THE INDIAN CONCRETE JOURNAL - icjonline.com · damage to the RC structures. ... beam testing, the authors feel it ... • “Manufactured Sand & its Behavioural Study” on 1st &

PUBLISHED BY ACC LIMITED

February 2018, Vol. 92, No. 2, 68 pages.

THE INDIAN CONCRETE JOURNAL

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Ninety-two papers have been divided into 13 categories, namely, Bridge construction, Bridges across major rivers, Rail and road bridges across Thane creek, Bridges on Konkan Railway, Design and optimization of bridges, Cable-stayed bridges, Pamban bridge, First-of-its-kind bridges, Condition monitoring and rehabilitation of bridges, Seismic design of bridges, Integral bridges, High performance concrete for bridges and Delhi Metro.

CD on Concrete Bridges An ICj Compilation

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3The Indian Concrete Journal February 2018

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CIRCULATION OFFICEThe Indian Concrete Journal

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Editor: Ashish Patil

Editorial Team:

Ulhas Fernandes

THE INDIAN CONCRETE JOURNALFebruary 2018, Volume 92, Number 2

TECHNICAL PAPERS

10Investigation of the behaviour of shear-critical reinforced concrete columnsKomathi Murugan and Amlan K. Sengupta

25

35

Rebar identification, cover thickness and diameter estimation in reinforced concrete members using cover meter and GPR techniquesBhaskar Sangoju, S. Vasanthakumar, K. Ramanjaneyulu and K. Sivasubramanian

Flexural toughness characterization of steel, polymer and glass fibre reinforced concrete based on the notched beam test Sujatha Jose, Ravindra Gettu and Shabari Indhuja

FEATURES

04 EDITORIAL

06 NEWS & EVENTS

51POINT OF VIEW: Suggestions to improve method of installation of cast-in-situ driven and bored concrete pilesKartik Chandra Ta

4 The Indian Concrete Journal February 2018

EDITORIAL

The next two years are crucial ! In the run up to the

General Elections in 2019, Government’s plan to

accelerate economic growth will unfold in the Union

Budget due this month. Execution of infrastructure projects

and allocation of funds for their timely completion are

likely to be the top priority for the Government to showcase

their achievements. This will surely create a lot of demand

for quality men and materials and auger well for the

construction industry!

This issue has four detailed papers…

Columns are most vulnerable during an earthquake.

Conventional design is based on the strength of a member.

But for the columns to resist lateral load, design based

on estimation of deformation and post-yield ductility

is preferred. A comparative study of the strength and

behaviour of a shear-critical cantilever column specimen

based on experimental, analytical and numerical

investigations is presented in this first paper.

Rebar corrosion is a major reason that causes extensive

damage to the RC structures. Providing adequate cover is

crucial especially in aggressive environments. To assess

deterioration in structures, it is essential to locate the

rebar, know the cover thickness and rebar diameter. This

study explores two techniques of cover meter and ground

penetrating radar used to determine the reliability in

gathering these details in RCC structure.

Using Fibres in Reinforced Concrete (FRC) provides

improved energy absorption capability, fatigue resistance,

fracture energy, ductility in tension and compression,

flexural strength and toughness, resistance against impact

loading, abrasion and durability. Researchers suggest

methods to measure the effectiveness of fibres in the

resistance offered by FRC against crack propagation.

With limited testing capability and experience in notched

beam testing, the authors feel it is necessary to establish

a database of results and propose test methodologies that

are appropriate.

Suggestions to improve method of installation of

cast-in-situ driven and bored concrete piles under

different sub-soil conditions, the load characteristics of a

structure, settlement, and any other special requirements

of the project is detailed by the author covered under Point

of View.

Thoughts come to my mind on what would interest you

every time a new issue is planned? We look forward to

your views and suggestions...

With Best Regards,

Ashish Patil

From the Editor’s Desk...

The Indian Concrete Journal February 20186

NEWS & EVENTSNEWS & EVENTS

AKC’s FEBRuARy 2018 PRogRAms

The February 2018 programs of Ambuja Knowledge Centre include the following.

AKC Andheri

• “Advance Concrete Mix Design Workshop on 7th & 8th Feb’ 2018

• “Basic course on fundamentals of concrete materials & technology” workshop on 19th – 23rd Feb’ 2018.

• Technical Lecture on “MasterEase- for making Low Viscosity for Concrete” on 23rd Feb’ 18 (Speaker-Swapnil Patil, Sales Manager, BASF India)

AKC Thane along with ‘Centre of Excellence’, ACC Thane

• “Manufactured Sand & its Behavioural Study” on 1st & 2nd Feb’ 2018.

• “Durability Parameters of Concrete” on 15th & 16th Feb’ 2018.

• Technical Lecture on “ Affordable Housing by green, Speedy RCB Construction Method” on 2nd Feb’ 18 (Speaker - Er. Ganesh Kamat, Technical Advisor, Ganaka Engineers & Architects)

AKC Navi Mumbai

• Technical Lecture on “Structural Waterproofing with Bituminous Membrane” on 16th Feb’ 18 (Speaker - Er. Swapnadeep Chaudhari, National Manager, Roofing, Sika India Pvt. Ltd)

Ambuja Cements Ltd., Mumbaip: ++91 22 40667620 / 8291885509e: [email protected]: www.foundationsakc.com

ExtREmE EnginEERing

The 11th IFHS Conference is being arranged in Singapore during February 26-28th 2019.

Engineering projects around the globe are getting more and more ambitious as major cities jostle to be known for signature landmark buildings and overcoming impossible engineering feats. Today, structures compete to dwarf anything that has come before. Records are broken, design and technical boundaries are challenged; and new concepts on liveability and sustainability have also emerged in the race for space. Plans beyond imagination are on the drawing board for

projects so huge - not only in scale but in their implications for society. Some examples include ‘Tunnelling under the Alps’, ‘Bridging the Bering Strait’, ‘Widening the Panama Canal’, ’Hong Kong- Zhuhai—Macao Bridge’, the ‘Burj Khalifa’ and the “Shanghai Tower”. Soon, we will be constructing beyond heights of 1000 meter, very long span bridges, very deep underground tunnels under the sea and structures in previously inhabitable spaces.

IHFS welcome research and case studies on these and many more exciting Extreme Engineering projects. IHFS welcomes researchers, architects, planners, builders, constructors, (consulting and structural) engineers, developers, contractors, safety and security experts, financiers, academia, suppliers and all professionals involved in the field of extreme engineering to meet in Singapore at the international Conference on 26-28 February 2019. Singapore is where the world meets and where passion is made possible.

International Federation for High Rise Structures (IFHS)p: 91-80-26614325e: [email protected]: www.ifhsassociation.org

ACCE(i)-AWARDs 2018 CAll FoR nominAtions

Association of Consulting Civil Engineers (India) has been conferring yearly Awards for outstanding work of persons / organisation / firms, for excellent / innovative design, excellence in the field of construction, significant contributions to civil engineering consultancy, best software package in civil engineering, best publication in civil engineering etc.

Some of the categories for the awards are for outstanding design of Industrial Plant/ Structure using special /advance /new Technique, specialized foundations; or moving structures or design of facilitating structure; for Innovative design of Structure other than Industrial structure; for Excellence

The Indian Concrete Journal February 2018 7


in construction of Industrial Structure; for Excellence in construction of High Rise Building; for Excellence in Construction of Highway Project; for Excellence in construction in the Field of Civil Engineering ; for Best Software Package for Civil Engineering Consultancy; for Best Publication (Book) in Civil Engineering (Useful to Consultants); for Effective Use of Pozolona or Blended Cements in Design & Construction of Civil Engineering Projects; for Best Construction by Budding Company of India; for Effective Use of Construction Materials/Systems in Construction Resulting In National Savings; for Best Use of Formwork In Civil Engineering; for Appropriate use of Construction Chemicals Epoxy for rehabilitation/ Retrofitting of Civil; Engineering Structure by Consultants; for Innovative Structural Design by Upcoming Structural Designer; for Excellence in the use of Rectangular/Square Hollow Sections In Steel Structures; for a person from the field of engineering, preferably, civil engineering; and more. For more details get in touch with ACCE

Association of Consulting Civil Engineers (ACCE)p: 080-26770365, 26770366e: [email protected]: www.acce.in

PCERF-FERRoCEmEnt AWARD ComPEtition 2018Ferrocement is a really old technology which emerged before cement concrete, i.e. in 1847. But due to ease in construction, reinforced concrete superseded and became popular. However when curved shapes and light weight construction is required engineers and architects have no other alternative than ferrocement. Ferrocement Society of India has its head office in Pune. Society is promoting this technology as this also gives solutions for earthquake resistant houses. Ferrocement is one of the components in GREEN housing concepts. PCERF had called for the best ferrocement structure entries from Indian constructors. The following entries were received for consideration.

1. Deep Ferrocement, Pune

2. Ferrotechnologies, Chenganassery, Kerala

3. P.J. Varghese, Kerala

4. Owen Waldschlagel, Intact Structures, USA

The bungalow (picture shown here) owned by P.J. Varghese and designed and constructed by Biji John, Ferrotechnologies, Chenganacherry, Kerala was selected by the juries for the PCERF-Ferrocement award 2018. The award will be given in a special function in Constro exhibition, on 20 January 2018. The participants will be felicitated by the Chief Guest

Dr Rajendra Pawar, Director General, MERI, Nashik. He said Government of Maharashtra is always open to new building materials and technologies which are directly helpful for mass housing, rural development and farmers. He stressed the need of skill developed labour in construction industry. Chandramohan Hangekar, President Ferrocement Society said the ferrocement technology is directly useful in rural area as the people are already using soil bamboo combination. After the overwhelming response to the Constro exhibition, Girish Sangle, Hon Secretary, Ferrocement Society said 25 architectural and engineering colleges will be tied up with this mission of ferrocement technology dissemination.

– Ferrocement Society Press Release

innovAtivE sustAinABlE stRAtEgiEs FoR ComPEtitivE mARKEting FoR CEmEnt AnD ConCREtE

NICE Consultants is organizing a two-day workshop for cement industry professionals on “Innovative strategies for Cost Reduction in Cement Industries” from 20- 21 February, 2018. The program will be held at Green Park Hotel, Begumpet, Hyderabad. The program will focus attention of participating professionals on cost reduction measures by adoption of new / emerging technologies, operation and maintenance of plant and equipment, maintenance strategies for improving uptime and productivity, energy audit / conservation, environmental issues, life cycle assessment and sustainable development. A unique feature of the program will be technical presentations by experts as well as from leading global technology suppliers.

New India Cement Engineering Consultantsp: 91-9958998059, 9958998065e: [email protected] , snmkhan@hotmail



EvEnt REPoRtThe following are the recommendations of the 22nd Annual Convention & Seminar on “Housing for All in Urban & Rural Areas” held by the Indian Buildings Congress at NDMC Auditorium on 22-23, December 2017

1. Pradhan Mantri Awas Yojna (PMAY) launched in June 2015 aims to build 20 million houses in urban areas for EWS, LIG and BPL categories by 2022. The progress so far upto March 2017 has only been 41,000 houses. This calls for revisiting the PMAY for removing the hurdles in implementation. It is desirable that all schemes district wise, are led preferably by the Collector of the district, with concerned central and state government officials working in close coordination. This could ensure speed-ier implementation of the projects.

2. Housing is a state subject which involves close interac-tion amongst Central, State and Local Governments and entails cross sectional co-ordination of social, economic, environmental and governance system. The important aspect to be dealt with include, access to lands, regula-tory and planning control and clearances, slum devel-opment and rehabilitation, housing upgradation and renewal, infrastructure services development, outsourc-ing processes and managing financial resources through loans, subsidies and grants.

3. Partnerships and Community Participation are new mod-els, which should be community led and participatory based on principle of equitable allocation of resources. The aim is to create inclusive housing which provides everyone with a house with linkages, livelihood, water, electricity, toilets and security. The community acts as the core actor in the shelter process, controls the money and resources, promotes a broader concept of upgradation and improvement of existing structures and optimizes uti-lization of urban land with focus on brown fields, capac-ity development, community empowerment including PPP mode of implementation.

4. According to TCPO estimate, the land requirement for 20 million houses is around 100,000 hectares of addi-tional land. Accordingly, a GIS based inventory and to-tal station survey of all potential lands suitable for hous-ing redevelopment and settlement should be prepared including the master plan of the concerned city. Land parcels in government possession, unused lands in SEZs, potential agricultural lands etc. should be explored for housing development, and digitized inventory prepared for all townships for long term planning and develop-ment.

5. Compulsory reservation of lands for social housing for BPL/EWS/LIG is required. For making social housing af-fordable and viable, there is need to subsidize the cost of land. New options of access to lands such as land pool-ing, transferable development right (TDR), land banking and mixed land use should be considered.

