Wastewater - Primary

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Unit CIV3264: Urban Water and Wastewater SystemsTopic 3 Wastewater

Department of Civil Engineering, Monash University

04-2012

MONASHU N I V E R S I T Y

1

TOPIC 3: WASTEWATER SYSTEMS

TABLE OF CONTENTS

PREVIEW ......................................................................................................................... 2Introduction ................................................................................................................ 2Objectives ................................................................................................................... 2Readings ..................................................................................................................... 2

3.1 COMPONENTS OF WASTEWATER SYSTEMS .................................................... 33.2 CHARACTERISTICS OF WASTEWATER .............................................................. 3

3.2.1 Quantity - Wastewater Flow Determination...................................................... 43.2.2 Water Quality .................................................................................................... 83.2.3 Variations in Wastewater Quantity and Quality ................................................ 9

3.3 OUTLINE OF A WASTEWATER TREATMENT PLANT .................................... 113.4 PRELIMINARY TREATMENT ............................................................................... 13

3.4.1 Screens ............................................................................................................. 133.4.2 Grit Chambers ................................................................................................. 223.4.3 Other ................................................................................................................ 35

3.5 PRIMARY TREATMENT ........................................................................................ 353.5.1 Primary Sedimentation Tanks ......................................................................... 35


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PREVIEW

Introduction

This topic introduces the main principles of wastewater systems focusing particularly on

technologies of waste water treatment systems.

Objectives

After completing this topic you will be able to:

understand functioning of wastewater systems, particularly wastewater

treatment plants;

conceptually outline several types of wastewater treatment plants:

conventional, modern and alternative;

design a conventional wastewater treatment plant; and

understand interaction between wastewater systems and other UWM systems

(e.g. water supply systems and stormwater management systems).

If you are not convinced that you can achieve each objective after studying the

materials, you should re-read the relevant parts of the Study Guide and associated

reading.

Readings

This summarises the suggested readings for the topic.

SUGGESTED

Tchobanoglous,Wastewater Engineering:

Treatment and Reuse, Metcalf & Eddy, 2002

Environmental Engineering, Gerard Kiely,McGrow Hill, 1997

Introduction to Environmental Engineering,

Mackenzie Davis & David Cornwell, McGrow

Hill, 1991

http://www.pub.gov.sg/research/Pages/default.

aspx?


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3.1 Components of Wastewater Systems

Municipal wastewater is collected from our homes and industrial plants via a sewage

collection system. In this context, it is defined as everything from the point of a

wastewater discharge in urban areas to the point of its disposal into a receiving water

body (e.g. the wastewater system of Melbourne is presented in Figure 3.1 to the left).

The system comprises of the following:

collection in households municipal sewage pipes and pumps, wastewater treatment plant, disposal of treated water into receiving water bodies, disposal of sewage sludge.

Figure 3.1: Diagram of wastewater system and separate stormwater system in

Melbourne.

In Australia, stormwater and municipal wastewater are collected in separate systems(Fig. 3.1), while in Europe combined systems are more common (stormwater and

wastewater are collected in one pipe).

3.2 Characteristics of Wastewater

Sewerage systems are designed to collect, transport, treat, and discharge sewage. These

requirements involve hydraulicprocesses and treatmentprocesses. Therefore, before

we start designing the systems, we have to determine the sewage quantity (flow rate)

and its water quality.


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In most cases, the hydraulics is reasonably basic. Nevertheless, a sound understanding

of the hydraulic processes is necessary to ensure efficient, economical, and safe

operation.

3.2.1 Quantity - Wastewater Flow DeterminationThe main components of municipal wastewater are:

Domestic wastewater - discharged from residences, commercial, institutional,and other facilities.

Industrial wastewaterdischarges from industrial plants. Inflow/infiltration - stormwater runoff which finds its way into the sewer, and

steady leakage from groundwater.

Average Dry Weather Flow (ADWF) in litres/day, that presents flow in pipes without

any infiltration, is calculated as:

ADWF=Domestic + Industri al [l /day] (3.1)

Domestic Wastewater

Principal sources are residential and commercial districts. Other sources include human

waste from institutional and recreational facilities.

Domestic discharge is usually expressed as:

Domestic = PE (F low discharged per capita) [l/day] (3.2)

where PE is Population Equivalent.

Currently, it is assumed that

Flow discharged per capita = 225 l/day/capita (3.3)

General PE values for different establishments are presented in Table 3.1

I ndustrial Wastewater

Industrial use varies widely, according to the nature of the manufacturing process. There

is no foolproof procedure for predicting industrial discharges. Consequently, careful

measurements are required.

Flow rates vary depending on:

type of industry,

size of industry,

degree of water re-use, and

on-site wastewater treatment methods.


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Table 3.1: Recommended PE values for different establishments.

Wastewater flow projections may be based on water use. However, it is usually

estimated according to some of the following assumptions:

the typical design value for industrial districts with no wet process industries is

30 m3/ha/day;

the same for light industry is 20 m

3

/ha/day;for industries without internal re-use programmes 85-90% of water used will

become wastewater; for large industries with internal water re-use programmes,

separate estimates must be made;

average domestic wastewater contributed from industrial activities may vary

from 30 to 95 l/capita/day;

special attention needs to be given to projections of future industrial flows; and

industrial flows are particularly troublesome in small sewage treatment plants

where there is limited capacity to absorb shock loadings.


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Infiltration

Sewers are not watertight. They may be damaged by ground movement, traffic & road

construction, tree roots, age, or illegal connections. Therefore, due to infiltration during

wet weather, flow rates in sewage pipes are much higher than the Average Dry Weather

Flows. As a consequence of the above, infiltration may be present even during dry

weather; it occurs mainly through seepage into sewers laid below the water table.

It is impossible to entirely avoid faulty joints, cracked sewer pipes, and damaged

manhole connections. However, infiltration is greatly reduced by use of proper

materials, construction techniques, supervision, and field testing.

Inflow/infiltration has a number of components as follows:

Surface infiltration rain water entering a sewer system from the ground

through defective pipes, pipe joints, connections, or manhole walls;

Steady inflow - water discharged from cellar and foundation drains, cooling-water discharges, and drains from springs and swampy areas. It is a steady flow

and is identified and measured along with infiltration;

Direct inflow - inflows that have a direct stormwater runoff connection to thesewer and cause an almost immediate increase in wastewater flows. Possible

sources are: roof leaders, yard drains, manhole covers or cross-connections from

storm drains.

Figure 3.2 presents different types of flow in sewage pipes. A part of the terms

described above flowrate also shows the following:

Total inflow - sum of the direct inflow at any point in the system plus any flowdischarged from the system upstream through overflows, pumping station

bypasses, etc.

