concluded as an expansive. A- LINE CHART [1Dakshanamurthy V. and Raman V.] As a convenience for comparing a variety of soils, Dr. A. Casagrande and V. Raman revised a plasticity chart.

Properties Range Katti’s range

Gravel % 0 - 10.0 -

Sand % 2.0 - 30.0 1.0 – 26.0

Silt+Clay % 70.0 - 100.0 -

Plastic Limit (Wp) % 20.0 - 35.0 20.0 – 50.0

Liquid Limit (WL) % 45.0 - 70.0 50.0 – 90.0

Plasticity Index (Ip) % 20.0 - 40.0 20.0 – 50.0

Dry Density g/cc 1.00 - 1.55 1.3 – 1.4

Field Moisture Content (%) 15.0 – 35.0 25.0 – 35.0

Specific Gravity (G) 2.5 - 2.9 2.7 – 2.9

Shrinkage Limit % 7.0 - 20.0 7.0 – 30.0

Free swell index 40.0 - 85.0 -

Shrinkage Index 30.0 - 60.0 -

Soil group CH,CI,MH CH

K S Berawala

Fig. 1 A-line chart From reference of chart given by Dr. A. Casagrande and V. Raman, the figure above show the A line chart in which test results range between 45% to 70% which gives the indication of medium to high plasticity and swelling of soil hence the soil is identified as high swelling and high plastic soil. DAMAGE TO FOUNDATIONS FROM EXPANSIVE SOILS [2Dr. Robert M. Brooks; 6S. Bhuvaneshwari, R. G. Robinson, S. R. Gandhi ] The most obvious way in which expansive soils damage foundations is by uplift as they swell with moisture increases. Swelling soils lift up and crack lightly-loaded, continuous strip footings, and frequently cause distress in floor slabs. Such differential movement of the foundation can also cause distress to the framing of a structure. To stabilize expansive soil, the waste product (bagasse) furnace ash is collected from boiler / furnace bed of sugarcane located at Sayan located near Surat, Gujarat. Following are the chemical properties of furnace ash. Table 2. Properties of Furnace Ash

Description Abbreviation Percentage (%) Silica Sio2 60.26 Iron Fe2o3 5.03

Calcium Cao 8.35 Magnesium Mgo 0.40

Sodium Na20 1.33 Potassium K2o 5.57 Chloride Cl 0.20 Sulphate So4 1.30

Phosphorus P2o5 2.69 Loss of Ignition - 3.39

Alumina Al2o3 10.73 Titanium Tio2 0.13

Manganese Mn 0.078 Wax Content - Nil

INDEX PROPERTIES OF THE STABILIZED SOIL: The soil samples were taken from 14 feet below the ground level of VESU area of Surat region,. Various tests like Liquid

Limit, Plastic Limit, Plasticity Index, Shrinkage Limit, Free Swell Index and Swelling Pressure performed. The percentage of furnace ash is kept 0 %, 3%, 5%, 7% and 10% respectively and all the tests are conducted. The results show that when the percentage of furnace ash is increased in the soil sample, all the properties decrease. The table below shows the laboratory test results of soil properties for region “VESU”. Table 3. Test Result

Where, Furnace ash (%) =Percentage of Furnace Ash F.S.I=Free Swell Index L.L=Liquid Limit P.L=Plastic Limit P.I=Plasticity Index S.L=Shrinkage Limit S.I=Shrinkage Index Sp=Swelling Pressure GRAPHICAL REPRESENTATION OF TEST RESULT: From the graph below, we analyze that Free Swell Index decreases as Percentage of Furnace Ash increases; and shows some linearity between them. Fig. 2 Relation between Free Swell Index & Percentage of Furnace Ash Swelling pressure is very important parameter of expansive soil. The factor affecting the swelling pressure of soil is percentage of clay mineral, percent of particle passing through 425 micron size, other basic properties such as LL, PL, FSI, SL, PI, water content etc. From the graph below, we analyze that Swelling Pressure decreases as Percentage of Furnace Ash increases; and shows some linearity between them.

Sr. No

Furnace ash %

L.L %

P.L %

P.I %

S.L %

F.S.I Sp Kg/Cm2

S.I %

1 0 72 30 42 21 143 0.120 21 2 3 67 29 38 19 127 0.099 19 3 5 63 28 35 17 103 0.084 18 4 7 58 26 32 15 85 0.074 17 5 10 52 25 27 12 80 0.046 15

(%)

Empirical correlation and stabilization of expansive soil using furnace ash for surat region

Fig. 3 Relation between Swelling Pressure & Percentage of Furnace Ash

REGRESSION ANALYSIS USING MATRICES IN EXCEL SHEET: LINEAR REGRESSION: [4K. S. Berawala and C. H. Solanki ;7Srirama A. Rao and Phani Kumar B.R.; 8Yusuf Erzin and Orhan Erol,] Based on experimental data obtained from the several test conducted on the samples from the 12 sources, correlation have been developed by Regression Analysis using matrices in Excel sheet. In regression analysis coefficient of an equation relating predicator variables (independent variable) with criterion variable (dependent variable) are evaluated. To obtain the correlation of swelling pressure with other properties step by step procedure was done using different combination of soil properties with % of furnace ash. And one final correlation is developed for the best combination of soil properties. RESULT: Using mathematical regression analysis, correlation between swelling pressure and other soil properties has been established for the Surat region as shown below. Sp (Kg/cm2) = C0+C1(LL)+C2(PI)+C3(SI)+C4(FSI)+C5(%)

(2) R2 = 0.97437

Where, C0 = 4.75; C1 = 0.01855; C2 = -0.0684; C3 = -0.1172; C4 = -0.004; C5 = -0.2109 CORRELATION COEFFICIENT (R): Correlation coefficient measures how well the least square regression line fits the sample data. If the total variation is all explained by the regression line i.e. R= ±1, we say that there is perfect linear relation and if the total variation is all unexplained R is zero i.e. R= 0.Correlation analysis provides a means of drawing interferences about the strength of the relationship between two or more variables i.e. it is a measure of a degree to which, values of these variables vary in a systematic manner. Thus it provides a quantitative index of the degree to which one or more variables can be used to predict the value of another variable. The correlation coefficient for a linear multivate model is calculated by the following equation.

(1) Where, Y = dependent variable X = Independent variable N = number of observation

CONCLUSION Soil stabilization method by applying waste product furnace ash was successfully applied to improve the existing poor and expansive sub grade soil. Furnace ash is free of cost and available locally, hence it proved economical also. Furnace ash effectively dries wet soils and provides an initial rapid strength gain, which is useful during construction in wet, unstable ground conditions. Furnace ash also decreases swell potential of expansive soils by replacing some of the volume previously held by expansive clay minerals and by cementing the soil particles together. The statistical analysis was carried out in order to obtain the most accurate and suitable relationship. The liquid limit, plasticity index, shrinkage index, free swell index and Percentage of Furnace Ash are correlated with swelling pressure excellently and have a considerable impact on predicting swelling pressure value. Also the equation of swelling pressure connecting different soil properties will be applicable to this region in future, unless there is extensive change in geological formation of the strata.

REFERENCE 1. Dakshanamurthy V and Raman V (1973), A Simple

Method of Identifying an Expansive Soil. Soils and Foundations, Jap. Soc of Soil Mech. and Foundation Engineering Vol. 13 No. 1 pp 97-104.

2. Dr. Robert M. Brooks, (2009), Soil Stabilization With Fly ash And Rice Husk Ash, International Journal of Research and Reviews in Applied Sciences, ISSN: 2076-734X, EISSN: 2076-7366, Volume 1, Issue 3.

3. Holtz W. G. and Gibbs H. J. (1956), Engineering properties of expansive clays” Transactions, ASCE pg. 121, 641-677.

4. K. S. Berawala and C. H. Solanki (2010), Empirical Correlations Of Surat Region Expansive Soils Parameter Based on Swelling Characteristics, Indian Geotechnical Conference , GEOtrendz.

5. Salma Tawfiq and Zalihe Nalbantoglu,(2009), Swell-shrink behavior of expansive clays” 28-30,pg. 336-340.

6. S. Bhuvaneshwari, R. G. Robinson, S. R. Gandhi (2005) , “Stabilization of Expansive Soils Using Flyash.” Fly Ash India, New Delhi

7. Srirama A. Rao and PhaniKumar B. R., “Correlation studies for swelling characteristics of expansive soils”, pg. 59-63.

8. Yusuf Erzin and Orhan Erol, “Correlations for Quick Prediction of Swell Pressures” EJGE paper 2004-0476.

Kg/cm2


CONSOLIDATION OF DOUBLE POROSITY CLAYS

Anurag Chafale, Postgraduate Student, Department of Civil Engineering, IITB, Email: [email protected] R Reddy Palle, UG Student, Department of Civil Engineering, G.P.R.E.C. Email: [email protected] Ashish Juneja, Associate Professor, Department of Civil Engineering, IITB, Email: [email protected]

ABSTRACT: Restoration of sites previously dumped with dredged clays and tailing sludge is a serious geotechnical challenge. Much of these soils exist as lumps suspended in slurry. Consolidation of this heterogeneous fill occurs because of expulsion of water from pores within the lumps and pores in between the lumps. Terzaghi's classical consolidation theory is unable to explain these complex characteristics of the "double-porosity" fill. In this paper, permeability results of clay lumps are discussed. The tests were conducted using flexible-wall permeameter. In this procedure, clay cubes were trimmed from a large block of clay. The cubes were then carefully arranged in a rubber membrane stretched over a split mould. The entire assembly was then placed in the permeameter and the lumps were allowed to consolidate at the predefined effective stress. The size and arrangement of the lumps, and the mean effective stress were the variables in the tests. Using the test results, the relationships between the effective stress, volume change and the permeability were derived. INTRODUCTION It is necessary to find an economical and environmentally friendly source of soils for land restoration use near coastal cities. In many of these projects, dredge clays and soil wastes from underground excavations have replaced the traditional sand based fill because of huge cost savings. Behaviour of the above "lumpy" fill is quite different from homogenous soils because the fill contains both intra-lump and inter-lump pores. Fig. 1 shows the lumps with double porosity. Consolidation of this heterogeneous fill occurs because of the expulsion of water from pores within the lumps and pores in between the lumps. Unfortunately, Terzaghi's classical consolidation theory is unable to explain these complex characteristics. This paper discusses the relationship between the mean effective stress and volumetric strain of clay lumps.

Fig.1 Double porosity model of lumpy clay

BACKGROUND Finite consolidation studies using clay lumps have traditionally been conducted using a rigid wall oedometer. Although these tests have the advantage of simplicity, yet they do not account for the rigid boundary effects [2]. In addition, the clay lumps are subjected to one-dimensional (1-

D) loading in oedometer, which does not truly represent the conditions of finite loaded area and when the boundary of the restored area is examined. Such boundary effects can be reduced by the use of an enlarged oedometer, but these effects cannot altogether be eliminated [3].More recently, some investigators have also observed that the stress transfer along the interface between the lumps and the rigid boundary in laboratory tests can be significant when compared to the inter-lump stresses measured inside the sample. Ignoring the above boundary effects can lead to serious errors especially since the porosity of the soil matrix is complex. The present study aims to address some of the above issues. The tests were conducted using the flexible wall permeameter. The combined volume-change of the intra-lump and the inter-lump void space was measured using the air-water interface which is commonly used in routine triaxial testing. Using this simplified procedure, an attempt was also made to approximately compare the results with Darcy's law of soil permeability. All dimensions in the figures are in mm unless and otherwise stated. EXPERIMENTAL SETUP Fig. 2 shows the experimental setup of the flexible wall permeameter used in this study. The apparatus consisted on a permeameter cell, volume change measurement, air-water interface, pore pressure transducers and data logger. Details of each of the apparatuses are briefly explained below. Permeability Cell Fig. 3 shows the schematic diagram of the permeability cell. It consisted of an aluminium cell base with a cylindrical cell made of Perspex. The top cap was originally made of aluminium. It was later replaced by a Perspex cap to reduce the dead weight on the clay lumps. It also helped to provide stability to the stack of lumps during the sample preparation.

Chafale A.,Palle R. R., Juneja A.

The top cap was then placed over the soil sample and the entire assembly mounted over the pedestal.

Fig.2 Setup of Flexible wall permeameter where, A is Permeability cell, B is Volume change measurement apparatus, C is Pressure control panel and D is Data logging system, respectively.

Four drainage lines were provided at the two ends of the sample. These "double drainage lines" helped provide flexibility in controlling the drainage and simultaneously measuring the pore pressure in the sample. In essence, using this procedure, a line from top as well as the bottom was connected to the pressure panel via the volume change apparatus. This helped to provide controlled pressure at both top and bottom of the sample and also create hydraulic gradient. The second set of drainage lines was attached to the pore pressure transducers to measure the pore pressures at the top and the bottom of the sample.

Fig.3 Schematic diagram of Flexible Wall Permeameter Cell

(After David Daniel et al. 1984)

Volume Change Measurement Apparatus: The volume change apparatus was a 100ml cylinder with piston assembly. The upward and downward flow could be regulated using this apparatus. Two such devices were used to measure the change in volume at the top and the bottom of soil sample. The first device was attached to the top of the sample using the top drainage line and second device was attached to the bottom of sample using the bottom drainage line. The inflow or the outflow was estimated using the change in volume given by the two devices. Air-Water Interface: The air-water interface helped to create a uniform and constant pressure in the cell as well as at the two ends of the soil sample throughout the test. Pore Pressure Transducer: The pore pressure transducers (PPTs) were used to measure the pore pressures at the top and the bottom of the sample. This helped to deduce the head loss between the two ends. An attempt was also made to estimate the hydraulic gradient after each stage of the consolidation. Data logger: Analogue signals from the PPTs were digitised using the commercial software Data-man. The sampling rate was set at 1 sample per second. No attempt was made to amplify or smooth the signals. METHODOLOGY Methodology of the test consist of two stages

1) Preparation of sample 2) Test procedure

Sample Preparation: All tests were conducted using commercially available Kaolin clay. Table 1 shows the physical properties of the clay used in this study. In this procedure, a cylindrical block of clay was consolidated from de-aired kaolin slurry using a pneumatic press. Small clay cubes of 10mm and 30mm side were then prepared by trimming the large clay block using a wire saw. Initially, a few samples were prepared by pressing and rubbing thin layers of clay near its plastic limit in the mould. The layers were smeared using the thumb. Although this procedure helped produced the sample within a short period, it also produced samples of unknown consistency. This method was therefore discontinued for all other tests. Nonetheless, the effect of sample preparation on its yield locus was not part of this investigation. Fig. 4 shows the photograph of cubes of 30mm side. The cubes were arranged in layers over the bottom pedestal and then enclosed by a membrane using a membrane stretcher. The number and orientation of the clay lumps was fixed to maintain the dimensions of the sample used in the permeameter. The

Effective stress and permeability of double porosity clays

details of the permeameter and other equipment used in this study are summarised in the following paragraphs. Figs. 5a and b show the photographs of the sample before and after the test.

Fig.4 Cubes of 30mm obtained by cutting with a wire saw

Table 1: Physical properties of Kaolin clay Sr. No. Properties (%)

1 Liquid limit 42.38 2 Plastic limit 32.06

Test Procedure: The normal procedure for testing the soil specimen consist of two stages

1) Consolidation Stage 2) Permeation Stage

Consolidation Stage: In this step, the average effective stress was obtained as the difference between the confining pressure and the back pressure. A confining pressure of 200kN/m2 and a back pressure of 100kN/m2at the top and the bottom of the sample, respectively, were used to consolidate the sample. During the consolidation process, the change in volume at top and bottom as well as the change in pore pressures were logged. Consolidation was deemed to be complete when there was no change in volume at top and bottom. Permeation Stage: After completion of the consolidation stage, ahead difference of about 40 kN/m2 was provided to permit the pore water to drain away from the sample. The head difference was applied between the top and the bottom of the sample. This stage was conducted as per ASTM D (5084). In essence, the following procedure was adopted to obtain the volumetric strains in the sample:

1) The cell pressure was increased from 200 to 600 kN/m2 while the back pressure was maintained constant at

100 kN/m2. This helped to generate an effective stress in between 100 and 500 kN/m2 in the samples

2) The change in volume of the sample was measured using the volume change device by maintaining a back pressure of 120kN/m2at top and 80kN/m2at bottom of the sample.

(a) (b)

Fig.5 (a) Small clay cubes before consolidation, (b) clay cubes after consolidation

3) The above step 2 was maintained until the change in volume of the sample reached a constant value. Volumetric strains were calculated throughout this step.

4) An attempt was made to estimate the combined permeability of the clay lumps using the volumetric strains obtained in the above step 3. These results were compared to the permeability measured in a 100mm diameter and 100mm long single kaolin clay lump for which the soil permeability was well established (Mir 2009). The following equation was used in the analysis

where, K is Coefficient of permeability, ΔV is Change in volume over a time t, Δh is head loss, L and A is post consolidation length and area of sample, respectively. RESULTS AND DISCUSSION Fig. 6 shows the variation of the average volume change measured at the top and the bottom of the sample with time in the sample prepared using 30mm size clay lumps. The mean effective stress was increased from 100 to 500 kN/m2. The figure shows that the gradient continuously reduced with the change in mean effective stress during the consolidation. This difference is likely due to the presence of inter-lump and intra-lump voids in the soil sample. As the mean effective stress increased, the voids between any two adjacent lumps decreased. Subsequently at higher mean effective stress, the voids within the lump became significant and had a

Chafale A.,Palle R. R., Juneja A.

dominating role in the volume change. Experiments were also conducted to compare the behaviour of single homogenous clay lump and a group of small clay lumps with inter-lump voids. It was similarly observed that rate of volume change in small clay lumps was more when compared to the single clay-lump. Fig. 7 shows the results between the void ratio and the effective stress.

Furthermore, an approximate relationship between the mean effective stress and the soil permeability was derived. It may be noted that not all assumptions of Darcy's law are truly applicable in our present study. Fig. 8 shows the plot of the results on the small clay lumps and single homogenous clay lump. The figure shows that the variation of the soil permeability using small clay lumps was high when compare to the single clay lump because of the presence of inter-lump voids. As inter-lump voids closed at high mean effective stress, the results of both the small clay lumps and of the large single block of clay are nearly the same. This occurred when the mean effective stress was about 300 kN/m2.

Fig. 6 Time vs Volume change for small lumps in

Consolidation stage

Fig.7 Void Ratio vs Effective Stress

Fig.8 Effective Stress vs Permeability

CONCLUSION: Consolidation characteristic of this material was investigated. The results were also compared to that of a homogenous clay sample for which Darcy's laws are well established. Only preliminary results have been presented in this paper and further investigation is still on-going. When there was no change in volume, it was assumed that the applied mean effective stress was fully utilised during the consolidation. It was also observed that the increase in mean effective stresses caused the soil permeability to reduce because of the reduction in the volume of the voids. Upon comparing the variation of the permeability of small clay lumps and single homogenous clay, it was observed that the change in permeability was significant in the case of sample made of small lumps. The permeability between the lumps and homogenous clay was similar when the mean effective stress was more than 300 kN/m2. However, this mean effective stress is usually higher than that observed in land restoration projects. REFERENCES

1. American Society for testing and Materials (ASTM).(2010), Standard Test Methods for Measurement of Hydraulic Conductivity of Saturated Porous Materials Using a Flexible Wall Permeameter. ASTM D 5084, Annual Book of ASTM Standards.

2. David E. Daniel, Stephen J. Trautwein, Stephen S. Boynton and David E. Foreman, Permeability testing with Flexible-Wall Permeameters, Goetechnical Testing Journal, GTJODJ, Vol.7, No. 3, Sept 1984, pp.113-122

3. Mir, B.A. and Juneja, A. (2009). "Some mechanical properties of reconstituted kaolin clays", 17th

Southeast Asian Geotechnical Conference, Taipei (Li and Lin eds.), Vol. 1, 145-148.

4. Robinson R. G., Dasari G. R. And Tan T. S. (2005), Experimental study of the Behaviour of a lumpy fill of soft clay, International journal of geomechanics (ASCE), Vol.5, No.2, pp125-137.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No. A 139)

EFFECT OF UNCERTAINTIES IN SITE CHARACTERIZATION USING SURFACE WAVE TECHNIQUES

Narayan Roy, Research Scholar, Dept. of Earthquake Engg., IIT Roorkee, India, Email: [email protected] Ravi Sankar Jakka, Assistant. Professor, Dept. of Earthquake Engg., IIT Roorkee, India, Email: [email protected] H.R. Wason, Professor, Dept. of Earthquake Engg., IIT Roorkee, India, Email: [email protected]

ABSTRACT: In geophysical site characterization using surface wave techniques, inversion is the most important step. In the analysis, experimental dispersion curve which is generated from the field measurements is matched with theoretical dispersion curves to get shear wave velocity profiles. The main uncertainty associated with surface wave techniques is in the inversion process which can provide several equivalent profiles and consequently it leads to different local seismic responses. In this paper, the effect of uncertainty in one-dimensional ground response analysis is studied by using a realistic ground motion input. Equivalent shear wave velocity profiles are selected from inversion using neighbourhood algorithm on the basis of low misfit value with respect to target dispersion curve. Then these equivalent profiles are subjected to conventional one-dimensional ground response analysis using software SHAKE2000 in the frequency range of engineering interest. Amplification and response spectra show significant difference and mean coefficient of variation of amplification spectra is as high as 20%. INTRODUCTION Surface wave method is used to geotechnical site characterization on the basis of shear-wave velocity profiles. Due to dispersive nature of Rayleigh wave, different frequency wave travel at different velocity in a layered medium and penetrates up to different soil thicknesses. As a result of the variation of the shear stiffness of the layers, waves with different wavelengths of frequency travel at different phase velocities. The applications of surface waves in engineering field started in the 1950s with the Steady State Rayleigh Method [1], but their revolution arrived only in the last two decades with the SASW method [2] and MASW [3]. In surface wave methods, experimental dispersion curve is developed from field data using different processing techniques. This experimental dispersion curve is then used for inverse problem solution to get shear wave velocity profiles. The solution of the inverse problem is non-unique and provides several equivalent velocity profiles. Its final model strongly depends on the initial one. The surface wave data measurement uncertainty has been studied by different researchers [4-6]. Uncertainty in measured shear-wave velocity profile has also been investigated in some literature [7,6,8]. Some research has been carried out on the effect of the inversion uncertainty on the seismic ground response analysis. Foti et al. [9] showed that the effect of the surface-wave inversion uncertainty is not significant in seismic ground response analysis. Later this study was extended by Boaga et al. [10] for different impedance contrast and found out that the equivalent profiles as a results of surface wave inversion are not equivalent in terms of seismic ground response analysis. In this paper, an attempt has been made to study the effect of inversion

uncertainty on seismic site response analysis using a realistic earthquake record for a low impedance contrast profile. SURFACE WAVE METHODS Different types of surface wave methods are used for constructing the dispersion curve. Active-source tests, in which waves are generated using a seismic source [11,3]. In passive-source tests, constant vibration of earth’s surface or microtremor [12,13] is used for the analysis. The main difference in active and passive-source tests is in frequency components which are directly related to the depth of investigation. Generally active-source tests are associated with high frequency components and in passive-source tests low frequency components are obtained. Sometimes both active-source and passive-source tests are used together for getting profile up to larger depths and better resolution at lower depth [14]. NEIGHBOURHOOD ALGORITHM IN SURFACE WAVE INVERSION The neighbourhood algorithm is a stochastic direct-search method for finding models of acceptable data fit inside a multidimensional parameter space [15]. A set of pseudo-random samples is generated after defining the variation of each model parameters (thickness and shear-wave velocity of each layer) in the parameter space. This set of samples is then processed to get the dispersion curves by using the forward problem algorithm for fundamental mode of Rayleigh wave propagation considering soil column as a stack of horizontal and homogeneous layers. Once the theoretical dispersion curve is developed from the random samples given by the neighbourhood algorithm, the misfit value is calculated. If the experimental dispersion curves

Narayan Roy, Ravi Sankar Jakka & H.R. Wason

are associated with an uncertainty estimate, the misfit is given by Eq. (1) as below

Misfit = (1)

where is the theoretical and is the experimental phase velocity of the calculated curve at frequency fi, is the uncertainty of the frequency samples considered and is the number of frequency samples considered. If no uncertainty is provided, is replaced by in above equation. The details about the procedure are described in the literature [16]. As a result of the inversion process, we get a set of shear-wave velocity profiles based on the misfit value. Higher the value of misfit, higher will be the set of selected profiles. SYNTHETIC ANALYSIS A synthetic study has been carried out to find out impact of non-uniqueness on seismic response of soil column using a realistic input motion. First we have taken a reference velocity profile consisting of three layers plus half-space with gradually increasing shear-wave velocity reported in Table 1. Theoretical dispersion curve is generated for the profile using the forward problem. Poisson ratio and unit weight are same for all layers (Poisson ratio- 0.33 and density-1950 kg/m3) because these parameters has a very little influence on Rayleigh wave dispersion. The neighbourhood of the dispersion curve is defined from previous study [6] so as to take an allowable range of standard deviation of the target dispersion curve. After the inversion 76 nos. of equivalent shear-wave velocity profiles are obtained (Fig. 1a). The dispersions curves developed from each equivalent profiles is shown in Fig. 1b which shows a good fit with the target dispersion curve. Within this 76 nos. of profiles, first 15 best fitting profiles having a misfit of less than 0.4 are

selected (Fig. 2) for one-dimensional ground response analysis. Table 1 Reference velocity profile

Thickness Shear-wave velocity (m/s) 5 180 7 240 12 300

Half-space 360 For shaking analysis, an earthquake record has been used as an input motion in the analysis. The earthquake data used, is taken from K-NET of magnitude 6.6 in Japan. The typical acceleration time history has shown in the Fig. 3. RESULTS The result of ground response analysis shows significant differences in the amplification spectrum in terms of amplification as well as in peak frequency also (Fig. 4a). Amplification varies from 7.2 to 9.2 and peak frequency varies from 1.4 to 2.6 Hz. For soil condition, when shear-wave velocity slowly increases with depth (i.e., low impedance contrast), it exhibits different ground response analysis. So, in this type of soil conditions, non-uniqueness of surface wave inversion may contribute significantly different ground motion. Large variation is also observed for response spectra (Fig. 4b). Variation in peak spectral acceleration is between 0.4g to 0.86g. Surface wave inversion uncertainty can ultimately lead to significant differences in the geotechnical site characterisation.