6. Out of total housing shortage of 20 millions, about 80% dwelling units comprise of dilapidated and congested houses under PMAY thus involving urban renewal, up-

gradation, regularization, redevelopment, rehabilitation and retrofitting. About 25% of the population in major cities of the country resides in unauthorized colonies, of-ten built on public land lacking basic infrastructure, open spaces and facilities. As such rather than greenfield de-velopment, it is necessary to focus on upgradation and redevelopment of old, dilapidated housing areas, urban-ized villages and unauthorized colonies.

7. Government needs to develop proper schemes for such rehabilitation and redevelopment of viable unauthorized schemes and implement them with a view to achieve the target by 2022. Additional Floor Area Ratio (FAR) and Floor space Index (FSI), can make slum development projects more financially viable. Slum rehabilitation projects can have a free sale component for market sale so as to cross-subsidize the project.

8. National Urban Rental Housing Policy (NURHP) 2015 was formulated with a view to provide social Rental Housing especially for BPL, EWS and LIG groups, and also shelters for hostels, PG, dormitories etc. Under the scheme, urban local bodies are to distribute rent vouch-ers for identified families eligible for such concession.

9. Government should construct large scale ‘Night Shelters’ on vacant Govt. lands along with community toilets for below poverty line population and pavement dwellers to provide them with roof over their head till such time they are provided housing under various schemes over a pe-riod of time. This could be done by State Govts. through their own funds in all townships.

10. A critical concern in redevelopment, re-densification, and slum rehabilitation projects relates to infrastructure services such as water supply, sewerage, power and oth-er services which are under severe stress. These require strategic interventions such as (i) Preparation of services plan of redevelopment, slum rehabilitation and regular-ization (ii) Water recycling, water conservation and en-ergy efficiency (iii) Checking of leakages, theft and trans-mission losses.

11. There is an urgent need to simplify procedure and pro-cesses for land acquisition and conversion of land use to encourage mass Housing Projects. Authorities should introduce single window online clearance mechanism which would facilitate speedy clearance of projects.



12. To encourage construction of appropriate infrastructure by private sector, Govt. should grant ‘infrastructure’ sta-tus to construction industry and accord further suitable tax incentives.

13. Lot of innovative Building materials and Technologies like GFRG etc. are coming up. There is need for Government to give financial support for their manufacturing units.

14. Government could finance and make available exten-sively semi developed land with roads, water and elec-tricity to prospective private builders to construct mass housing projects.

15. Developed countries solved their housing problem, by industrializing the building industry. Industrialization of building involves the mass manufacture of components in factory and their erection at site using mechanical ap-pliances. It needs total integration of production, trans-portation, erection through competent management us-ing mechanization.

16. Functional design, rationalization of specifications, op-timal structural design of buildings with integrated ap-proach of mass housing construction would optimize cost. Mass housing technologies like Monolithic concrete construction system using plastic/aluminum formwork, Expanded polystyrene core panel system, Industrialized 3-S system using cellular light weight concrete slabs and precast columns, light gauge steel framed structures, fac-tory made fast track modular building system, etc. are feasible solutions for mass Housing.

17. Public Private Partnership (PPP) mode is in nascent stage as far as low cost housing projects are concerned. The Govt. could consider creating a robust PPP model to at-tract private capital for low cost housing.

18. Housing finance is a long term investment and asset li-ability mismatch is a major problem for housing finance. Such financing of the social housing besides government grants and private sector funds can be supplemented by mortgage guarantee fund, social housing fund, mi-cro financing, pension funds, real estate mutual funds and the like. National Housing Bank could be allowed to raise funds through capital gains bonds. Housing Fi-nance companies could provide funds at subsidized in-terest rates. Real Estate Investment Trusts (REITs) and In-frastructure Investment Trusts could be alternate sources of funding.

19. Training Institute may be opened by State Govt./PWD’s at District Headquarters under skill development schemes of GOI. On the job training may be given by PWD to artisans at their construction sites free of cost at various locations to augment availability of skilled labour for ur-ban and rural housing.

20. In respect of Rural Housing: Pradhan Mantri Awas Yo-jna – Gramin caters for providing financial assistance to one crore dwelling units from 2016-17 to 2018-19. Implementation mechanism of Gramin Yojna needs to be streamlined and strengthened, including proper in-put from National Technical Support Agency at national level.

21. Provision of Urban Amenities to Rural Areas (PURA) ca-ters to providing rural hubs physical connectivity by pro-viding roads, energy self sufficiency, drinking water and communication network. This has to be carried out in an integral way on a war footing so that economic activities including housing get properly facilitated.

22. There is need to enhance the availability of financial re-sources for procuring lands for BPL as well as APL house-holds through additional instruments – grants as well as interest subsidy based, to address different needs of rural households.

23. State Govts. should make available adequate number of home plots in terms of locality and size to meet basic livelihood needs of the family in rural areas, preferably to be made available on annual lease hold basis.

24. Strengthening of Panchayati Raj Institution through ap-pointment of secretary cum CEO, and engineers through State Public Service Commission as suggested by Expert Panel headed by Sumit Bose needs to be carried out.

25. NGO’s have been playing an important role in facili-tating rural housing through promotion of community based process and alternative technologies. NGO’s could be professionally engaged to facilitate infrastruc-ture and housing development.

26. Central Govt. could support the States to prepare State Housing and Habitat Plan and schemes giving a road-map of actions required. A well equipped PMU is need-ed to be set up at the state and district level to lead the process of implementation of rural habitat schemes.

27. In conclusion, for realizing the goal of ‘Housing for All’, public sector and private sector both, need to work to-gether. Central and State Governments should proac-tively develop and make adequate land and infrastruc-ture available for housing at a fast pace. The building industry must be modernized, technologically upgraded, made environmentally conscious and globally competi-tive, to meet the housing challenge in the country.

IBC Press Release


TECHNICAL PAPER

Investigation of the behaviour of shear-critical reinforced concrete columns

Komathi Murugan and Amlan K. Sengupta

INTRODUCTION Damage to structural members is inevitable during an earthquake. The vulnerability of a primary gravity load carrying member such as a column in a multi–storeyed building, may even lead to collapse of the entire structure. The design of such a member should be done with more accuracy, in addition to adoption of a good construction practice. Conventional design is based on the strength of a member. But for members of a lateral load resisting system, design based on estimation of deformation and post–yield ductility is preferred. This needs modelling of the behaviour of each member and subsequent non–linear analysis of the building. A pushover analysis provides the modes of failure and sequence of failure of the members. The present study demonstrates the modelling of the behaviour of a tested shear–critical short column designed for gravity loads,

The Indian Concrete Journal, February 2018, Vol. 92, Issue 2, pp. 10-24.

A comparative study of the strength and behaviour of a shear–critical cantilever column specimen based on experimental, analytical and numerical investigations is presented in this paper. The specimen was tested under monotonic lateral load. Its strength was predicted using codal provisions and a strut–and–tie model. Its lateral load versus drift behaviour was analytically predicted using a piecewise linear model based on truss analogy, and numerically simulated using three–dimensional (3D) and one–dimensional (1D) finite element (FE) models. Correlation between the tested, predicted and simulated results is discussed in this paper. The study verifies whether the selected methods of analysis give reasonable estimates of strength and behaviour. It was found that the codal expressions under–estimate the shear strength of a short column failing under diagonal crushing. A strut–and–tie model can predict the strength accurately, provided the strut width and softening of concrete are considered appropriately. The behaviour was accurately predicted by a 3D FE model. The analytical model based on a linear truss analogy and the corresponding 1D FE model could trace the behaviour accurately up to cracking. They provided reasonable estimates of the drift at ultimate. These models are suitable in professional practice for a pushover analysis of a building.

Keywords: Column; reinforced concrete; strut–and–tie model; shear hinge property; truss analogy.

similar to those present in a reduced height open ground storey or as captive columns in a typical multi–storeyed residential building in urban India.

A comparative study of the strength and behaviour of a reinforced concrete (RC) column specimen based on experimental, analytical and numerical investigations is presented in this paper. The analyses check how well the predictions match with the actual strength and behaviour.

The vertical cantilever specimen was tested under monotonic lateral load generating single curvature bending. Its strength was predicted using codal provisions (based on Indian standard and selected international standards) and a strut–and–tie model. The lateral load versus drift behaviour was approximately predicted using a piecewise linear model based on truss analogy, suitable for professional practice. The steps of calculations are explained. Further, numerical simulations were done using three–dimensional


TECHNICAL PAPER

(3D) and one–dimensional (1D) finite element (FE) models. Correlations between the tested, predicted and simulated results are discussed. The study verifies whether the selected methods of analysis give reasonable estimates of strength and behaviour. The methods can be applied in the pushover analysis of a multi–storeyed building, with shear–critical columns.

LITERATURE REVIEW Selected studies related to testing and modelling of shear–critical RC column specimens are summarized as follows.

Experimental Investigations

Capacity of concrete to resist shear before the formation of diagonal crack, was found to be a major factor of the shear capacity of a column [1]. It was observed that, not just the spacing, but proper configuration of the transverse reinforcement (ties) governs the ductility in the behaviour under lateral load [2]. The shear strength was found to be related to the displacement ductility demand in columns with light transverse reinforcement [3]. At large drifts, column specimens with axial loads collapsed even under decreasing lateral loads [4]. Shear deformations were found to be the major fractions of the total lateral deformations in short columns [5].

Analytical Models

An equation to compute the shear strength of a column considering flexural ductility demand was proposed [6]. However, the shear strength was found to be mildly sensitive to the ductility demand [7]. A new model for shear strength of columns based on truss analogy, gave better predictions if about two–thirds of the strength contributed by the ties was considered effective [8].

The models for predicting the shear responses of RC members are as follows.

1. Truss analogy: This approach provides an estimate of the shear deformation of a frame member after diagonal cracking [9].

2. Strut–and–tie model: The model was developed for the ultimate state to predict the shear strength of a short member or disturbed (D) region subjected to patch loads [10]. This model has been extended to predict deformation also.

3. Softened truss model: The model was proposed for wall type members such as shear walls and

large girders [11]. Subsequently, the approach was extended to incorporate the Poisson’s effect under two−dimensional state of stresses and was termed the softened membrane model [12].

4. Modified Compression Field Theory (MCFT): This model was also proposed for wall type members [13]. Later, the model was extended for finite element analysis under cyclic loads (disturbed stress field model [14]).

An analytical procedure developed based on a two–dimensional nonlinear model provided conservative estimates of the lateral strength, deformation and failure modes of columns than those obtained from experiments [15]. An approach to quantify lateral drift was used to develop an idealized backbone model for the key damage states of an existing column [16]. A simplified monotonic shear model, independent of flexural yielding of column, was developed based on MCFT [17].

Numerical Simulations

Finite element analysis of RC columns of single storey height was conducted to study their non–linear behaviour [18]. A similar analysis showed that the column capacity increased by 80 percent with increase in the column size by two inches on all sides [19]. Pushover analyses of columns using hinge properties developed based on experimented data were corroborated with test results [20].

RESEARCH SIGNIFICANCEFrom the literature review, it was noted that experimental data and corresponding analytical models or numerical simulation to predict the behaviour of shear–critical RC columns under lateral load are limited. Most of the previous studies were on columns showing ductile behaviour. The importance of the present study lies in the investigation of short non–ductile columns failing in shear. The models available for prediction of the response of columns failing under shear need to be applied to develop shear hinge properties for columns in a pushover analysis of a building.

The analytical and numerical methods reported in this paper demonstrate the application of the selected models. The analytical model and the 1D FE model are based on the simplified approach of modelling piecewise linear behaviour. This is suitable for professional practice in the pushover analysis of a building with many columns in each storey. A rigorous 3D FE model is also presented for demonstration of analysing an individual column.


TECHNICAL PAPER

DETAILS OF COLUMN SPECIMEN

Specimen Geometry and Reinforcement

A typical column in an open ground storey of a building on raft foundation has approximately a fixed–fixed configuration, with the point of contra–flexure under lateral load at its mid–height. The geometry of the selected specimen was derived considering the bottom half of such a column. The specimen consisted of a cantilever column

Figure 1. Details of the specimen (dimensions in mm)

of rectangular section, with a foundation block to generate fixity at base (Figure 1).

The specimen was made shear–critical by selecting a suitable height–to–depth ratio and designing the amounts of lateral and transverse reinforcements in the test region. The detailing of reinforcement aimed to control any premature failure. The details are as follows.

The height was decided based on the bolt holes of the reaction wall available in the laboratory. The width of the column section was selected as 230 mm, a common practice in Indian construction to flush the column with walls above. The depth of the column section was then arrived to maintain the shear span–to–depth ratio at around 3.0. The plan dimensions of the foundation block were selected based on the spacing of the anchor holes of the strong floor in the laboratory. The thickness of the block ensured adequate anchorage of the column longitudinal bars, to avoid any slip of the bars.