Delayed inflow - Stormwater runoff which my require several days or more todrain through the sewer system. This can include the discharge of sump pumps

from cellar drainage as well as the slowed entry of surface water through

manholes in ponded areas.

Infiltration may vary during the year in response to groundwater levels. Normally, it is

estimated during the early morning hours when water use is at a minimum and the flow

consists essentially only of infiltration. This is particularly suitable for small systems

where travel times through the system are short, however not so suitable for largesystems where travel times may be long.

Inflow rates are determined by using a network of continuous flow meters operating

before and during a significant storm event. It is determined from the flow hydrograph

by subtracting the normal dry-weather domestic and industrial flow and the infiltration

from the measured flow rate.

For design purposes, the maximum infiltration rate for well constructed sewers is

I nfi ltration = 50 l/mm/km/day D L [l /day] (3.4)


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whereD is the pipe diameter, andL is the pipe length. Although design values are given

in the standards, there are many factors which can affect infiltration: length of sewers,

area served, soil and topographic conditions, population density (which affects the

number and total length of house connections).

Sewers first built in a district usually followed water courses in the bottom of valleys.

They are often close to (or even below) the bed of the stream. As a result, these old

sewers can receive large infiltration flows. Newer sewers are often built at higher

elevations and receive comparatively less infiltration

It is usually found that:

only a small part of the collection system contributes most of theinflow/infiltration;

typically about 75% of the inflow comes from 20 - 30% of the system; andtypically about 75% of the infiltration comes from 40% of the area.

There is an attempt to produce watertight sewer systems. They are sometimes known as

Smart sewers. The benefits of a leak-free or tight system include:

no overloaded or surcharged sewers and the associated problems of wastewater

backups and overflows,

more efficient operation of wastewater treatment facilities, and

use of sewer hydraulic capacity for wastewater instead of inflow/infiltration.

Figure 3.2: Components of wastewater flow


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3.2.2 Water Quality

NOTE: Water quality depends on the source!

Domestic Wastewater

Typical data on the individual constituents found in domestic water are reported in

Table 3.2. Main water quality characteristics of domestic waters are: BOD5

(biochemical oxygen demand) and SS (suspended solids). Some other constituents

that are important for biological treatment should be also determined (Ca, Cu, Fe, Mg,

Zn, Mn, Sulphates). If the inflow from industrial wastewater is high the characteristics

of municipal waters (collected by the drainage system) could be different from those

presented in Table 3.2.

Table 3.2: Main water quality characteristics of domestic wastewater

Concentration

Constituent Strong Medium Weak

Solids, total :

Dissolved, total

Fixed

Volatile

Suspended, total

Fixed

Volatile

Settleable solids, mL/LBiochemical oxygen demand, 5-day, 20C

(BOD5, 20C)

Total organic carbon (TOC)

Chemical oxygen demand (COD)

Nitrogen (total as N) :

Organic

Free ammonia

Nitrites

Nitrates

Phosphorus (total as P) :Organic

Inorganic

Chloridesb

Alkalinity (as CaCO3)b

1200

850

525

325

350

75

275

20400

290

1000

85

35

50

0

0

15

510

100

200

150

720

500

300

200

220

55

165

10220

160

500

40

15

25

0

0

8

35

50

100

100

350

250

145

105

100

20

80

5110

80

250

20

8

12

0

0

4

13

30

50

50a

mg/L = g/m3

bValues should be increased by amount in domestic water supply

I ndustrial Wastewater

Numerous materials can be found in industrial waste that can cause pollution. The most

common are: anorganic salts, acids or alkalis, organic matters, suspended solids,


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floating solids and liquids, colour, heated water, toxic chemicals, radioactive materials,

foam-producing matter, and microorganisms.

Quality of industrial wastewatervaries widely with the type of industry and it usually

includes:process waste from manufacturing,

water from heating and cooling,

wash water, and

employees sanitary waste products.

Typical quantities for process wastewater from four different manufacturing industries

are presented in Table 3.3. The wastewater is more specific for each industry and can

range from wastewaters with high BOD5 (meat, food industry) to inorganic and toxic

waste (textile, chemical and plating industry).

Table 3.3: Industrial process wastewater quality

Parameter Food Meat Plating Textile

Volume (L)/capita/day - - - -/tonne prod. 10,000 12,000 - 100,000% runoff - - - -

MPN (106

/100mL) 0 - 0 0

BOD5

1,200 640 0 400

COD - - - -TOC - - - -Susp Solids 700 300 0 100Diss Solids - 200 - 1,900Total N 0 3 0 0Total P 0 - 0 0pH - 7.0 4 or 10 10Copper 0.29 0.09 6 0.31Cadmium 0.006 0.011 1 0.03Chromium 0.15 0.15 11 0.82Nickel 0.11 0.07 12 0.25Lead - - - -Zinc 1.08 0.43 9 0.47

Food: canning factory (pickles, beets tomatoes, pears);Meat: meat processing (poultry plant with no manure or blood recovery);

Plating: wastes are acidic with chromate baths, alkaline with cyanide baths;

Textile: textile mill (spun cotton yarn processed into cotton goods)

3.2.3 Variations in Wastewater Quantity and Quality

The principle factors for loading variations are the following:

(i) established daily people habits (short-term variations);

(ii) seasonal activities (longer-term variations); and

(iii) industrial activities (both short and long term, plus shock loads).


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Daily variations due to established people habits are presented in Figure 3.3. A double-

peaked diurnal pattern is common, depending on the mix of commercial, industrial, and

domestic connections. Diurnal patterns are regular because of:

the same daily pattern for Monday to Friday flows;the weekend patterns differ from weekdays; and

Saturdays and Sundays may be different from each other.

Figure 3.3: Daily diurnal patterns

Both quantity and quality

of industrial wastewater

may vary significantly

throughout the day.

Problems with short-term

loading from one

particular part of the

manufacturing processcan cause problems to

treatment plants. Seasonal

variations are typical for

food industry, as shown

in Figure 3.4.

When industrial wastes

are to be accommodated in municipal collection and treatment facilities, special

attention should be given to developing adequate characterisations and projections.

Figure 3.4: Seasonal variations of flow in a small town

with a large peaches cannery plant.