Fig. 1(a) Equivalent shear-wave velocity profiles (b) Equivalent dispersion curves with target dispersion curve (black)

Effect of uncertainties in site characterization using surface wave techniques

Fig. 2 First 15 best fitting shear-wave velocity profiles

Fig. 3 Input-motion (Magnitude: 6.6, Date: 2012/03/27-20:00:00.00, Latitude-39.80N, Longitude-142.33E)

Fig. 4(a) Comparison of amplification spectra from equivalent velocity profiles (b) Comparison of response spectra from equivalent velocity profiles

Fig. 5(a) Coefficient of variation of amplification spectra with respect to frequency (b) Coefficient of variation of response spectra with respect to frequency To quantify the uncertainty, a statistical study has been carried out to show the relative variation of different

spectra with frequency. Very high value of COV (mean 20%) is observed for the amplification spectrum (Fig. 5a)

Narayan Roy, Ravi Sankar Jakka & H.R. Wason

derived from best fitting 15 equivalent profiles and high data scatter is found out up to 10 Hz frequency. For response spectra, the COV observed (Fig. 5b) is little bit lower (Mean COV is 17%) and it also shows high data scatter up to 5Hz frequency. DISCUSSION AND CONCLUSIONS Surface wave method becoming very much popular from geophysical and geotechnical point of view but a very few research has been carried out to find out the uncertainty associated with method. In this paper, an effort has been made to quantify the inversion uncertainty on site characterisation. The findings obtained from the entire study can be summarized as follows:

Equivalent profiles, results of surface wave inversion are not equivalent in terms of seismic ground response analysis. While the one recent study claim that amplification spectra shows variation in terms of peak frequency but our study shows significant variation is observed in amplification amplitude also.

Statistical analysis shows very high value of mean of COV for amplification spectra (20%) and a little low values of mean COV (17%) is observed for response spectra.

Surface wave inversion uncertainty has significant impact on site characterisation and it can mislead the calculation of design ground motion.

REFERENCES 1. Jones, R.B. (1958), In-situ measurement of the

dynamic properties of soil by vibration methods, Geotechnique, 8 (1), 1-21.

2. Nazarian, S., Stokoe, K.H.II and Hudson, W. R. (1983), Use of spectral analysis of surface waves method for determination of moduli and thicknesses of pavement systems, Transp. Res. Rec., 930, 38–45.

3. Park, C.B., Miller, R.D. and Xia, J. (1999), Multi-channel analysis of surface waves, Geophysics, 64(3), 800-808.

4. Xia, J., Miller, R.D., Park, C.B., Hunter, J.A., Harris, J.B., and Ivanov, J. (2002), Comparing shear-wave velocity profiles inverted from multichannel surface wave with borehole measurements, Soil Dyn. Earthquake Eng., 22, 181–190.

5. Marosi, K.T. and Hiltunen, D.R. (2004a), Characterization of SASW phase angle and phase velocity measurement uncertainty, Geotech. Test. J., 27(2), 205–213.

6. Lai, C.G., Foti, S. and Rix, G.J. (2005), Propagation of data uncertainty in surface wave inversion. J. Environ. Eng. Geophys., 10(2), 219–228.

7. Marosi, K.T. and Hiltunen, D.R. (2004b), Characterization of SASW shear wave velocity measurement uncertainty, J. Geotech. Geoenviron. Eng., 130(10), 1034–1041.

8. Strobbia, C. and Foti, S. (2006), Multi-offset phase analysis of surface wave data (MOPA), J. Appl. Geophys, 59, 300–13

9. Foti, S., Comina, C., Boiero, D., and Socco, L.V. (2009), Non uniqueness in surface wave inversion and consequences on seismic site response analyses, Soil Dyn. Earthquake Eng., 29(6), 982–993.

10. Boaga, J., Vignoli, G. and Cassiani, G. (2011), Shear wave profiles from surface wave inversion: the impact of uncertainty on seismic site response analysis, J. Geophys. Eng., 8, 162–174.

11. Stokoe, K.H.II., Wright, S.G., Bay, J.A., and Roesset, J.M. (1994), Characterization of geotechnical sites by SASW method, Geophysical characterization of sites, R. D. Woods, ed., Oxford & IBH Publishing, New Delhi, India15–25.

12. Horike, M. (1985), Inversion of phase velocity of long-period microtremors to the S-wave velocity structure down to the basement in urbanized areas: Journal of Physics of the Earth, 33, 59-96.

13. Tokimatsu, K. (1995), Geotechnical site characterization using surface waves.” Proc., First Int. Conf. on Earthquake Geotechnical Engineering, IS-Tokyo ’95, Japanese Geotechnical Society, Balkema, Rotterdam, Netherlands, 1333–1368.

14. Rix, G.J., Hebeler, G.L., and Orozco, M.C. (2002), Near surface vs profiling in the New Madrid Seismic Zone using surface wave methods, Seismol. Res. Lett., 73(3), 380–392.

15. Sambridge, M. (1999), Geophysical inversion with a neighbourhood algorithm –II. Appraising the ensemble, Geophys. J. Int., 138, 727-746.

16. Wathelet, M., Jongmans, D., and Ohrnberger, M. (2004). “Surface-wave inversion using a direct search algorithm and its application to ambient vibration measurements.” Near Surf. Geophys., 2(4), 211–221.


HOLISTIC STUDY ON SITE CHARACTERIZATION FOR HABITAT DEVELOPMENT A. Ghosh, Chief Scientist, Central Building research institute [email protected]

P. K. S. Chauhan Principal Scientist, Central Building Research Institute, [email protected] S. K. Jain, Principal Scientist, Central Building Research Institute, [email protected] Dalip Kumar, Principal Technical officer, Central Building Research Institute, @yahoo.com

Zameer Ahmed, Senior Technical officer, Central Building Research Institute, @yahoo.com

ABSTRACT: A large area at the foothills of Himalaya has been acquired for developing habitation wherein multistorey towers comprising of flats are proposed to be constructed and some part of the area will be delivered to individual clients as plots for their own construction. In view of the constructions proposed, for holistic site characterisation, geological study, geotechnical investigation, electrical resistivity survey, seismic survey using Engineering Seismograph, ambient vibration study using Strong Motion Accelerograph were carried out. INTRODUCTION A large infrastructure development company proposed to develop a real estate site on the foothills of Himalaya. In this connection the company has acquired large area where multistorey towers will be constructed and some part of the area will be delivered to individual clients as plots for their own construction The company approached Central Building Research Institute, Roorkee for the holistic site characterisation. Through discussion it was felt that for such type of construction in the seismic zone V, Geotechnical, Geological, Geophysical, and Ambient vibration study was required so that complete details of the site could be obtained. The data so generated will also help the designer to design the towers. The company decided to sponsor the study and consequently study has been taken up for the site. Under this study, general geological study, geotechnical investigation, electrical resistivity survey, seismic survey using Engineering Seismograph, ambient vibration study using Strong Motion Accelerograph were carried out. GEOLOGICAL STUDY Old Doon Gravel Old Doon Gravel near the foot hills of Lesser Himalaya and Siwalik is the old remnant of fan deposits of Post Siwalik age. Lithologically it consists of big boulders, pebbles, gravels, embedded in reddish yellow brown clayey matrix lying unconformably over the Middle and Upper Siwaliks, showing poor stratification and forming massive hilly topography elongated in north-south direction. The clasts are angular to sub rounded, made up of quartzite and sandstone, indicating provenance from Lesser Himalaya and Siwalik range. These isolated hillocks are characterized by thick vegetative cover mostly of sal trees as observed around near by Reserved Forest in the southeast and Reserve Forest in the north.

The proposed site is situated on a terrace on the right bank of a river. The geological formation is Old Doon Gravel is an unconsolidated strata with rock particles of shales, lime stone and quartzite of the size of cobble, pebble and gravel in a sandy and clayey matrix. Doon Fan Gravel The unit is comprised of boulders of smaller size, pebble, gravel, sand, silt, clay and chips of phyllites and slates ranging in age from Late Pleistocene to Early Holocene. Materials composition consists of quartzite, schist, phyllite and slate of Pre-Tertiary formations and sandstone, conglomerate and shale of Siwalik Group. Reddish and yellow clay patches are also observed in this formation. The local conglomerates which are considered to be the characteristic of Doon Fan Gravel found in Dehradun city. River Alluvium This unit includes river terrace which is of sub recent in age comprising of boulder, gravels, pebbles, sand, silt and clay. It is developed on the lower parts of the piedmont zone. In Doon valley this unit has occupied an extensive part of the different rivers such as Tons, Gauna, Suwarna and Yamuna. GEOPHYSICAL SURVEY

To know the thickness of the gravel deposit at the site geophysical survey has been carried out. However, the detailed stratification of the area could be unfolded after the drilling. Resistivity Survey Electrical Resistivity method is a highly effective geophysical techniques used for subsurface investigations. The resistivity of a geological formation depends mostly on porosity, amount of water contained in the pores, salinity and geomaterial quality. In general hard and compact materials give high resistivity than the weathered formations.

Ghosh. A, Sarkar. S, Kanungo. D.P, Chauhan, P.K.S , Jain,, S.K Sharma Shaifaly, Kumar Dalip and AhmedZameer

The resistivity sounding was conducted at four sites in the proposed site area using Schlumberger configuration. In Schlumberger configuration the current and potential electrodes pairs are placed at equidistance from the centre in a straight line on the surface and the distance between current electrodes is always more than five times of the distance between potential electrodes. In both the configurations the potential electrodes are usually in between the current electrodes. The analysis of the data suggests that the resistivity values are decreasing with depth. The top 12 to 19m layers have higher resistivity values as compared to the soil strata below. The interpreted curves are shown in figure 1.

Fig 1. Typical resistivity curve of the site.

SEISMIC SURVEY

Twelve channel engineering seismograph was used at site for data collection in order to delineate the subsurface layers. 12 numbers of 6 Hz vertical geophones were used for the detection of the longitudinal waves. Hammer and strike plates were used as energy source. The refraction survey carried out at the site broadly detected two layers. The upper layer is compacted Doon gravel. Thickness of the layer varies from 42m to 52 m from the surface in the plotted area of the site. Beyond this layer the second layer detected is hard and compacted material which may be limestone or hard consolidated gravely material. In the proposed constructed area of the multi-storey flats, three layers were detected. Thickness of the 1st layer i.e. the compacted alluvium mixed with gravel known as Doon Gravel varies between 40m to 80 m. Second layer could be limestone thickness varies between 25m to 80m below the first layer. The third layer may be hard sedimentary rock. Thickness of the third layer could not be ascertained because the energy imparted could not reach up to the bottom of this layer.

AMBIENT VIBRATION STUDIES Ambient vibration monitoring of any site is performed to obtain the natural ground frequency of the area. Ambient vibration data is collected using either seismograph or accelerograph. In the present case Digital Triaxial Strong Motion Accelerograph (SMA) was used to collect ambient vibration data. A digital triaxial Strong Motion Accelerograph (Altus K2, Kinemetrics, USA) was kept at various places in the premises of ATS site area for some time. The SMA was kept in trigger threshold mode for recording acceleration time histories of ground motion in digital form. The threshold was set very low to get the even cultural noise generated inside the ground and transmitted to the structure. SMAs have full-scale range of 2.0 g with sampling rate of 100 samples per second (sps). The data has been collected from many different locations. The collected data was processed for the computation of the natural frequency of the site area. Results of the study for few locations are shown in the Table 1 and figures 2 & 3. The natural ground frequency varies from 0.5 to 0.8 Hz. Table 1 Natural ground frequency of the site

Fig 2. Ambient vibration record of the site.

Site No. Natural Frequency (Hz)

1 0.7 2 0.8 3 0.6 4 0.7 5 0.5

Holistic study on site characterization for habitat development

Fig 3. FFT of Time-Acceleration history of the site. GEOTECHNICAL INVESTIGATION Geotechnical tests were carried out to determine the engineering properties of soil. The tests carried out in the field are as follows:

o Plate load test - at 1.0m and 1.5m depth from the ground surface

o Boring with standard penetration tests - up to a depth of 10.5 each in three locations

o Dynamic cone penetration tests - at 8 locations FIELD AND LABORATORY TESTS Plate Load Test Two plate load tests were conducted on 30 cm x 30 cm size plate at 1.0 m and 1.5 m depth below the existing ground level in a pit of size 1.5 m x 1.5 m. The tests were conducted as per Indian Standard Specification (IS: 18881982). The gravity loading was applied by girders, sleepers and sand bags. The load was applied in the increment of 2.5 t/m

2 to 5.0

t/m2. Fig. 4 shows the pressure settlement curve.

Boring with Standard Penetration Test Three bore holes B

1, B

2 and B

3 were made. Boring was

advanced up to 10.5 m depth at the proposed locations. Standard penetration tests were conducted at 1.5 m, 2.25 m, 3.0 m, 4.5 m, 6.0 m, 7.5 m, 9 m and 10.5 m depths as per IS: 2131-1981. The observed numbers of blows for the hole B1 is shown in Fig. 5. Dynamic Cone Penetration Tests Eight dynamic cone penetration tests were conducted. These tests were planned to conduct up to 10 m depth or refusal

whichever is earlier. The dynamic cone penetration resistance i.e., the number of blows for 30 cm penetration for C

5 & C

6 are shown in Fig. 6. D.C.P.T. values show that the soil mass is loose up to about 1 m to 1.5 m, then it is medium to dense compacted. Refusal was observed at 3 m to 7.5 m depth. Laboratory analysis of the soil samples Samples collected from the field were transported in the laboratory for evaluation of physical and engineering characteristics. Following tests were carried out in the laboratory:

o Grain size analysis o Proctor density test o Direct shear test o Oedometer test and permeability test

The values obtained are given in the table 2 to 5. RECOMMENDATIONS & CONCLUSION On the basis of the field and laboratory tests and analysis of the test data the following recommendations and conclusions are drawn:

1. Sub-surface strata consist of coarse grained and fine grained soil mass. In-situ strata is silty of low compressibility (ML) mixed with sand and silty sand (SM) mixed with gravel upto about 2 m depth. It is followed by poorly/well graded gravel (GP/GW) mixed with sand, cobbles and boulders upto the depth of investigations i.e. 10 m.

2. Penetration test results show that soil mass is loose upto about 1 m to 1.5 m, then it is medium to dense compacted.

3. Net allowable soil pressure of intensity 15 t/m2

may be adopted for the design of strip/isolated foundation at 1.2 m, 1.5 m and 2.2 m depth.

4. If during excavation for laying the foundation any loose/soft pocket/cavity is observed, it should be cleaned and filled-up with lean concrete.

5. As far as possible, the entire foundation of one

building should rest on uniform sub-surface strata, however if it is unavoidable, settlement joints shall be provided in the building at places of abrupt change. Suitably designed R.C.C. plinth beam are also to be provided to check the tendency for differential settlement.

6. R.C.C. bands at lintel are preferred in seismic zone.

The foundation should normally be so designed that no tension is created at the foundation plane. In

Ghosh. A, Sarkar. S, Kanungo. D.P, Chauhan, P.K.S , Jain,, S.K Sharma Shaifaly, Kumar Dalip and AhmedZameer

seismic zone foundations should be designed for earthquake resistant as per IS: 1893-2002.

Table 2. Grain size distribution

Table 3. Proctor density test

Table 4. Direct Shear Test

Condition of test

Cohesion C (kg/cm2)

Friction φ (deg)

Sample 1

At OMC

0.62 20

At Saturation 0.36 21

Sample 2 At OMC

0.47 40

At Saturation 0.35 37

Sample 3

At OMC

0.41 36

At Saturation 0.35 36 Table 5. Consolidation and Permeability properties

Parameter Values Coefficient of consolidation Cv (cm2/Sec)

1.07 X 10 -3

Compression Index (Cc) 0.066Permeability K (cm / sec) 1.16X 10-4

REFERENCES

1. Bureau of Indian Standards. (1982), Method of Load Test for Soils. IS: 1888

2. Bureau of Indian Standards. (1976) , Indian Standard Code of Method for sub-surface Sounding for Soils, Part 1 Dynamic Method Using 50 mm cone without Bentonite Slurry. IS: 4968

3. Bureau of Indian Standards. (1981), Method for Standard Penetration Test for Soils. IS: 2131

4. Bureau of Indian Standards. (1970), Classification and Identification of Soil for General Engineering Purposes. IS: 1498

5. Bureau of Indian Standards. (1978), Code of Practice for Structural Safety of Buildings Shallow Foundations. IS: 1904.

6. Bureau of Indian Standards (2002), Indian Standard Code on Criteria for Earthquake Resistant Design of Structures. IS: 1893

Fig. 4 Load settlement curve

Fig. 5. Typical SPT curve at Location B1

Fig. 6 DCPT curves at location C5 & C6

Size Sample 1 Sample 2 Sample 3 Gravel % 29.0 36.0 28.5 Sand % 55.5 50.5 49.0 Silt % 13.5 11.5 20.0 Clay % 1.5 2.0 2.5

Parameter Sample 1 Sample 2 Sample 3 OMC % 13.0 13.8 16.3 MDD gm/cc 1.88 1.87 1.75

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi, Paper No.A145

AN EXPERIMENTAL STUDY ON THE RESPONSE OF MODEL FOOTINGS RESTING ON REINFORCED FLYASH BEDS UNDER REPEATED LOAD

S. Gangadhara 1, Associate Professor, UVCE, Bangalore University, Bangalore, E-mail- [email protected] H. C. Muddaraju 2, Assistant Professor, UVCE, Bangalore University, Bangalore, E-mail [email protected] ABSTRACT: Fly ash generated by Thermal power plant, an industrial waste slag, possessing good engineering properties, can be used in the construction of embankment. An important engineering property that is necessary for using non-Pozzolonic fly ash in any geotechnical applications is its strength. To achieve a higher strength, inclusion of geogrid reinforcement into compacted flyash is one of the most effective techniques. Several attempts have been made in the laboratory to study the behavior of reinforced sand beds that are exposed to different type of loadings. All such efforts are mainly focused on understanding the Bearing capacity and settlement behavior of the Reinforced sand beds. However, efforts to study the behaviour of reinforced Fly ash beds when subjected to either the monotonous loading or the repeated loading are scanty. Hence in the present investigation attempt has been made to study the behaviour of reinforced fly ash beds under repeated loads by conducting carefully designed experiments. As no unique standard equipment is available for the application of repeated loads and to measure the response, different researchers have designed and fabricated different types of equipments for the testing. The repeated load of known intensity with waveform type and frequency is applied on the surface and embedded footing in unreinforced and reinforced flyash beds. The response of the flyash beds, in the form of settlement is measured using LVDTs. The experimental results clearly demonstrated that the provision of reinforcement in the flyash beds is effective in improving the performance of both surface and embedded footing under repeated loads. Key words: Reinforced, fly ash bed, Repeated loads, Cyclic résistance ratio, Settlement ratio. 1. INTRODUCTION The simple mechanism of reinforced earth and the economy in cost and time have made it instant success with research workers and construction engineers alike for temporary as well as permanent structures. The performance of reinforced earth when exposed to dynamic loading is a critical issue, since the behavior of these structures under monotonic loading has been understood to some fair degree of satisfaction. However, understanding the dynamic behavior of soils is essential in areas where the seismic or man-made dynamic loading occurs. The performance and behavior of reinforced soil structure both in the field and the laboratory are well documented. Many researchers have contributed immensely to the better understanding of the concepts, design procedures and construction methods of Reinforced Soil Structures through laboratory studies, field investigations and monitoring of constructed structures (Choudhary A. K., et al.,(2010), Ayyappan S., et.al.,(2010), Prasad D.S.V. et al., (2010), Ashis Kumar Bera., et al., (2009)). Efforts to study the behaviour of reinforced earth when exposed to dynamic loading are very scanty. In this investigation, it is intended to investigate the potential benefits of using reinforced earth to improve the performance of surface and embedded footings in flyash beds under repeated loading. The experiments are conducted on model footing resting on/in reinforced and unreinforced flyash beds subjected to repeated loads. The flyash beds were reinforced with geogrids. Model footing used for this investigation is

Mild steel square footing of 100mm size and 4mm thick. The experimental results presented earlier had shown that the performance of footing is strongly influenced by the pattern of reinforcement distribution and the excitation parameters Nagaraja .P.S (2006). Hence in the present investigation, it is intend to study the performance of both surface and embedded footings as influenced by the number of reinforcement layers in the flyash beds. Reinforcement location and spacing were based on optimization of previous research results. In the present study the repeated load tests are performed in an Automated Dynamic Testing Apparatus (ADTA) specially designed, fabricated and calibrated for the purpose. 2. MATERIALS AND METHODS 2.1 Fly ash The fly ash used in the study is collected from Raichur thermal power plant, Karnataka, India. It is a non-pozzolanic fly ash belonging to ASTM classification “C”. This fly ash is directly collected from open dry dumps. The property of flyash is given in Table 1. 2.2 Reinforcement Polyethylene reinforcement in the form of Geogrid is used in the present investigation. Table 2 presents the properties of geogrid used. 2.3 Preparation of Fly ash beds Fly ash bed is prepared by manual compaction at its optimum moisture content, to maximum dry density. Unreinforced

S. Gangadhara, H.C. Muddaraju

sample is compacted up to a height of 360mm in 3 equal layers of 120mm thick. For reinforced sample, the geogrid reinforcements are placed at predetermined spacing in between fly ash layers from the bottom of footing, and by the same procedure remaining height of the tank is compacted. The reinforcements are provided in the shape of circular discs. A clearance of 5mm is provided to ensure that no friction was generated between the reinforcement and the walls of the tank. Table. 1: Properties of fly ash

Physical properties Test

Results

Colour Light grey

Specific gravity 2.07 Grain size distribution Sand size fraction (%) 15 Silt and clay size fraction (%) 85 Atterberg’s limits: Liquid Limit (%) 31.8 Plastic Limit (%) --

Plasticity Index (%) Non

plastic Compaction characteristics Optimum moisture Content (%) 23

Maximum Dry Density(kN/m3) 12.7 Unconfined Compressive Strength at MDD(kPa) 51.4 Table. 2 Properties of Reinforcement Physical properties Unit Test Results Aperture size MD mm 34 CD mm 32 Ultimate tensile strength MD kN/m 33.2 CD kN/m 31.1 Strain at ultimate MD % 14.4 CD % 6.9 2.4 Method of Testing The reinforced and unreinforced fly ash beds are subjected to repeated loading in the Automated Dynamic Testing Apparatus. The excitation values, viz., cyclic pressure (repeated load) and frequency are selected and fed in to the computer. The load is applied on to the model footing and the settlements are measured through three different LVDT’s placed orthogonal to each other. The load cell and the LVDT’s are in turn connected to the control unit, where the analog to digital conversion takes place, and is recorded in the data acquisition system. The measured settlements after each cycle of loading are recorded in the data acquisition system, which is then recovered through the computer.