The amount of longitudinal reinforcement in the column ensured high flexural strength. The lateral load corresponding to flexural yielding was higher than the load corresponding to shear failure. The ties were spaced at 230 mm in the central test region as per the codal provision for conventional design for gravity loads. However, closely spaced ties were provided at the potential plastic hinge region above the foundation block to avoid flexural failure. The close tie spacing was continued inside the foundation block to reduce the effect of any yield penetration of the longitudinal bars. Closely spaced ties were also provided at the top of the column near the loading zone to avoid any local crushing of the concrete.

For economy, the reinforcement in the foundation block was limited to the minimum amount of skin reinforcement in each surface. Four vertical holes were made in the block by inserting plastic pipes of suitable diameter at the required locations of the rods for anchoring. Spiral reinforcements were provided around the holes to avoid any local failure in these regions due to the bearing of the anchoring rods.

Properties of Materials

Ordinary portland cement, fine aggregate (river sand), coarse aggregate of nominal sizes 12 mm and 20 mm, and water were the constituents of standard concrete used to cast the specimen. Mineral or chemical admixtures were not added. High yield strength deformed steel bars were used as reinforcement. The cubes cast during specimen preparation were tested simultaneously on the day of testing of the specimen, to ascertain the compressive strengths of concrete. The yield strengths of reinforcements were obtained from


TECHNICAL PAPER

tension tests of coupon specimens. The properties of the materials used in the specimen are shown in Table 1.

EXPERIMENTAL INVESTIGATION

Preparation of Specimen

After bonding the strain gauges, the reinforcement cages were fabricated for the column and foundation block, and the formwork was prepared. The concrete was cast in two stages, with one batch for each stage. First, the concrete in the foundation block was cast. After its initial setting, the column concrete was cast. After 24 hours, the specimen was demoulded and subsequently cured using wet burlap. For every batch of casting, 150 mm cubes were cast for evaluating the compressive strength, and these were cured alongside the specimen.

Details of Test Set–Up and Instrumentation

The specimen was anchored to the strong floor using threaded rods through the holes provided in the foundation block (Figure 2a). During the initial loading, the foundation block cracked. It was then prestressed externally both in the horizontal and vertical directions using stiffening beams and high strength threaded rods, to arrest further cracking. Linear variable displacement transducers (LVDTs) were used to measure deformations or displacements (Figure 2b). Two rosettes of diagonal LVDTs (1 and 2, 3 and 4) were placed to measure the average shear strains across the depth of the specimen. LVDTs 5 and 6 measured axial strains in the potential plastic hinge region. LVDTs 7 to 10 monitored the sliding, uplift and rotation of the foundation block. LVDT 11 measured the drift.

Grid lines were marked in the test region at 50 mm spacing, both horizontally and vertically, to monitor the growth and measure the inclination of diagonal cracks. Strain gauges

Table 1. Strengths of materials

Material Measured strength (N/mm2)

Compressive strength of concrete cubes

Column 29.7

Foundation block 40.5

Yield strength of reinforcing bars

25 mm φ longitudinal bars 411.5

6 mm φ transverse ties 371.3

8 mm φ confining bars in foundation block 417.8

Figure 2a. Details of test set-upFigure 2b. Positions of LVDTs and strain gauges 1–11 → LVDTsi–x → strain gauges (dimensions in mm)


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appearance of shear cracks. There was limited widening of the shear cracks with increase in load. The specimen failed suddenly with crushing of the concrete strut. The load versus drift curve for the column is shown in Figure 3a. This curve was corrected for the effects of cracking (during the initial loading) and minor sliding of the foundation block. The behaviour was not ductile as the yielding of ties was

were installed at desired locations of reinforcement to monitor their yield. Gauges i to viii measured the strains in the ties. Pairs of gauges (ii and iii, v and vi) were provided for both the legs of the respective ties.

The column was subjected to bending about the major axis. The lateral load was applied over a patch of size 300 mm × 300 mm, using an actuator of 100 T capacity fixed to the reaction wall. Axial load was not applied to avoid increasing the shear strength of concrete in the column. The test protocol was displacement–controlled. Crack widths were measured using a hand–held microscope. The test was terminated after substantial drift beyond the peak load.

Behaviour of Specimen

The specimen failed in shear with the formation of diagonal tension cracks (Figures 2c and 2d). The sequence of behaviour is as follows. When the specimen was loaded laterally, thin flexural cracks formed at the tension face of the column. With continued loading, the flexural cracks did not widen or propagate. On further loading, shear cracks appeared rapidly in the test region, which led to the formation of a diagonal strut. The drift increased at a faster rate after the Figure 3a. Lateral load versus drift

Figure 2c. Formation of shear cracks at lateral load = 124.8 kN Figure 2d. Crushing of concrete at failure


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delayed due to dowel action and aggregate interlock. The variation of calculated average shear stress versus average shear strain in the test region is shown in Figure 3b. The reduction in stiffness after the formation of shear crack is evident. The local variations of strains in the ties as recorded by the strain gauges are shown in Figure 3c, with respect to the applied lateral load. The strain in each tie varies along its length. The strain gauge marked SG vii showed yielding of the respective tie closer to peak, based on its proximity to a crack.

ANALYTICAL INVESTIGATIONThe analytical prediction of strength and behaviour of the specimen using existing methodologies is discussed below. The methods were selected based on their applicability to a shear–critical column specimen.

Analysis of Shear Strength

A cantilever column with a concentrated lateral load applied at the top, has uniform distribution of shear force throughout its height. For a shear–critical specimen, the ultimate lateral load gives the shear strength of the column. The shear strength was predicted using two methods: codal provisions and a strut–and–tie model.

Codal ProvisionsThe shear strength is composed of two components: concrete contribution and tie contribution. Expressions available in selected standards such as IS 456 : 2000 (SP : 24 − 1983), ACI 318 : 2014, EN 1992–1–1 : 2004, CSA A23.3 : 2004 and NZS 3101–Part 1 : 2006 were used to calculate the shear strength [21–26]. The material safety factors or resistance factors were not considered in the computations.

Strut–and–Tie ModelThis method is applicable in this study considering the prominent formation of a single strut beyond shear cracking. The model was developed based on the observed cracks and the strut formed in–between them, in the specimen. The idealized truss with the diagonal struts (DS), vertical struts (VS), horizontal ties (HT), vertical ties (VT) and nodes (N) is shown in Figure 4a.

Figure 3b. Average shear stress versus average shear strain

Figure 3c. Lateral load versus strain in ties

Figure 4a. Idealized truss – assembly of struts and ties(dimensions in mm)


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The idealizations involved the positions, widths and inclination of struts, positions of the ties and nodes. The struts marked VS1 and VS2 were placed along the compression face of the column, with the corresponding nodes at the centroid of compression longitudinal reinforcement. The strut marked DS2 is the critical strut in the test region, which dictates the non–linear shear behaviour of the column. The inclination of this strut was selected based on the experimental observation. The struts marked DS1 and DS3 were placed in the regions of closely spaced transverse bars. These struts were considered to be linear elastic. The ties marked VT1, VT2 and VT3 were placed along the centroid of tensile longitudinal reinforcement. The ties marked HT1, HT2 and HT3 were placed along the transverse bars located at the nodes of the diagonal struts. The intermediate transverse bars were grouped in the modelled ties.

The shear strength of the specimen was computed based on the compression capacity of Strut DS2. This was selected based on the observation of diagonal crushing of concrete along the strut. The width of Strut DS2 is highlighted in Figure 4a. A softening coefficient ‘βs’ was introduced to

consider reduced compression capacity of the concrete in the strut, based on the observed bulging of the strut due to orthogonal tensile strain. The value of βs = 0.2 was found to be suitable in predicting the strength. The value is substantially low for the strut width selected based on the observed crushing.

Corresponding to the failure load, the calculated force in Tie HT2 based on nodal equilibrium was more than that to yield the tie. This is because part of the compression in the strut is balanced by the dowel action of the high amount of longitudinal reinforcement at the crack, and the friction due to aggregate interlock (Figure 4b), which were not considered in the model. The capacity of the concrete in the nodal zones around the nodes N3 and N4 was found to be higher than the force corresponding to the failure load.

Analysis of Lateral Load versus Drift Behaviour

The behaviour plot for the specimen is shown schematically in Figure 5. The critical points of the behaviour plot as observed in the specimen are onset of flexural cracking (‘ΔMcr’, ‘PMcr’), onset of diagonal shear cracking (‘ΔVcr’, ‘PVcr’) and ultimate strength (‘Δu’, ‘Pu’). These can be connected to form an approximate piecewise linear model of the actual behaviour. Since the tensile longitudinal bars did not yield, the point of flexural yielding was not considered. Due to the dowel action and friction, the yielding of ties in the test region was delayed till the crushing of the strut. Hence, the point of yielding of the ties is not considered separately.

The loads corresponding to the critical points were computed based on linear elastic theory for cracking, and using the strut–and–tie model at ultimate as explained earlier in the analysis of shear strength. The drifts corresponding

Figure 4b. Mechanisms of shear resistance after diagonal cracking

Figure 5. Piecewise linear model showing critical points of behaviour curve


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to these loads were computed using linear elastic theory up to cracking, and using curvature diagram for flexural component and truss analogy for shear component, at ultimate. The flexural and shear components were added to obtain the total drift. The steps involved in determining the loads and drifts corresponding to the critical points, are explained further.

Onset of Flexural Cracking 1. Find the moment corresponding to flexural cracking of the section ‘Mcr’ (Eq. 1).

...(1)

where, ‘fcr’ = Modulus of rupture of concrete (= 0.7√fcm, as per IS 456 : 2000)‘fcm’ = Mean compressive strength of concrete in the column‘fv’ = Compressive stress due to axial load (P/Ag = 0, for the studied column specimen)‘P’ = Axial compressive load‘Ag’ = Gross cross−sectional area of the column'Ig’ = Gross moment of inertia of the section ‘yt’ = Distance of the extreme tension fibre from the neutral axis

2. Find the curvature at cracking ‘ΦMcr’ using ‘Ig’ (Eq. 2). The modulus of elasticity of concrete ‘Ec’ can be taken as 5000√fcm as per IS 456 : 2000, in absence of any measured data. This is typically lower than the actual modulus of the concrete in a specimen.

...(2)

3. Find the lateral force corresponding to flexural cracking ‘PMcr’. For a vertical cantilever column, the flexural cracking initiates at the base. The cracking force is equal to the cracking moment divided by the column height ‘H’.

4. Calculate the flexural component of the drift ‘Δf ’ corresponding to ‘ΦMcr’ based on the expression for horizontal drift of a column under lateral load.

5. Find the shear component of the drift ‘Δv ’ corresponding to ‘PMcr’ using the uncracked shear stiffness ‘kv1’ of the section (Eq. 3). Here, ‘bw’ is the width of the section at neutral axis, ‘D’ is the depth of the section and ‘f ’ is a factor to consider non–uniform distribution of shear stress across the depth of the section. For the present column specimen of rectangular section, f = 1.2.

...(3)

6. Find the total drift corresponding to flexural cracking ‘ΔMcr’ by adding the flexural and shear components.

Onset of Shear Cracking1. Find the shear force corresponding to shear cracking of the section ‘Vcr’ (Eq. 4). This is estimated based on the maximum cracking shear stress ‘τcr’ corresponding to the principal tensile stress at neutral axis reaching the cracking value (Eq. 5).

...(4)

...(5)

The tensile strength of concrete ‘fct’ can be taken as 0.2√fcm, in absence of any measured data. This is a lower estimate considering the variability of strength of concrete under direct tension. A is the fractional area of the section adjacent to neutral axis. is the first moment of that area about the neutral axis. For a vertical cantilever column with a lateral force at the top, the corresponding lateral force ‘PVcr’ is equal to the shear force ‘Vcr ‘.

2. Calculate the flexural component of the drift ‘Δf ’ corresponding to ‘PVcr’ based on the curvature diagram along the height of the column, and using the second moment–area theorem or conjugate beam method.

3. Find the shear component of the drift ‘Δv ’ corresponding to ‘PVcr’ using the uncracked shear stiffness ‘kv1’ as before. A reduced sectional depth shall be considered (equal to depth of shear area ‘jd ’). The total drift ‘ΔVcr’ is the sum of the two components.

At Ultimate Strength1. Find the lateral force ‘Pu’ corresponding to the shear strength of column, which is calculated using a strut–and–tie model.

2. Calculate the flexural component of the drift ‘Δf ’ corresponding to ‘Pu’ based on the curvature diagram as before.

3. Find the shear component of the drift ‘Δv’ corresponding to ‘Pu’ using the post–cracking shear stiffness ‘kv2’ (Eq. 6).