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Peak Flows

Peak hourly flow must be used in design of sewers, pumping stations, and treatment

plant components. It is determined as:

Peak Flow = PFF ADWF (3.5)

where ADWF is the Averaged Dry Weather Flow (domestic + industry) and PFF is the

Peak Flow Factor. The PFF depends on the size of the development and is generally

calculated as:

1000

7.4 11.0

PEPwhere

PPFF

(3.6)

where PE is the Population Equivalent (Table 3.1) and P is the population equivalent inunits of 1,000. Thus, the PFF for different sizes of a development is:

PE = 1,000 PFF = 4.7

PE = 10,000 PFF = 3.6

PE = 300 PFF = 5.4

If commercial, institutional, and/or industrial wastewaters make up > 25% of all flows

(excluding infiltration) consideration should be given to estimating PFFs for each flow

category.

For industrial wastewater the PFF is estimated on the basis of:

average water use;

number of shifts worked; and

pertinent details of plant operation.

3.3 Outline of a Wastewater Treatment Plant

The products of a wastewater treatment plant are:

an effluent of accepted quality in relation to a receiving water course, and

a sewage sludge that contains all pollutants.

There are the following major categories for municipal wastewater treatment:

(1)Preliminary (or pre-treatment);(2)Primary treatment;(3)Secondary treatment;(4)Tertiary (or advanced treatment), and(5)Sludge treatment.

Efficiencies of the main categories are listed below:


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Primary treatment: Removal of 60% of TSS and 35% of BOD5. It removes the

pollutants that either settle or float by screening and sedimentation methods.

Secondary treatment: Removal of 85% of BOD5 and more than 85% of TSS. Itremoves BOD5 and SS by biological processes.

Tertiary treatment: 99% of BOD5 and phosphorus, all TSS, and bacteria, 95%of nitrogen are now removed. The pollutants are removed by chemical treatment

and filtration, or by land infiltration treatment.

Traditional treatment plants have no tertiary treatment as shown in the case of the

Melbourne Eastern Treatment Plant in Figure 3.5. They treat wastewater usually to

TSS=30 mg/l and BOD5 = 20 mg/l. However, modern treatment plants have a tertiary

treatment for further removal of pollutants as shown in Figure 3.6.

Figure 3.5: Diagram of a typical secondary treatment plant (Melbourne).


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Figure 3.6: Diagram of a tertiary treatment plant.

3.4 Preliminary treatment

Preliminary or pre-treatment is aimed to provide protection to the wastewater treatment

plant, but it has little effect on the reduction of BOD5 and nutrients. The main devices

used in the preliminary treatment are discussed below. However, be aware that some of

these devises could also be used at other stages of the treatment process chain.

The most common devices used in preliminary treatment are screens and grit chambers.

3.4.1 Screens

Screening of sewage is one of the oldest treatment processes. Although used mostly in

the preliminary treatment they are also used in other stages of treatment. Therefore,

screens are classified as:

primary screens,

secondary screens, and

microstrainers.

In this section, primary and secondary screens are discussed, while microstainers are

discussed later (they are used in the secondary treatment stage).


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Primary Screens

Primary screens (Fig. 3.7) are typically located

at the inlet to sewage treatment plants and also

at the inlet to pumping stations. They are

designed to remove coarse debris such as rags,

solids, and sticks which could cause damage by

damaging pump impellers or interfering with the

downstream performance in sewage treatment

plants. They have to be cleaned either manually

or mechanically.

Primary screens are normally classified as:

Coarsewith openings of 50-150 mm;

Mediumwith openings 20-50 mm.

The following factors need to be taken into account in screen design:

the strength of the screen material and its resistance to corrosion,

the clear screen area (this is related to cleaning),

the maximum flow velocity through the screen to prevent dislodging ofscreenings,

the minimum velocity in the approach channel to prevent sedimentation of

suspended matter, and

the head loss through the screen.

Secondary Screens

Secondary screens have smaller openings than primary screens and are installed after

the pumping section and ahead of the grit chamber. Their purpose is to remove material

such as paper, plastic, cloth, and other particles which may affect the treatment process

downstream; and to minimise blockages in sludge handling and treatment facilities.

Secondary screens are analysed and designed in the same way as primary screens. The

only difference is in the maximum clear spacing of bars. This is typically around 12

mm, although openings as small as 6 mm have been used in practice.

Hydraul ics of Screens

The analysis of a screen involves the determination of the head loss across it. The head

loss is primarily a function of the flow velocity and the screen openings, but may also

be dependent on bar size, bar spacing, and the angle of the screen from the vertical.

Several equations have been developed, but only those most widely used are considered

here.

Figure 3.7: A primary screen.


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Figure 3.8: Schematic of Sloping Bar Screen

Figure 3.8 shows a schematic of a sloping bar screen. Application of Bernoullis

equation yields:

lossesg

vhg

vh 22

2

2

2

2

1

1 (3.7)

where

h1 is the upstream depth of flow

h2 is the downstream depth of flow

g is the acceleration due to gravity

v1 is the upstream velocity

v2 is the downstream velocity

For a clean or partially blocked screen, the losses are usually incorporated into a

coefficient and Equation (3.7) is expressed as:

losses h h hgC

v vd

sc1 2 2

2

1

21

2(3.8)

where vsc is the velocity through the screen

Cd is a discharge coefficient with a typical value of 0.84.

Alternatively, an orifice equation may be applied in the form:2

2

2

2

1

2 AC

Q

ggC

vh

dodo

sc (3.9)

where Q is the flow rate,

A is the effective open area of the submerged screen,Cdo is a discharge coefficient.

It should be noted that the discharge coefficient in Equation (3.9), Cdo is different from

that in Equation (3.8). In the latter equation, the value of Cdo is dependent on screen

design parameters and is supplied by the screen manufacturer or by experimentation.

If the screens are to be manually cleaned, the effective open area should be taken as 50

% of the actual open area, representing the half-clogged condition. The head loss should

be estimated under conditions of maximum flow.

v1


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If the bar screen is clean, Kirschmers equation may be used for estimating the head loss

as follows:

hW

bhv

1 33.

sin (3.10)

where is a bar shape factor, as given in Table 3.4

W is the total transverse width of the screen

b is the total transverse clear spacing between bars

hv is the upstream velocity headv

g

1

2

2

is the angle of the bars to the horizontal

Table 3.4: Bar Shape Factor for Kirschmers Equation

Bar TypeSharp-edged rectangular 2.42

Rectangular with semicircular upstream face 1.83

Circular 1.79

Rectangular with semicircular upstream and

downstream face

1.67

Tear shape 0.76

It should be noted that Kirschmers equation is a general form of the standard head loss

equation:

h Kv

g

2

2 (3.11)

where v is identified as v1

K is given by KW

b

1 33.

sin

It should be noted that the expressions developed above are of use in determining the

minimum energy losses through screens, but are of little value in determining the energy

loss once material begins to accumulate behind the screen.