3. RESULTS AND DISCUSSIONS To bring out the effect of inclusion of geogrid reinforcement on the performance of surface and embedded square footing resting on/in flyash beds, experiments are conducted both on unreinforced and reinforced flyash beds under repeated loading. Fig 1 presents the number of load cycles V/s settlement curves for square surface footing resting on unreinforced and reinforced flyash beds with two and three layers of reinforcement layers subjected to a repeated loading of 250 kPa. It can be seen from the Figure that at any level of settlement, the footing resting on reinforced fly ash bed, both two layer and three layer, resisted more number of load cycles compared to their counterpart resting on unreinforced flyash bed. For example at a settlement of 10mm, the footing on unreinforced flyash bed resisted 8 repetitive loading whereas its counterpart resting on two and three layer reinforced flyash beds resisted about 300 and 9000 repetitions. This clearly demonstrate the effectiveness of inclusion of geogrid reinforcement in flyash beds. Further it is to be observed from Fig 1 that among the footing on reinforced flyash beds, footing on three layer reinforced flyash bed performed better than footing on two layer reinforced flyash bed. This result clearly confirms the trend of results of earlier researcher which established the fact that the optimum number of reinforcement layer is three. Fig 2 presents the number of load cycles v/s settlement curves for embedded footing in unreinforced and reinforced fly ash beds with two and three layers of reinforcement subjected to a repeated loading of 250kPa. It can be observed from Fig 2 that the trend of results in case of embedded footing is same as that for surface footing (Fig 1). Fig. 2 demonstrates that even for the embedded footing, • The reinforced flyash beds experience less settlement

compared to the unreinforced flyash beds at the same number of repeated loads.

• The three layer of reinforcement is optimum, which induces minimum settlement of the embedded footing at any number of load cycles

A comparative study of Fig 1 and Fig 2 indicates that, both in the unreinforced and reinforced flyash beds, the embedded footing experiences relatively higher value of settlement compared to the surface footing. This is true irrespective of number of layer of reinforcement. Similarly in case of reinforced flyash beds, the surface footing in three layer reinforced flyash bed experienced a settlement of 2mm where as its counterpart embedded in three layer flyash bed experienced a settlement of 7mm for the same 50 number of repetitions. This clearly demonstrates that the surface footing resting on unreinforced/reinforced flyash beds perform better than the embedded footing in unreinforced/reinforced flyash beds. This is further analyzed in terms of cyclic resistance ratio and settlement ratio.

An experimental study on the response of model footings resting on reinforced flyash beds under repeated load

Fig. 1 Effect of number reinforcement (S=0.3B) on the performance of surface footing resting on flyash beds under a repeated load of 250 kPa,

Fig. 2 Effect of number reinforcement (S=0.3B) on the performance of embedded footing resting in flyash beds under a repeated load of 250 kPa, 3.1 Cyclic Resistance Ratio and Settlement Ratio. The results of the experiments performed to understand the behavior of surface and embedded footing are analyzed in terms of cyclic resistance ratio (CRR) and the Settlement ratio (SR) using the definitions given by Nagaraja.P.S (2006). (i) Cyclic Resistance Ratio (CRR) �

Fig. 3 and Fig. 4 presents the cyclic resistance ratio curves for square footing and embedded footing resting on/in two layer and three layer reinforced flyash beds. It is seen from these figures that the cyclic resistance ratio for both surface and embedded footing increases as the settlement of the footing increases. At lesser value of settlement, the cyclic resistance ratio is less and increases exponentially at higher values of settlement. For example for surface footing at a settlement of 20mm the cyclic resistance ratio is 40 and it increases to about 100 at settlement of 30mm. Similarly for embedded footing these values are 5 and 42 at 20mm and 30mm settlement, respectively. This clearly demonstrates that the

provision of reinforcement is effective in increasing the repeated load carrying capacity both in surface footing and embedded footing. Further it can be inferred that providing the reinforcement is more effective or becomes more useful at higher values of settlement (as indicated by higher values of Cyclic Resistance Ratio at higher settlement level). Further a comparison of these curves indicate that at any settlement level, the surface footing exhibit higher values of cyclic resistance ratio compared to its counterpart embedded in flyash beds. For example in two layered flyash beds, the surface footing at a settlement of 20mm exhibits a cyclic resistance ratio of 38 whereas the embedded footing at the same settlement level exhibits a cyclic resistance ration of about 5 only (Fig-3). Similarly at 30mm settlement level, the cyclic resistance ratio values of surface and embedded footing are 90 and 40 respectively (Fig.3). Similar trend of results is observed in case of three layer reinforced flyash beds also (Fig 4) i.e., the surface footing exhibited higher values of cyclic resistance ratio compared to the embedded footings. It is further interesting to observe from Fig. 3 and Fig. 4 that the footings, both surface and embedded, in three layer reinforced flyash beds (Fig. 4) exhibited higher values of cyclic resistance ratio compared to their counterparts in two layer reinforced flyash beds (Fig. 3). The surface footing on three layer reinforced flyash bed exhibited a cyclic resistance ratio as high as 1600 at about 25mm settlement level whereas the same value reduces to about 60 for two layer reinforced flyash beds. The two layer reinforced flyash beds showed a maximum cyclic resistance ratio value of 100 only (at about 30mm settlement). From the above discussion it can be inferred, beyond doubt, that the optimum reinforcement number is three, for both surface and embedded footing under repeated loads.

Fig. 3 Cyclic Resistance Ratio curves for footing in two layer reinforced flyash beds. (ii). Settlement Ratio(SR) To bring out the effect of reinforcement on the settlement behavior of footings in the term flyash beds, Settlement Ratio is calculated from the experimental results. In the present study, Settlement Ratio (SR) is defined as,

S. Gangadhara, H.C. Muddaraju

Fig. 4 Cyclic Resistance Ratio curves for footing in three layer reinforced flyash beds. Fig. 5 and Fig. 6 presents the settlement ratio curves for surface and embedded footing resting on/in reinforced flyash beds with two and three layer of reinforcement, respectively under repeated load of 250kPa. It is to be observed in these two figures that the Settlement Ratio is calculated to a maximum of 100 number of load cycles, through the experiments were conducted up to 20000 load cycles for reinforced flyash beds as the experiments are terminated after 100 number of load cycles for the unreinforced flyash beds. The footing in unreinforced flyash beds experienced the limiting settlement of 40mm at 100 numbers of load cycles itself. Fig. 5 and Fig. 6 shows that, for all the testing conditions, at any number of load cycles the embedded footing exhibits, higher value of Settlement Ratio compared to the surface footing resting on reinforced flyash beds experiences less settlement compared to the embedded footing in reinforced flyash beds. The same trend of results is observed in case of both two layered and three layered reinforced flyash beds. 4. CONCLUSIONS Based on the results of experiments conducted, the following conclusions are drawn. • Both the surface and embedded footing perform better

when they are placed on/in reinforced flyash beds. Footings in reinforced flyash beds exhibit resistance for more number of repeated loads and show less settlement for a given number of repetitions compared to the unreinforced condition.

• The performance of footings is calculated in terms of cyclic resistance ratio and settlement ratio. The surface footing exhibited higher values of cyclic resistance ratio and lower values of settlement ratio under all testing conditions, indicating that they perform better than the embedded footing.

• The cyclic resistance ratio curves and the settlement ratio curves indicate that the optimum number of reinforcement layer is three.

Fig. 5 Settlement Ratio curves for footings in two layer reinforced flyash beds.

Fig. 6 Settlement Ratio curves for footings in three layer reinforced flyash beds.

5. REFERENCES 1. Ashis Kumar Bera., Sowmendra Nath Chandra.,

Amalendu Ghosh., Ambarish Ghosh., (2009), “Unconfined compressive strength of fly ash reinforced with jute geotextiles”, Geotextiles and Geomembranes, Vol. 27, No 4, PP 391-398.

2. Ayyappan, S., Hemalatha, K., Sundaram, M., (2010), “Investigation of engineering behaviour of soil, polypropylene fibers and flyash mixtures for road construction”, International journal of Environmental Science and Development, Vol.01, No.2, PP 171-175.

3. Choudhary, A.K., Jha, J.N., Gill. K.S., (2010), “Laboratory investigation of bearing capacity behaviour of strip footing on reinforced flyash slope”, Geotextiles and Geomembranes, Vol. 28, No 4, PP 393-402.

4. Prasad, D.S.V., Anjan Kumar, M., Prasada Raju, G.V.R., Kondayya, V., (2010),“ Behaviour of Flyash Reinforced Sub-bases on Expansive Soil Subgrades under Cyclic Loading”, Indian Geotechnical Conference-2010 pp 131 I.I.T., Mumbai (GEOtrendz)

5. Nagaraja.P.S. (2006) Behavior of model footing resting on geogrid reinforced Sand beds under Monotonic, Cyclic and dynamic loads, Ph.D Thesis, Bangalore University, Bangalore.


GEOPHYSICAL INVESTIGATIONS FOR A HYDROELECTRIC PROJECT

M.S. Chaudhari, Senior Research Officer, Central Water and Power Research Station, Pune, [email protected]

ABSTRACT: Engineering geophysical methods have gained wide acceptance for deciphering subsurface stratigraphy including rock topography. This information is required for deciding the level and for designing the foundations of the civil structures of a hydroelectric project economically. Continuous seismic refraction survey and cross-hole seismic studies were conducted at a Rupsiabagar Khasiabara hydroelectric project in Pithoragarh district, Uttarakhand. The project area falls in the main central crystalline belt which consists of mylonite gneisses, phylites, garnetiferous schist, calcareous silicate rock and quartzites with associated syntectonic granite gneiss. Continuous seismic refraction survey revealed that the subsurface in general comprises three/ four layers including weathered rock/ good quality rock. The subsurface layers depending on compressional wave velocity values were interpreted as loose overburden (velocity values ranging from 250 m/sec to 750 m/sec), overburden (750 m/sec to 1200 m/sec), compact overburden (1200 m/sec to 2000 m/sec), weathered rock/ bouldery bed (2000 m/sec to 3500 m/sec) and good quality rock (above 3500 m/sec). At power house site where rock depth was large, because of limited spread length available, the minimum depth to the rock was evaluated after making certain assumptions. The cross-hole seismic studies at power house site from 1.5 m depth up to 27 m depth in East and South directions yielded compressional (P-) wave velocities varying from 820 m/sec to 2620 m/sec and shear (S-) wave velocities from 330 m/s to 1130 m/s. Lower P- wave velocity of 1540 m/sec to 1950 m/sec obtained from 16.5 m to 27 m depth was attributed to change of strata from bouldery bed to sandy layer. This inference was corroborated by the geological log of the source hole and the results of seismic refraction 3 m away from cross-hole test. P- and S- wave velocities determined in South direction at a particular depth were of the same order as in East direction. The in situ dynamic Young’s and Shear moduli with depth varied from 0.07 X 105 kg/cm2 to 0.87 X 105 kg/cm2 and 0.02 X 105 kg/cm2 and 0.31 X 105 kg/cm2 respectively. INTRODUCTION Engineering geophysical methods have gained wide acceptance as a cost effective and rapid means for deciding the levels and designing the foundations of civil structures of a hydroelectric project. Since dam in a hydroelectric project is a massive structure, the foundation should be geotechnicaly sound to sustain high stresses those are expected to be developed due to self weight of the structure, water pressure of the reservoir and earthquake vibration induced forces. The geological and geotechnical investigation is to be directed towards determining the subsurface geological structure, stratigraphy, faulting, foliation and jointing and to establish ground water conditions adjacent to the dam site, including the abutments. This paper deals with the utility of seismic refraction technique for delineating subsurface stratigraphy including rock topography and cross-hole seismic technique for evaluating dynamic elastic properties with depth for a hydroelectric project in Uttarakhand.

Rupsiabagar Khasiabara Hydroelectric Project is a run-off river scheme proposed to be constructed on river Goriganga a tributary of river Mahakali in Ganga basin that will generate 260 MW hydroelectric power on completion. The project is situated in Munsiari tehsil of Pithoragarh district and the dam is located at Lat 300 09’ 56.45" and Long 800 15’ 06". The

project includes a 60 m high concrete gravity dam, surface power house, head race tunnel, desilting chambers, surge tank, surface penstock and tail race tunnel. Tail race tunnel will discharge the flows back into Goriganga river after power generation. For economic and safe design of foundation of these structures, subsurface stratigraphy including faults and shear zones and rock topography and its quality are important input parameters. This information can be obtained at discrete locations by drilling boreholes but the procedure of drilling boreholes is both time consuming and expensive. Also, in complex geological settings like Himalayan region where depth to rock varies from place to place, interpolation of geological information between holes may be misleading and the detailed project report prepared only on the basis of borehole data may not be the true representative of the site conditions. Shear wave velocities of the subsurface strata are needed both for seismic hazard analysis and for siting of critical structures of a hydroelectric project. Shear wave velocities vary spatially as well as with depth and wide range of values for strata particularly in overburden are available. Therefore, which velocity value to adopt for a particular site is a very difficult decision to make. It is always advisable to carry out site specific surveys to measure the precise shear wave velocity. The site specific in situ P- and S- wave velocities are best evaluated by cross-hole seismic technique. Though the technique needs special arrangements and equipment for generation and recording of shear waves with

M.S. Chaudhari

depth but provides accurate shear wave velocities. With the above in view, continuous seismic refraction survey was conducted to establish the rock topography and to evaluate its quality. Cross-hole seismic studies in two mutually perpendicular directions at power house site was conducted to find out the preferred direction of orientation of in homogeneities i.e. joints and fractures. Cross-hole studies were also aimed at to evaluate dynamic moduli as well as to delineate the weak zones with depth. GEOLOGY The project area consists of mylonite gneisses, phylites, garnetiferous schist, calcareous silicate rock and quartzites with associated syntectonic granite gneiss and late to post tectonic tourmaline granite. The terrace where power house is proposed to be constructed is occupied with ill assorted boulders of gneissic schist quartizites and few granite pieces with sand. CONTINUOUS SEISMIC REFRACTION SURVEY Out of the 14 profiles taken at different locations for Rupsiabeger Khasiabara Hydroelectric Project (RKHEP), four seismic refraction profiles of 84 m to 138 m length and cross-hole seismic test up to 27 m depth at one location were taken on the proposed power house site (Fig. 1).

The refraction data were collected using continuous refraction technique while data were interpreted using ‘Reciprocal Technique’ deploying Winsism software. Typical depth

section along profile 12 of length 138 m detailing subsurface layers along with their velocities is shown in Figure 2.

The subsurface along this profile comprises three layers. The first layer having compressional (P-) wave velocity of between 350 m/s and 450 m/s is inferred as loose overburden and the second layer having P- wave velocity between 950 m/s and 1200 m/s is inferred as overburden. The thickness of first 2 layers is evaluated below shot point at ch 3.0 m, ch 45.0 m, ch 87.0 m and ch 135.0 m have been joined to make it look continuous. The depth to compact overburden from ground surface, which is addition of first and second layer thicknesses varied from 11.3 m to 22.8 m. The RL’s of the compact overburden along this traverse varied between RL 1247 m and RL 1256 m. The compact overburden velocity varied from 1650 m/s to 2000 m/s. The cross-hole test is taken at about 3 m away from 54 m chainage of profile 12 and the source borehole log is plotted on the same depth section. It is seen from the depth section that velocity increases up to 16.5 m depth. This result matches with the borehole log up to 16.5 m where the strata is sand with boulders. Because of limited spread length available and the rock being deeper, the depth to rock could not be ascertained. However, after making certain assumption, the minimum depth to rock was evaluated. CROSS-HOLE SEISMIC STUDIES Shear wave velocity governs the transmission of seismic signal of earthquake from site to structures and also controls the site response itself. All geotechnical tests provide information of shear wave velocity from point to point and the values are interpreted in between places. These tests grossly under sample the subsurface and frequently inadequate [1].

Fig.1 Layout of seismic refraction profiles and boreholes for cross-hole seismic test

P-9 P-10

51700 51800 51900 52000

44800

44900

45000

45100

North

ing

44800

44900

45000

45100

51700 51800 51900 52000Easting

P-11

P-12

P-12 Profile - 12 with chainage

INDEX

0 m

0 m

0 m

0 m

138 m

138 m84 m

90 m

Borehole

DH-06-19

DH-06-17

DL-06-05

Power House

Set of boreholes for cross-hole studies

Source borehole CH-06-02

3.0 m

FIG.2 Depth-Section along Profile - 12 ( 0-138 m ) at Power House Site

INDEX

Shot Point with Chainage

P-wave velocity

Horizontal Scale

-20 m 0 20 m

Overburden

Surface Topography

350 m/s

3.0 m Loose overburden

45.0 m 87.0 m 135.0 m

1220

1230

1240

1250

1260

1270

Red

uced

Lev

el (m

)Compact overburden

Minimum depth to good quality rock

1650 m/s2000 m/s

450 m/s350 m/s400 m/s360 m/s

1200 m/s1050 m/s 950 m/s

1200 m/s

Geophysical Investigations for a Hydroelectric Project

Cross-hole technique provides greater radius of investigation compared to up-hole and down-hole methods and provides greater measuring accuracy. The cross-hole seismic measurements were carried out up to 27 m depth at power house site where a set of three boreholes 5 m apart in east and south directions were rotary drilled. All the holes were PVC cased and the annular space between casing and the borehole wall was pressure grouted with bentonite clay and water mixture. Borehole CH-06-02 was used as source hole and the remaining two holes were used to record P- and S- wave arrivals where tri-axial geophones were lowered. The logging interval were kept 1.5 m by moving the source hammer and recording geophone step by step in respective holes up to the depth logged. He criteria for selection of distance between boreholes is the borehole should be far enough to give discernible difference in arrival times between P- and S- waves and borehole should be close enough to yield sharp arrivals avoiding picking up of refracted arrivals from adjacent layers [2]. If ‘Vp’ and ‘Vs’ are the compressional and shear wave velocities in the medium , ‘ρ’ the density and ‘σ ’ Poisson’s ratio then dynamic Young’s modulus of elasticity ‘Ed’ and Shear modulus of elasticity ‘Gd’ are calculated using [3]

)1()21)(1(2

σσσρ

−−+

= pd VE (1)

Gd = ρ Vs

2 (2) where

22

225.0

sp

sp

VVVV

−

−=σ (3)

The densities for the bouldery bed and the sandy strata were assumed to be 2.4 gm/cc and 2.2 gm/cc respectively. The corresponding values of dynamic Young’s and shear moduli determined using equations (1), (2) and (3). Figures 3 and 4 show plot of P- and S- wave velocities and Young’s and Shear moduli with depth in east and south directions in the power house site respectively. It is seen from the figure that shear wave velocity with depth varied from 330 m/s to 1130 m/s. P- wave velocity with depth increases up to 16.5 m up to which borehole log reveal bouldery bed. From 18.0 m onwards a decrease in P- wave velocity with depth is noticed. Through unusual this was attributed to change of strata from bouldery bed to sand.

Shear

Shear wave velocity also starts decreasing from 18 m depth confirming the beginning of sandy layer.The in situ dynamic Young’s modulus varied from 0.07 X 105 kg/cm2 to 0.87 X 105 kg/cm2. The shear modulus values were between 0.02 X 105 kg/cm2 and 0.31 X 105 kg/cm2. Dynamic Poisson’s ratio varied from 0.35 to 0.44. It is also seen that velocities in two mutually perpendicular directions are the same. It was inferred from this that the inhomogenities have no preferential direction of orientation. Therefore average of two sets of values of velocities at any level can be adopted as representative velocity for the strata at that depth. CONCLUSIONS From continuous seismic refraction survey the subsurface layers depending on P- wave velocity values were interpreted as loose overburden (velocity values ranging from 250 m/sec to 750 m/sec), overburden (750 m/sec to 1200 m/sec), compact overburden (1200 m/sec to 2000 m/sec), weathered rock/ bouldery bed (2000 m/sec to 3500 m/sec) and good quality rock (above 3500 m/sec). The results of the survey did not indicate any undesirable subsurface feature like a major shear zone, or a geological fault which might have posed foundation

0 1000 2000 3000

Vp / Vs m/sec

30

20

10

0

Dept

h (m

)

Fig. 3 Wave Velocities and Elastic Moduli in east direction along with source borehole log

30

20

10

0

Receiver Borehole

a)b)

Sand with boulders

Sand

0 0.4 0.8

Young's / Shear modulus x 106 Kg/cm2

30

20

10

0

c)d)

Sand with boulders

Source Borehole CH06 - 02

a) b) c) d)

Compressional wave velocity

Young's Modulus

Shear wave velocity

Shear Modulus

INDEX

Fig. 4 Wave Velocities and Elastic Moduli in south direction along with source borehole log

30

20

10

0

Receiver Borehole

Source Borehole CH06 - 02

a) b)

Sand with boulders

Sand

0 1000 2000 3000

Vp / Vs m/sec

30

20

10

0

Dep

th (m

)

Sand with boulders

0 0.4 0.8

Young's / Shear modulus x 106 Kg/cm2

30

20

10

0

c) d)

a)

b)

c)

d)

Shear Wave Velocity

Compressional Wave Velocity

Shear Modulus

Young's Modulus

M.S. Chaudhari

problems. The cross-hole seismic studies at power house site from 1.5m depth up to 27m depth revealed that the P- wave and S- wave velocity varies from 820 m/s to 2620 m/s and 330 m/s to 1130 m/s respectively. The values of P- wave velocities obtained from refraction and cross-hole survey matched well with the source borehole log. The dynamic Young’s and Shear moduli with depth varied from 0.07 X 105 kg/cm2 to 0.87 X 105 kg/cm2 and 0.02 X 105 kg/cm2 and 0.31 X 105 kg/cm2 respectively. The evaluated P- and S- wave velocities and Young’s and Shear moduli will help in deciding the level and in designing the foundations of civil structures so as to safeguard them against earthquake forces. ACKNOWLEDGEMENT Author is thankful to Dr.I.D.Gupta, Director, for constant encouragement and for according permission to publish this work. REFERENCES 1. Sarman, R., and D.F.Palmer, 1990, Engineering

Geophysics, The need for its development and application, h International IAEG Congress, Rotterdam, pp. 1017- 1023.