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...(6)

...(7)

where,‘ρv’ = Ratio of transverse reinforcement (= Asv/sv.bw)‘Asv’ = Total area of the legs of ties‘sv’ = Vertical spacing of the ties ‘α ’ = Angle between the concrete strut and the longitudinal axis of the column‘m’ = Short–term modular ratio (= Es/Ecu)‘Es’ = Modulus of elasticity of reinforcement‘Ecu’ = Secant modulus for concrete at strength (2fc’/ε0)

Here, ‘fc’’ is the compressive strength of concrete cylinder,

which was considered to be 0.8 times the measured cube strength and ‘ε0’ is the compressive strain corresponding to peak stress, which is around 0.002. The angle ‘α’ can be found using Equation 7, based on the level of axial stress. For the present study, it was taken as the actual crack angle observed from the experiment.

The above expression of shear stiffness is applicable before the yielding of the ties. Since the ties yielded beyond the crushing of the strut, the above linear expression was used to compute the drift at ultimate. To have a precise estimate of the drift in a column where the ties yield substantially, the non−linear stress versus strain behaviour of the strut under compression, needs to be considered. Note that a reduced secant modulus of concrete should be considered at failure.

4. Find the total drift 'Δu’ as the sum of the flexural and shear components.

NUMERICAL INVESTIGATIONSThe column specimen was numerically simulated to predict its non–linear lateral load versus drift behaviour using two following FE models with available commercial softwares.

1. A 3D model

2. A 1D model

3D Model

A 3D model of the column is necessary to consider the effects of variation of shear stress across the column section, softening of concrete under compression due to diagonal

tensile strain, tension stiffening of cracked concrete and the yielding of the ties. However, such a model is not suitable in an analysis of a building with many columns in each storey.

The concrete was modelled with 8–noded elements. The continuum model however cannot capture the discrete cracking and the aggregate interlock. The longitudinal and transverse bars were modelled with 2–noded elements. It is to be noted that, the link elements for the vertical bars cannot model the dowel action. An optimum mesh density was used to arrive at adequately converging results. 8–node elements were used to model the steel loading plate. This was necessary to avoid stress concentration and subsequent local crushing of concrete due to application of load (Figure 6a). The base of the column was considered to be fixed.

Figure 6a. Numerical simulation – three-dimensional model Isometric view


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The non–linearity in the behaviour was incorporated through the material stress–strain curves. The constitutive relationships used to model the ascending and descending branches of the stress–strain ( σc−εc ) curves for concrete under compression are given by Equations 8 and 9, respectively [27].

For εc ≤ ζ ε0 : ...(8)

For εc > ζ ε0 : ...(9)

The descending branch was continued up to a level of one–fifth of the peak stress and then maintained constant to avoid numerical instability. It is to be noted that the value of softening coefficient ‘ζ ’ needs to be specified a priori, and it stays constant throughout the analysis. Two models were used for concrete in compression (Figure 6b). The concrete in the test region with widely spaced ties was considered to soften. A value of ζ = 0.5 was found to be suitable based on trials. The concrete in the regions with closely spaced ties was considered to be intact. A value of ζ = 0.9 was used to consider the size effect. Any confinement of concrete due to rectangular ties was neglected.

The constitutive relationships used to model the ascending and descending branches of the stress–strain ( σt−εt ) curves for concrete under tension are given by Equations 10 and 11, respectively [27].

For εt ≤ εcr : ...(10)

For εt > εcr : ...(11)

The descending branch was continued up to a level of one-half of the peak stress. Here, ‘σcr’ and ‘εcr’ are the tensile cracking stress and strain respectively, of concrete. This model (Figure 6c) was used for concrete in tension, along with both the compression models mentioned above. The pre-cracking portion is linear and the post–cracking is concave with a drastic drop of strength. For the reinforcements,

Figure 6b. Numerical simulation – three–dimensional model. Material models for concrete in compression

Figure 6c. Material model for concrete in tension

Figure 6d. Contour of principal compressive stress at peak load of 195.8 kN (values in N/mm2)


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bi-linear elasto–plastic curves were used, with the measured yield strengths.

The contours of the principal compressive stress and principal tensile strain are shown in Figures 6d and 6e, respectively. The respective vector plots are shown in Figures 6f and 6g. The plots reveal the formation of strut, with substantial diagonal tension. The lateral load versus drift curve was obtained.

Figure 6e. Contour of principal tensile strain at peak load of 195.8 kN (values in mm/mm)

Figure 6f. (Left) Vector plot showing distribution of principal compressive stress at peak load of 195.8 kN. Figure 6g. (Right) Vector plot showing distribution of principal tensile strain at peak load of 195.8 kN

Figure 7a. Numerical simulation – one–dimensional model. Locations of hinges in the column model

Figure 7b. Moment–curvature values used to develop flexural hinge properties


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1D Model

The 1D model of the column is suitable in an analysis of a building with many columns in each storey. A single frame element is used to model the column. The non–linearity in the behaviour was incorporated through the properties of lumped plastic hinges.

Two types of hinges were assigned at the potential plastic regions: a flexural (or moment) hinge at the fixed base and a shear hinge at the mid–height of the frame element (Figure 7a).

The moment versus curvature and shear force versus shear deformation curves used to define the properties of the hinges are shown in Figures 7b and 7c, respectively. These curves were based on the piecewise linear analysis of the

lateral load behaviour of the column using truss analogy, as explained earlier.

A displacement–controlled non–linear static pushover analysis was carried out with incrementing the drift in steps. The progression of hinge formation was observed with increase in lateral push (Figure 7d). First, a flexural hinge formed at the base, which did not progress beyond the level of flexural cracking. Next, a shear hinge formed at mid–height of the element. The output pushover curve showed the simulated lateral load versus drift behaviour.

RESULTS AND DISCUSSIONSThe results obtained from the experimental, analytical and numerical investigations are presented under two headings.

1. Comparison of shear strength of the column section

2. Comparison of lateral load versus drift behaviour of the column

The data measured from the various instruments are not provided for brevity.

Table 2. Comparison of prediction of shear strengthMethod Code or model Shear strength (kN)

Experimental 200.9

Analytical

IS 456 : 2000 151.5ACI 318 : 2014 153.7EN 1992–1–1 : 2004 41.8CSA A23.3 : 2004 166.6NZS 3101–Part 1 : 2006 160.9Strut–and–Tie Model 199.5

Numerical3D FE model 195.81D FE model 199.5

Figure 7c. Shear force–shear deformation values used to develop shear hinge properties

Figure 7d. Progression of hinge formation with increase in lateral push Figure 8a. Comparison of shear strength


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Table 3. Comparison of critical values of behaviour of the column

MethodLateral load (kN) Drift (mm)

Onset of flexural cracking

Onset of shear cracking Ultimate Onset of flexural

crackingOnset of shear

cracking Ultimate

Experimental 23.0 124.8 200.9 0.5 5.5 18.7

Analytical model 24.4 69.0 199.5 0.5 2.0 15.8

3D FE model 21.4 137.5 195.8 0.7 7.0 13.0

1D FE model 24.4 69.0 199.5 0.5 2.2 15.5

Comparison of Shear Strength

The values of shear strength as per the methods described in the previous section are compared in Table 2 and Figure 8a. The observations are summarised below.

1. The strengths predicted using codal provisions (IS, ACI, EN, CSA, NZS) were lesser than those obtained from the experiment. This is because, the codal expressions for shear strength are conservative and based on flexural–shear failure. However, the failure observed in the specimen was due to diagonal tension cracking throughout the depth of the member, followed by diagonal crushing of the concrete strut. The prediction due to EN code is substantially less as it neglects any concrete contribution (including dowel action and aggregate interlock) when the failure is governed by yielding of ties.

2. The strength predicted using the strut–and–tie model (STM), was very close to the test result. This demonstrates that this method is suitable for a short column dominated by shear behaviour.

3. The strength obtained from the 3D FE model was close to the experimental strength. However, the softening coefficient in the concrete compressive stress−strain curve needs to be precisely evaluated.

4. The strength by the 1D model is similar to that obtained from the strut–and–tie model. This is because it is governed by the shear hinge property calculated based on the strut–and–tie model.

Comparison of Lateral Load versus Drift Behaviour

The lateral load versus drift behaviour of the column, traced based on the analytical method and numerical simulations described in the previous sections, are compared in Table 3 and Figure 8b. The values of drift obtained from the test were corrected for the minor base rotation and sliding. The observations are as follows.

1. At flexural cracking, the values of load and drift predicted by the analytical model and 1D model were comparable to that obtained from the experiment. The 3D model exhibited a closer cracking load, but at a slightly higher drift due to the lower estimate of the modulus of concrete.

2. At shear cracking, the values of load were under-predicted by the analytical and 1D models due to the lower estimate of the tensile strength of concrete. The 3D model showed a higher load. The drift was under–estimated by the analytical and 1D models. These models neglected precise modelling of concrete deformation. The 3D model demonstrated a higher drift.

3. At ultimate, the predictions of the strength by all the models were close. The drift was underestimated by all the models. The analytical and 1D models did not consider the yielding of the ties. The analysis by Figure 8b. Comparison of lateral load versus drift


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the 3D model terminated due to failure from stress concentration.

SUMMARY AND CONCLUSIONSA comparative study of the strength and behaviour of a shear–critical cantilever column based on experimental, analytical and numerical investigations is presented in this paper. The specimen was tested under monotonic lateral load. Its lateral strength was predicted using the codal expressions and a strut–and–tie model. The lateral load versus drift behaviour was predicted using a piecewise linear analytical model and two types of numerical models. The analytical model, based on truss analogy and explicit expressions of stiffness, provides results without any need of iteration.

It is observed that the shear strength of a short column with dominant shear behaviour is under–estimated by the codal expressions. The strut–and–tie method provides a better estimate. The prediction of behaviour is reasonable using a three−dimensional finite element model. The analytical model and the one−dimensional finite element model under–estimated the drift. However, these models are suitable in professional practice for a pushover analysis of a building with many columns in each storey.

References

1. Woodward K.A. and Jirsa J.O., Influence of reinforcement on RC short column lateral resistance, ASCE Journal of Structural Engineering, January 1984, Vol. 110, No. 1, pp. 90–104.

2. Saatcioglu M. and Ozcebe G., Response of reinforced concrete columns to simulated seismic loading, ACI Structural Journal, January 1989, Vol. 86, No. 1, pp. 3–12.

3. Lynn A.C., Moehle J.P., Mahin S.A. and Holmes W.T., Seismic evaluation of existing reinforced concrete columns, EERI Earthquake Spectra, November 1996, Vol. 12, No. 4, pp. 715–739.

4. Nakamura T. and Yoshimura M., Gravity load collapse of reinforced concrete columns with brittle failure modes, Journal of Asian Architecture and Building Engineering, March 2002, Vol. 1, No. 1, pp. 21–27.

5. Li Y.A., Huang Y.T. and Hwang S.J., Seismic response of reinforced concrete short columns failed in shear, ACI Structural Journal, July 2014, Vol. 111, No. 4, pp. 945–954.

6. Priestley M.J.N., Verma R. and Xiao Y., Seismic shear strength of reinforced concrete columns, ASCE Journal of Structural Engineering, August 1994, Vol. 120, No. 8, pp. 2310–2329.

7. Sezen H. and Moehle J.P., Shear strength model for lightly reinforced concrete columns, ASCE Journal of Structural Engineering, November 2004, Vol. 130, No. 11, pp. 1692–1703.

8. Sasani M., Shear strength and deformation capacity models for RC columns, Proceedings of 13th World Conference on Earthquake Engineering, August 2004, Paper No. 1838.

9. Park R. and Paulay T., Reinforced Concrete Structures, A

Wiley–Interscience Publication, New Jersey, 1975, pp. 270–300.

10. Schlaich J., Schafer K. and Jennewein M., Towards a consistent design of structural concrete, PCI Journal Proceedings, May 1987, Vol. 32, No. 3, pp. 74–150.

11. Hsu T.T.C., Softened truss model theory for shear and torsion, ACI Structural Journal, November 1988, Vol. 85, No. 6, pp. 624–635.

12. Hsu T.T.C. and Zhu R.R.H., Softened membrane model for reinforced concrete elements in shear, ACI Structural Journal, July 2002, Vol. 99, No. 4, pp. 460–469.

13. Vecchio F.J. and Collins M.P., The Modified Compression–Field Theory for reinforced concrete elements subjected to shear, ACI Journal Proceedings, March 1986, Vol. 83, No. 2, pp. 219–231.

14. Vecchio F.J., Disturbed stress field model for reinforced concrete: formulation, ASCE Journal of Structural Engineering, September 2000, Vol. 126, No. 9, pp. 1070–1077.