The screen design should take into account the maximum increase in head loss likely tooccur under the conditions of maximum flow rate and minimum cleaning frequency. It

is especially important with manually raked screens that sufficient freeboard is provided

in the upstream channel to avoid the danger of spills at high flows.

ExampleA mechanically cleaned wastewater bar screen is constructed using 6.5 mm wide bars

with a clear spacing of 5.0 cm. The wastewater flow velocity in the channel

immediately upstream of the screen will vary from 0.4 m/sec to 0.9 m/sec.

Determine the design head loss for the screen at the two extremes of flow. Assume that

the discharge coefficient Cdhas a value of 0.84.


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Solution:

Energy Equation: Head Loss1

2 22

1

2

gCv v

d

sc

Ifv1 is given, vsc can be calculated, knowing the screen geometry.

Continuity Equation:)(1111 clearscsc whvwhv

where w1 is channel width in front of the screen,

wsc is the total width of clear openings of the screen.

w

wsc clear

1

( )

bar spacing + bar width

bar spacing

50 6550

. = 1.13

vsc = 1.13v1

h v v1

2 9 81 084113

2

2

1

2

1

2

x xx

. .. = 0.02v1

2

v1 = 0.4 m/sec h = 3.2 mm

v1 = 0.9 m/sec h = 16.2 mm

Design of Screens

The velocity in the approach channel is normally kept between about 0.3 m/sec and 1m/sec. The lower limit is designed to prevent the settling of coarse matter while the

upper limit is designed to prevent the screens being carried away by the flow.

Screens may be manually cleaned or mechanically raked. Manually cleaned screens are

only fitted in small treatment plants, typically servicing a population equivalent, PE of

less than 5,000. Mechanically raked screens are recommended for all plants servicing a

PE greater than 2,000.

Figure 3.9 shows a schematic of a manually raked screen. The maximum clear spacing

between bars is typically set at 25 mm, although American practice permits spacings up

to 50 mm. To facilitate cleaning, the bars are normally set at 30 45 from the vertical.

The screenings are manually raked on to a perforated plate where they drain, prior to

removal. Cleaning must be frequent to avoid clogging. Infrequent cleaning may result in

significant upstream backwater caused by the buildup of solids. When cleaning is

carried out, the sudden release of the ponded water leads to flow surges.


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Figure 3.9: Schematic of Manually Raked Screen

A schematic of a mechanically raked bar screen is shown in Figure 3.10. Typically, the

maximum clear spacing between bars is 25 mm, although American practice permitsspacings up to 38 mm. A spacing of 18 mm is considered satisfactory for the protection

of any downstream equipment.

Figure 3.10: Schematic of Mechanically Raked Bar Screen

Mechanically raked screens are normally set at between 0 and 45 from the vertical. The

use of such screens leads to reduced labour costs, improved flow conditions, and

improved capture of screenings. A large number of proprietary screens with mechanicalrakes are available. Manufacturers will normally provide design charts to facilitate

selection of the correct screen size for a particular service.

Figure 3.11 shows a schematic of another type of screen a drum screen. Screenings

naturally fall from the screen as it rotates above the hopper. A water spray assists in

removing screenings.


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Figure 3.11: Schematic of Drum Screen

An example illustrating the design technique for a screen and screen chamber is

presented in an example below..

ExampleDesign a screen and screen chamber and determine its hydraulic characteristics for a

loading of 10,000 PE. All material larger than 12 mm is to be screened out. The screen

is a bar screen with rectangular bars of 5 mm transverse dimension. At the peak design

flow, the velocity through the screen should be 0.9 m/sec

The water level downstream of the screen is controlled by a downstream long-throated

flume which gives a depth of 400 mm at the peak design flow and 175 mm at ADWF.

In particular, a) determine the head loss across screen;

b) determine the screen chamber width;

c) check the velocities; and

d) if the screen is 50 % blocked, calculate the head loss across it.

Solution:

Estimate loads

ADWF = 225l/day/PE

Peak flow factor (PFF) = 4.7 (PE)

-0.11

(PE in thousands)


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Load = 10,000 PE

ADWF = 2.25Ml/day = 26l/sec

Peak flow factor = 4.7 10-0.11

= 3.65

Peak flow = 3.65 26 = 95l/sec

Choose the screen:

Bar spacing = 12mm (will screen out all larger material)

Bar thickness = 5mm

If screen velocity is 0.9m/sec for peak flow, calculate v1

bar widthspacingbar

spacingbar1 scvv 0 9

12

17. = 0.64m/sec

a) Determine head loss

2

1

2

22

1vv

gCh sc

d

1

2 9 81 0 840 9 0 642

2 2

. .. .

m029.0h

Depth upstream of screen mhg

v

g

vhh 43.0

22

2

1

2

221

b) Determine screen chamber width.

From continuity, required clear screen width ( Wsc clear( ) ) is

Q h W vsc clear sc1 ( )

Wsc clear

0095

0 429 09

.

. .= 0.246m

Required screen chamber width


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w 0 24617

12. 0.349m or 350mm

(CHECK against approach velocity)

429.0349.0

095.0

1

1hw

Qv =0.64m/sec

c) Check velocities

ADWF = 0.026m3/sec

Associated h2=175mm

350.0175.0

026.02v 0 426. m / sec

Now, because the flow is lower, we would expect a reduced head loss as well.

The upstream depth will be less than 0.175 + 0.029 < 0.204m

v1

0026

0 204 0 349 0365

.

. . . m / sec >0.3m/sec

O.K. (we could calculate v1exactly, but the above argument removes

the need to do so)

d) Head loss with screen half blocked

Energy equation: hv

g

hv

g

hL11

2

2

2

2

2 2

For peak flow Q = 0.095m3/sec

Upstream from the screen (Section 1):

11

1

271.0

35.0 hh

Qv

On the screen:

Wsc (clear) = * 0.246m = 0123 m


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11

77.0

123.0

095.0

hhvsc

After the screen (Section 2):

h2 0 4. m smwh

Qv /679.0

35.04.0

095.0

2

2

Losses: h v vL sc1

2 9 81 0 842

2

1

2

. .

Substitute for v h v vsc1 2 2, , , in energy equation

2

1

2

2

1

2

2

2

2

1

2

1

271.077.0

84.06.19

1

6.19

679.0

4.06.19

271.0

hhhh

hh1 1

2

0 003750 4235 0

..

Solve by trial

h1 0539. m

Head loss = 539400 139mm

vQ

hsc 0124

0 095

0124 0 5391.

.

. .=1.42m/sec

vh1 1

0 271 0 271

05390503

. .