2. Butler D.K and Curro R. (Jr), 1981, Cross-hole seismic testing-procedures and pitfalls; Geophysics V.46, No.1, pp. 23-29

3. Dobrin and Savit, 1988, Introduction to Geophysical

Prospecting, McGraw Hill Book Company, Fourth edition, pp. 25-26

4. C.W.P.R.S. 2007, Report on Seismic Refraction survey for

Rupsiabagar Khasiabara Hydroelectric Project, Uttarakhand. Technical Report No. 4455

5. C.W.P.R.S. 2007, Report on Cross-hole seismic studies for

Rupsiabagar Khasiabara Hydroelectric Power Project, Uttarakhand. Technical Report No.4458

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No.A-149.)

RESILIENT MODULUS OF COHESIVE SOIL

Vidya Dodwad, Postgraduate Student, Dept. of Civil Engineering, SIT, Tumkur-572103, [email protected] Muttana .S. Balreddy Asst. Professor, Dept. of Civil Engineering, SIT, Tumkur-572103, India, [email protected] T S Umesha, Associate Professor, Dept. of Civil Engineering, SIT, Tumkur-572103, India, [email protected] S.V. Dinesh, Professor, Dept. of Civil Engineering, SIT, Tumkur-572103, India, [email protected] ABSTRACT: Resilient modulus is an important parameter and is widely used in the design of pavement system. AASHTO design guidelines recommended the use of resilient modulus (Mr) for characterizing base and subgrade soils and for designing flexible pavements. There are many correlation equations based on various soil properties for predicting Mr of different soils and these models have been developed based on extensive testing of different soil and there is need to verify the prediction capability of these models of soils of some other region. In the present investigation an attempt is made to evaluate the applicability of some of the model to predict Mr Values of Black cotton soil. Unconfined compressive strength tests have been carried out for both soaked and unsoaked condition on samples prepared at proctor condition. Repeated load triaxial test was also carried as per AASHTO guidelines. A comparison of resilient modulus values from experimental and those predicted by various model is reported. Results indicate none of these models are satisfactory and there is large variation in experimental and predicted values. INTRODUCTION Pavement life depends on the performance and condition of the pavement system, which consists of a bituminous overlay, base, subbase and subgrade. During the life of the system, the subgrade is subjected to variations in moisture content, and depending on the soil type of the subgrade, could result in variations of the moduli. In optimum conditions, the subgrade would be compacted to 99% of dry unit weight and at optimum moisture content. During seasonal changes, storm and groundwater may infiltrate the subgrade, changing the moisture content and, therefore, changing the resilient modulus. Temperature fluctuations (freezing and thawing) in the subgrade, depending on the depth, may also affect the performance of the subgrade resilient modulus. Most of the methods which use California Bearing Ratio (CBR) and Soil Support Value (SSV) do not represent the conditions of a pavement subjected to repeated traffic loading. Recognizing this deficiency, the 1986 and the subsequent 1993 American Association of State Highway and Transportation Officials (AASHTO) design guides recommended the use of resilient modulus (Mr) for characterizing base and subgrade soils and for designing flexible pavements. Several researchers have suggested correlation equations based on various soil properties (i.e [1], [2], [3], [4] and [5]) such as deviator stress, confining pressure, proctor values, moisture contents, CBR and unconfined compressive strength, percent compaction, degree of saturation, % sand, %clay, liquid limit, plasticity index. The Standard Test Method for determining the Resilient Modulus of Soils and Aggregate Materials, AASHTO Designation T 307-99 has been adopted as the universal laboratory testing procedure to determine resilient modulus of subgrade soils. The resilient modulus is defined as the ratio of the maximum cyclic (deviatoric) stress (σd) to the recoverable resilient strain (εr)

Mr = σd / εr

where, σd = Applied deviator stress εr = Resilient strain Several factors which affect resilient modulus of subgrade soil are stress state which includes the deviator stress and confining stress, soil type and its structure, degree of saturation and compactive effort (density), specimen size and preparation, temperature effects, effects of end conditions. The objective of this study is to conduct Unconfined Compressive Strength and repeated load triaxial shear tests on Black Cotton Soil samples at proctor conditions to evaluate Unconfined Compressive Strength and Resilient Modulus under both soaked and unsoaked conditions for different percentages of lime and curing period. An attempt was also made to evaluate a predictive equation for resilient modulus based on soil index properties including the Unconfined Compressive Strength MATERIALS Materials used in the present study are Black Cotton soil and lime. Black cotton soil samples were collected from Bagalkot District, Karnataka. The geotechnical properties of the B.C Soil are presented in Table 1. Grain size indicate that the soil is classified as Silty-Clay soil with the group symbol MH-CH as per IS classification. Commercially available lime was used in the present investigation.

Vidya Dodwad, Muttana S Balreddy, T S Umesha & S V Dinesh

Table 1 Geotechnical properties of Black Cotton soil

Sl.no

Property B.C Soil

1 Specific Gravity 2.70

2

Grain Size Distribution (%) a Gravel 1.20 b Sand 13.70 c Silt & Clay 85.1

3 Soil Classification a I.S Soil classification MH-CH

4

Atterbergh׳s Limit (%) a Liquid Limit 65 b Plastic Limit 37 c Plasticity Index 28 d Shrinkage Limit 22

5 Compaction Characteristics (std proctor test) a OMC (%) 22.69 b Max dry density (gm/cc) 1.62

6 Free Swell Index (%) 50

7

California Bearing Ratio (CBR) Test (%) I.S Light compaction-Unsoaked condition

3

-Soaked condition

2

EXPERIMENTAL INVESTIGATIONS Specimen Preparation for Resilient Modulus Soil specimens were prepared according to the procedure described by [6]. Moulds of diameter 50mm and height 100mm were used to prepare soil specimens by static compaction. Specimens were prepared at maximum dry unit weight and optimum moisture content, repeated load triaxial tests were performed on specimens as per AASHTO T-307 procedure. After the soil specimen was prepared under a specified unit weight and moisture content, it was placed in a membrane and mounted on the base of the triaxial cell. Porous stones were placed at the top and bottom of the specimen. The triaxial cell was sealed and mounted on the base of the dynamic materials test system frame. All connections were tightened and checked. Cell pressure, LVTD’s, load cell, and all other required setup were connected and checked. The loading sequence and the combinations are presented in Table 3. Axial deformation of the specimen was recorded by two externally mounted Linear Variable Differential Transducers (LVDT). The average of the resilient modulus values of the last five loading cycles of the 100 cycle sequence was used to compute resilient modulus.

Specimen Preparation for Unconfined Compressive strength Soil specimens were prepared based on proctor conditions for untreated and treated soil. Samples were prepared in a static mould of diameter 3.8cm and height of 8.0cm. Curing: The UCS test was carried out after curing the specimen for required curing period in a dessicator at 100% humidity under a constant strain rate of 0.254mm/min. For 1 day cured sample testing, samples were soaked for one hour before testing and for 7, 14 days curing, samples were soaked for 1 day before testing for soaked condition. Lime Stabilization Lime Fixation The lime content required to improve the physical properties of soils is called lime fixation point, which varies between 3 to 10%, depending on the soil. [7]. Optimum lime content is determined by pH method (plotting a graph, Lime v/s pH). This test will identify the lime content required to satisfy immediate lime-soil reactions and still provide the proper conditions for the long-term pozzolanic reaction. In the present study optimum lime content obtained by the above method is 7%.

Table 2 Testing Sequence for Subgrade soils as per [6] Guidelines

Sequence No

Confining Pressure, σ1 (kPa)

Deviator Stress, σd (kPa)

No of load applications

0 * 41 28 500-1000 1

41

14 100 2 28 100 3 41 100 4 55 100 5 69 100 6

28

14 100 7 28 100 8 41 100 9 55 100 10 69 100 11

13

14 100 12 28 100 13 41 100 14 55 100 15 69 100

*(Preconditioning) Experimental Programme The following experimental program was deviced for carrying out UCS and Resilient Modulus test on Black cotton soil with lime as additive for different curing periods.

Resilient Modulus of Cohesive Soil

Table 3.0 Experimental Programme for UCS Pr

oper

ty Initial state Test

condition % lime

Curing period (days) DD

(gm/cc) w.c (%)

(soaked/ unsoaked)

UCS

1.62 22.69 Unsoaked 0 0

1.62 22.69 Soaked 0 1 1.62 22.69 Unsoaked 8,10,

12,14 7, 14

1.62 22.69 Soaked 8,10, 12,14

7, 14

Table 4 Experimental Programme for Resilient Modulus

Prop

erty

Initial state Test condition (soaked/ unsoaked)

Curing period (days)

Dry Density (gm/cc)

Water Content (%)

Dry/ OMC / wet

RM 1.62 22.69 OMC Unsoaked 28

RESULTS AND DISCUSSIONS Unconfined Compressive Strength Figures 1 and 2 shows the UCS for unsoaked and soaked condition that the addition of lime beyond optimum lime content results in increase in UCS value for both unsoaked and soaked condition for different curing period. The Unconfined compressive strength values for different test condition, % lime and curing period are shown in Table 5

Fig 1 Unconfined Compression Strength results for Unsoaked condition

Fig 2 Unconfined Compression Strength results for soaked condition

Table 5 Unconfined Compression Strength Results

Sl. no

UCS Test

condition %

lime

Curing period

(kPa) (days)

1 433.68 Unsoaked 0 0 2 61.79 Soaked 0 1 3 538.72

Unsoaked

8

7 4 458.33 10 5 623.80 12 6 692.23 14 7 717.24

Unsoaked

8

14 8 626.45 10 9 817.04 12

10 1283.0 14 11 109.19

Soaked

8

7 12 84.64 10 13 267.49 12 14 407.36 14 15 321.28

Soaked

8

14 16 222.68 10 17 694.98 12 18 784.52 14

Resilient Modulus Figure 3 shows the resilient modulus versus deviator stress for B.C Soil compacted at proctor conditions. The resilient modulus increases with increase in deviator stress under constant confining pressure. Figure 4 shows resilient modulus versus confining pressure for B.C Soil compacted at proctor conditions. The resilient modulus increases with increase in confining pressure under

Vidya Dodwad, Muttana S Balreddy, T S Umesha & S V Dinesh

deviator stresses of 41.4, 55.2 and 68.9 kPa. For the 27.6 and 13.8 kPa there is decrease. However the resilient modulus values are higher at confining pressure 27.6 kPa.

Fig 3 Resilient Modulus Vs Deviator Stress

Fig 4 Resilient Modulus Vs Confining Pressure Table 4 shows the comparison of Mr values obtained by repetitive load triaxial tests and the predicted values from various models. There is wide variation between experimental and predicted values. The constitutive model which uses UCS does not reflect the Mr behaviour of real soils used in the present investigation. Table 4 Comparison of Laboratory Mr value (AASHTO 307)

with various models

Tes

t Seq

uenc

e

Con

finin

g pr

essu

re

D

evia

tor

stre

ss

A

ASH

TO

T

- 307

USD

A M

odel

[3

]

TD

OT

Mod

el

[8]

Con

stitu

tive

mod

el [8

]

kPa kPa MPa MPa MPa MPa 1

41.4

13.8 83.0 110.61 81.80 1.37 2 27.6 97.77 106.12 80.16 1.18 3 41.4 127.32 101.64 78.52 1.09 4 55.2 140.70 97.16 75.25 1.03 5 68.9 155.09 92.68 75.25 0.99 6 27.6 13.8 125.26 108.13 78.08 1.26

7 27.6 122.20 103.65 76.52 1.10 8 41.4 137.16 99.17 74.95 1.02 9 55.2 150.57 94.69 73.39 0.96 10 68.9 156.90 90.21 71.83 0.93 11

13.8

13.8 127.30 105.66 74.36 1.12 12 27.6 120.16 101.18 72.87 0.99 13 41.4 123.24 96.70 71.39 0.93 14 55.2 131.54 92.22 69.90 0.89 15 68.9 136.48 87.73 68.41 0.86 CONCLUSIONS The following conclusions are drawn 1. Addition of lime beyond fixation point increases both

soaked and unsoaked compressive strength. 2. The experimental data indicates increased Mr with

confining pressure and deviator stress. 3. Experimental models fail to estimate the Mr values for

the local soil and there is need to revise these empirical models by incorporating additional terms to model the real behaviour.

REFERENCES

1. Ramesh. B. Malla and Shraddha Joshi. “Resilient Modulus Prediction Models Based on Analysis of LTTP data for Subgrade Soils and Experimental Verification”.Journal of Transportation Engg ASCE- Vol 133 (2007).

2. Drumm, B. C., Boateng-Poku, Y., and Pierce, T. J. (1990). "Estimation of subgrade resilient modulus from standard tests:' Journal Geotech. Engg., ASCE, 116(5), 774-789.

3. Carmichael III, R. F., and Stuart, B. (1985). "Predicting resilient modulus: a study to determine the mechanical properties of subgrade soils:' Transp. Res. Rec., 1043, 145-148.

4. Farrar, M. J., and Turner, J. P. (1991). "Resilient modulus of Wyoming subgrade soils." Rep. No. 91·1, Mountain-Plains Consortium, Fargo, N. Dak.

5. Yau, A.,and Von Quintus (2004). “Predicting Elastic Response Characteristics of Unbound Materials and Soils,” Transportation Research Record No. 1874, Transportation Research Board, National Research Council, Washington, D. C., pp. 47-56.

6. AASHTO Guide for Design of Pavement Structures, American Association of State Highway and Transportation Officials, Washington, DC, 1986

7. Dallas N. Little. “Evaluation of Structural properties of Lime Stabilized Soils and Aggregates”. Prepared for the National Lime Association, January 1999.

8. Dong-Gyou Kim, M.S. “Development of a Constitutive Model for Resilient Modulus of Cohesive Soils”. A Dissertation presented in partial fulfillment of the requirements for the Degree Doctor of Philosophy in the Graduate School of the Ohio State University (2004).

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No. A 150.)

AN EXPERIMENTAL STUDY ON STRENGTH BEHAVOIUR OF REINFORCED JOINTED ROCK MASS

Lok Priya Srivastava, Research Scholar, Department of Civil Engineering, IIT Roorkee, Email:[email protected] Mahendra Singh, Professor, Department of Civil Engineering, IIT Roorkee, Email:[email protected] ABSTRACT: Rock bolts are used in the field to enhance the self supporting capacity of jointed rock mass. The main aim of this study is to investigate the strength enhancement in rock mass due to provision of rock bolts. For this purpose large size direct shear tests were conducted on jointed rock mass specimens with and without reinforcement. Normal stress during tests was varied between 0 to 2 MPa. The results of the study indicate substantial strength enhancement at low level of normal stresses on shearing plane. As the normal stresses increases, strength enhancement has been found to be reduced. There is a limiting value of normal stress beyond which no strength enhancement occurs due to provision of rock bolts. 1 INTRODUCTION Rock masses encountered in the field are invariably jointed. The strength of jointed rock mass depends upon many factors. Most important of these are the strength of parent intact rock, number of joint sets, joint orientation, and joint surface characteristics. Joints introduce or form a plane of weakness in the rock mass. Hence, to strengthen a rock mass, it is necessary to prevent shearing/sliding along the plane of weakness. Generally, reinforcement has been used for this purpose. One of the most widely used reinforcement techniques is rock bolting. Many researchers have conducted studies on reinforced rock joints in past however studies on rock masses are very few. It has been shown that the strength of reinforced rock joint depends upon various factors like strength of parent rock [1,2, 3], joint orientation, joint condition, inclination between joint and bolts [3,4], type of bolt, strength of bolt, diameter of bolts and pretensioning of bolts [2,3,5,6]. Literature available has generally focused on the reinforcement of joint and not the rock mass as a whole. In case of a rock mass, complex interaction between blocks occurs, and the response of the mass cannot be explained solely by understanding joint behaviour. The failure modes of the rock mass as a whole play important role in governing the rock mass behaviour [7]. The presence of rock bolts influences the likely mode of failure and this has considerable impact on rock mass strength. 2 EXPERIMENTAL PROGRAMME As mentioned above, several studies have been conducted in past to understand the strength behavior of joints, however the studies on rock masses are very few. The main aim of the present research work is to investigate the effect of rock bolts on the strength enhancement of jointed rock mass. To

understand the strength behavoiur of reinforced jointed rock mass, a laboratory investigation was planned in which large sized direct shear tests were conducted on the rock mass specimens with and without rock bolts. Concrete has been used to simulate intact rock material. The tests have been conducted at normal stress levels ranging between 0 to 2 MPa. A servo controlled large size direct shear test machine has been used for conducting the tests. The size of shear box is 75 cm x 75 cm x 100 cm (height). The machine has loading capacity of 2000 kN in shear loading and 1500 kN in normal loading. Concrete cubes of size 15 cm x 15 cm x 15 cm were used for simulating rock. The ratio of different constituents of concrete is given in Table 1. The cube strength of the model material has been found to be ranging between 25 to 40 MPa. About 150 cubes of size 15 cm x 15 cm x 15 cm were required for conducting one test. The formed jointed rock mass consisted of vertical and horizontal joint sets having average spacing of 13 cm to 15 cm. For reinforcing jointed specimen, steel bars of 6 mm diameter simulating rock bolts were installed in the mass at an orientation perpendicular to shearing direction. The tensile strength of steel bars is about 550 MPa. Three rock bolts were installed in the specimen at 30 cm centre-to-centre spacing. For this purpose, special concrete cubes of size 15 cm x 15 cm x 15 cm, with a central hole of 10 mm diameter were prepared. Steel bars were grouted with 4 mm thick cement mortar. No pretension was applied to the rock bolts. The elevation and plan of large direct shear test is shown in Fig 1. Wooden plates of 5 cm thickness were placed at top and bottom of the shear box to fill the reaming gap in the shear box. The tests were done in strain-controlled mode. The shearing rate was maintained at 1.25 mm per minute.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No. A 150.)

Fig. 1 Elevation and plan of large direct shear test on reinforced jointed rock mass

Table 1 Constituent of concrete used as model material

Material Ratio (by weight) Cement 1 Sand 1.57 Coarse aggregate 2.91 Water 0.48

Four LVDTs were used to measure the vertical displacement while two were used to measure the horizontal displacements. Shear load, horizontal displacement, and vertical displacement were recorded during the test. 3 RESULTS AND DISSCUSSION The jointed rock mass specimens with and without reinforcement were tested at 0, 0.5, 1 and 2 MPa of normal stress respectively. Figure 2 shows the shear stress vs horizontal- shear- displacement plots of unreinforced and reinforced rock mass for normal stress equal to zero. In case of unreinforced rock mass, the shear stress increases with shear displacement, reaches it peak value of 0.05 MPa at a horizontal displacement of 25.6 mm, and after that more or less remains constant. The test was stopped at a horizontal displacement equal to 47.7 mm. In case of reinforced rock mass, the peak stress was 0.17MPa, which occurs at a horizontal displacement of 10.9 mm. After this peak, there was a small drop in shear stress and the test was stopped. The percent increase in the shear strength due to reinforcement is about 240%. Inspection of the reinforcing bars, after the

conduct of the test, indicated no yielding of bars or failure of intact material around the bars. The reinforcement has played role in interlocking the rock mass rather than actually acting as a structural members. It is the self supporting capacity of the rock mass, that is enhanced, and the strength of the rock bolt does not play significant role, in mobilising the strength of the rock mass. The shear stresses at the shearing plane are too low to induce yielding of bolts.

Fig. 2 Shear stress vs horizontal shear displacement plot for σn = 0.0 MPa Figure 3 through 5 show the shear stress vs horizontal shear displacement plots of unreinforced and reinforced rock masses at 0.5, 1.0 and 2 MPa normal stresses. Tests

An experimental study on strength behavoiur of reinforced jointed rock mass

conducted at normal stress of 0.5 MPa indicate increasing shear stress with horizontal shear displacement. In case of unreinforced rock mass, the shear stress remains almost constant after reaching a peak value of 0.36 MPa at 33 mm of displacement. Whereas the reinforced rock mass exhibits sudden drop after the peak stress of 0.70 MPa at 16.4 mm of displacement. For normal stress of 0.5 MPa, the percent increase in peak shear stress due to provision of reinforcement, was found to be 94%.

Fig. 3 Shear stress vs horizontal shear displacement plot for σn = 0.5 MPa

Fig. 4 Shear stress vs horizontal shear displacement plot for σn = 1.0 MPa In case of tests at normal stress equal to 1 MPa, the peak stresses for unreinforced and reinforced rock mass occur at 42.9 mm and 18.6 mm respectively. The peak shear stresses were 0.78 MPa and 0.88 MPa for unreinforced and reinforced cases. The strength of reinforced mass was 13% higher than that of unreinforced mass. When the normal stress increases to 2 MPa, the peak shear stress exhibited by unreinforced and reinforced rock masses are found to be almost same. Though the displacement was higher (40.6 mm), in case of

unreinforced mass compared to that of reinforced rock mass (16.3 mm).

Fig. 5 Shear stress vs horizontal shear displacement plot for σn = 2.0 MPa 4 SHEAR STRENGTH ENHANCEMENT DUE TO REINFORCEMENT Enhancement observed in shear strength of rock mass due to provision of reinforcement is presented in Table 2. Percent increase in shear strength of the rock mass with respect to the strength of unreinforced rock mass, is given in this table. Figure 6 shows the variation of percent increase in shear strength due to reinforcement with normal stress. It is seen that substantial strength enhancement occurs due to provision of rock bolts at low normal stress levels. This is due to interlocking introduced by reinforcement at low normal stress levels. With increasing normal stress level, the strength enhancement reduces, and becomes negligibly small beyond a limiting value of normal stress. At higher normal stress, the joints are tight and interlocked; provision of rock bolts brings no further improvement in interlocking. In present case, the limiting value has been observed to about 2.0 MPa. For a given geometry of rock mass and pattern of reinforcement, there will be a different value of limiting normal stress. The normal stress level is therefore an important parameter for grouted rock bolts. Further research is required on this aspect. Failure envelopes for the two conditions of the rock mass are shown in Fig 7. Mohr–Coulomb shear strength parameters cmass and ϕmass as obtained from these plots are given in Table 3. One can see substantial increase in cmass due to provision of rock bolts. It is observed that when the rock mass is reinforced ϕmass value reduces. The reduction in ϕmass value is due to curvilinear nature of the failure envelope which indicates that, at higher normal stress level the behavoiur of reinforced and unreinforced rock mass tend to be similar.

Lok Priya Srivastava & Mahendra Singh

Fig. 6 Effect of normal stress on strength enhancement of jointed rock mass Table 2 Shear strength enhancement due to reinforcement

Normal stress (MPa)

Peak shear stress (MPa) Percent increase in

strength Unreinforced

mass Reinforced

mass

0.0 0.05 0.17 240 0.5 0.36 0.70 94.4 1.0 0.78 0.88 12.8 2.0 1.56 1.58 1.3

Fig. 7 Normal vs shear stress plot of unreinforced and reinforced rock mass Table 3 Mohr-Coulombs shear strength parameters

5 CONCLUDING REMARKS An experimental study has been conducted by performing large size direct shear tests (75 cm x 75 cm x 90 cm (height))

on rock mass specimen. The tests have been conducted for two conditions of the mass i.e. unreinforced and reinforced with a particular pattern of grouted rock bolts. The following conclusions have been drawn from this study i. There is substantial enhancement in shear strength of rock

mass when the shearing plane is subjected to very low normal stress. Strength enhancement is due to interlocking induced by the rock bolts, and not by failure of bolt material.

ii. As normal stress level increases the strength enhancement reduces.

iii. Beyond a limiting normal stress level there is negligible strength enhancement due to provision of rock bolts.

iv. If Mohr-Coulomb criterion is used, substantial increase is observed in cohesion cmass whereas the friction angle ϕmass reduces due to non-linear shape of failure envelope.