15. Pincheira J.A., Dotiwala F.S. and D’Souza J.T., Seismic analysis of older reinforced concrete columns, EERI Earthquake Spectra, May 1999, Vol. 15, No. 2, pp. 245–272.

16. Elwood K.J. and Moehle J.P., Evaluation of existing reinforced concrete columns, Proceedings of 13th World Conference on Earthquake Engineering, August 2004, Paper No. 579.

17. Sezen H., Shear deformation model for reinforced concrete columns, Structural Engineering and Mechanics, January 2008, Vol. 28, No. 1, pp. 39–52.

18. Amin S.A. and Ahsan R., Finite element modeling of reinforced concrete column under monotonic lateral loads, International Journal of Computer Applications, March 2012, Vol. 42, No. 3, pp. 53–58.

19. Ahmed F.S. and Khan M.A., Finite element analysis of reinforced concrete column under lateral load, International Journal of Engineering Research and Applications, July 2013, Vol. 3, No. 4, pp. 228–231.

20. Sung Y.C., A study on pushover analyses of reinforced concrete columns, Structural Engineering and Mechanics, September 2005, Vol. 21, No. 1, pp. 35–52.

21. ______Indian standard code of practice for plain and reinforced concrete, IS 456 : 2000. Bureau of Indian Standards, New Delhi.

22. ______Explanatory handbook on Indian standard code of practice for plain and reinforced concrete IS : 456 − 1978, SP : 24 − 1983. Bureau of Indian Standards, New Delhi.

23. Building code requirements for structural concrete and commentary, ACI 318 : 2014. American Concrete Institute, Michigan.

24. Eurocode 2: Design of concrete structures – part 1–1: general rules and rules for buildings, EN 1992–1–1 : 2004. European Committee for Standardization, Belgium.

25. Design of concrete structures, CSA A23.3 : 2004. Canadian Standards Association, Ontario.

26. Concrete structures standard part 1 – the design of concrete structures, NZS 3101–Part 1 : 2006. New Zealand Standard, Wellington.

27. Hsu T.T.C., Unified Theory of Reinforced Concrete, CRC Press, U.S.A., 1993.


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Komathi Murugan holds a B.E. in Civil Engineering from SRM Engineering College (affiliated to Anna University Chennai); M.E. in Structural Engineering from College of Engineering Guindy; pursuing her PhD at Indian Institute of Technology Madras. Her research interests include seismic retrofit of reinforced concrete buildings, non-linear modelling and earthquake resistant design.

Amlan K. Sengupta is a Professor at the Department of Civil Engineering, Indian Institute of Technology Madras. His research interests include the behaviour of reinforced and pre-stressed concrete members, seismic analysis and retrofit of reinforced concrete buildings. He is a member of the Sectional Committees on National Building Code and Special Structures, under Bureau of Indian Standards.


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Rebar identification, cover thickness and diameter estimation in reinforced concrete members using

cover meter and GPR techniques

Bhaskar Sangoju, S. Vasanthakumar, K. Ramanjaneyulu and K. Sivasubramanian

INTRODUCTIONConcrete is the most popular and widely used construction material due to its relatively low cost, versatility and adaptability. Experience in the recent past shows that structures made of reinforced concrete (RC) do not provide adequate resistance to aggressive environments, mainly due to inadequate cover to the reinforcement bar (rebar). Rebar corrosion is a major concern that causes extensive damage to the RC structures. For new or freshly constructed RC structure, adequate concrete cover to rebar is essential to protect it from external aggressive chloride or carbon-di-oxide induced environments. For the condition assessment of deteriorated reinforced concrete (RC) structures, it becomes essential to locate the rebars, know the cover thickness and rebar diameter. Also, for core sampling and extraction, rebar location is important to avoid cutting of the reinforcement. A reliable estimation of rebar location, concrete cover


To explore the reliable applicability of Cover meter and Ground Penetrating Radar (GPR), parametric studies have been carried out on the laboratory cast test specimens for the identification of rebar, estimation of cover thickness and rebar diameter. Studies are also carried out on field members for the rebar identification and cover thickness estimation. Laboratory test results reveal that (i) Cover meter is effective in identifying the rebar, estimating cover thickness and diameter for shallow covers up to 40 to 50 mm, (ii) GPR is good in identifying rebar and cover thickness estimation for shallow as well as deep covers even beyond 150 mm. Field studies further reveal that GPR is good and effective for rebar identification and cover thickness estimation in the case of closely spaced/heavily reinforced concrete structural members, when compared to that of Cover meter.

Keywords: Non-destructive testing (NDT); Cover meter; Ground Penetration Radar (GPR); Concrete cover; Rebar diameter.

thickness and rebar diameter, non-destructively, is very much important for the quality assurance of new and for the condition assessment of existing RC structures.

Various non-destructive testing (NDT) techniques such as Rebound hammer, Ultrasonic pulse velocity, Cover meter, Impact echo (IE), Pulse echo (PE), Infrared thermography (IRT), Ground penetrating radar (GPR), Linear polarization resistance (LPR), etc., are being used for the condition assessment of reinforced concrete structures [1-2]. Techniques such as IE, PE, IRT and GPR are sometimes called as advanced NDT techniques, because, they greatly increase the speed, utility and allow the near continuous acquisition of data along any given path over structure and also play an important role when accessibility is from one side only [3-5]. Among various NDT techniques, two prominent techniques that are most commonly used in the field for the rebar identification, cover thickness and rebar diameter estimation are Cover meter and GPR techniques. It is to be noted that these two techniques have their own


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merits and demerits, which depend on the depth of cover, spacing between rebars and diameter of rebars, etc. [6-10].

The commercially available Cover meter (Profoscope+), working principle, application, limitation, etc., are described in the manufacturer manual [7]. Advanced versions of Profoscopes are being released by the manufacturer. It is to be noted here that Cover meter probe setting (detectable range), bar diameter setting, etc., shall affect the cover thickness measurement. Limited research is reported on the reliable applicability of Cover meter for cover thickness and rebar diameter estimation. The laboratory study on concrete blocks with single embedded rebar reported that bar diameter can be estimated with an error percent between 1% to 35% for bar diameters 16 mm and above for cover thickness up to 50 mm [11]. The study carried out by Reuben et al. [10] reported that for cover thickness measurement, bar diameter setting and range setting (low or high) has greater influence on the cover thickness estimation. It is reported in the manufacturer manual that Cover meter survey is not affected by the presence of non-conductive materials such as wood, plastics, etc., however, any kind of conductive materials within the magnetic field will have an influence on the measurement [7]. The applicability of Cover meter in case of deep cover structural members such as footings, dams, etc. and closely spaced members such as strong floor reaction walls, turbo generator foundations, retaining walls, etc. is not known.

GPR is a new technique being used for subsurface investigation, rebar identification and mapping [12-14]. GPR uses electromagnetic pulses to image the subsurface. The commercially available GPR and its working principle is described in GSSI manual [15]. Accurate estimation of depth, or cover requires a reliable knowledge of the dielectric properties of the concrete [15, 16]. It is to be noted that rebar size cannot be estimated directly through GPR image. However, qualitatively, rebar size can be commented on a relatively comparative basis based on the strength of the signal [9].

The working principle, application and limitations of Cover meter and GPR are described in literature [8, 9, 12, 13, 17]. Many times during field investigations, difficulty is experienced in the (i) rebar identification, (ii) cover thickness, and (iii) diameter estimation using the Cover meter. This could be due to uneven/rough surface, deep concrete cover, congestion of rebars, etc. Hence, a laboratory parametric study is proposed to explore the reliable applicability of Cover meter and GPR for the rebar identification, estimation of cover thickness and estimation of rebar diameter in concrete structural members considering different cover

thicknesses and rebar diameters. The present study is also extended to identify the rebar position and cover thickness in field concrete walk-way and strong reaction floor concrete wall, available at CSIR-SERC, Chennai.

EXPERIMENTAL PROGRAMME

Specimen Details

As already mentioned that the purpose of present study is to explore the applicability and limitations of Cover meter and GPR on the estimation of cover thickness and rebar diameter, non-destructively. Hence, number of test specimens are exclusively cast with different cover thicknesses and rebar diameters for the data generation and reliable interpretation of the above two techniques. Five concrete blocks (B1 to B5) and two slabs (S1 and S2) with different rebar diameters and cover thicknesses were cast. The concrete block is of 150 mm thickness with a size of 400×250 mm. For easy interpretation and also to avoid interference effect from the adjacent rebars, single bar is placed in each block. For easy casting, an effective cover of 50 mm from top is maintained for all blocks. Rebars with different diameters viz., 32 mm, 25 mm, 20 mm, 16 mm, and 12 mm are used to get different clear covers. To identify the location and spacing between embedded rebars and also to estimate the rebar diameter and cover thickness, a slab S1 of size 1500×1000 mm with 150 mm thick; embedded with three rebars at sufficient spacing (i.e., more than 200 mm) between them to avoid interference effect from the adjacent rebars was cast [7, 18]. The diameters of rebars are 6 mm, 8 mm and 10 mm; placed at 40 mm effective cover from top surface (with a clear cover of 37 mm, 36 mm and 35 mm respectively, for 6 mm, 8 mm and 10 mm diameter rebars). To represent deep cover to the rebars in situations such as isolated foundations, irrigation dams, etc., a slab, S2 with a square of size 1500×1500 mm and 500 mm thick is cast. Three rebars of diameter 16 mm, 20 mm and 25 mm are placed with an effective cover of 160 mm from the top surface. Sufficient spacing i.e., more than 200 mm between the rebars is maintained to avoid interference

Table 1. Details of different specimensConcrete block /

slab

Diameter of rebar used

(mm)

Effective cover from the top

(mm)

Effective cover from the bottom ( mm)

B1 12 50 100B2 16 50 100B3 20 50 100B4 25 50 100B5 32 50 100

S110 40 1108 40 1106 40 110

S225 160 34020 160 34016 160 340


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effect from the adjacent rebars and also to get reliable GPR radagram [18]. As it is already mentioned that the main idea of present study is to understand the applicability of Cover meter and GPR in identifying the embedded rebars, reliable estimation of rebar diameters and concrete cover thickness in practically challenging situations such as footings, retaining walls, underground metro stations, etc. Therefore, concrete specimens with embedded rebars of shallow and large cover thicknesses are intentionally cast. Table 1 illustrates the details of different test specimens. Figure 1 shows the scanning of a typical concrete block using Cover meter and GPR.

Diameter Estimation using Cover meter

Commercially available Cover meter (Profoscope+) is used in the present study [7]. Rebar diameter estimation is made on all blocks (B1 to B5) and also on slab S1. The procedure for estimation of rebar diameter is: (i) identify the rebar position and (ii) place the centreline of the Cover meter over the rebar and click the function key to get the rebar diameter. For reliability and repeatability of diameter estimation using

Figure 2. Diameter estimation in blocks (B1 to B5) from top face

Table 2. Diameter estimation on block B1 and slab S1 using cover meter

Trails Estimated diameter from top face ( mm)

Block B1 12 mm

Slab S16 mm 8 mm 10 mm

1 11 9 12 12

2 12 8 8 8

3 9 8 9 10

4 11 8 11 8

Cover meter, several trials are made on each block. Figure 2 presents the estimated diameters using Cover meter for different trials. For blocks B2 to B5, the rebar diameters have been estimated correctly for all the seven trials. It can be said that the diameters have been estimated correctly with 100% repeatability, except for block B1. However, from the bottom face (clear cover thickness ranging from 84 mm to 94 mm), the diameter could not be estimated. This could be due to the limitation of Cover meter device, beyond 64 mm and more cover thickness, the diameter cannot be estimated [7]. Also, diameters are estimated for rebars embedded in slab S1. Table 2 presents the results of diameter estimation for Block B1 and slab S1 for four trials. It is observed that consistent and reliable diameter could not be estimated when the rebar diameter is equal to 12 mm and lower.

Figure 1. Scanning on a typical block specimen. (left) using Cover meter. (right) using GPR


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the rebar diameter using Cover meter before going for cover thickness estimation/measurement.

Cover thickness estimations are also made for three different embedded rebars (6 mm, 8 mm and 10 mm diameter) of slab S1 with an effective cover of 40 mm. Table 4 presents the cover thickness estimations for different assumed input/set diameters. It can be observed that, as the input/set diameter is increasing, the estimated cover thickness is also increasing, as observed in the case of block specimens earlier. However, when the set/input diameter is equal to rebar diameter, the estimated cover thickness is marginally higher than the actual cover thickness and the error in estimation is 8.5% to 13.5%. Similar observation has been made for block B1

Cover Thickness Estimation using Cover meter

To determine the cover thickness, scanning is carried out on the concrete blocks B1 to B5. As it is already discussed that, to get a correct cover thickness, the rebar diameter needs to be known apriori and set accordingly in the Cover meter device [7]. In the previous section, rebar diameter has been estimated for different blocks and slabs. In general, many times in the field, rebar details may not be available, especially for old structures and sometimes not possible to estimate the rebar diameter due to reasons such as congestion of rebars, thick concrete cover, etc. In such cases, an appropriate diameter is to be assumed based on the history of the structure and experience. Keeping that as back ground, in the present study, a parametric study has been carried out for cover thickness estimation by assuming different input diameters (four diameters/trials used). Again, to understand more about the applicability of Cover meter, both top and bottom covers of block specimens are proposed to be estimated.