.. / secm

3.4.2 Grit ChambersWithin sewage treatment plants, gritcomprising sand, egg shells, coffee grounds and

other non-putrescible materialmay cause severe problems in pumps, sludge digestion

facilities, and de-watering facilities. In addition, it may settle out in downstream pipes

and processes.

The grit removal process is carried out in grit chambers (Figure 3.12) at an early stage

of treatment because the grit particles cannot be broken down by biological processes

and the particles are abrasive and wear down the equipment. Because the grit material is

non-putrescible, it requires no further treatment following removal from the sewage

treatment process and ultimate disposal.


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It should be noted, however, that the location of grit chambers upstream of the sewage

pumps at the entrance to the sewage treatment plant would normally involve placing

them at a considerable depth involving substantial expense. Therefore, it is usually more

economical to pump the sewage, including the grit, to grit chambers located at a

convenient position upstream of the treatment plant units. It is recognised that thepumps may require greater maintenance as a result.

Figure 3.12: A grit chamber

Grit chambers are designed to remove inorganic solids of sizes greater than about 2 mm.

Removal is commonly effected using settlement, separation using a vortex, or

settlement in the presence of aeration (in the latter process, aeration keeps the lighter

organic particles in suspension). There are important hydraulic principles associated

with each of these three processes.

In this section, the choice of grit removal process is first discussed. The three main

types of grit chamber are then described and the hydraulic aspects of the operation of

each are described qualitatively and, where appropriate, quantitatively. Design aspects

are also discussed.

Choice of Gr it Removal Process

The choice of grit removal process depends largely on the size of the sewage treatment

plant:

For a PE < 5,000 (small treatment plants), a horizontal flow (constant velocity)

settling chamber is commonly used.

For PE of between 5,000 and 10,000 (medium-sized treatment plants), a vortextype grit chamber is commonly used.


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For PE > 10,000 (large treatment plants), an aerated grit chamber is often

specified, although the vortex type chamber may also be used.

Whichever type is used, it is vital that the unit must operate effectively over the full

range of expected flows.

Other non-hydraulic considerations include grit removal from the unit, which may be

manual or mechanical; handling, storage, and disposal of grit, and the provision of

standby or bypass facilities.

Horizontal F low (or Constant Velocity) Gri t Chamber

The horizontal flow grit chamber is basically an open channel with a detention time

sufficient to allow design particles to settle. Additionally, the velocity must be

sufficiently high that organic materials are scoured so that they pass through the gritchamber for subsequent biological treatment.

The Camp-Shields equation is commonly used to estimate the scour velocity required to

re-suspend settled organic material. This equation is expressed as:

vkgd

fs

p8(3.11)

where vs is the velocity of scour

d is the particle diameterk is an empirical constant (typically 0.040.06)

f is the Darcy-Weisbach friction factor (typically 0.02)

p is the particle density

is the fluid density

Typically, this equation yields a required horizontal flow velocity of 0.15 0.3 m/sec.

This compares well with the Malaysian design standard of 0.2 m/sec.

The primary hydraulic design issue for the horizontal flow grit chamber is the

maintenance of the constant velocity in the channel, despite large variations in the flow

rate, based on a typical diurnal flow pattern.

The problem is illustrated in the following.

Consider a rectangular chamber with the

flow passing over a low rectangular weir

placed at its end (Figure 3.13).

The discharge relationship for the weir is:

Q C B gH d 23

2 (3.12)

where

Figure 3.13: A chamber with a weir


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Cd is a discharge coefficient,

B is the channel width,

H is approx = the channel depth (if the weir height is very small).

(The derivation of Equation 3.12 is presented in the Study Guide of CIV2262).

Now, the horizontal velocity, vh, is related to the flow rate, Q, and channel geometry by:

vQ

BH

C B g H

BHC g Hh

dd

22

32 1

2 (3.13)

Substituting forH1

2 from Equation (3.12) yields:

v C gQ

C gBh d d2

2

13

(3.14)

v

v

Q

Q

h

h

max

min

max

min

13

(3.15)

Now, a typical value for the ratio of maximum to minimum flow rates (Qmax/Qmin) is

about 5. Substitution of this ratio into Equation (3.15) yields a corresponding value for

the ratio of maximum to minimum velocities (vh(max)/vh(min))of:

71.15

3/1

min

max

h

h

v

v

If 0.2 m/sec is chosen for the value ofvh(min), the corresponding value forvh(max) would

be 0.342 m/sec, which would be unacceptably large.

Therefore, the shape of either the channel or the weir must be modified to

maintain a constant and satisfactory horizontal velocity.

Modification of Channel Shape

The issue to be resolved is whether or not it is possible to develop a channel shape such

that the horizontal velocity remains constant for all flow rates. It is assumed that the

channel discharges into a rectangular control section, such as a long-throated or Parshall

flume. Such a device acts as a water level control and a flow measurement device (see

CIV2262 Study Guide; a flume is shown in Figure 3.14 to the right).

The analysis that follows is generally applicable to any rectangular cross-section. The

analysis specifically makes use of the properties of a long-throated flume because it is

widely used in practice.

The flow through a long-throated flume may be expressed in the form:


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Q g b H c2

3

2

31

32 (3.16)

where bc is the throat width,

H1 is the upstream head.

Differentiation of Equation (3.16) yields:

dQ gb H dH c2

31

12

1 (3.17)

Now, within the channel, the horizontal velocity, vh, is given by:

vQ

wHh

1

(3.18)

or:Q v wH h 1 (3.19)

where w is the channel width.

Differentiation of Equation (3.19) yields the flow through an elemental horizontal strip

of width w in the channel in the form:

dQ v wdHh 1 (3.20)

Equating the right hand sides of Equations (3.17) and (3.20) yields:

2

31

12

1 1gb H dH v wdHc h (3.21)

Solving Equation (3.21) forw yields:

w gb

vHc

h

2

31

12 (3.22)

If we want vh to be constant for any depth in the channel (i.e. any flow rate), the channelshould have the width determined as a function of its height:

w Hconstant x 11

2 (3.23)

Equation (3.23) describes a parabola, indicating that a parabolic shape for the channel

cross-section will ensure a constant value of velocity, vh, regardless of flow rate. To

reduce construction costs, the parabolic shape is normally approximated with a

trapezoid.


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Design Aspects

Figure 3.14 presents a layout of a typical design grit chamber with a flume for depth

control at its downstream end. The individual chambers have a trapezoidal shape with a

grit storage channel at their bottom for collection of deposited grit.

Figure 3.14: Schematic of Channel-modified Horizontal Constant Velocity Grit

Chamber

As a minimum, one channel and a bypass should be installed. However, usually at least

two are designed for operational reasons (maintenance, etc). When the number of

channels is determined, the maximum, average, and minimum flows in an individual

channel can be determined (i.e. Qemergency, Qmaximum, Qaverage, and Qminimum, are used to

design the shape and length of the grit channel).