REFERENCES 1. Sakurai, S. (2010), Modeling strategy for jointed rock

masses reinforced by rock bolts in tunneling practice, Acta Geotechnica, 5,121–126.

2. Dight, P. M. (1983), Improvement to the stability of rock walls in open pit mine, PhD thesis, Department of Civil Engineering, Monash University, Australia.

3. Hartman, W. and Hebblewhite, B. (2003), Understanding the performance of rock reinforcement elements under shear loading through laboratory testing — A 30-year history, 1st Agcm Conference, 10-13 November 2003.

4. Grasselli, G. (2005), 3D Behaviour of bolted rock joints: experimental and numerical study, International Journal of Rock Mechanics & Mining Sciences, 42, 13–24.

5. Ferreroq, A. M. (1995), The shear strength of reinforced rock joints, Int. J. Rock Mech. Min. Sci. & Geomech, Vol. 32, No. 6, 595-605.

6. Jalalifar, H. and Aziz, E. N. (2010), Experimental and 3D numerical simulation of reinforced shear joints. Rock Mech Rock Engg, 43, 95–103.

7. Singh, M. (1997), Engineering behaviour of jointed model materials. PhD thesis IIT Delhi, India.

Case Cmass

(MPa) ϕmass°

Unreinforced mass 0.01 38 Reinforced mass 0.23 34


SLURRY CONSOLIDATION METHOD FOR LARGE CLAY BED PREPARATION Yashwant Apparao Kolekar, PhD student, IIT Bombay, [email protected] Dasaka S. Murthy, Assistant Professor, IIT Bombay, [email protected] Owasis Mir, PhD scholar, IIT Bombay, [email protected] ABSTRACT: This paper focuses on evaluation of degree of uniformity of clay bed prepared using slurry consolidation method. In this paper two series of tests are conducted to ascertain the uniformity and consistency of clay samples prepared. In the first set clay slurry is subjected to 18 kPa consolidation pressure and in another set 36 kPa consolidation pressure. The water content and vane shear strength of the clay bed after consolidation were measured horizontally at the surface and vertically down along the centre of the clay bed where stone columns are installed. The results from the study revealed that the coefficients of variation of water content of the clay bed horizontally and vertically are 1.43% and 3.82% for 18 kPa & 36 kPa respectively. Similarly, coefficient of variation of vane shear strength of the clay bed is 6.58% & 2.83% for 18 kPa & 36 kPa respectively. Based on the above results, it can be concluded that the samples prepared by slurry consolidation method produces uniform and consistent clay beds. Key words: Slurry, consolidation, unit cell, uniformity, consistency INTRODUCTION With scarcity of sites with good soils, the soft soils are being put to use by improving them by various techniques. Marine clay though found in abundance, is available only along the sea shores which are the regions of high economic growth. These marine clays are formed due to deposition of marine sediments with each tidal wave and undergo consolidation over a geological time scale. Earthen embankments, liquid storage tanks, silos, raft foundations etc are constructed on these soft soils along the coastal belt. These soft soils have a tendency to undergo large settlements when subjected to loading. To have a better understanding of the behavior of these soft soils, laboratory model studies become a necessity. LITERATURE [1, 3, and 4] have conducted studies on clay by adding water to clay in required quantity and mixing it thoroughly and leaving for 2-3 days for allowing the soil to reach a saturation stage, which does not simulate the process of consolidation. [5, 7] have conducted their research studies by consolidating clay in the laboratory. For consolidating clay in the laboratory, [7] have adopted the hydraulic consolidation method over a very low effective stress range, whereas [5] adopted the dead weight method of consolidation. Dead weight method though effective has an inherent problem of applying a huge dead weight on the tank for consolidation of clay soil, especially for large samples which can be reduced by adopting the lever arm technology. This paper discusses the preparation of large clay beds by consolidation of marine clay from slurry state to plastic state and the uniformity and consistency of the prepared clay beds. SET-UP In this study, the unit cell concept is adopted to consolidate the clay slurry for preparing the clay beds to perform laboratory studies on stone columns as shown in Fig. 1. The

unit cell has dimensions of 350 mm internal diameter and a height of 520 mm along with a collar of same diameter and height 250 mm. Unit cell has 3 outlets provided at the base at 120° to allow the excess pore water to exit. The marine clay required for preparing clay beds is obtained from Uran, near

Fig. 1 Unit cell along with collar Navi Mumbai. The soil properties are shown in Table 1. The ratio of L.L. of oven dried soil to L.L. of soil (natural state) is less than 0.75 hence the soil is organic. The clay slurry is prepared at a water content of 1.5 – 1.6 times liquid limit of marine clay [6]. A steel loading plate of diameter 330 mm and 12 mm thick with perforations is placed over the sand bed along with a geotextile provided at the interface of sand bed and the clay slurry as shown in Fig.2. The perforations in steel plate permit excess pore water to escape from the top, thus providing drainage boundary at the top. The diameter of

Yashwant Apparao Kolekar, Owais Mir, Dasaka S. Murty

loading plate is 20 mm less than the internal diameter of the unit cell and collar, to prevent any build up of friction at the boundary and at the same time sufficient enough to prevent any clay slurry coming out through the

Table 1 Properties of marine clay

Properties Values Natural moisture content (%) Field density (kN/m3) In-situ vane shear strength (kPa) Specific gravity of soil solids (Gs)

84 14.96 8 - 9 2.74

Liquid limit (Natural) (%) Liquid limit (Oven dried) (%)

101 51

Plastic limit (%) 40 Plasticity index (%) USCS classification

61 OH

Pre-consolidation pressure (kPa) Free swell index (%)

18 31

gap between unit cell and loading plate. Also, a layer of geotextile is placed over the aggregates at the bottom of the unit cell to act as a drainage layer, thus simulating a double drainage condition. On the inner surface of unit cell a thin plastic sheet is placed above which geotextile is laid, which acts as a drainage path for the pore water to flow towards the outlet provided at the bottom of unit cell and assist in expediting the rate of consolidation.

Fig. 2 Loading arrangement

TEST In the present studies two series of tests are carried out with 4 tests in each series. In first series of tests (1 – 4), the clay slurry is subjected to a consolidation pressure of 18 kPa, and in the second series of tests (5 – 8) to 36 kPa. Clay is soaked in water for a month’s time and then slurry is prepared using mechanical mixer for an hour to ensure homogeneous mixing. Slurry has a water content of 1.5 - 1.6 times the natural liquid limit. The high water content in the slurry

enhances the workability and minimizes the possibility of presence of entrapped air. The slurry is then placed in the unit cell. Furthermore during placing of slurry, the sides of unit cell assembly is slowly tamped on the exterior surface with the help of a wooden mallet, so that any air bubble present in the clay slurry is removed. An initial consolidation pressure of 1 kPa and 2 kPa is applied to first and second series of test respectively for 3 days under which the consolidation progresses. This is done so that, the marine clay slurry does not come out from the gap between loading plate and collar under sudden loading. After 3 days, consolidation pressure is increased to 4.5 kPa and 9 kPa for the first and second series respectively, through the lever arm mechanism which is widely used for conducting the odeometer tests. On completion of 6 days the entire consolidation pressure of 18 and 36 kPa is applied to the respective series of tests till the rate of settlement is less than 1 mm/day. Settlements are recorded in 3 days interval till the settlement is equal to or less than 3 mm for 3 days. [5] stated the criteria for consolidation to be complete for a given loading, when settlement is equal to or less than 1 mm/day is reached. It requires a period of 25 - 30 days to reach this stage of settlement. During the initial stage of consolidation the total settlements are greater than 50 mm, the range up to which the dial gauges are available. Hence, a water proof plastic tape having a least count of 1mm is fixed on the inner side of the unit cell for recording the settlements. [2] specified that the compressible silts and clay having undrained shear strength in the range of 15 - 50 kPa are suitable for better performance of stone columns. The clay beds consolidated at 18 kPa and 36 kPa yielded undrained strengths of 8 - 9 kPa and 20-22 kPa respectively. The water content and vane shear strength for these clay bed specimens are recorded on the surface at a distance of 25 mm, 100 mm, 175 mm, 250 mm and 325 mm from edge of the unit cell and water contents are also recorded along the core made at the centre of clay bed at depths 0 mm, 100 mm, 200 mm, 300 mm, and 400 mm from top as shown in Fig. 3.

(a) (b)

Fig. 3 Layout (a) horizontal & (b) vertical sampling locations RESULTS AND DISCUSSION Water content-Horizontal profile

Slurry consolidation method for clay bed preparation

The variation of water content along top surface of clay bed in horizontal direction is shown in Fig. 4. For 18 kPa and 36 kPa consolidation pressure the water content is higher in the

0 50 100 150 200 250 300 35075

80

85

90

95

100

Test 1 - 4 at 18 kPaTest 5 - 8 at 36 kPa

WA

TER

CO

NTE

NT

(%)

DISTANCE FROM EDGE OF UNIT CELL (mm)

1 2 3 4 5 6 7 8

Fig. 4 Water content variations at top layer

central part of clay bed and reduces laterally towards at the boundaries, this is due to the presence of drainage boundary at the periphery in the form of geotextile and may be because of non-uniform contact pressure at the interface of the sand and clay bed with the higher pressures at the edges than at the center. The COVs for 18 kPa and 36 kPa consolidation pressures are 0.85% and 1.76% respectively.

80 85 90 95 100 105

400

300

200

100

0

DIS

TAN

CE

FR

OM

TO

P (m

m)

WATER CONTENT (%)

1 2 3 4 5 6 7 8

Test 1 - 4 at 18 kPaTest 5 - 8 at 36 kPa

Fig. 5 Water content variations along the centre of clay bed

Water content-Vertical profile The vertical water content profile is also prepared along the central core of clay bed as shown in Fig. 5. For tests 1 - 4 the water contents are nearly uniform over the depth and for tests 5 - 8 it shows a slight variation in the profile. The water content at the top surface is least and increases downward and then decreases at the base. This is due to the proximity of drainage boundary at the base and top. It is observed that the variation in the values of water content is very minimal; indicating uniform sample with good accuracy can be obtained with the given set up. The COVs for 18 kPa and 36 kPa are 1.43% and 3.82% respectively.

Shear strength Fig 6 shows the variation of shear strength with the horizontal distance along the top surface of clay bed. For test series 1 - 4, the shear strength varies in the range 8 – 9.5 kPa and test series 5 - 8, it varies in the range 20 - 22 kPa. The variation of shear strength is in tandem with the water content values recorded. The COVs of shear strength for 18 kPa and 36 kPa consolidation pressures are 6.58% and 2.83% respectively. Greater values of shear strength are recorded at the edges than that at the centre. It is observed that variation is quite minimal; hence the samples obtained indicate good repeatability of clay samples and the same can be prepared for laboratory testing. It is noted from literature that the COV of 10% is well accepted in geotechnical engineering, in view of the large uncertainties involved in the associated process.

0 2 4 6 80

10

20

30

VA

NE

SH

EA

R S

TRE

NG

TH (

kPa)

TEST (N)

Test 1 - 4 at 18 kPa Test 5 - 8 at 36 kPa

Fig. 6 Average shear strength at top layer

CONCLUSION The slurry method of consolidation is more appropriate for preparing the clay beds for laboratory model studies and with the use of lever arm mechanism a better control over the uniformity of the clay bed is achieved. 1. The Coefficients of variation of water content at the

surface of clay bed are 0.85% & 1.76%. 2. The Coefficients of variation of water content along the

centre of clay bed are 1.43% and 3.82%. 3. The Coefficients of variation of average shear strength of

clay bed are 6.58% and 2.83%.

REFERENCES 1. Ambily, A. P., and Gandhi, S.R. (2007), Behaviour of

stone columns based on experimental and FEM analysis, Journal of Geotechnical and Geoenvironmental Engineering (ASCE), Vol. 133, No. 4, 405–415.

2. Barksdale, R. D., and Bachus, R. C. (1983), Design and construction of stone columns, Report No. FHWA/RD-83/026, Office of Engineering and Highway Operations Research and Development, Federal Highway Administration, Washington, DC, USA.

Yashwant Apparao Kolekar, Owais Mir, Dasaka S. Murty

3. Deb, K., Samadhiya, N., and Jagtap, N, B., (2011), Laboratory model studies on unreinforced and geogrid reinforced sand bed over stone column improved soft clay, Geotextiles and Geomembranes, Vol. 24, No. 6, 190-196.

4. Malarvizhi, S. N, and Ilamparuthi, K. (2004), Load versus settlement of clay bed stabilised with stone and reinforced stone columns, Proceedings of GeoAsia-2004, Seoul, Korea, pp. 322-329.

5. Murugesan, S., and Rajagopal, K. (2007), Model tests on geosynthetic encased stone columns, Geosynthetics International, 14, No. 6, 346–354.

6. Prakash, K. (2011), Seepage induced consolidation test for the evaluation of compressibility characteristics of soft clays, Proceedings of IGC, Dec. 15-17, 2011, Kochi (Invited Talk-12).

7. Sridharan, A. and Prakash, K. (1999), Simplified seepage consolidation test for soft sediments, Geotechnical Testing Journal, ASTM, 22(3), 235-244.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No A153)

FROM GEOACOUSTICS TO GEOTECHNICS: EXTRACTING SOIL PARAMETERS USING ACOUSTIC METHODS

Gopu R. Potty, Department of Ocean Engineering, University of Rhode Island, Narragansett, RI, USA, [email protected] Jeanette Greene, Department of Ocean Eng., University of RI, Narragansett, RI, USA, [email protected] James H. Miller, Department of Ocean Engineering, University of Rhode Island, Narragansett, RI, USA, [email protected] Christopher D.P. Baxter, Department of Ocean Engineering, University of RI, Narragansett, RI, USA, [email protected] ABSTRACT: Acoustic propagation in a shallow water waveguide is sensitive to properties of the bottom such as the compressional and shear speeds and attenuation. Many methods have been developed during the past to estimate geoacoustic properties of shallow water sediments using acoustic approaches. The University of Rhode Island is in the forefront of geoacoustic sensing and has developed techniques to estimate compressional wave speed and attenuation in the past. A shear measurement system capable of estimating the shear speed profile in shallow water sediments will be presented in this paper. The shear measurement system consists of an array of geophones which collects Scholte wave data which will be inverted to obtain the shear speed profile in the sediment. The shear speed is strongly correlated to soil parameters and a geoacoustic model will be used to extract soil parameters based on the estimated shear speed profile. Data from sea tests will be presented and estimates of the shear speed and soil parameters will be compared to ground truth data from the test locations. [Work supported by U.S Office of Naval Research] INTRODUCTION Geoacoustic parameters such as compressional and shear wave speeds and attenuation are important in acoustic propagation modelling in shallow water. University of Rhode Island has been in the forefront of developing acoustic based techniques for the inversion of sediment geoacoustic parameters. Inversion techniques for estimating compressional wave speeds and attenuation has been developed and validated using field data [1-3]. Recently a new shear measurement system was developed with the objective of estimating the shear wave speeds in the sediment. The system collects interface (Scholte) wave data which is then used to estimate the shear speed. The Scholte wave propagates at the boundary between elastic (sediment) and a fluid medium and its speed is closely related to the shear wave speed in the sediment. The properties of a Scholte wave can be summarized as follows [4]:

• It has a rotational particle movement. • It is dispersive if the shear speed varies with depth. • It propagates with a velocity of approx. 0.9 times the

shear speed. Extraction of shear wave speed in the sediment based on Scholte wave speed measurements is a well-established technique in the ocean acoustics [4] and the geotechnical communities [5-8]. In the present study we extend this technique to estimate physical properties of the sediment using the shear wave speed and the Biot-Stoll model. The Scholte wave data will be used to estimate a shear wave speed profile in the sediment and some of the sediment properties will be then estimated to fit this shear wave speed profile. The multi-step procedure is summarized as follows:

1. Calculate the phase speed of the Scholte wave arrivals from the geophone data.

2. Estimate the shear speed profile in the sediment by fitting the dispersed Scholte wave arrival data to modelled arrivals calculated using Chapman-Godin approach. This model requires that the shear speed is expressed as power law profile.

3. Calculate the shear speed as a function of depth for a given set of sediment parameters (porosity, grain size and grain density) using Biot-Stoll model. By iteratively varying the parameter values fit this shear speed profile to that estimated in step 2.

The present study is documented as follows: The shear measurement system and its components are described in the next section. A brief description of data processing employed is also included in this section. A concise review of sediments in Narragansett Bay is provided in the following section. The Chapman-Godin approach to model Scholte wave dispersion and Biot-Stoll model to model the shear wave speed are summarized in the subsequent sections. Discussion of the results is followed by conclusions of this study. URI SHEAR MEASUREMENT SYSTEM At University of Rhode Island, Miller and Potty have developed a geophone/hydrophone array capable of collecting interface wave data. This system consists of a number of geophones and hydrophones, Several Hydrophone Receive Units (SHRUs) built by Woods Hole Oceanographic Institution (WHOI) as data acquisition system, and a cage which houses the SHRUs to which the hydrophone/geophone array will be connected [9]. We excited Scholte waves by means of direct impact by dropping a weight. A field test was conducted north of the R/V Endeavor Pier in the Narragansett Bay Campus of the University of Rhode Island on March, 2011 [9]. One SHRU and a four geophone

Potty, G. R., Greene, J., Miller, J. H and Baxter, C.

array were deployed off the stern of R/V Endeavor in approximately 6 m of water. The geophones were properly placed, with a spacing of 5 m underwater, with the help of divers. A 135 kg weight was released from just below the surface of the water off the stern of R/V Endeavor using the ship’s capstan (Fig. 1). The signals received on the four geophones were used to calculate the phase velocity as a function of frequency. The sea test yielded time series of geophone recorded seabed motions due to the weight drops. Fig. 2 shows an example of such a time series collected by four geophones separated by 5 m each.

Fig. 1 Locations of the source (weight drop), and the geophone receivers [5]. The data collected by the geophones were used to calculate the phase speed of the interface waves [9]. The phase difference between interface wave arrivals between pairs of geophones as a function of frequency is calculated from the cross-spectral density between the geophone data. This phase difference is used to calculate the phase speed (as a function of frequency) knowing the distance between the geophones. Fig. 3 shows the phase speed data calculated for different geophone pairs for various source events.

Fig. 2 Interface wave signal received at the four geophones due to one of the weight drops. The geophone closest to the source is A and farthest from it is D. NARRAGANSETT BAY SEDIMENTS Leblanc et al., [10] McMaster, [11] Richardson [12] and Baxter et al. [13] have provided description of sediment characteristics in Narragansett Bay. Approximately 20,000 years ago, melting ice formed a fresh water lake covering an area larger than the current size of Narragansett Bay. The

soils found in Narragansett Bay today that were deposited during this time consist of sands and inorganic silts and are

Fig. 3 Phase speed computed from the geophone data. The dashed line is the modeled phased speed calculated using Chapman-Godin model for a shear speed profile in the sediment as shown in Figure 5. commonly referred to as outwash deposits. These deposits consist of layers of silts and sand and they have been observed to be as much as 15 to 50 m thick in the vicinity of the Jamestown Verrazzano Bridge [12-13]. McMaster [11], based on a nearby seismic sections, reported the thickness of the sediments near the western shore of the bay as 45–50 m. This seismic section is near the above mentioned bridge which is approximately 3 to 4 km north of the location where acoustic data were collected. Leblanc et al. [10] reported compressional wave speed as approximately 1650 to 1700 m/s for nearby sediments. Based on the seismic sections and bore logs presented in Richardson [12] a sediment profile as shown in Fig. 4 was assumed. It should be noted that this sediment description is approximate and is based on published data from nearby location. This sediment model was used to estimate the shear speed in these layers using the Biot-Stoll model.

Fig. 4 The sediment profile assumed for the Narragansett Bay. Numbers on the left indicate the depth of each layers. MODELING SCHOLTE WAVE DISPERSION A simple approach developed by Chapman and Godin [14] will be used to model the dispersion (frequency dependent phase speeds) of Scholte (interface) wave data. This method models the shear speeds in the sediment as a smooth depth dependent function using a power law variation. Based on this power law variation of shear speed in the sediment, the

From Geoacoustics to Geology: Extracting soil parameters using acoustic methods

phase speeds of the Scholte wave can be calculated using the Godin-Chapman approach [14] as described below: Let us assume the shear profile in the form, ( ) (1) 0

νzczcs = Where, cs is the shear speed, z is the depth below the water-sediment boundary, and c0 and ν are constants. c0 is the shear speed at unit depth and the parameter ν, which must be in the range 0 < ν <1, governs the rate of increase of shear speed with depth. Exact solutions for the phase speed (Vn) of the nth mode Scholte wave exists for ν=0.5 (square root profile) and is given as [14]:

( ) ( )

( )

1,2,3,..n

0n 1

21

20

20

⎪⎪⎩

⎪⎪⎨

⎧

=

=+

=⎟⎠⎞

⎜⎝⎛ =

fc

Rfc

Vn

π

πν

(2)

Where, f is the frequency in Hz and R is the ratio of the water density to sediment bulk density. The power law type variation is a reasonably good approximation to the layered sediment shear speed depth dependence as demonstrated by Chapman and Godin [14] (based on the analysis of a number of published data sets).

Fig. 5 Shear speed profile (continuous line) that will fit the phase speed data in Fig. 3. The stair case profile is the Biot-Stoll model predictions using the sediment parameters shown in Table 1. GEOACOUSTIC MODEL In order to model acoustic propagation, sediments are modeled as a fluid or equivalent fluid medium, or as a visco-elastic solid with associated compressional and shear waves (p and s waves) or as a poro-elastic medium which considers the soil medium as a porous frame with fluid filled pore spaces. Even though there has been considerable debate on the validity of these models, it has been generally accepted that poro-elastic models provide a realistic description of the energy propagation and dissipation in porous sediments. Biot-Stoll model will be used in this study to model the shear speed in the poro-elastic sediment. There have been many papers discussing the modeling aspects of Biot theory. Some of the key expressions from one such paper by Badiey et al. [15] are summarized here. Since

the main focus of this study is shear waves, only the expressions for shear speeds will be highlighted, but all the other wave parameters are described in detail by Badiey et al., [15]. Following Badiey et al., [15] the shear speed can be expressed as:

( )

frametheofmodulusshear -G porosity -

density fluid densitygrain

-1bygiven is density bulk thewhere

(3) '

'

f

21

2

φρρ

φρρφρρ

ρρ

−−

+=

⎟⎟⎠

⎞⎜⎜⎝

⎛

−=

s

fs

fs m

Gmc

In the above equation m’ is given by the following expression [15]:

( )

frequencyangular - typermeabili intrinsic -k

viscositydynamic fluidtcoefficien mass madded -

(4) 1'

f

ω

μ

ωμ

φρ

−

−+=

cwhere

kF

icm ff

F is a viscosity correction factor and Badiey et al., [15] has described the method to calculate this parameter. It can be seen from these equations that the shear speed is a function of a number of sediment parameters such as the permeability, density, porosity and shear modulus of the frame. The shear modulus of ocean sediments, in turn, is a function of its porosity and overburden pressure. In summary, it should be possible to estimate many important sediment parameters by inverting the shear speed in the sediment. Many investigators have compiled field data and developed empirical models to calculate the wave parameters in the sediments. For example, Jensen et al. [16] provide empirical relationships to calculate shear wave speed which, for sand, silt and clay, are reproduced below:

(clay) / 100(silt) 80

(sand) 1103.0

s

3.0

smVzV

zV

s

s

<=

= (5)

Where, Vs is the shear speed in m/s and z is the depth below water-sediment interface in meters. RESULTS AND DISCUSSION The results of the present study are summarized in Table 1 and Fig. 5. The continuous line in Fig. 5 is the shear speed profile (ν=0.5) which fits well to the Scholte wave dispersion data which is shown in Fig. 3. The broken stair step line indicates the shear speeds calculated using Biot-Stoll model. The dotted and dash-dotted lines in Fig. 5 show the approximate shear speed profiles for sand and silt given in

Potty, G. R., Greene, J., Miller, J. H and Baxter, C.