Table 3 presents the details of four diameters/trials used for cover thickness estimation. Among the four input diameters, two are less than the actual, one is equal and the other is higher than the actual rebar diameter. The clear cover thicknesses are estimated and Figure 3 shows the plot of effective cover thicknesses for different input diameters. It is observed from Figure 3 that for the rebar diameters between 16 mm and 32 mm, the cover thickness is estimated correctly, only when setting the input diameter as actual bar diameter. It is also observed that when the input diameter is set as a value higher than the actual, the estimated cover is higher and vice versa. This has been observed for both top and bottom covers. From Figure 3, it can be inferred that the accuracy of cover thickness greatly depends on the set value of input diameter in the Cover meter device. The maximum error observed in the present study is about 21% for different incorrect input diameters. Therefore, in order to have reliable and correct cover thickness estimations, it is desirable to collect the rebar diameter details first, if available or estimate

Table 3. Different blocks-Actual and assumed input/set diameters

Concrete block

Actual rebar diameter ( mm)

Assumed input diameters ( mm)

B1 12 8 10 12 16B2 16 10 12 16 20B3 20 12 16 20 25B4 25 16 20 25 32B5 32 20 25 32 36

Table 4. Cover thickness estimation for Slab S1

Trail

Set / input

diameter (mm)

Actual rebar

diameter (mm)

Actualclear cover( mm)

Estimated clear cover

from top (mm)

Error in estimation

(%)

1 6

10 35

34 2.82 8 37 5.73 10 38 8.54 12 42 20.01 6

8 3638 5.50

2 8 41 12.53 10 43 19.41 6

6 3742 13.5

2 8 45 21.63 10 48 29.7

Figure 3. Cover thickness variation for different input diameters


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with rebar diameter 12 mm, where the estimated clear cover thickness is 47 mm instead of 44 mm (an error of 6.8%), even after setting the correct input diameter. Though the error in estimation is not significant, based on the results of rebar diameter and cover thickness estimations, it can only be said at this point of time that the cover thickness estimation may not be consistent and reliable for rebar diameters less than or equal to 12 mm. For slab specimen S2, cover thickness could not be estimated due to the embedment of rebars at 150 mm and above. However, Proceq manual states that the measuring range is dependent on the bar size and for 25 mm rebar diameter, the thickness measuring range is up to 160 mm [7].

Cover Thickness Estimation using GPR

A portable GPR with inbuilt antenna (2.6GHz frequency) is used in the present study [9]. Appropriate dielectric constant value has been evaluated by following methods [15].

• On site calibration over the specimens

• Hyperbolic fitting using the migration processing method.

GPR scanning is carried out on each of the concrete block specimens across the rebar position. Typical radargram images are shown in Figure 4. The hyperbola peak in radargram indicate the location/position of rebar. To get the cover thickness to the rebar, the radargram is processed for time zero correction using a processing software. The vertical position of the rebar i.e., the clear cover to the rebar can also be observed by studying the hyperbola along with its O-scope [9]. It can be seen that there is a change in the

signal polarity in the O-Scope corresponding to the peak of the hyperbola. This represent the position of rebar from the surface. The cover thickness results estimated using GPR are closely matching with the actual cover thicknesses. Similarly, for slab S1 and S2, GPR scanning is carried out. Figure 5 shows the processed radargram image for slab S2. It is clear that the three hyperbola peaks indicate the position of three embedded rebars. The cover thickness is estimated by studying the hyperbola along with its O-scope and the estimated cover thicknesses are closely matching with that of the actual cover thickness corresponding to the rebar. Table 5 presents the comparison of cover thickness estimations by using Cover meter and GPR. The error in estimation of the cover thickness is calculated with reference to actual clear cover thickness. For smaller or shallow cover depths upto 40 mm by using the Cover meter, it can be observed that, the error in estimated cover thickness is about 1% to 4% for embedded rebars with diameters more than 16 mm and is about 8% to 15% for embedded rebars with diameters 12 mm

Figure 5. Radargram with hyperbolas for slab S2

Figure 4. Radargram for typical block specimen B2; top and bottom face

(a) Top face (b) Bottom face


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and less. Also, using Cover meter, cover thickness could not be estimated for embedded rebars in slab S2, wherein the rebars are placed at about 160 mm. Nevertheless, with GPR, the cover thickness estimation is possible for shallow as well as deeper cover ranges and also the the error in estimation is less. The additional advantage of GPR is that the radargram images can be used to study the internal features of concrete, rebars, mapping/arrangement of rebars, etc. [3,14,19].

Diameter Estimation using GPR

For diameter estimation using GPR, slab S2 is considered. As discussed earlier, the collected radargram is processed for time zero correction and migration analysis has been carried out using RADAN Software. Radar does not estimate/measure the diameter of rebar, directly. For studying the hyperbolic signatures of radar images, different researchers adopted different methodologies for estimating the rebar diameter/radius quantitatively. Shaw et al. [20] proposed a methodology, which uses neural network to automate and facilitate the post-processing of radargram. Shihab et al. [21] attempted to mathematically model the GPR signatures to evaluate different parameters from the shape of the detected radargrams (hyperbola), by subjecting to a series of image processing stages followed by a curve-fitting procedure. Zhan et al. [22] developed a method for estimation of rebar radius by studying the hyperbolic signatures using stationary wavelet transform. Chang et al. [18] proposed a methodology in evaluating the diameter of embedded rebar using the following equations.

...(1)

...(2)

where, E is energy radius; λ is wave length; H is vertical position; L is scan length; ε is relative dielectric constant and R is radius of rebar

It is to be noted here that, for estimating rebar radius, energy radius of the cone (E) and length of the scan (L) are required. The energy radius, E depends on the wavelength (λ) of the penetrating radiation and the vertical position (H) of the rebar. Using a software, the radargram is post-processed by applying time zero correction and migration analysis. The vertical depth (H) or the rebar cover thickness and relative dielectric constant ( ε ) of steel concrete interface is estimated. However, the procedure to estimate the scan length is complex, lengthy and cumbersome and needs knowledge and understanding of the image processing. Hence, a simple methodology which is in similar lines as reported in literature is proposed [23]. In this procedure, in order to get the scan length, L, the collected radargram is converted into ASCII file that contains the amplitude values of the reflected signals at every interface. The key function of the process is converting the analog signals into digital signals. The digital signal is represented by numerically encoded values corresponding to the amplitude of the reflected electromagnetic waves. The amplitude values corresponding to steel concrete interface are traced out based on the significant difference in the numerical values. From the start and end position of the amplitude values of steel concrete interface, the number of scans recorded for a particular rebar is calculated. It is observed that sufficient spacing is required between rebars for easy selection of start and end position of steel concrete interface. From the obtained number of scans, the scan length, L (cm) is determined by dividing number of scans recorded with the scan density (for the present study, the scan density, i.e., the number of scans/cm is equal to 8). Eq. (2) is used to estimate

Table 5. Error in estimation of cover thickness using cover meter and GPR

Block/Slab

Diameter of rebar (mm)

Cover meter GPR

Error (%) in estimated cover from top face

Error (%) in estimated cover from bottom face

Error (%) in estimated cover from top face

Error (%) in estimated cover from bottom face

B1 12 11.4 14.9 2.2 2.0B2 16 2.4 3.3 -2.3 3.0B3 20 2.5 3.3 -2.5 2.2B4 25 1.3 3.3 -1.3 0.6B5 32 0.0 1.2 0.5 -1.2

S16 13.5

-----4.7

-----8 12.5 6.310 8.5 5.7

S216 Could not estimate due to

higher cover thickness (150 mm and above)

-----2.2

-----20 1.525 2.5


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the rebar radius. The radius of the rebar and its diameter is estimated using the above procedure for slab S2. The results are encouraging and further studies are in progress.

FIELD STUDIES ON PAVEMENT WALK-WAY AND STRONG FLOOR REACTION WALL Field studies are carried out to identify the rebar position and estimate the cover thickness of newly constructed walk-way (nominally reinforced) under the space grid roof and strong floor reaction wall (heavily reinforced) in the CSIR-SERC campus.

Pavement Walk-way

Figure 6 shows the concrete pavement walk-way under the space grid of CSIR-SERC. It is reinforced with 10 mm diameter bars; spaced at 25 mm from bottom surface and 200 mm c/c in both ways. The two way reinforcement photograph is shown in Figure 6(b). Longer direction bars are placed over shorter direction bars. That means, clear cover from bottom to the shorter direction bars is 25 mm. Therefore, clear cover from top is 65 mm for shorter direction bars (from bottom it is 25 mm) and 55 mm for longer direction bars (from bottom it is 25+10 = 35 mm). Cover meter and GPR survey is carried out to identify the position/location of rebars. Diameter of rebar is explored using Cover meter alone. Number of trials have been carried out to estimate the cover and diameter of rebars. Diameter of rebar could not be estimated using Cover meter. This could be due to higher cover (in this case it 55 to 65 mm) and lower size of rebar. It is worthy to note that during laboratory studies, it has been observed that the rebar diameter estimation is not reliable and consistent when the bar size is 12 mm and lower. However,

(a) Walk-way

(b) Rebar position

Figure 6. Concrete walk-way with rebar position

Figure 7. Radargram with hyperbolas - walk-way


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in the profoscope manual, it is reported that diameter can be estimated for cover depths up to 64 mm [7]. Nevertheless, the cover thickness estimated is 56 to 57 mm and 66 to 69 mm, respectively for shorter direction and longer direction bars, closely matching with the actual values. Figure 7 shows the line scan image of GPR radargram with hyperbolas. The hyperbola peaks in the radargram indicate rebar locations. The cover thickness is estimated by studying the hyperbola along with its O-scope. The cover thickness to the rebar and spacing between rebars is estimated and are closely matching with actual values. It can be said that Cover meter and GPR, both are effective in identifying the rebars, which include cover thickness and the spacing between them.

Strong Floor Reaction Wall

GPR survey and Cover meter survey is carried out at selected locations to identify the position/location of rebars in the reaction wall. Figure 8 (left) shows the concrete reaction wall with inserts. The reinforcement is placed in two layers on both faces of the wall. Some of the internal details are as follows: thickness of concrete wall, 750 mm; horizontal reinforcement, 16 mm diameter and vertical (main) reinforcement, 32 mm diameter, spaced at 150 to 200 mm; inserts @ 1000 mm c/c; clear cover to the 32 mm dia. main bars is 40 mm and 16 mm dia. bars is 72 mm (40+32=72 mm).

Initially, Cover meter survey is carried out to identify the position of rebars. The rebar identification is found difficult with Cover meter, which could be due to the presence of metal inserts, congestion of reinforcement, more cover thickness (72 mm), etc. Figure 8 (right) shows the GPR radargram with

hyperbolas. As it is reported earlier, the hyperbola peaks in the radargram indicate rebar locations between four inserts. It is observed that the horizontal and vertical reinforcement could be identified using GPR and is matching closely with the available drawings. This shows the limitation of Cover meter in identifying the rebars in closely/heavily reinforced large concrete structures.

CONCLUSIONSFollowing conclusions can be drawn from the limited experimental study:

• Cover meter and GPR techniques can be used for rebar identification, cover thickness and rebar diameter estimation. In general, Cover meter technique is simple to use and easy to interpret the results when compared to that of GPR technique.

• In laboratory cast single rebar specimens, rebar diameter estimation using Cover meter is effective and reliable for cover depths up to 40 to 50 mm. Consistent and reliable diameter could not be estimated when the rebar diameter is 12 mm and less.

• Using Cover meter, cover thickness estimation is effective and reliable for embedded rebar diameters greater than 12 mm for depths up to 50 to 60 mm. The cover thickness could not be estimated for cover thicknesses beyond 150 mm and more.

• To get more reliable and correct cover thickness estimations using Cover meter, it is desirable to get the

Figure 8. Strong floor reaction wall. (Left) Reaction wall. (Right) Radargram with hyperbolas


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rebar diameter details, if available or estimate the rebar diameter before going for cover thickness estimation/measurement.

• Structural members with deep cover such as foundations and closely spaced rebars sections with metal inserts such as reaction floor walls, rebar identification using Cover meter is difficult. In such cases, GPR is effective in identifying rebar locations and cover thickness estimations.