The system should be designed such that, when one channel is out of service, its flow is

diverted to the other channels. The resulting emergency flow for each channel is based

on the maximum flow into the set of grit chambers with one out of service.

Other practical aspects are associated with the turbulence which occurs in the inlet and

outlet zones of the chamber. These zones are similar as in any sedimentation tank, and

are illustrated schematically in Figure 3.15.

Turbulence occurs in the inlet zone as the flow is established. A similar phenomenon

occurs in the outlet zone as the flow streamlines turn upwards. To allow for this

disturbance, a 2550 % increase in the calculated settling length is applied.

Typical design criteria for a channel-modified horizontal grit chamber are presented in

Table 3.5.


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In the case of a flume being

used for flow regulation in the

chamber, the cross-section area

of the chamber is defined

according the followingprocedure. Equation 3.16 for the

flume is:

23

1

23

3

2HbgQ c

where bc is the

considered

width, and H1

the water depth.

Table 3.5: Typical design criteria for Channel-Modified Grit Chambers

Design Parameter Typical Values Comments

Water depth (m) 0.61.5 Dependent on channel area and

flow rate

Length (m) 325 Function of channel depth and

grit settling velocity

Extra for inlet and outlet 2550 % Based on theoretical length

Detention time at peak flow

(seconds)

1590 Function of velocity and

channel lengthHorizontal velocity (m/sec.) 0.150.3 0.2 m/sec is Malaysian

Standard

Equation 3.22 expresses the chamber width at the surface, w, as:

21

13

2H

v

bgw

h

c or

21

13

2Hg

wvb hc

Combining Eqs. 3.16 and 3.22 the flow can be expressed as:

hvwHQ 13

2

Since Q=Avh, the cross-sectional area of the chamber, A is:

13

2wHA (3.24)

The design procedure for a channel-modified grit chamber is illustrated in the following

Example.

Figure 3.15: Schematic of Settling Process in Grit

Chamber


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Example:Design a horizontal/constant velocity grit chamber for a hydraulic load of 2,000 PE.

Consider only the ADWF and the peak flow. The water level within the chamber iscontrolled by a downstream long-throated flume which gives a depth of 205 mm at the

peak design flow and 80 mm at ADWF. The following should apply:

Maximum horizontal velocity is v=0.2 m/sec

Channel length > 18 times maximum water depth

Grit quantity is estimated as 0.03 m3/ML of wastewater

Grit collection channel to be cleaned out twice per week

Solution

Average dry weather flow

ADWF = 225 2,000 = 0.45 ML/day

ADWF = 5.2l/sec

Peak flow factor PFF 4 7 2 0 11. . = 4.35

Peak flow PF= 4.35 5.2

PF= 23 l/sec

The long-throated flume gives depth of

205 mm at peak flow and 80 mm at ADWF

Therefore, cross-sectional areas of the trapezoidal cross-section should be:

ADWF :0.2

0.0052

v

ADWF=Area 0026

2. m

Peak: Area =0.023

0.2 0115. m2

Surface widths at each flow are now calculated.


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At average dry weather flow

Surface width12

3

HA

0 026 3

2 0 08

.

.

049. m

At Peak Flow

Surface width0115 3

2 0 205

.

.

=0.84m

Length of chamber:

> 18 max. depth

> 18 0.205

Use 3.7m

Grit quantity is based on ADWF

Grit quantity = 0.45 0.03 = 0.014m3

/day

At twice weekly cleanout, grit accumulation

0 014 4. ~ 0056. m3

Required cross-sectional area of grit collection channel

0056

37

.

.0015. m2

Use grit collection channel 150mm wide 110mm deep

(gives some margin). Allow for freeboard (say, 200mm)

Parabolic section to be approximated by trapezoid.

Modification of Downstream Control Weir

The chamber can have a rectangular cross-section, only if the flow is controlled by a

special device at the downstream end.


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For a rectangular grit chamber, the flow rate is given by:

BhvQ h (3.25)

where B is the chamber widthh is the flow depth in the chamber

The form of Equation (3.25) indicates that forvh to be constant, regardless of flow rate,

the flow rate should be linearly proportional to the depth, h. This may be assured by

using a downstream control weir characterised by a linear relationship between flow

rate and head on the weir crest.

Such a weir is the Sutro weir which is described and analysed in CIV2262 (see

CIV2262 Study Guide).

Sutro weir

Vortex Grit Chamber

A schematic of a typical vortex grit chamber is shown in Figure 3.16.

With reference to this figure, grit-laden flow enters the unit tangentially at the top. The

resulting spiral flow pattern tends to lift the lighter organic particles while the

mechanically induced vortex captures grit at the centre. The grit is then removed by air-lift or through a hopper. It should be noted that the grit sump has a tendency to become

compacted and will potentially clog. Sometimes provision is made for the use of high-

pressure agitation water or air to clear the sump.

The adjustable rotating paddles maintain the proper circulation within the unit for all

flows. Attention should be paid to the tendency for these paddles to collect rags.


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Figure 3.16: Schematic of Typical Vortex Grit Chambers (a) PISTA Unit (b) TeacupUnit

Vortex grit chambers are highly energy-efficient. The head loss across the unit is

minimal when operating correctly and unclogged. American practice indicates a value

of 6 mm, although an allowance of 100 mm is recommended.

Vortex grit chambers have the great advantage that they are very compact. Their design

is usually proprietary so that manufacturers will usually produce a suitable unit to

accommodate stated performance specifications. Manufacturers specifications will

provide information on the maximum water depth within the chamber.

Aerated Grit Chamber

Aerated grit chambers are commonly used in medium to large sewage treatment plants.

The introduction of air through a diffuser, located on one side of the tank, induces a

spiral flow pattern in the sewage as it moves through the tank, as shown in Figure 3.17.

Correct positioning of the tank inlet and outlet directs the flow perpendicular to the

spiral roll pattern. Inlet and outlet baffles are normally installed to dissipate energy and

minimise short-circuiting. Head loss across the chamber is minimal.


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Figure 3.17: Helicoidal Flow Pattern in an Aerated Grit Chamber

The roll velocity is set so that it is sufficient to maintain lighter organic particles in

suspension while allowing heavier grit particles to settle. Because conditions change

with flow rate, the air supply is adjustable to provide the optimum roll velocity.

A further advantage of the introduction of air is that the sewage is freshened, leading to

a notable reduction in odour. If desired, the chamber can be used for chemical addition,

mixing, and/or flocculation ahead of primary treatment. Grease removal may be

achieved with a skimmer.