Eq. 5 [16]. The estimated shear speed profile lies between the shear speeds corresponding to sands and silts. The values for the near-surface sediments are lower than 100 m/s which indicate the presence of softer sediments near the sea-bottom. This is reflected in the phase speeds of the Scholte wave also (Fig. 3). The higher frequencies (lower wavelengths) have lower phase speeds compared to lower frequencies. The inputs to the Biot-Stoll model, which produced the stair-step shear speed profile in Fig. 5, are shown in Table 1. Table 1: Inputs to the Biot - Stoll model. The sediment type (last column) is the approximate sediment description based on Hamilton's classification [17]. Layer (m)

Grain density (g/cm3)

Grain size (mm)

Porosity (%)

Sediment type

0-5 1.7 0.02 62 Silt/sand-silt-clay

5-15 1.86 0.10 48 Very fine sand 15-20 1.90 0.125 47 Fine sand/very

fine sand 20-35 1.95 0.4 41.5 Fine sand/coarse

sand 35-50 2.03 0.53 38.6 Coarse sand Based on the inputs to the Biot-Stoll model which provides the best fit to the shear speed profile estimated using Chapman-Godin approach, the sediments can be classified as shown in the last column of Table 1. This description is based on Hamilton’s classification [17] of continental margin sediments. The types of sediments in the layers vary from very fine sand and silt to coarse sand. They compare reasonably well with the previously reported sediment description in the Narragansett Bay which is shown in Fig. 4. CONCLUSIONS The shear speed profile at a location in Narragansett Bay very close to the shore is estimated using Scholte wave data collected using a geophone array. Biot-Stoll model is then used to back calculate some of the physical properties of the sediments which fit the estimated shear speed profile. If developed further this approach has the potential to provide a tool for the rapid estimation of range averaged and depth dependent sediment parameters such as porosity, density and grain size. REFERENCES 1. Potty, G., Miller, J. H., Lynch, J. F., and Smith, K. B.,

(2000) Tomographic Inversion for sediment parameters in shallow water, J. Acoust. Soc. Am. 108(3), 973-986.

2. Potty, G., Miller, J. H., Dahl, P. H., and Lazauski C. J., (2004) Geoacoustic inversion results from the ASIAEX East China Sea Experiment, IEEE J. Oceanic Eng., 29(4), 1000-1010.

3. Potty, G., Miller, J. H., and Lynch, J. F., Newhall, A., Wilson, P., (2008) Geoacoustic inversions using combustive sound source signals, J. Acoust. Soc. Am. 124 EL146.

4. Frivik. S. A., and Hovem, J. M., (1998) Determination of Shear Wave Properties in the Upper Seafloor Using Seismo- Acoustic Interface Waves, Proc. of the OCEAN’S 98, Vol. 2, 682 - 686.

5. Luke, B. A. and Stokoe, K. H., (1998) Application of SASW Method Underwater, J. Geotech. and Geoenv. Eng., 124 (6), 523.

6. Park, C.B., Miller, R.D., Xia, J., Ivanov, J., Hunter, J.A., Good, R.L. and Burns, R. A., (2000) Multi-channel Analysis of Underwater Surface Waves Near Vancouver B.C., Canada, Technical Program with Biographies, SEG, 70th Annual Meeting, Canada, 1303-1306.

7. Rosenblad, B.L., Kalinski, M., Stokoe, K. H., and Kavazanjian, E. (2003), Application of Surface Wave Methods for Measuring Very Near Surface Seafloor Stiffness Properties, Proc., 13th Int. Offshore and Polar Engineering Conference, Vol. 1, May. 594-600.

8. Kaufmann, R.D., Xia, J., Benson, R.C., Yuhr, L.B., Castro, D.W., and Park, C. B., (2005) Evaluation of MASW Data Acquired with a Hydrophone Streamer in a Shallow Marine Environment, J. of Env. and Eng. Geophysics, 10 (2), 87-98.

9. Greene, J., Potty, G. R., and Miller, J. H., (2011) A Measurement System for Shear Speed Using Interface Wave Dispersion, Proceedings of the International Symposium on Ocean Electronics (SYMPOL), Cochin, India, 211-216.

10. LeBlanc, L. R., Mayer, L., Rufino. M., Schock, S. G and King, J. (1992), Marine Sediment Classification using the Chirp Sonar, J. Acoust. Soc. Am., 91(1), 107-115.

11. McMaster, R. L., (1979), Bedrock Surface and Overburden at the Jamestown Bridge, West Passage, Narragansett Bay, Rhode Island, Based on Seismic Reflection Profiling, Technical Report, University of Rhode Island, USA.

12. Richardson, B., (2011) A Case Study on Pile Relaxation in Dilative Silts, Master’s Thesis, University of Rhode Island.

13. Baxter, C. D. P, Page, M. J., Bradshaw, A. S., and Sherrill, M., (2005) Guidelines for Geotechnical Site Investigations in Rhode Island, Final Report, FHWA-RIDOT-RTD-05-1S, RI Department of Transportation.

14. Chapman, D. M. F., and Godin, O. A., (2001) Dispersion of interface waves in sediments with power-law shear speed profiles. II. Experimental observations and seismo-acoustic inversions, J. Acoust. Soc. Am. 110 (4).

15. Badiey, M., Cheng, A. H-D., and Mu, Y., (1998) From geology to geoacoustics—Evaluation of Biot–Stoll sound speed and attenuation for shallow water acoustics, J. Acoust. Soc. Am. 103, 309.

16. Jensen, F. B., Kuperman, W. A., Porter, M. B., and Schmidt, H., (2000) Computational Ocean Acoustics, Springer- Verlag, New York.

17. Hamilton, E. L., (1980) Geoacoustic modeling of the seafloor, J. Acoust. Soc. Am. 68(5).

Proceedings of Indian Geotechnical Conference December 13-15,2012, Delhi (Paper No: A 156)

MODEL STUDIES ON GEOSYNTHETIC REINFORCED UNPAVED ROAD SECTIONS M. N. Asha, Ph.D student, Indian Institute of Science, Bangalore, India, [email protected] G. Madhavi Latha, Associate Professor, Indian Institute of Science, Bangalore, India, [email protected] ABSTRACT: This paper studies the behaviour of unreinforced and geosynthetic reinforced unpaved road sections through laboratory model tests. Repeated loading tests were carried out on unpaved road sections built in a steel test tank. A clay subgrade was used in the tests with aggregate sub base. In model tests on reinforced sections, geosynthetic layers are placed at the interface of subgrade – sub base. Different geosynthetics, namely, geotextile, geogrid and geocell are used in the experiments. The development of elastic and plastic strains in the systems with load cycles and the secant modulus of the test sections were analyzed. It is observed that reinforcement has generally improved the elastic strains and decreased the cumulative plastic strains. Significant reduction in the settlements was observed with the reinforcement, which was visible from the deformation profiles plotted on the sections. Comparison of different geosynthetic reinforced sections revealed that the geocell reinforced section showed highest secant modulus compared to other sections. INTRODUCTION Road network of any country is indicative of the development of the country. Pavements are always subjected to repeated loads and the life of the pavement essentially depends on the resistance of the component layers in resisting the superimposed load. The application of geosynthetics in pavement construction has increased drastically in the recent past. Suitability of different geosynthetic materials like geogrids and geotextiles in reducing the rut depth and increasing the service life of the pavements has been studied by many researchers by carrying out static load tests [1, 2, 3] or repeated/cyclic load tests [1, 3, 4, 5] on model pavement sections. Use of geocells to improve the bearing capacity of sand fills underlain by weak clayey subgrade has been investigated by few researchers by carrying out static load tests [6, 7, 8] and cyclic load tests [8]. Repeated load tests on reinforced road sections were carried out by few researchers and there are many grey areas in these studies. Especially the investigations on elastic and plastic strains and their implications on overall behaviour of reinforced road sections subjected to repeated loading are very limited. This study is aimed at understanding the repeated load behaviour of reinforced test sections in terms of elastic and plastic strain development. Reinforcement in different forms like grid and geocell is used in this study to explore the effect of these forms on the cyclic deformation behaviour and modulus of the unpaved road sections. MATERIALS USED The subgrade soil used in the experiments is the locally available red soil. The soil used is classified as CI according to IS: 1498, 1987. The properties of the subgrade soil used in the experiments are summarized in Table 1. Aggregate from a nearby quarry is used as the sub-base material. Granular material of various size ranges are collected and sampled such that it conforms to Grading III of granular sub-base design as given by Ministry of Rural Development, Specifications for rural roads, 2004. The grain

size distribution curves for the subgrade soil and the sub-base course material used are shown in Fig. 1. Table.1. Soil Properties

Fig. 1. Grain size distributions of the subgrade soil and aggregate used for sub-base in the experiments Geosynthetics used for reinforcement in the experiments are strong geogrid, weak geogrid and geocells. To prevent the mixing of subgrade soil and sub-base layer, the geogrid and layer are underlain by geotextile, which functions as a separator as well as additional reinforcement. The photograph of geogrid and geocell reinforcement (underlain

Colour Reddish brown Specific gravity 2.71 Soil classification CI Liquid limit & Plastic limit, % 36 & 24 Shrinkage limit, % 21 Maximum dry unit weight, kN/m3 18.24 Optimum moisture content, % 15.5 Unsoaked CBR value, % 19

M.N.Asha & G.Madhavi Latha

by geotextile) used in the experiments is shown in Fig. 2. Properties of geosynthetics are given in Table 2.

Fig. 2. Reinforcing materials used in the experiments (a) Geogrid layer and (b) Geocell layer Table.2. Properties of geosynthetics used Property Geotextile Strong

geogrid Weak geogrid

Aperture size (mm)

- 30×30 35×35

Ultimate tensile strength (kN/m)

55 38 7.6

Failure strain (%) 38 17 18 Secant modulus at 2% strain (kN/m)

151 588 152

Mass per unit area (g/m2)

230 530 220

Geocell Geocell Depth (mm)

75

Weld distance (mm)

400

Each cell opening size (mm×mm)

289×256

Seam strength (N)

1050

Table.3. Details of experiments carried out Systems Description

UR No reinforcement WG Weak geogrid underlain by geotextile SG Strong geogrid underlain by geotextile GC_75 Geocell of 75 mm height underlain by

geotextile TEST SET-UP Experimental studies are carried out in a steel tank of 750 mm × 730 mm cross section and 600 mm height. Load is applied through a steel plate of 147 mm diameter and 10 mm thickness. To minimise the boundary effect, diameter of the plate is maintained as 1/5th of the size of the mould [5, 9]. Four model tests were carried out on the unpaved road sections and the details of the tests are summarized in Table 3.

SAMPLE PREPARTION Within the tank, model sections of reinforced and unreinforced unpaved roads are constructed. The height of soil subgrade is maintained as 400 mm and that of granular sub-base layer is maintained as 200 mm in all tests. The beneficial effect of geosynthetics will be evident at high water contents. Hence, the subgrade soil is compacted to a water content of 17 % (slightly to wet of OMC) at a density of 12 kN/m3 in four lifts. Each lift is given 50 blows using a drop hammer of 5 kg falling from a height of 450 mm on a square base plate of 150 mm × 150 mm in size.

The granular sub-base layer of the required gradation is compacted in a single lift for the remaining 200 mm height of the tank at a density of 16 kN/m3. A total of 75 blows using the drop hammer were given for the compaction of sub-base layer.

In tests with reinforcement, the reinforcing layer viz., geogrid layer or geocell layer are placed at the interface. The schematic sketch of the reinforced soil-aggregate system with geocell layer is shown in Fig. 3. For geocell reinforced samples, initially the geocells are stretched within the tank using C-clamps as shown in Fig. 2. For filling the sub-base material in geocells, starting from one end, material is filled in each cell to 1/2 height. Once all the cells are 1/2 filled, material is filled up to the top of the geocell. The photograph of aggregate being filled in geocell pockets is shown in Fig. 4.

(1) Granular sub-base layer (2) Subgrade soil (3) geocell layer (4) Load cell (5) Hydraulic jack (6) Oil pump

All dimensions are in mm

Fig. 3. Schematic diagram of the test set-up with geocell layer TESTING PROCEDURE A load of 5 kN is applied repeatedly on the unreinforced and reinforced model pavement sections for a total of 100 cycles . The loading is applied in increments upto a maximum load of 5 kN and unloading is done gradually. Two dial gauges are placed on the loading plate to measure its deformations. For measuring surface deformations of the granular sub-base layer, three dial gauges are placed on one side of the tank and one dial gauge on the other side of the tank. The dial gauge readings of the plate are recorded at regular intervals during

Model Studies on Geosynthetic Reinforced Unpaved Road Section

loading and unloading. The deformation profile is recorded for every 10 cycles. A typical loading-unloading cycle takes around 2.5-3 minute.

Fig. 4. Photograph of aggregate being filled within geocell pockets RESULTS AND DISCUSSIONS All the systems (both unreinforced and reinforced systems) exhibited a punching failure. The experimental results are analysed to understand the relative benefit of different types of geosynthetic reinforcement. In all the reinforced cases, the geotextile could prevent the intermixing of granular sub-base and subgrade soil. Pressure-settlement response The pressure-settlement plots of unreinforced and reinforced model sections are compared to understand the relative performance of various geosynthetic reinforcements. The unreinforced section experienced large deformations beyond a load magnitude of 4 kN. Hence, repeated loading was carried out for a load magnitude of 4.25 kN for unreinforced sections, whereas for reinforced sections the load magnitude was 5 kN. The pressure-settlement plot in the loading stage of first cycle for various systems is shown in Fig. 5.

Fig. 5. Pressure-settlement response for various systems

From the figure it is clear that the reinforced systems settled less in comparison to unreinforced system and the performance of various reinforced systems is more or less the same. The unreinforced system settled by an amount of 80

mm in the loading stage of first cycle with a lesser bearing capacity. Plastic settlement and Elastic settlement The elastic and plastic settlements developed in the various systems are computed for each cycle and are summarised in Fig. 6. From Fig. 6a it is seen that unreinforced systems developed a plastic settlement of 110 mm which is very huge when compared to that of unreinforced systems. The settlement developed in strong and weak geogrid is more or less the same. The plastic settlement of geocell reinforced system is less in comparison to that of geogrid reinforced systemGeocell layer worked like a basal mat for the sub-base with increased stiffness due to the apparent cohesion imparted by all round confinement effect of the geocells filled with granular sub-base.

Fig. 6. Variation of (a) cumulative plastic settlement and (b) elastic settlement with number of cycles for various

systems

Fig. 6b compares elastic settlements developed in various systems. The deformations are divided into two parts, elastic deformations, which were recovered after the load reversal and plastic deformations, which were not recoverable with the release of load. It is beneficial for the system to have more elastic deformations and less plastic deformations, as it represents the resilience of the system. More resilient the system is, higher would be its sustainability to load with recoverable deformations. Elastic settlement developed in

(a)

(b)

M.N.Asha & G.Madhavi Latha

geocell reinforced system and that of strong geogrid reinforced system are the highest and they are more or less comparable. Weak geogrid has less elastic settlement than that of other reinforced sections whereas unreinforced section has the least elastic settlement. Hence, it is clear that elastic settlements developed are in proportion to the stiffness of the systems. Deformation Profiles The surface deformation profile for all the model test sections at the end of 100 cycles is shown in Fig. 7. From the figure it is clear that all the test sections are subjected to punching failure. The beneficial effect of geosynthetic reinforcement is evident in reducing the displacement of the centre plate. However, the adjacent surface has been more or less unaltered in all these tests, as it happens in punching shear failure.

Fig. 7. Deformation profiles of unreinforced and reinforced sections at the end of 100 cycles

Secant Modulus Secant modulus is defined as the slope of the pressure vs. plastic settlement response at any load cycle. Secant modulus can be used to quantify the plastic settlement within the system. Fig. 8 compares the secant modulus of the various model sections. From the figure it is clear that the geocell reinforced section has the highest modulus and unreinforced section has the least modulus. The geogrid reinforced sections have exhibited more or less same moduli. CONCLUSIONS Based on the repeated load tests carried out on model unreinforced and reinforced unpaved road systems, the following conclusions are drawn. • Important benefit of including reinforcement in unpaved

roads is the reduction in plastic settlements. • Stiffer reinforcement provided better resistance to

repeated loads and the computed secant modulus of the test sections increased with the stiffer reinforcement at any load cycle.

• Geocell reinforcement showed much resilient behaviour compared to geogrid reinforcement, as significant amount of deformations observed were elastic and were recovered after the removal of load.

Fig. 8. Variation of secant moduli with load cycles for

unreinforced and reinforced unpaved road systems REFERENCES 1. Subaida, E.A., Chandrakaran, S. and Sankar, N. (2009),

Laboratory performance of unpaved roads reinforced with woven coir geotextiles, Geotextiles and Geomembranes, 27 (3), 204–210.

2. Milligan, G. W. E., Fannin, J. and Farrar, D. M. (1986), Model and full-scale tests of granular layers reinforced with a geogrid, 3rd Intl. Conf. on Geotextiles, April. 7-11, 1986, Vienna, Austria, 1A/8, 61-66.

3. Jarrett, P. M. (1986), Load tests of on geogrid reinforced gravel fills constructed on peat subgrades, 3rd Intl. Conf. on Geotextiles, April. 7-11, 1986, Vienna, Austria, 1A/8, 87-92.

4. Bhosale, S. S. and Kambale, B. R. (2008), Laboratory study for evaluation of membrane effect of geotextile in unpaved road, 12th Intl. Conf. of Intl. Association for Computer Methods and Advances in Geomechanics (IACMAG), Oct. 1-6, 2008, Goa, India, 4385-4391.

5. Leng. (2002), Characteristics and Behavior of Geogrid Reinforced Aggregate under Cyclic Load. Ph.D Thesis Submitted to Graduate Faculty of North Carolina State University, Raleigh, pp. 1-152

6. Sireesh S., (2005), Behaviour of geocell reinforced foundation beds. PhD thesis, Indian Institute of Science, Bangalore, India.

7. Mhaiskar, S. Y. and and J. N. Mandal, J. N. (1996), Investigations on soft caly subgrade strengthening using geocells, Construction and Building Materials, 10 (4), 281-286.

8. Tafreshi, S. N. M. and Dawson, A. R. (2012), A comparison of static and cyclic loading responses of foundations on geocell-reinforced sand, Geotextiles and Geomembranes, 32, 55-68.

9. IRC: 2004. Specification for Rural Roads, March 2009, Ministry of Rural Development, Indian Roads Congress, New Delhi, India.

10. Perkins, S. W. (1999), Mechanical Response of Geosynthetic-Reinforced Flexible Pavements, Geosynthetics International, 6 (5), 347-382.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No.A.158)

EXPERIMENTAL AND NUMERICAL STUDIES ON SAND-GEOTEXTILE INTERFACES V.Prashanth1, Research Student, Indian Institute of Science, Bangalore-500612, India, [email protected] A. Murali Krishna2, Assistant Professor, Indian Institute of Technology Guwahati, Assam, India, [email protected] G. Madhavi Latha3, Associate Professor, Indian Institute of Science, Bangalore-500612, India, [email protected] ABSTRACT: This paper presents the results of large direct shear tests on unreinforced and reinforced sand to determine the interface friction properties. Numerical modeling of these experiments is also presented. A modified large direct shear test setup (300 mm × 300 mm) is used for the geotextile interface tests. Woven and non-woven geotextiles with large variation in the surface roughness are used in the tests. The influence of test parameters like normal stress and type of geotextile on the interface frictional properties is analyzed. Laboratory large direct shear tests are simulated in a numerical code using FLAC2D and validated with the test results of experiments. Sand-sand and sand-geotextile interfaces are created in the models using the interface elements. Mohr-coulomb model is used to represent the sand in the model and the interface is represented using a linear Coulomb-shear strength interface. It is observed that the interface friction properties or shear stress-shear strain behavior obtained from the model are in close agreement with the experimental results. INTRODUCTION The mechanisms of soil-geosynthetic interaction are very complex. The interfacial friction characteristics drastically vary based on the type and properties of the geosynthetic material and soil, as observed from the laboratory test results presented by many authors [1, 2]. In all the cases, the peak interfacial frictional angle is less than the peak frictional angle of the soil. Therefore the possible weakest portion of the soil-geosynthetic composite system lies at the soil-geosynthetic interface. The bonding resistance and the lateral displacements caused by destabilizing forces in the reinforced structures are significantly influenced by interface frictional properties. It is established that the safe and economical design of geosynthetic reinforced soil structures is mainly influenced by the interaction properties between geosynthetic material and its neighboring soil. Design of reinforced soil structures based on numerical simulations incorporate such interface frictional properties directly. Thus, numerical and experimental studies of soil-geosynthetic interfaces are very important and the results from such tests have direct bearing on the designs. Interfacial friction behavior of soil-geosynthetic material is usually obtained by traditional direct shear tests with some modification. Several other methods like tilt table tests [3], inclined plane test [4] and pullout test [1, 5] were also used by many authors to investigate the interfacial friction properties. Modified direct shear testing is the best common method especially for the flexible geotextiles and has been used by several investigators [6, 7]. In this paper, the effect of shear stress-shear strain behaviour of geosynthetic reinforced sand and unreinforced sand under different normal stresses at a relative density of 70% was investigated through experimental and numerical simulations. A large scale modified direct shear test setup (300 mm× 30 mm × 185 mm) was used for this study. Laboratory experiments were simulated in a numerical model using

software FLAC2D (Fast Lagrangian Analysis of Continua) [8]. Earlier few researchers have developed numerical models to simulate direct shear test using FLAC3D to determine the shear strength of a rock joint [9] and [10]. Joint in the model is modelled with an interface element which satisfies the Mohr-coulomb shear-strength criterion. Plane strain properties of sandy soil were investigated using FLAC2D in few studies [11]. The numerical model succeeded to closely predict the actual measurements in experiments. Similar attempt has been made in the present study to simulate the interfacial shear tests in numerical model using FLAC2D. EXPERIMENTAL PROGRAM In the modified large scale direct shear box (Fig. 1), interface shear tests were carried out on sand-sand and sand-geosynthetic interfaces. Sand-sand interface tests were carried out by filling sand in both the halves of the direct shear box. In tests with sand-geosynthetic interfaces, one face of the geosynthetic layer is glued to a wooden plank and it is fixed to a wooden rigid block placed in the lower half of the direct shear box. Sand is filled in the upper half of the direct shear box and the interface is sheared through the movement on bottom box at a constant applied strain rate. Rigid block in the lower half is used for avoiding the sagging effect of the geosynthetic layer caused by normal load application. The testing program of sand–sand and sand-geotextile shearing test is shown in Table.1. Materials Sand Sand used in this study is locally available river sand. This sand is classified as poorly graded sand (SP) according to unified soil classification system. The maximum and minimum unit weights of the sand were determined as 16.58 kN/m3 and 14.02 kN/m3 respectively. Modified direct shear tests were performed on samples prepared by sand pluviation technique at a relative density (Rd) of 70% and the average

Prashanth,v., A, Murali Krishna., G, Madhavi Latha.

friction angle value obtained was 45º. Other index properties measured are presented in Table 2.