• GPR is very good in estimating the cover thickness for shallow as well as deep covers, however, diameter estimation is tedious and complex.

AcknowledgementsAuthors would like to acknowledge the help rendered by the staff of the Advanced Concrete Testing and Evaluation Laboratory (ACTEL) at the CSIR-Structural Engineering Research Centre (SERC), Taramani, Chennai.

References1. Malhotra, V. M., and Carino, N. J. (Eds.), Handbook on

Nondestructive testing of concrete, CRS Press, Washington DC, 2004.

2. Bungey, J. H., Millard, S. G. and Grantham, M. G., “Testing of concrete in structures”, 4th edition, 2006, Taylor & Francis, NY, USA.

3. McCann, D.M. and Forde, M.C., Review of NDT methods in the assessment of concrete and masonry structures, NDT and E International, 34, 2001, pp. 71–84.

4. Bhaskar, S., and Wiggenhauser, H., Advanced NDT for the evaluation of quality, durability and integrity of concrete/RC structural components, CEL-OLP131-RR-04, December, 2008.

5. Bhaskar, S., Srinivasan, P., Murthy, S. G. N., Nagesh R. Iyer and Ravisankar, K., “Evaluation of thickness and defects in concrete using Impact-echo technique”, The Indian Concrete Journal, January 2013, Vol. 87, No 1, 19-27.

6. Helal, J., Sofi, M. and Mendis, P. “Non-destructive testing of concrete A review of methods”, Special Issue: Electronic Journal of Structural Engineering, 14 (1), 2015, pp. 97-105.

7. Proceq “Profoscope operating instructions”, www.proceq.com, Switzerland, 2013.

8. British Standard BS1881 Part 204 “Reco mmendations on the use of electromagnetic Cover meters”. British Standards Institution, London, 1998.

9. “GSSI Handbook for RADAR Inspection of concrete”, Geophysical Survey Systems, Inc.Salem, New Hampshire, USA, August 2006.

10. Reuben BARNES., Tony ZHENG. “Research on Factors Affecting Concrete Cover Measurement”, www.ndt.net; The e-Journal of Non-destructive Testing, December 2008.

11. Sivasubramanian K., Jaya K.P. and Neelemegam M., “Cover meter for identifying cover depth and rebar diameter in high strength concrete”, International Journal of Civil and Structural Engineering, 3 (3), 2013.

12. ASTM Standard D 4748-10, “Standard Test Method for Determining the thickness of bound pavement layers using Short-Pulse Radar”, Annual book of ASTM standards, American Society of Testing and Materials, USA, 2010

13. ASTM D 6432 - Standard Guide for Using the Surface Ground Penetrating Radar Method for Subsurface Investigation, Annual book of ASTM standards, American Society of Testing and Materials, USA, 2011.

14. Srinivasan, P., Ravisankar, K., and Thirugnanasambandam, S., “Non-destructive evaluation of concrete structures with ground penetrating radar and influencing parameters”, IUP Journal of Structural Engineering, V (4), 2012, pp. 43-52.

15. GSSI “Manual for RADARTM 7”, Geophysical Survey Systems, Inc.Salem, New Hampshire, USA, 2013.

16. Bhaskar S., and K. Ramanjaneyulu., “Estimation of rebar diameter in concrete structural elements using ground penetrating radar.” 25th National Seminar & International Exhibition on Non-Destructive Evaluation (NDE 2015), Hyderabad, 2015 (CD format).

17. ACI 228.2R (2013) Report on Nondestructive test methods for evaluation of concrete in structures, American Concrete Institute, Farmington Hills, USA.

18. Chang, Che Way, Chen Hua Lin, and Hung Sheng Lien, “Measurement radius of reinforcing steel bar in concrete using digital image GPR.” Construction and Building Materials, 23 (2), 2009, pp. 1057-1063.

19. Bungey, J. H., and S. G. Millard, “Radar inspection of structures”, Proceedings of the International in Civil Engineering Structures and Buildings, 1993, pp. 173-178.

20. Shaw, M. R., S. G. Millard., T. C. K. Molyneaux, M. J. Taylor. and J. H. Bungey, “Location of steel reinforcement in concrete using ground penetrating radar and neural networks.” NDT & E International 38, 2005, pp. 203-212.

21. Shihab, S. and Al-Nuaimy, W., “Radius estimation for cylindrical objects detected by ground penetrating radar”, Subsurface Sensing Technologies and Applications, 6 (2), 2005, pp. 1-16.

22. Zhan, Runtao, Xie and Huicai., “GPR measurement of the diameter of steel bars in concrete specimens based on the stationary wavelet transform”, Insight- Non-destructive Testing and Condition Monitoring, 51 (3), 2009, pp. 151-155.

23. Ambika, L. K., Bhaskar, S., Vasanthakumar, S. and Alfa, R., “Estimation of diameter using Ground Penetrating Radar”, National Conference on Challenges of Civil Engineering Innovations (NCCCEI-2016), 4-5 May 2016, Bengaluru (CD Format).


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Dr. Bhaskar Sangoju holds a PhD degree from IIT Madras, Chennai. He is a Principal Scientist at CSIR-Structural Engineering Research Centre (SERC), Chennai. He is working at CSIR-SERC in the area of reinforced concrete (RC) durability and condition assessment of RC structures using non-destructive testing (NDT) and advanced NDT. His areas of interest are durability and service life enhancement of RC structures, condition assessment of RC structures using NDT, repair and rehabilitation of RC structures.

S. Vasanthakumar holds B.E. Civil Engineering from Anna University and was working as Project Assistant at Advanced Concrete Testing and Evaluation Laboratory, CSIR-SERC, Chennai.

Dr. K. Ramanjaneyulu, obtained doctoral degree in Structural Engineering from the Indian Institute of Technology Madras. He is heading the Advanced Concrete Testing and Evaluation Laboratory. His areas of research interests include condition assessment and performance evaluation of reinforced and prestressed concrete structures, analysis and design of special structures such as, bridges and cooling tower shells, seismic upgradation of gravity load designed structures, structural health assessment and management of bridges, repair and rehabilitation of structures.

Dr. K. Sivasubramanian is a Senior Scientist in CSIR-Structural engineering research centre, Chennai. He has been working in this organisation for over 15 years. He completed his PhD from Anna university, Chennai. He has about 50 technical publications. He has been working in the frontier research areas like ‘condition assessment of concrete structures’, ‘precast concrete structures’ and ‘numerical methods’. He is involved in the condition assessment of several TG foundations, marine structures, cooling towers and chimneys.


TECHNICAL PAPER

Flexural toughness characterization of steel, polymer and glass fibre reinforced concrete

based on the notched beam test

Sujatha Jose, Ravindra Gettu and Shabari Indhuja

IntRoductIonResearch on fibre reinforced concrete (FRC) has resulted in many applications due to its superior post-crack load-carrying capacity, which occurs because of the stress transfer by the fibres across the crack. Moreover, compared to plain concrete, FRC has improved energy absorption capability, fatigue resistance, fracture energy, ductility in tension and compression, flexural strength and toughness, resistance against impact loading, abrasion and durability

[1-7]. Although steel fibres are the most widely used, recently polymer and glass fibres are also being employed as reinforcement in various applications [8]. However, knowledge regarding the mechanical characteristics of polymer and glass fibre reinforced concrete is limited. The enhanced performance of fibre reinforced concrete is reflected by higher energy absorption capacity during fracture, which can be termed as ‘toughness’. The main


In the present work, the flexural behavior of concrete with different kinds of fibres have been characterized by center point loading (CPL) notched beam testing according to the EN 14651:2005 (E) and RILEM TC 162-TDF methods. The load-crack mouth opening (CMOD) and load-deflection curves were used to determine various toughness parameters, viz., flexural strength, residual flexural strength and equivalent flexural strength. A good correlation was found to exist between equivalent (feq) and residual flexural tensile strength (fR,j) parameters for steel and polypropylene fibre reinforced concretes. It was observed that hooked-ended steel fibres are more efficient in retaining the post-crack load-carrying capacity in comparison with polypropylene and glass fibres for the same volume fraction (0.4% Vf). The present study shows that the relationship between CMOD and midpoint deflection (δ) is almost the same for steel and macro polypropylene FRC at different dosages and, therefore, can be used to calculate CMOD or deflection when only one of them is known. The classification of FRC using the fib Model Code is also illustrated, which enables the designers to choose and specify FRC, based on its performance rather than by a type of fibre.

Keywords: Flexural toughness; notched beam test; crack mouth opening displacement (CMOD); equivalent flexural strength; fracture

objective of toughness characterization is to measure the effectiveness of fibres in the resistance offered by FRC against crack propagation. Researchers have long suggested the flexural characterization of FRC using a notched beam test done in a servo-controlled testing machine using Crack Mouth Opening Displacement (CMOD) control [4,9]. The notched beam test with centre point loading (CPL) was consequently recommended by the EN 14651:2005 (E) [10] standard and the RILEM TC 162-TDF Report [11]. However, in India, very limited testing capability and experience exists in the area of notched beam testing. Therefore, it is necessary to establish a database of results and propose test methodologies that are appropriate.

In this study, different dosages of hooked-ended steel, polypropylene and glass fibres were incorporated in a typical M40 grade concrete, and the corresponding toughness parameters have been determined. The flexural toughness parameters assessed are based on the load-crack mouth opening displacement (CMOD) and load-deflection curves


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obtained experimentally. The residual strength parameters are obtained from the load-CMOD curve as per EN 14651:2005 [10] and the equivalent flexural strength parameters are obtained from the load-deflection curve based on RILEM TC 162-TDF [11] method. A detailed analysis was also done to examine correlations between the equivalent and residual flexural tensile strength parameters for steel and polymer fibre reinforced concretes. The parameters obtained are used in the fib Model Code 2010 [12] classification to illustrate the specifications of toughness parameters that respond to the design requirements.

Figure 1. Test configuration of the notched beam test

Figure 2. Position of the notch

FlexuRal touGhneSS chaRacteRIzatIon by the notched beam teStThe notched beam test is performed as per EN 14651:2005 with the test configuration given in Figure 1; the specimen dimensions are 150 × 150 × 700 mm, and the span is 500 mm. A notch is cut across the width of specimen at mid-span on a face that is perpendicular to the casting direction, as indicated in Figure 2.

The tests have been performed here in a 1 MN closed-loop servo-controlled Controls testing system (Figure 3), and the control and data acquisition were done through a Controls Advantest interface. For measuring the CMOD (see Figure 4), a clip gauge is mounted on a pair of 3 mm thick knife edges across the notch at the mid-width of the specimen. Since the crack opening is measured at a distance of y (= 3 mm) from the bottom face of the specimen, the value of CMOD has to be corrected as: CMOD = CMODy where h is the depth of the specimen (i.e., 150 mm) and CMODy is the measured value. For measuring the deflection, two LVDTs are mounted on rigid frames fixed to the either side of the

Figure 3. Experimental setup of notched beam test (CPL)


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specimen and the tips rest on 1 mm thick steel plates placed across the notch (see Figure 4).

The prism is loaded such that the direction of casting is perpendicular to the loading direction and the notched face is at the bottom. The deflection of the specimen is taken as the average of the measurements of the two LVDTs (of 10 mm range). The test is performed initially by increasing the load at a constant rate of 100 N/s up to about 40% of the estimated peak load, and then by changing to CMOD control. The ranges of CMOD rates for the concretes reinforced with steel, polypropylene and glass fibres were 0.8-3.3 µm/s, 0.5-2.5 µm/s and 0.3-1 µm/s, respectively, all within the range mentioned in EN standard. Slightly slower rates had to be used for the concretes with polypropylene and glass fibres to ensure a stable post peak response in the testing system employed here. The time to peak in all cases was 2-3 minutes. The test is performed up to a CMOD of at least

4 mm and the time taken for the entire test for the concretes with steel, polymer and glass fibers were about 30, 45 and 60 minutes, respectively.

The limit of proportionality (LOP) is the flexural strength obtained using the first peak load (EN 14651:2005 [10] ) given by:

....(1)

where, is the LOP, in MPa;FL is the load corresponding to the LOP, in N, taken as the highest load in the CMOD interval of 0-0.05 mm (see Figure 5);l is the span length, in mm;b is the width of the specimen, in mm; hsp is the ligament length, in mm; (Figure 2).

The residual flexural tensile strength, fR,j, is an estimate of the flexural strength retained by FRC after cracking up to a particular crack width, calculated from the load-CMOD curve as (EN 14651:2005 [10] ):

...(2)

where,fR,j is the residual flexural tensile strength, in MPa, corresponding to CMOD = CMODj ( j = 1,2,3,4) for CMOD values of 0.5, 1.5, 2.5 and 3.5 mm as shown in Figure 5Fj is the load corresponding to CMODj, in N;l is the span length, in mm; b is the width of the specimen, in mm; hsp is the ligament length, in mm.