If correctly designed, an aerated grit chamber with a minimum hydraulic detention time

of 3 minutes will capture about 95% of grit larger than 0.2 mm when operating at its

peak flow. The usual range of design specifications is given in Table 3.6.

Table 3.6: Typical Design Specifications for an Aerated Grit Chamber

Design Parameter Range of Values Comments

Depth 25 m Varies widelyLength 820 m

Width 2.57 m

Width:Depth Ratio 1:15:1 2:1 typical

Length:Width Ratio 3:15:1

Minimum Detention Time 25 minutes 3 minutes typical

Air Supply 0.250.75 m /min/m 0.45 m /min/m typical

Diffuser Distance from

Bottom

0.61.0 m

Transverse Roll Velocity 0.60.75 m/sec


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Example:Design an aerated grit chamber for a hydraulic load of 20,000 PE. The following is

chosen:

The minimum detention time at peak flow is 3 minutes,

The width to depth ratio is 2:1,

The length to width ratio is 2:1,

Grit quantity is estimated as 0.03 m3/ML of wastewater,

The aeration requirement is 10 litres/sec/m length of tank.

Solution

ADWF = 20,000 225l/day= 4,500m3/day=52l/sec

PFF= 4.7 20-0.11= 3.38

Peak flow = 52 3.38 = 176 l/sec

Grit chamber volume:

Minimum detention time at peak flow = 3minutes

= 3 60 = 180seconds

Required volume = 0.176 180 = 31.7 = 32m3

W

D

L

W2 2,

Volume = D W L = 32

W = 2D, L = 2W = 4D

D 2D 4D = 32

Dimensions D = 1.6m

W = 3.2m

L = 6.4m

Aeration requirement

10l/sec/m length = 10 6.4 = 64 l/sec

Means should be provided to vary the air flow rate to control grit removal

rate and grit cleanliness

Grit quantity

Based on average flow rate,

= 4.5ML/day 0.03m3/ML = 135 l/day


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3.4.3 Other

Comminutors

Comminutors mince wastewater solids (rags, paper, plastic, etc) by revolving cuttingbars. They are used instead of fine bar racks (after grit chambers).

EqualizationThe purpose of equalization is to dampen flow variations so that wastewater can be

treated at a nearly constant flow rate. It is achieved by constructing large basins.

3.5 Primary Treatment

After preliminary treatment, the wastewater contains light organic solids, some of which

can be removed by gravity or screening.

3.5.1 Primary Sedimentation TanksSedimentation tanks (also known as clarifiers) are used as a part of both primary

treatment and secondary treatment processes. Here only preliminary sedimentation

tanks will be discussed.

Sedimentation tanks may be rectangular, square, or circular in shape. A schematic of a

typical rectangular clarifier is shown in Figure 3.18, of a circular clarifier in Figure 3.19,

and a square one in Figure 3.20.

Rectangular tanks (Fig. 31.18) are commonly used for primary sedimentation. They

occupy less space than circular tanks and can be economically built side by side with

common walls.

Figure 3.18: Schematic of Rectangular Sedimentation Tank

In the case of a circular, the flow enters at the centre and settlement takes place as the

flow moves outwards and rises. The effluent is collected in a channel or launder, which


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then conveys the flow to an exit channel or pipe. Circular tanks require a careful design

of the inlet stilling well to achieve a stable radial flow pattern without causing excessive

turbulence in the vicinity of the central sludge hopper. Inlet design is considered in

subsequent paragraphs.

Figure 3.19: Schematic of Circular Clarifier

Square or upflow tanks (Fig. 3.20) typically have deep hopper bottoms and are common

in small treatment plants. Their primary advantage is that sludge removal is carried out

entirely by gravity. The steeply sloping sides typically 60concentrate the sludge atthe bottom of the hopper. A significant disadvantage is that hydraulic overloading may

cause major problems because any particles with a settling velocity less than the surface

loading rate will not be removed, but will escape with the effluent.

This section emphasises the hydraulic aspects of the design of clarifiers. The basic

design procedure is reviewed and design guidelines are presented. The important

procedure for the design of the launder is then discussed. Finally, a design example is

presented to aid understanding.


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where Q is inflow;

A is surface area;

Detention Time

The detention time is given by the equation:

SLR

H

Q

Volumet (3.25)

where H is depth of the tank.

Following the specification of these parameters, the dimensioning of the tank then

proceeds as follows:

SLR

QAArea,SurfaceTank (3.26)

ADL orDiameter,orLengthTank (3.27)

whereL

4for rectangular tanks

4for circular tanks

Clarifiers are normally designed to provide a detention time of between 1.5 and 2.5

hours, based on the peak flow rate. It is noted that the design criteria for Malaysian

systems incorporate a time of 2 hours based on the peak flow rate.

The Forward Velocity

The forward velocity is also an important aspect of the design of rectangular tanks. If

this is excessive, scouring and re-suspension of the sludge will result.

The forward velocity is given by:

vQ

WHh (3.28)

Incorporating Equation (3.26) for the detention time,

vL

th (3.29)

It is evident from Equations (3.28) and (3.29) that the forward velocity influences the

choice of length to width ratio. The maximum forward velocity to avoid the risk of

scouring settled sludge is 10 to 15 mm/sec, indicating that the ratio of length to width

should preferably be about 3:1.


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Values of L/W in practice range between 3 and 6. The Malaysian Draft Guidelines

specify a value of 3.

Weir Loading RateThe weir loading rate is defined as

WLR = Q/Lw (3.30)

where Lw is the length of the outlet weir.

If this value is too high, the approach current generated by the weir will extend

upstream into the settling zone, creating a potential disruption of the flow pattern. A

weir loading rate of between 100 and 200 m3/m/day is typically specified. Achieving

this value is a particular problem for rectangular tanks which is usually overcome by

utilising multiple suspended weir troughs.

In circular tanks, the weir loading rate associated with a perimeter weir is normally

satisfactory at high flows. At low flows, however, difficulties may arise from a weir

loading rate which is too small because the consequent very small flow depths over the

weir make the tank flow pattern very sensitive to errors in weir levelling. This problem

may be overcome by constructing the perimeter weir as a saw-tooth weiror multiple

V-notchto increase the flow depth.

Design Guideli nes

Design guidelines for clarifiers vary significantly from country to country. Typical

guidelines from American practice are presented in Table 3.7.