Fig.1 Schematic view of modified direct shear test Table 1 Testing Program

No Purpose Sample Normal Stress (σn) kPa

Rd(%)

1 Normal stress

WG-sand and NWG -

sand

27 70 52

77 2 Normal

stress

Sand-Sand 27 70 52

77 Geosynthetics A woven geotextile (WG) and a non-woven geotextile (NWG) were used in this study. The tensile properties of geosynthetics were determined using wide width tension strength test as per ASTM D4595 [12] (Fig.3). The tensile strength of woven and non-woven geotextiles are 35 kN/m (elongation at failure is 22%) and 38.8 kN/m (elongation at failure is 30%) respectively.

Table 2 Properties of sand used in the study D60

(mm) D30

(mm) D10

(mm) Cu Cc

0.45 0.22 0.16 2.81 0.67

NUMERICAL SIMULATION A two dimensional explicit finite difference program FLAC2D

was used to model all direct shear tests. The dimensions of the soil sample are 300 mm × 300 mm × 185 mm, which is identical to the laboratory soil sample in the large scale direct shear box. The numerical model is divided into upper half and lower half. The contact between these two halves is simulated using interface elements.

Fig. 2 Geosynthetic materials used in study Both sides of upper half and lower half are fixed only in horizontal direction to allow the sample to undergo vertical displacements. Testing procedure in numerical model is very similar to experimental procedure. At first, normal stress (σn) is applied on the sample and then lower half is allowed shear at a velocity corresponding to the experimental shearing. Displacement vectors observed in the model are shown in Fig. 3. Sand in the upper half of all direct shear tests (sand-sand & geotextile- sand) was represented using Mohr-Coulomb model. However in modified direct shear tests the wooden block fitted in the lower half is represented by an elastic model. The mechanical response of given soil model is normally visualized in two different phases, elastic and plastic phases. The elastic phase is normally characterized by the soil shear modulus G, and bulk modulus, K, whereas the plastic phase in Mohr-coulomb model is characterized by the frictional angle, φ and dilation angle, ψ. In all the cases the interface elements are represented by normal and shear stiffness between two planes which may contact each other. For either side of the interface, FLAC uses contact logic. Normal force (Fn) and shear force (Fs) at contact nodes were calculated using the given normal stiffness kn and shear stiffness ks. In this numerical model the interface element is imposed with a coulomb shear-strength condition. This results in adjustment of the contact forces. The coulomb shear-strength criterion limits the shear force by the relation.

Fsmax = cL+tan (φ)Fn (1) Where Fsmax= Maximum shear force at interface nodes, c=cohesion along the interface, L=effective contact length and φ=friction angle of the interface. Actual interface properties like interface frictional angle,φ, dilation angle, ψ were determined from direct shear test. Whereas normal stiffness Kn (normal stress, kPa/normal displacement, m), shear stiffness Ks (shear stress, kPa/shear displacement, m) were estimated from direct shear stress-strain response at 50% of peak shear stress. The numerical calibration exercise was carried out by adjusting the soil shear modulus, G and bulk modulus, K until the expected variations found in simulated results for the unreinforced sand and the same values were used for modeling the reinforced sands.

Experimental and Numerical studies on Sand-Geotextile interfaces

Fig. 3 Displacement vectors of numerical model (sand-non woven geotextile) Results and Discussion Effect of confining pressure The results from direct shear test on sand-sand at different normal stresses at a relative density of 70% are presented in Fig. 4. Increase in normal stress (σn) from 27 kPa to 77 kPa resulted in significant increases in initial stiffness, peak shear stress and peak shear strain. The frictional angle determined from the test is 45.5°. Fig. 4 also shows the comparison of results from numerical studies with the experimental observations. The results obtained from modified direct shear tests on sand-NWG and sand- WG under different normal stresses (σn) are shown in Figs. 5 and 6 respectively. These Figures show a distinguishable interfacial peak (τip) and interfacial post peak (τipp) shear stress behavior. The peak interfacial shear stress (τip) and the shear strain of both cases increase with an increase in normal stress (σn) (Table. 3 and Fig. 7). The variation of peak interfacial shear stress of both the cases and peak shear stress of sand-sand with the change of normal stress is shown Fig. 7. Increased peak shear stains (%) with increase in normal stress (σn) in all cases are presented Table 3. The average interfacial frictional angle (δ) of NWG-sand (42°) and WG-sand (40.5°) has been observed to be less than the average friction angle (φ) of sand-sand (45.5°). Increase in normal stress (σn) from 27 to 77 kPa at relative density of 70% caused an increase of peak shear stress to 37.49%, 45.30% and 44.75% in sand-non woven geotextile, sand-woven geotextile and sand-sand samples respectively.

One interesting observation is, nonwoven geotextile-sand interface exhibits perfectly plastic behavior after peak interfacial shear stress (τip) (Fig. 5), while the sand-WG interface exhibits strain softening behavior after the peak interfacial shear stress (τip) (Fig. 6).The slippage of soil particles could be the reason behind strain softening behavior of interfacial shear stress with shear strain immediately after the peak shear strength in sand-WG. Whereas in sand-NWG, the impinging of soil particles with NWG fibber give rise to perfectly plastic behavior after the peak interfacial shear stress (τip). This difference in the response of these two different interfaces could not be captures with a simple Mohr-

coulomb model and hence the plots show a perfect elasto-plastic response of numerical models irrespective of the type of geosynthetic material used in the tests. From Figs. 4 to 6 it is clear that there is close agreement between the results from laboratory tests and numerical simulations in terms of the peak strength and constitutive behaviour prior to the peak. Table. 3 peak shear strain (%) of woven and non woven geotextile with sand and sand-sand at different normal stress

No

σn ,kPa Shear strain (%) Sand-NWG Sand-WG Sand-Sand

1 77 5.9 4.1 2.8 2 52 4.4 3.8 2.2 3 27 3.04 3.05 2.1

Fig. 4 Sand-sand shear stress-shear strain response from experimental and numerical simulations

Fig. 5 Sand-NWG shear stress-shear strain response from experimental and numerical simulations Post peak difference in the behaviour for sand-WG interface could only be captured using advanced numerical models

Prashanth,v., A, Murali Krishna., G, Madhavi Latha.

including strain softening behaviour, which is not attempted in this study. There is significant difference in the stiffness of experimental and numerical stress-strain plots, as observed from Figs. 5 and 6. In case of unreinforced sand, this difference is not significant. Few studies in literature [13] reported that, a difference of 10-20% in the initial stiffness of the soil when using finite element model to simulate direct shear test can be possible. Peak shear stresses at different normal stress in numerical model of sand-sand have occurred at higher strains than the experimental models. In case of sand-WG and sand-NWG, the peak shear stresses at different normal stresses fairly match with the experimental results.

Fig. 6 Sand-WG shear stress-shear strain response from experimental and numerical simulations

Fig. 7 Peak shear stress versus normal stress of sand-sand, sand-WG and sand-NWG CONCLUSIONS The numerical models used in this study to represent the behaviour of unreinforced and reinforced sand in interface shear tests were successful in predicting the shear stress-shear strain responses of these interfaces very well. Limitations of the numerical models were the inability to capture the post peak response of sand-woven geotextile and relatively higher initial stiffness compared to experimental models. Given the

simplicity of the constitutive models used in the study, the comparisons are encouraging. REFERENCES 1. Juran, I., Knochenmus, G., Acar, Y. B. and Arman, A.

(1988), Pull-out response of geotextiles and geogrids (Synthesis of available experimental data). Proceedings of Symposium on Geotextiles for Soil Improvement, ASCE, Geotech. Special Publication 18, 92-11

2. Koerner, R. M. (2006), Designing with Geosynthetics, 5th Ed., Prentice Hall, Englewood Cliffs, NJ.

3. Wu, W., Wick, H., Ferstl, F., Aschauer, F. (2008), A tilt table device for testing geosyntheticinterfaces in centrifuge, Geotextiles and Geomembranes 26 (1),31–38.

4. Lopes, P.C., Lopes, M.L. and Lopes, M.P. (2001), Shear behaviour of geosynthetics in the inclined plane test – influence of soil particle size and geosynthetic structure, Geosynthetics International, 8(4), 327-342.

5. Lopes, M.L. & Ladeira, M. (1996), Influence of the confinement, soil density and displacement ratio on soil-geogrid interaction, Geotextiles and Geomembranes, 14(10), 543-554.

6. Lee K.M. and Manjunath V.R. (2000), Soil-geotextile interface friction by direct shear tests. Can.Geotech.J, 37(1), 238-252.

7. Lopes, M. L., Silvano. R. (2010), Soil/geotextile interface behavior in direct shear and pullout movements. Geotech Geol Eng, 28(6), 791-804.

8. FLAC User’s Guide (2005), FLAC - Fast Lagrangian Analysis of Continua. v 5.00, Itasca Consulting Group Inc, Minneapolis, MN, USA.

9. ZHOU Yong., LIU Jian-xin., LIU Qun-yi. (2009), 3- Dimensional modeling for direct shear test of structure plane in rock mass, Intl. conf on Information Engineering and Computer Science, ICIECS, Dec. 19-20, 2009.

10. Hang Lin., Ping Cao., Yong Zhou. (2010), Numerical simulation for direct shear test of joint in rock mass, I.J Image, Graphics and signal processing, 1(2), 39-45.

11. Magdi M.El-Emam., Mousa F.Attom. and Zahid H.Khan. (2012), Numerical prediction of plane strain properties of sandy soil from direct shear test, International Journal of Geotechnical Engineering, 6(1), 79-90.

12. ASTM D4595 – 09. Standard test method for tensile properties of geotextiles by the wide-width strip method. American Society for Testing and Materials, Philadelphia.

13. Potts, D.M., Dounias, G.T., and Vaughan, P.R. (1987), Finite element analysis of direct shear box test. Géotechnique, 37(1), 11-23.


ELECTRICAL PROPERTIES OF SOIL AND GEOTHECNICAL ENGINEERING

Author: Dennis M. Anderson, P.E., USA, email: [email protected]

ABSTRACT: Every unique soil type has a unique geo-electric signature. When the geo-electric signature is researched and established for a given soil type and integrated with physical, chemical and soil solution properties, then that data may be used to determine geotechnical field characteristics of the soil under test. This paper presents engineering principals that relate the investigation of soil electrical properties with geotechnical engineering. In 1992 the author and a co-researcher, William J. Ehni, theorized that the relationship between soil electrical resistivity and geotechnical engineering could be developed for applications in soil engineering. By 1999Anderson and Ehni successfully patented technologies that use electrical geophysics for civil engineering applications. In 2000 the Electrical Density Gauge Limited Liability Company was formed and capitalized to commercialize the new density technology. Mr. John Lundstrom was one of the principals in the newly formed company. His technology contributions and expertise in electrical engineering lead to a quantum shift in the technology that has become known as the “Electrical Density Gauge.” The research and development moved from measuring direct current resistance to the alternating current measurements of current, voltage, phase and temperature to calculate soil material density and moisture content. Dr. Darrell Word, PhD in electrical engineering was engaged the research and development of the electrical density gauge and added greatly to our understanding of the theory and applied science. In year 2008 United States Patent 6,963,205 titled Electrically Measuring Soil Dry Density was issued to the electrical density gauge research team. INTRODUCTION The technical aspects of this paper provide an expanded understanding of soils science and metamorphosis of the abstract nature of geophysics to a definitive and clear application in geotechnical engineering. In soil science, this paper provides a new understanding of soil electrical properties as measured and then used in geotechnical engineering. Separate soil properties, including physical, chemical, mineralogical, and soil mechanics when analyzed, both separately and collectively, with electrical properties of soil materials provides the basis for the use of electrical measurements in geotechnical engineering. The measurement soil electrical properties support the interrelated nature of soil properties. This paper explains the theory and practice of using electrical geophysical measurements of soil as combined with conventional geotechnical engineering to derive the physical characteristics of a soil material. Geophysics has historically been an interpretive science that has used creative investigative means such as electrical, nuclear, and seismic techniques to assess gross geological features deep within the earth. Even nuclear geophysics as employed in the density gauge, that has become an industry standard for soil geotechnical engineering, is not well understood by most soil engineers. The nuclear density gauge essentially counts hydrogen atoms and then correlates the count with a curve fitting program to approximate the density of the soil under test. The density and moisture approximations generated by a nuclear density gauge are within an allowable engineering tolerance for soil investigation. The electrical geophysical methods and apparatus for determining the in-situ soil

properties of density, moisture, and hydraulic conductivity represents the next generation in geotechnical engineering. MODERN SOIL MECHANICS & GEOTECHNICAL ENGINEERING Since these early researches investigated soil engineering properties, a wealth of knowledge has been amassed to help present day soil engineers safely design foundations and footing for civil construction and building. Standardized testing procedures have been developed to ensure public safety and uniformity in building practices. Most of the test procedures involve physical measurements and testing procedures to assess soil strengths and reactions to imposed loads by constructed structures. Knowing soil density and moisture content is of major importance in the construction of roads and foundations. Proper density and moisture are necessary to prevent premature failure of these constructions. To enable the construction of engineered foundations to meet civil construction specifications soils engineers must conduct geotechnical investigations to determine the character of the soil materials that will be used in the design and construction of the foundations. To properly engineer and design a foundation, the soil characterization is done by both laboratory and field tests that provide strength data that is used in the design calculations for that subject foundation. The routine laboratory test of evaluation of a soil material density is known as the proctor test. The ASTM D-698-00a "Standard Test Methods for Laboratory Compaction Characteristics of Soil Using Standard Effort" or ASTM D-1557-00 "Standard Test Methods for Laboratory Compaction Characteristics of Soil Using Modified Effort " are used to determine a soil materials maximum density at an optimum

Dennis M. Anderson, P.E., USA

moisture content. The engineering soil specification will require that the foundation is built and tested to meet some design density criterion at a design moisture content. Field soil density measurements are made physically by a process involving a replacement of a known weight of soil with a measured amount of sand of known and repeatable density. This test is commonly known as the Sand Cone Test. ASTM D-1556-00 "Standard Test Method for Density and Unit Weight of Soil in Place by Sand Cone Method" provides a detailed procedure and protocol of conducting the test. Soil moisture content is measured by determining the weight loss after oven drying. ASTM has several test procedures for determining the moisture content of soil. ASTM D-2216-98 "Standard Test Methods for Laboratory Determination of Water (Moisture) Content of Soil and Rock by Mass"; ASTM 4643-00 "Standard Test Methods for Laboratory Determination of Water (Moisture) Content of Soil by Microwave Oven Heating"; and ASTM 4959-00 "Standard Test Methods for Laboratory Determination of Water (Moisture) Content of Soil by Direct Heating" are three of the geotechnical industry standards of measuring soil moisture. The results of these measurements are used to determine the Wet Density, the Moisture Content, and the Dry Density of the soil. These are some of the engineering parameters necessary to determine that the soil construction is adequate for the intended use. To permit more rapid field measurements of the physical character of the soil, the Nuclear Density Gauge is commonly used in conjunction with the sand cone test. The nuclear density gauge ASTM standard is 2922-96 "Standard Test Method for Density and Soil-Aggregate in Place by Nuclear Method (Shallow Depth". The sand cone test is used to standardize the nuclear gauge for a specific type of soil, which allows the nuclear gauge to be used repeatedly in the same area on the same type of soil. This permits many more measurements on each site to be made quickly and without continual resort to the cumbersome sand cone test. Nuclear gauges are quite expensive and very costly to have repaired due to the nuclear source they contain. Principals of soil mechanics are important to understand the relationship of physical soil properties to measured electrical properties. Soil mechanics relates to the study of the response of masses composed of soil, water and air to imposed loads. By employing principals of soil mechanics which provides the analytical tools required for foundation engineering, retaining wall design, highway and railway subbase design, tunneling, earth dam design, mine excavations design engineers are able to design civil structures with appropriate factors of safety for public use. Soil consists of a multiphase aggregation of solid particles, water and air. This fundamental composition gives rise to unique engineering properties as evaluated with both physical characteristics measurements and electrical measurements. The description of the mechanical behavior of soil requires some of the most sophisticated principles of engineering mechanics. The present invention integrates the complex science of soil

mechanics with electrical engineering to provide efficient evaluation of soil materials that are used in civil construction. MODERN ELECTRICAL SOIL ANALYSIS AS USED IN SOIL COMPACTION AND SOIL MOISTURE GEOTECHNICAL ENGINEERING Currently three emerging electrical technologies are being used in civil construction for compaction analysis. The electrical density gauge (EDG) is one of the three technologies and is the topic of this paper. The other two technologies include the time domain reflectometry for compaction control (TDR) and a coil impedance technology. All three technologies measure dielectric properties of soil material and calibrate against a standard sample of the soil material that is being evaluated. The TDR and the coil impedance technology work in the gigahertz radio frequency range where the wave length is measured in millimeters. The EDG uses a 3 megahertz radio frequency as the operating signal with a wave length of the 100 meters. Dr. Darrell Word, E.E., the University of Texas, and John Lundstrom MSc. E.E., Principal with the Electrical Density Gauge Company, along with the electrical density gauge research and development team developed the theory and electrical methodology for the analysis of soil density and soil moisture. This research was conducted mainly between 2000 and 2004 and resulted in the beta unit equipment of the electrical density gauge. The Electrical Density Gauge Technology For Soil Material Compaction And Soil Moisture Testing In Geotechnical Engineering The Electrical Density Gauge (EDG) invention provides maintenance free, and rapid measurement of the electrical dielectric properties of soil that can be related to soil wet density, and dry density. Because soils have such a wide variety of characteristics that affect the electrical dielectric properties, it is necessary to employ the sand cone test as a calibration means for the various specific types of soils. The frequency of the sand cone tests required for good accuracy and correlation is the same as that already practiced when using the nuclear density gauge. The EDG provides advantages over the nuclear gauge and other radio frequency means, because it is low cost, light and portable, battery powered, and not subject to calibration degradation over time. Once standardized against a sand cone test, the EDG will provide results that are as good as, or better than the nuclear gauge. The measurement circuit of the EDG contains a 3 MHz radio frequency source that is applied to the soil under test by spike type probe electrodes that are pushed into the soil to a prescribed depth and distance apart. The volume of soil that is measured is controlled by the depth and spacing of the measurement probes. The radio frequency current that is

Electrical properties of soil and geotechnical engineering

passing through the probes into the soil and the voltage that appears across the probes are measured electronically. Additionally the electrical phase relationship between the soil current and the probe-to-probe voltage is determined. These parameters are termed Is for soil current, Vs for probe-to-probe voltage, and Ps for their phase relationship. The electrical dielectric parameters of the soil are then calculated using well known electrical engineering equations that use current (Is), voltage (Vs), and phase (Ps) to determine the equivalent values of soil resistance (Rs) and soil capacitance (Cs). Calibration of the EDG invention for a specific soil type involves first taking a series of electrical measurements in a chosen spot of the site under test. These measurements are then stored. A geotechnical (sand cone) test is then performed on the center of the spot where the electrical measurements were taken. This assures that the soil that is used for physical measurements is the same for which electrical values were taken. When calibrating the EDG for a specific soil type, two equations are generated that relate the soil physical properties to the calculated electrical dielectric parameters. First an equation is derived that represents the relationship between soil wet density and soil electrical real impedance. Then a second equation is derived that relates the soil unit weight of water to the quotient Cs/Rs. It has been found through considerable testing of various soil types and calculations with synthetic modeled soil physical parameters that these relationships provide a good prediction of the soil physical properties using the equivalent electrical parameters calculated from measured electrical dielectric properties. In practice, linear regression equations are used to relate the measured electrical parameters to the measured physical parameters. Over a wide range of density and moisture it may be desirable to employ non-linear equations for better accuracy. After calibration for a specific soil type, the equivalent electrical values (Rs, Cs) of the soil under test are computed, then used to calculate the Real Impedance and the Cs/Rs quotient. These values are then applied to the soil specific equations generated during calibration. The resulting values of wet density and unit weight of water are then used to calculate soil dry density using equations that are well known by geotechnical engineers. Measurements of the soil electrical parameters (Is, Vs, Ps) are made with circuitry that is well known in the field of electronic engineering, and implementation of these measurement means can take a variety of forms. The frequency source used in the EDG operates at 3.0 mHz. The EDG invention is capable of making the described measurements at other radio frequencies without any change to the substance of the invention. The EDG uses a micro-processor based mini-computer module that contains memory, processing, A/D converters,

keypad entry, RS-232 I/O ports, LCD display, and other features that are useful in the computation and handling of electrical signals. This computer utilizes proprietary programming to convert the electrical signals Is, Vs, and Ps to representative digital form, then the computes actual soil electrical values measured each time the soil is tested. From the electrical soil measurements, the software then calculates Rs and Cs, the quotient Cs/Rs, and real impedance (Zs). Cs/Rs is then applied to the equation that was generated when the EDG was calibrated on soil for which the physical parameters were determined by the sand cone test or any other standard test for assessing the physical soil density and moisture. The result is the unit weight of water in the soil under test. Zs is applied to its respective equation resulting in the wet density of the soil under test. These measured physical parameters are used to determine dry density. These final values are displayed on the computer display and stored for future download as required. The soils engineers expect also to see the percent of maximum compaction that the measured dry density would provide. This additional output requires another physical test on the soil that is typically performed under ASTM D-698-00a "Standard Test Methods for Laboratory Compaction Characteristics of Soil Using Standard Effort " or ASTM D-1557-00 "Standard Test Methods for Laboratory Compaction Characteristics of Soil Using Modified Effort." The procedure specifies that a several kilogram sample be prepared at increasing moisture contents and then the various prepared samples are compacted in a standard proctor mold using a specified compactive effort. The test results in determining a value of maximum compaction and optimum moisture content for the soil that is being tested. The EDG will accept the results of this test as an input, calculate, and display the percent of maximum compaction that results from the estimated value of dry density. The Electrical Density Gauge Operating Practices For Soil Material Compaction And Soil Moisture Testing The Electrical Density Gauge consists of electrode means for electrically connecting to an in-place test spot of compacted construction material (soil). The soil electrical equivalent parallel resistance and equivalent parallel capacitance of soil is measured both in the calibration phase of the testing and in the field measurement and determination of the in-place soil density and the in place soil moisture by performing the necessary computations and display of results. The procedure requires characterization of the soil products that are being tested by first measuring a plurality of in-place test spots in a field of soil construction material to determine values of said equivalent resistance and capacitance. Then determining with the use of geotechnical means, the in-place wet density and in-place weight of water of the same plurality of test spots where electrical measurements were made, and with the use of said electronic computational means. The real impedance of the measured test samples is determined, the best fit regression equation between the physical wet density

Dennis M. Anderson, P.E., USA

data points and real impedance data points is calculated with the use of on-board electronic computational means. The ratio of measured capacitance and measured resistance is determined for all data points, then the best fit regression equation between the unit weight of water data points and capacitance / resistance ratio data points is calculated. This initial procedure established the geo-electrical signature for the soil product that is being tested. Then the Electrical Density Gauge is used to measure field test spots of the same type of construction material with previously unknown electrical characteristics and with the newly determined values of real impedance and capacitance / resistance ratio applied to the aforementioned regression equations. The on-board computer is used to compute a value of wet density and a value of unit weight of water for said unknown test spot, and using the aforementioned electronic computational means, the dry density of the constructional material at the unknown test spot is calculated from the wet density and unit eight of water. The Electrical Density Gauge calculated value of maximum dry density of the tested constructional material as determined by geotechnical means is entered into the aforementioned computational means, and used with the newly determined value of dry density to compute the percent of maximum compaction of the constructional material at each field test spot. APPLIED RESEARCH DATA Our electrical research on soil products shows that each soil material contains a unique geo-electric signature that relates the physical, chemical, mineralogical components of the soil and soil wetting agent to an electrical signature. When the geo-electric signature is researched and established for a given soil material and correlated with geotechnical soil properties, then that data may be used to determine geotechnical field characteristics of the soil material. The test data presented in this paper is contains research that has been conducted by independent researchers that are interested in non-nuclear density technologies as well as our contractors and internal research and development programs. The chart below shows a graphic comparison of the EDG as verses the nuclear density gauge and uses the nuclear density gauge as the baseline standard. This research was conducted by Dr. Hudson Jackson, P.E. under the direction of Dr. Tomas Bennert, Senior Research Engineer, Civil and Environmental Engineering/CAIT, Rutgers University, and was sponsored by the New Jersey Department Transportation. The Electrical Density Gauge Company conducted a research project during the summer of 2007 Transportation. These government agencies were solicited to participate in the volunteer that included the New Hampshire Department of Transportation, the New York Department of Transportation, the Rhode Island Department of Transportation, and the Vermont Agency of program and were provided with an

EDG and user instructions. The chart below shows data that was collected and compiled the Vermont Agency of Transportation.