Table 3.7: Typical Design Guidelines for Circular Primary Clarifiers

Parameter Value

Detention time

For average weather flow 1.5 and 2.5 hours

Surface loading rate = Q/Surface area

For average dry weather flow

For peak flow conditions

32 - 49 m3/m2/day

49 - 122 m3/m2/daySidewater depth 2.15 m

Weir loading rate = Q/Weir length 125500 m /m/day

Primary clarifiers are designed more conservatively if sedimentation is the only

treatment and if activated sludge is being returned to the primary clarifier. Rectangular

clarifiers are generally designed under the same criteria as circular clarifiers. Typical

length to width ratios for rectangular primary clarifiers range from 3:1 to 5:1, although

many existing tanks are characterised by ratios of between 1.5:1 and 15:1.

A well-designed and operated primary clarifier should be capable of removing between

50 and 65% of the total influent suspended solids.


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The issues of surface loading rate, detention time, and weir loading rate are illustrated

by the example below.

ExampleTwo primary clarifiers are 26 m in diameter with a 2.1 m side water depth. Single

effluent weirs are located on the peripheries of the tanks. For a wastewater flow of

26,000 m3/day, calculate:

a) the surface loading rate,

b) the detention time, and

c) the weir loading rate.

Solution

Surface area of each clarifierD2 2

4

26

4530 2m

Total surface area = 530 2 = 1,060m2

Total volume = 1,060 2.1 = 2,230m3

a) Surface loading rateQ

A

26 000

1060

,

, 245. m / m / day3 2

b) Detention timeVolume

Flow rate

2 230

26 00024

,

,2 06. hours

c) Weir loading rateflow rate

weir length

26 000

2

,

D

26 000

2 26

,

159m / m / day3

Tank I nlets

Sedimentation tank inlets must be designed to distribute the flow as uniformly aspossible so that the best possible flow pattern is maintained. The influent jet has a high

amount of kinetic energy that must be dissipated.

For rectangular tanks, various baffled inlet arrangements have been used which are

effective for energy dissipation and flow distribution. Typical arrangements are shown

schematically in Figure 3.21.

With circular tanks, the radial flow from the inlet is inherently less stable than the

horizontal flow in a rectangular basin. Careful design is needed to achieve a stable

radial flow pattern. Typical arrangements are shown in Figure 3.22 for (a) side feed, (b)


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vertical pipe feed, and (c) slotted vertical pipe feed. In all cases, the primary design

principles are that energy must be dissipated and the flow distribution must be uniform.

Figure 3.21: Schematics of Typical Rectangular Sedimentation Tank Inlets

Figure 3.22: Centre-feed Inlets for Circular Clarifiers: (a) Side Feed, (b)

Vertical Pipe Feed, (c) Slotted Vertical Pipe Feed

Ef fl uent L aunder Design

Rising wastewater in a clarifier flows over a weir into a channel or launder which, in

turn, conveys the collected effluent to the exit channel. Flow in the launder is classified

as spatially varied because the flow rate increases with distance along the launder. This


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characteristic requires the use of the momentum equation for its analysis, rather than the

energy equation.

The basic flow condition is illustrated schematically in Figure 3.23 which shows the

flow spilling over the multiple V-notch weir into the launder. A full momentumanalysis, including the effects of friction are needed. A simplified approach is usually

adequate and is presented herein.

The first issue is the size of V-notch weir required. The individual V-notches are

typically set out with a centre-to-centre spacing of between 150 and 300 mm. With the

number of V-notches consequently established, the flow through each notch can be

determined from:

QQ

NperV notch

(3.31)

where N is the number of V-notches.

The maximum height, h, over the weir is then determined from the standard V-notch

weir equation (see Study Guide CIV2262):

25

2tan2

15

8hgCQ dnotchperV (3.32)

The discharge coefficient, Cd, is a function of the notch angle, . For = 90

0

, Cd has avalue of 0.58.

The head over the weir, calculated from Equation (3.32), should be increased by a

safety factor of 15%.

The next stage in the hydraulic design is to determine the maximum depth in the

launder. First, the critical depth at the discharge point of the launder is calculated from:

yqL

b gc

2

2

13

4(3.33)

Figure 3.23: Definition Sketch for Flow in a Launder and image of a completed design.


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where LwQq / is discharge per unit weir length,

Lw is the length of weir,

L is the length of the launder (circumference of the tank),

b is the width of the launder.

The depth at the upstream end of the launder is then calculated from:

H yq x

gb yc

c

22 2

2

122

(3.34)

where x=L/2 for a circular basin.

The depth, H, calculated from Equation (3.34) should be increased by a factor of safety

of 50% to allow for friction loss, freeboard, and a free fall allowance.

The design of a launder is illustrated by the example below.

ExampleDesign the overflow weirs and launders (collection channels) for two identical circular

clarifiers that treat a design flow of 20,000 m3/day, and a peak hourly flow of 32,000

m3/day. They are 18 m in diameter each. The critical condition is when the peak flow

occurs with one clarifier out of service. The launder must be able to cope with the

corresponding flow.

Solution

Weir design

One clarifier must handle peak flow.

Peak weir loading rate32,000m / day

2 18m

3

where: 2 represents the inflow on both sides

18 represents the diameter

WOR /m/daym2833

Assume that weir comprises V-notches with spacing of 25cm centre to centre (this may

need adjusting).

Total number of V-notches 20 25

D

.

Take D as 18m, even though it will be less for the inner ring and more for the outer.

Total number of V-notches 218

0 25.= 452


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Flow per notch /sm00082.0452

1

24600,3

000,32 3

Now, for each V-notch, notch angle is =900 and Cd=0.58.

25

2tan2

15

8hgCQ d

mh 051.06.1958.08

00082.015 52

A safety factor of 15% is normally appropriate.

Allow for water depth over notch of 1.15 0.051= 0.059m

= 60mm

Width of V-notch at the top = 60mm 2 = 120mm.

Weir design as follows:

Launder design

q is discharge/unit weir length = weir loading rate 2

(because launder is fed from both sides)

q283

3 600 24

2

,

m / m / sec3

0 0066. m / m / sec3

Assume a launder width

Try 500mm

Calculate depth at launder discharge point


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45

m0.24381.95.04

180066.0

4

31

2

231

2

2

gb

qLyc

Calculate maximum depth in launder at upstream end

H yq x

gb y

2

2 2

2

0 52

.

(Note: xD

2)

mH 419.0243.05.081.9

2

180066.02

243.0

21

2

2

2

2

Increase this depth by 50% to allow for friction loss in the launder, freeboard, and free-

fall allowance.

Total depth to be provided in launder

= 0.419 1.5

= 0.629, say 0.65m

Launder depth below vertex of V-notch weirs = 0.65m

Launder width = 0.50m

Documents

Wastewater - Primary