CONCLUSION: The electrical density gauge has the potential to become an industry standard for investigating in-situ density on civil construction projects. The Model B of the electrical density gauge is currently being used globally on a limited basis. As the technological advances enhance the performance and operational efficiencies, the electrical density gauge will gain a wider acceptance and use. Model C will have several improvements over the 2004 Model B, making the instrument much easier to use. A new laboratory calibration technique is being developed that will allow the user to develop a soil model for the EDG in a soils lab, upload the soil model and deploy an EDG to a construction site for quality control testing. The electrical density gauge testing technology has substantial advantages over existing technology. The new electrical technology's primary advantage over existing technology is that the apparatus and method do not use a nuclear source, but still offer a high level of efficiency for QC/QA work and testing related to soil in-situ density. ACKNOWLEDGEMENT, REFERENCES, APPENDICES & END NOTES Several people, agencies and institutions contributed to the research and development of the Electrical Density Gauge. This paper submitted to IGC-Delhi 2012 is a concise form of the original paper “Electrical Properties Of Soil And Geotechnical Engineering”, submitted by the author to the Civil Engineering Department, The University of Nevada, Reno, Nevada on December 12, 2007. Readers are advised to contact the author for the acknowledgments, references, appendices, end notes and the full paper.


SHEAR STRENGTH OF SOILS SUBJECTED TO RELOADING Mohammed Aslam A. K, Research Scholar, Indian Institute of Technology Bombay, email: [email protected] Ashish Juneja, Associate Professor, Indian Institute of Technology Bombay, email: [email protected] ABSTRACT: Shear strength of soil is an important parameter that governs any foundation design. However, many a time soils subjected to loading and unloading show different shear strength values. To better understand the effects of repeated loading on soils, triaxial compression tests were performed on graded soil samples and the shear strength exhibited was assessed. Samples that were loaded to a fraction of the shear strength and allowed to drain were also subjected to triaxial compression tests. The changes in shear strength and stress strain behaviour due to past loading and drainage were studied. INTRODUCTION Shear strength of any given soil depends on the consolidation state of the soil and the confining stresses acting on it. Classical design of foundations assumes a constant strength of soils. However, it has been observed that soils subjected to loading and unloading shows different shear strength. In this context, acquiring a better understanding of this phenomenon gains importance as any increase in shear strength of a soil subjected to previous loading would translate to an increase in its load bearing capacity. BACKGROUND Bishop and Henkel [1] noticed that after shearing of clay samples, a fraction of the pore pressure remained in the soil even after unloading. This residual pore pressure is proportional to the applied stress and the stress history of the soil. A soil sample subjected to repeated or cyclic loading was found to exhibit a change in residual pore water pressure with each cycle. If the sample is allowed to drain after unloading it would result in a decrease in the void ratio of the soil. In the past, researchers have reported changes in properties of soil subjected to cyclic loading followed by drainage in terms of reduction of pore pressures in subsequent cycles [e.g. 2, 3], reduction in cyclic strains [e.g. 2, 4] and increase in cyclic strength of soil [e.g. 5]. Studies on increase in undrained shear strength due to dissipation of residual pore pressures are limited. The change in the strength of soil under these circumstances can be explained using the Modified Cam-Clay model proposed by Roscoe and Burland [6]. The parameters of the model are illustrated in Fig. 1. When a normally consolidated soil is subjected to undrained compression followed by unloading, it does not attain its original state A (Fig. 2). Due to some excess pore pressure, the soil attains state B. Any drainage would ultimately cause the soil to attain state C. C corresponds to an increased size of Modified Cam-Clay yield surface and hence an increase in the shear strength.

If the soil has a mean effective stress of p1’, a specific volume of ν1 prior to the commencement of loading (state A) and a excess pore pressure of ur after removal of load, the shear strength at initial state A, q, is given by

)''(' 11,1 ppMMpq f Δ−== (1)

where, M is the slope of the critical state line in stress space and p1’is also the initial size of the Modified Cam-Clay yield surface. Mean effective stress at state B, p2’is given by

r'-u'=pp 12 (2)

At state C after the dissipation of excess pore pressure, the soil has an effective stress of p1’. The specific volume at this point, ν2 is given by

ννν Δ-= 12 (3)

The change in specific volume, ∆ν, can be related to the volume change obtained from experimental results or from the relationship between pore pressure dissipation and volume change [7]. When the section BC is extended to meet the normal compression line at point D, the mean effective stress at D is given by

⎟⎠⎞

⎜⎝⎛

−−

= κλνν

e'pκλ

3 (4)

where, vλ is the specific volume of the isotropic normal compression line (NCL) at p' = 1kN/m2, λ is the slope of the normal compression line in the v-lnp' space, vκ is the specific volume of the swelling line for p' = 1 kN/m2 and κ is the slope of the recompression line in the v-lnp' space. p3’ becomes the size of the new Modified Cam-Clay yield surface.

'ln 21 p= κννκ + (5)

or

'ln 12 p= κννκ + (6)

The shear strength at state C, qnew, is given by )''(' 33,3 ppMMpq fnew Δ−== (7)

Mohammed Aslam A.K. & Juneja, Ashish

Now,

'''' 1133 pppp Δ−>Δ− (8)

Hence,

qqnew > (9)

In this concept the shear strength of the normally consolidated soil improves by the dissipation of the excess pore pressure.

Fig. 1 Critical state parameters

Fig. 2 Different states of soil in compression space during

loading, unloading and draining EXPERIMENTAL PROGRAM All tests were done on soil obtained from Jhansi region. This soil was chosen because it represents a wide variation of particle size. The soil has a high fines content (greater than 40%). The particle size distribution is represented by Fig. 3. When subjected to the standard Proctor test, the soil exhibited a maximum dry unit weight of 21.5 kN/m3 at an optimum moisture content of 10%.

All triaxial samples were of 100 mm diameter and 200 mm height. Test specimens were prepared in 10 layers by dry tamping with a mass of 175 g falling over a height of 144 mm with a 50 mm circular tamper. The average unit weight achieved by this method was between 15.0 and 16.2 kN/m3. The samples were saturated by back pressure saturation after carbon dioxide flushing. Two test series were conducted. In the first series A the samples were sheared using conventional consolidated undrained triaxial tests. In the test series B, after consolidation, the sample was loaded to a fraction of the ultimate load and then unloaded. The pore pressure remaining in the soil was then allowed to dissipate to the past effective stress. This was followed by undrained loading to ultimate load. The aim of conducting these tests were to study the change in shear strength behaviour after load-unload and drainage phase.

Fig. 3 Particle size distribution of Jhansi soil The details of the test performed in series A are recorded in Table 1 and those in series B are recorded in Table 2. Table 1 Test details of series A

Test Cell pressure σ3

Back pressure ub

kPa kPa A1 250 150 A2 350 250 A3 150 50 A4 200 100 A5 250 100

Table 2 Test details of series B

Test Cell pressure σ3

Back pressure ub

First Strain

kPa kPa % B1 200 100 0.45 B2 200 100 0.07

Shear strength of soils subjected to reloading

RESULTS AND DISCUSSIONS The strength of normally consolidated soil was obtained from the test series A. This step was important because critical state parameters of the sample were obtained from these tests. An excess pore pressure of 42 kPa was generated under a strain of 0.45% in sample B1. When unloaded, the excess pore pressure was equal to 32 kPa. This sample, when allowed to drain to expel about 2.8 ml of water, which corresponded to a decrease of 0.003 in the void ratio of the sample. The difference in stress-strain behaviour over large strain values is shown in Fig. 4 for tests A1, A2 and B1. The stress path for test B1 is given in Fig. 5. Fig. 6 gives the excess pore pressures developed in the series of tests.

0 10 20 30Axial Strain, axial (%)

0

20

40

60

80

Dev

iato

ric S

tress

, q (k

N/m

2 )

A1A2B1

Fig. 4 Stress-strain behaviour in tests A1, A2, B1

0 20 40 60 80 100Mean effective stress, p' (kPa)

0

20

40

60

80

100

Dev

iato

ric st

ress

, q (k

N/m

2 ) A1A2B1CSL

Fig. 5 Stress path in test A1, A2, B1

0 10 20 30 40Axial Strain, axial (%)

-0.4

0

0.4

0.8

1.2

Exce

ss p

ore

pres

sure

ratio

, u/

3'

A1A2A3

A4

A5

B1

B2

Fig. 6 Excess pore pressure ratio in tests A1-A5, B1, B2 A relationship between volume of water drained and pore pressure released was also obtained from these tests (Fig. 7). The compression space plots during the loading-unloading-draining-reloading phase of test B1 is given in Fig. 8. The major states during the test are listed in Table 3. Manual calculations for the determination of shear strength based on critical state soil mechanics were also done for test B1. The critical state parameters used for the calculations are given in Table 4. Size of the Modified Cam-Clay yield surface for the initial state in test B1 was 94 kPa and the predicted mean effective stress at failure, pf’ is 36 kPa. This corresponds to a deviatoric stress at failure, qf, of 52 kPa. When the sample was subjected to loading-unloading and draining, the predicted size of the new of the Modified Cam-Clay yield surface was 95.6 kPa and the predicted mean effective stress at failure, p’f was 37 kPa. This corresponds to a deviatoric stress at failure, qf, of 54 kPa which corresponds to a 4% increase in the shear strength.

y = 3.3592x ‐ 195.58

70

80

90

100

110

120

130

140

150

80 85 90 95 100 105

Vol

ume

of W

ater

dra

ined

, VW

(cc)

Pore pressure released, Δu (kPa) Fig. 7 Relationship between volume of water drained and

pore pressure released

Mohammed Aslam A.K. & Juneja, Ashish

20 30 40 50 60 70 80 90 1001.432

1.434

1.436

Spe

cific

Vol

ume,

v

Mean Effective Stress, p' (kPa)

AB C

DE F

Fig. 8 Compression space plot in test B1 during loading-

unloading-draining-reloading phases Table 3 Points on the compression space during loading-

unloading-draining-reloading phases for test B1 Points States

A End of isotropic consolidation, Start of first loading

B End of first loading, Start of unloading C End of unloading, Start of draining D End of draining, Start of second loading E Minimum effective stress during second

loading F End of test

Table 4 Critical state parameters for Jhansi soil

Parameter Value λ 0.15 M 1.46 Γ 1.973 νλ 2.201 νκ 1.5029 κ (assumed) 0.015

If a soil is subjected to load-unload-drain cycles, there can be a further increase in the strength of the soil. In addition, if the loading is cyclic in nature, the excess pore pressures increase with number of cycles [e.g. 5, 8] and this would correspond to a larger decrease in the void ratio during draining and hence result in higher shear strength.

CONCLUSIONS It was observed that there was excess pore pressure in the soil even after the removal of load. In addition, loading-unloading resulted in a change in the stress path. The dissipation of this excess pore pressure effected a marginal improvement in the shear strength of the soil. It therefore can be concluded that when a soil is exposed to similar cyclic loading with multiple loading and drainage periods, an appreciable increase in strength can be expected. However the gain in strength may peak after a limited cycles followed by drainage phase. This needs to be examined. ACKNOWLEDGEMENT The first author would like to acknowledge the financial support from the Indian Institute of Technology Bombay (IITB) in the form of research scholarships. REFERENCES 1. Bishop, A. W. and Henkel, D. J. (1953), Pore pressure

changes during shear in two undisturbed clays, Proceedings of the 3rd International Conference on Soil Mechanics and Foundation Engineering, 1, Zurich, Switzerland, 94–99.

2. Yasuhara, K. and Andersen, K. H. (1991), Recompression of Normally Consolidated Clay after Cyclic Loading, Soils and Foundations, Japanese Society for Soil Mechanics and Foundation Engineering, 31(1), 83-94 .

3. Yıldırım, H. and Ersan, H. (2007), Settlements under consecutive series of cyclic loading, Soil Dynamics and Earthquake Engineering, 27, 577–585.

4. O'Reilly, M. P., Brown, S. F., and Overy, R. F. (1991), Cyclic loading of silty clay with drainage periods, Journal of Geotechnical Engineering, ASCE, 117(2), 354-362.

5. Hyde, A. F. L., Higuchi, T. and Yasuhara, K. (2007), Postcyclic Recompression, Stiffness, and Consolidated Cyclic Strength of Silt, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 133(4), 416-423.

6. Roscoe, K .H. and Burland, J.B. (1968), On the generalised stress-strain behaviour of 'wet' clay, in Engineering plasticity (eds. J. Heyman and F.A. Leckie) Cambridge University Press, 535-609.

7. Juneja, A., Raghunandan M.E., Chatterjee, D. (2011), Post-cyclic Shear Strength of Sands, Proceedings of Indian Geotechnical Conference, December 15-17, 2011, Kochi.

8. Yang, J. and Sze, H. Y. (2011), Cyclic behaviour and resistance of saturated sand under non-symmetrical loading conditions, Geotechnique 61(1), 59–73.

Proceedings of Indian Geotechnical Conference December 13-15, 2012, Delhi (Paper No A 161.)

DEVELOPMENT OF HANDHELD STEP FREQUENCY GPR USING FIELDFOX NETWORK ANALYZER AND ITS APPLICATIONS

F.N. Kong (Consultant, NGI Oslo, Norway, [email protected]) Rajinder Bhasin (Regional Manager , NGI Oslo, Norway, [email protected]) Neeraj Chadha (Complete Instrumentation Solutions Pvt Ltd, [email protected])

Abstract— NGI’s classical handheld Step Frequency GPR has recently been developed for carrying out geotechnical investigations in the field. The investigations can be performed in a simple and quick manner. The GPR is made light in weight using the FieldFox handheld network analyser which Agilent introduced in 2008. The FieldFox has many attractive features for GPR engineers, such as wide bandwidth, low power consumption, light weight & compact. To investigate the performance of using FieldFox as a GPR system, we have made following tests: (1) data acquisition speed tests (2) laboratory model tests (3) field tests. Test results show that a GPR based on Fieldfox is able to make high-resolution detections at different frequency bands.The highest frequency of Fieldfox is 6 GHz. This is a distinct feature comparing with other commercially available GPR systems. The Fieldfox based GPR can be very much suited for non-destructive testing of roads, bridges, dams, river dykes, and testing inside buildings, tunnels etc. Key wo rds-step frequency GPR, vector network analysor, handheld GPR, non-destructive testing

INTRODUCTION Vector Network Analyzers (VNA) can make Step-Frequency measurements with high precision, high dynamic range & wide bandwidth. Hence they are sometimes used as GPR transmitters and receivers (Kong & By, 1995 and Sato et. al. 2008). However, a VNA for laboratory use e.g., agilent E5062a is high weight (13.5 kg), large size (43 x 22 x 31 cm) and large power consumption (350 VA), which needs a generator for electrical power. Such a vna based ground penetrating radar can only be used where a car can reach. The E5062a VNA provides Vector Network Analysis at frequency band 300 khz to 3 GHz with easy-to-use feature and low cost, and may be used for building GPR systems. We therefore use E5062a for performance comparison in our discussions. For GPR engineers, following features of Field Fox are attractive: • It can make vector (phase and magnitude)

measurements over a wide bandwidth from 2MHz to 6GHz.

• It consumes only 14W electrical power. Therefore the battery operation mode can be easily implemented whch lasts for 3.5 hrs.

• Light weight (only 2.7 kg including battery) with a compact size: 188x 72mm.

• Has a LAN port for fast data transfer and SCPI programming to communicate to an acquisition computer.

• Commercial product.

To evaluate the possibility of using FieldFox as a GPR system, the following investigations were carried out - Measurement speed - Dynamic range and noise level - Ability to detect targets in field measurements at different frequency ranges. It is reported that a VNA has also been developed by Tohoku University, Japan which is compact, light weight and with small power consumption.

Measurement Time and System Dynamic Range It is well known that the noise power at the output of a receiver is proportional to the receiver bandwidth. The advantage of using a step frequency radar system is that the receiver bandwidth, specified as IFBW (Intermediate Frequency Bandwidth), can be made rather small. For instance, one can choose IFBW 300, 1000, 3000 Hz for Fieldfox. It is also well known that the measurement time (or sweep time) is inversely proportional to IFBW. This is because the transient time for the receiver from measuring one frequency to another frequency is inversely proportional to IFBW – a principle of a linear system. Thus increasing measurement time means allowing to use narrower IFBW to receive signals, resulting in smaller receiver noise power. As we will discuss later, Fieldfox takes longer time to acquire data than e.g., E5062a. We will use the

FN.Kong,Rajinder Bhasin &Neeraj Chadha

above discussion to quantify the increased acquisition time as to the increasing of the noise power or the decreasing of the system dynamic range. The total data acquisition time is the sum of the sweep time and the data transferring time. We have made a data acquisition software in C# through LAN communication, and tested the data acquisition time (‘time’ in the table) and the sweep time for IFBW: 3000, 1000, 300 Hz and frequency numbers (points) 51 and 101. The results are listed in Table 1. IFBW (Hz)

3000 3000 1000 1000 300 300

Points 51 101 51 101 51 101 Time (s)

0.74 0.86 0.87 1.14 1.2 1.85

Sweep time

0.17 0.27 0.33 0.58 0.66 1.23

Table 1 Data acquisition time and sweep time of using Fieldfox The data transferring time is the difference between the data acquisition time and the sweep time. Table 1 shows the data transferring time is about 0.55 s, which doesn’t vary much with IFBW and frequency points, and is somewhat longer than the data transferring time (0.2 s for 201 points) through GPIB communication used at E5062a. When using E5062a at IFBW 300 Hz and frequency points 101, it takes about 0.6 s total acquisition time. So the acquisition time of Fieldfox is about 3 times (5 dB) of the E5062a acquisition time, or equivalently, Fieldfox receiver noise level is about 5 dB higher than E5062a. Consider that the maximum output power of E5062a is 10 dBmW, while it is 5 dBmW for Fieldfox. We can therefore say that the dynamic range of the Fieldfox is about 10 dB lower than E5062a. Noting that the dynamic range of E5062a is 115 dB, the dynamic range of Fieldfox is therefore 95 dB, at IFBW 300 Hz. Here we briefly discuss what a dynamic range of 95 dB means for target detection. Assume that the directly coupled signal from the transmitter antenna to the receiver is -40 dB of the transmitted signal, and that the weakest signal one can detect is -40 dB of the directly coupled signal, i.e., -80 dB below the transmitted signal. In that case, the system dynamic range required is 80 dB. In many applications, the GPR detection is limited by the ability of subtracting the weak reflections from the background of strong directly coupled signals. A dynamic range of 95 dB can be quite enough for

those applications. We use following real tests to support this statement. Laboratory Model Tests Three laboratory models are made to test the wide bandwidth performance of a GPR based on FieldFox. Figure 1 shows a photo of the sand model (Model 1) with a saw tooth shaped target, together with Fieldfox and the antennas used for the model tests. Both the transmitter dipole and the receiver dipole are built inside the tiny metal box. Hence antennas are shielded. The frequency band 1 GHz to 6 GHz is chosen for the model tests. For all the measurements in this article, we use 101 frequency points and IFBW 3000 Hz.

Fig. 1: Test Model 1, Fieldfox NA, and antenna box

Figure 2 shows the photo of Model 2: a piece of rock with three holes and Model 3: round shaped (anticline) target in a sand box.

Fig. 2: Model 2 (Top): a granite block (30cm length, 12cm

depth, 12cm height) with three holes. Model 3 (Bottom): a sand model with a round target


Figure 3-5 show the GPR test results for Model 1-3. The results shown are not much processed, only 60% of the ‘parallel background’ is subtracted. The targets: the saw-tooth shaped layer in Model 1 and the round shaped layer in Model 3 can be clearly seen. The results for Model 2 show clearly three hyperbolic curves at different depths corresponding to the reflection of three holes. One can also see the corner reflections in all three measurements. The detection resolution is really good, owing to the wide bandwidth of Fieldfox and antennas used for test.

Fig. 3: GPR results of Model 1



Indoor Re-Bar Test These tests are performed inside a garage. Figure 6 shows the test layout for testing the re-bars inside a concrete wall. Figure 7 shows a very clear rebar image. The frequency band used for this test is from 700 MHz to 3000 MHz.

Fig. 6: Fieldfox, acquisition computer and antenna box for testing re-bars in a concrete wall

Fig. 7 Test result of the concrete wall

Fig. 8 The antenna with 1.2m in length is used in Turag river dike monitoring in Bangladesh

FN.Kong,Rajinder Bhasin &Neeraj Chadha

Fig. 9 Test result from the dike showing weak spots

Fig. 10 Handheld GPR system in use at CRRI, New Delhi

Fig. 11 Result of the dike-side test at another location

River Dike And Pavement Thickness Tests The newly developed hand held GPR has been used used in several countries including China, India, Bangladesh and Vietnam. In China, NGI’s hand-held GPR was successfully used to detect the structure of Qian Tang River Dike in Hang Zhou. The purpose of the investigations was to identify the different stone layers used in the construction of the dike and to detect the interface between stone and soil (Kong et al, 2012). In Bangladesh

and Vietnam GPR investigations (Fig 8-9) were successfully performed on existing river dikes. The purpose was to locate weak spots in dikes so that these could be strengthened before the onset of the next monsoon season. Figure 8 shows the system in use in Dhaka Bangladesh using 1.2 m long antennas. Figure 9 shows the test results for the red river dike in Vietnam where weak spots were detected in the sub-surface. It was subsequently observed that there were some sink holes on the sides of the dyke beneath the detected weak spots. In India the hand-held GPR was successfully used to check the thickness of the road pavement in the Central Road Research Institute (CRRI) in New Delhi. Figure 10 shows the hand-held GPR system in use at CRRI. The results from the GPR tests at CRRI are shown in Figure 11 where the bitumen surfacing thickness is about 40mm followed by Bituminous course of around 120mm. The reflections in the radargram in Figure 11 shows the different pavement layers and the depth.

CONCLUSION The GPR made from FieldFox is a step frequency radar system. Its frequency band can be adjusted by software according to the application needs, within a very wide frequency band: 2MHz to 6 GHz. Fieldfox itself is a good instrument for measuring the characteristics of antennas and other hardware of the GPR system. Low power consumption, lightweight and compact size of the system can make the mobilization and field survey very convenient. The highest frequency of Fieldfox is 6 GHz. This is a distinct feature comparing with commercially available GPR systems. The Fieldfox based GPR can be very useful when detections with very high resolution, e.g., non-destructive testing of roads and bridges, testing inside buildings, tunnels and platforms etc., are needed. Tests for low frequency bands using long antennas are also made, and show that a GPR based on Fieldfox can very well be used for low frequency applications. REFERENCES

1. Kong, F.N., and T.L. By, 1995, Journal of Applied Geophysics, Vol. 33, 1995, pp 15-26.

2. Kong, F.N., Bhasin, R., Wang, L.Z. and Zhang, Y.G. (2012) Proc. 14th International Conf. on Ground Penetrating Radar, June 4-8 2012, Shanghia, China pp.151-156

Detected weak spots


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