The equivalent flexural strengths have been calculated as per RILEM TC 162 TDF [11] from the load deflection curve based on the average load in the post peak region. The area under the load-deflection curve DBZ,2 (or DBZ,3) up to a deflection δ2 (or δ3), is taken to be the sum of the contribution of the plain concrete and that of the fibres (Df

BZ,2 or DfBZ,3) (see Figure 6).

The two contributions are separated by drawing a straight line connecting the point on the curve corresponding to FL and the point on the abscissa equal to “δL +0.3”, where δL is the deflection at FL.

The deflection limits δ2 and δ3 are in turn defined as:

...(3)

...(4)

Figure 4. Position of the clip gauge and LVDT

Figure 5. Points at which residual flexural strengths are determined


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and F2 and F3 are calculated from DfBZ,2 and Df

BZ,3 as follows:

...(5)

...(6)

The equivalent flexural tensile strengths feq.2 and feq.3 can consequently be determined (RILEM TC 162 TDF [11] ) as:

...(7)

...(8)

expeRImental pRoGRammeThe flexural behavior of concrete, having 40 MPa design compressive strength (denoted as M40), reinforced with hooked-ended steel, polypropylene and glass fibres was characterized. The details of the fibres are given in Table 1.

In all the mixes, portland pozzolana cement (PPC), river sand (with grain size range of 0-5 mm, corresponding to zone 2 of IS 456) and crushed granite coarse aggregates (in the fractions of 5-10 mm and 10-20 mm) were used, and the water/cement ratio was 0.45. A polycarboxylate (PCE) based superplasticizer (with a density of 1080 kg/m3 and solid content of 33%), was used to attain the desired workability. The fibre dosages incorporated are given in Table 1. The nominal mix proportions for the concretes are given below in Table 2. The aggregates were considered to be in the saturated surface-dry state; the humidity of the aggregates

Figure 6. Contribution of fibres for the estimation of equivalent flexural strength

Table 1. Details of the fibres used

Material Type Specific gravity*(g/cc)

Length*(mm)

Diameter*(mm)

Tensile strength*(MPa)

Fibre dosage in kg/m3

Steel Hooked-ended steel 7.80 60 0.75 1225 10,15,20,30,45

Polymer Polypropylene macro fibres 0.92 40 0.44 620 2.5,3.75,5

Glass Glass 2.68 36 0.54 1700 5,10,15

* specified by the manufacturer

Table 2. Nominal mix proportions, in kg/m3

Material Quantity

Cement 380

Fine aggregates 760

5-10 mm Coarse aggregates 390

10-20 mm Coarse aggregates 700

Water 171


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was checked before the fabrication of each batch of concrete and the corresponding adjustments were made in the water added.

The mixing of the concrete was done in a forced-action vertical axis pan mixer of 250 litres capacity. The sequence of mixing was: initial dry mixing of aggregates and cement; adding of water and wet mixing for few minutes; and then addition of superplasticizer to attain workability. In the case of steel and polypropylene fibres, the concrete was mixed for about 3 minutes after fibre addition, whereas the mixing was limited to 45 seconds for concretes with glass fibres (as per the instructions of the manufacturer) to prevent any degradation of the fibres.

Cubes of 150 × 150 ×150 mm and prisms of 150 × 150 × 700 mm were cast from all the concretes. A high frequency vibrating table was used for compaction and the specimens were vibrated for about 15 seconds. In the case of prisms, the concrete was poured into the middle of the mould and allowed to flow to the ends; few scoops of concrete were placed at the ends and at top the mould. This method was followed so as to avoid weak planes where failure is expected to occur. The specimens were left in the moulds for 24 hours after casting, then demoulded and cured for the next 27 days in a mist room. Three days prior to the testing, a notch of 25 mm length was cut at mid-span across the width and

the specimen was kept again in the mist room for curing. Flexural tests were performed at the age of 28 days.

ReSultS and dIScuSSIonS

Fresh Properties

There was a loss in workability due to the addition of fibres, irrespective of the type of fibre, as expected. The superplasticizer dosage, by weight of cementitious materials, was increased (e.g., from 0.19% in the plain concrete to 0.3% for 30 kg/m3 of steel fibres, 0.75% for 5 kg/m3 of polypropylene fibres and to 0.8% for 15 kg/m3 of glass fibres) in order to obtain a slump of 100±20 mm for concretes with steel, 60±20 mm for polypropylene and 70±20 mm for glass fibres. In general, all mixes showed similar trends of workability reduction with an increase in fibre dosage. In all cases, the mixes were uniform and convenient for placing. The fresh properties of the different concretes are given in Table 3 along with the mix design. It can be seen that the unit weight is in the range of 2430–2490 kg/m3, indicating that all the mixes attained uniform degree of compaction. Note that in the mix design, M40 denotes the concrete grades, SF, PF and GF denotes the use of steel, polypropylene and glass fibres, and the number at the end denotes the fibre dosage in kg/m3. Reference concrete without any fibres were cast separately for the three series of SFRC, PFRC and GFRC due to extended duration of

Table 3. Fresh properties and compressive strengths of the different mixes

Concrete Type of FibreFibre dosage (kg/m3) and

Vf

Unit weight (kg/m3)

Slump (mm)

Compressive strength (MPa)(mean ± standard deviation)

at 3 days at 7 days at 28 days

M40SF0

Steel

- 2450 90 22.3±1.2 33.7±0.5 47.1±0.3M40SF10 10;0.13% 2450 95 22.5±0.8 34.4±0.9 48.4±0.8M40SF15 15;0.20% 2470 95 23.5±0.7 35.6±1.5 49.2±0.6M40SF20 20;0.26% 2490 100 26.7±0.8 36.7±1.6 50.4±1.3M40SF30 30;0.40% 2475 110 27.9±1.3 37.3±0.7 51.5±1.7M40SF45 45;0.58% 2490 100 28.6±0.4 38.0±0.9 52.8±0.7

M40PF0

Polypropylene

- 2480 90 21.1±0.3 29.9±1.7 46.5±1.6M40PF2.5 2.5;0.27% 2450 70 24.1±1.4 31.1±0.8 46.5±1.6

M40PF3.75 3.75;0.41% 2470 65 22.2±0.6 30.4±2.6 49.3±2.8 M40PF5 5;0.54% 2460 45 22.5±0.2 31.8±0.8 47.3±3.3 M40GF0

Glass

- 2440 105 20.6±1.3 27.9±3.4 44.8±0.5 M40GF5 5;0.19% 2430 80 22.9±2.2 32.8±1.9 45.8±1.2

M40GF10 10;0.38% 2430 70 23.3±1.7 34.5±0.8 44.5±1.2M40GF15 15;0.57% 2400 60 20.7±0.3 30.1±2.5 46.5±1.0


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the experimental programme, and to account for variations in the materials used in fabricating the concrete.

Compressive strength

The uniaxial compressive strength was assessed with cubes tested at the ages of 3, 7 and 28 days using a 3 MN capacity Controls machine; a minimum of three specimens were tested at each age. The mean values along with the standard deviations for concretes with different fibre types and volume fractions are given in Table 3.

The results indicate that incorporation of steel fibres resulted in a slight increase in the compressive strength compared to plain concrete, from 3% to 12% for the fibre dosages of 10 kg/m3 to 45 kg/m3. This increase in compressive strength is attributed to the crack restraining effect of the steel fibres and consequent delay in crack growth rate. However, the results also indicate that there is no significant change in compressive strength due to the addition of polypropylene and glass fibres, for the dosages considered here. The variability of compressive strengths is within the usual range and not influenced by the type or amount of fibres.

FlexuRal behavIoR baSed on load-cmod and load-deFlectIon cuRveS Typical load-CMOD curves for all the mixes are presented in Figures 7 to 9 and the corresponding load-deflection curves are shown in Figures 10 to 12. The plots given in 7(a), 8(a), and 9(a), shows the behaviour up to a CMOD or deflection of 0.5 mm and the plots given in 7(b), 8(b), and 9(b), shows the behaviour of up to a CMOD or deflection of 4 mm. It is clear from the curves that, in the pre-peak region, there is no significant influence of the incorporation of steel fibres (say until a CMOD of about 0.18 mm), whereas in the post crack region, is considerably dependent on the fibre dosage. However, for the mixes with polypropylene and glass fibres,

Figure 7(a). Typical load-CMOD curve of M40SF mixes, up to 0.5 mm

Figure 7(b). Typical load-CMOD curves of M40SF mixes

Figure 8(a). Typical load-CMOD curves of M40PF mixes, up to 0.5 mm

Figure 8(b). Typical load-CMOD curves of M40PF mixes


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a marginal increase in peak load is observed. For all the mixes, the post-peak load-carrying capacity and the area under the curve have a direct relation with the fibre dosage and as expected, increase with dosage.

In general, the response of most mixes exhibits a softening type behavior, except for the concrete with steel fibres (Figure 7). For the SFRC mixes there is a gradual change from softening-type response to hardening-type response with an increase in fibre dosage (from 10 to 45 kg/m3), especially at larger deflections (after a CMOD of 0.2 mm). In fact, for the M40SF45 mix, the post-crack load-carrying capacity was even higher than the peak load after a CMOD of 1 mm. Such hardening-type behavior for higher dosages of steel fibres have been reported by other researchers [9,13-14]. For lower dosages of steel fibres (10 and 15 kg/m3), the

softening branch is followed by a region of constant residual load.

A sudden drop in the curve beyond a CMOD of about 0.5 mm was observed for all dosages of polypropylene fibres, beyond which there is considerable retention of post-cracking capacity. Also, it is observed that the post-cracking load carrying capacity increases as the dosage increases from 2.5 to 3.75 kg/m3, whereas the increase is not to the same order when the dosage increases from 3.75 to 5 kg/m3.

The curves for GFRC (Figure 9) show a sudden drop in the load-carrying capacity after the peak, for all dosages, with little load retention beyond 1.5 mm. An increase in toughness as the dosage increases is noticed only until the crack opening of about 1.5 mm, beyond which there is no significant improvement. Similar behavior regarding GFRC

Figure 9(a). Typical Load-CMOD curves of M40GF, mixes up to 0.5 mm

Figure 9(b). Typical load-CMOD curves of M40GF mixes

Figure 10(a). Typical load-deflection curves of M40SF mixes, up to 0.5 mm

Figure 10(b). Typical load-deflection curves of M40SF


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was also observed in the load-deflection response in the unnotched beam test with the same dosages considered here [15].

RelatIonShIp between dISplacement and cmod In the notched beam teSt It can be seen in the Figures 7 to 12 that the load-CMOD and load-deflection curves has similar shapes for all the concretes tested. This promotes the possibility of obtaining a relationship between deflection (δ) and CMOD, which

could lead to the measurement of CMOD alone and thus save time and cost for the setup of deflection measurement. By converting CMOD into equivalent mid span deflection, it is possible to evaluate the design parameters (feq,2 and feq,3) of RILEM TC-162-TDF [11] from load-CMOD curves which would otherwise need the load-deflection (δ) curves.

It can be seen in Figures 13 and 14 for plain and M40SF45 concretes that the response in the pre-cracked region is

Figure 11(a). Typical load-deflection curves of M40PF mixes, up to 0.5 mm

Figure 11(b). Typical load-deflection curves of M40PF

Figure 12(a). Typical load-deflection curves of M40GF mixes, up to 0.5 mm

Figure 12(b). Typical load-deflection curves of M40GF

Figure 13. Typical plot of d against CMOD for Plain concrete

Figure 14. Typical plot of d against CMOD for SFRC (45kg dosage)


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nonlinear, though the major part of the response justifies a linear relation between CMOD and deflection with a constant δ/CMOD ratio in the post crack phase, as suggested by the rigid body model. Jamet et al. [16] previously observed a linear relationship between deflection (δ) and CMOD for concrete with 1% Vf of hooked-ended steel fibres, and reported that the ratio of deflection δ/CMOD was approximately 0.65. Oh et al. [17] also reported a linear relationship between CMOD and deflection (δ) for macro synthetic FRC for Vf of 1% and 1.5%. Other researchers have assumed that when the specimen cracks at the mid span, the dominant deformation resembles the rigid body motion of the two halves rotating about a ‘plastic hinge’ [16,18]. Consequently, in the post-cracked phase of the notched beam test, a linear relation arises between deflection (δ) and CMOD as shown in Figure 15. The displacement at the mid span of the prism is then given by δ = (L/2) θ, for the crack rotation of 2θ, and the relation between deflection (δ) and CMODy is:

...(9)

where L is the span of the beam; d is the apparent depth about which the beam rotates. Based on a large series of tests at different laboratories, Vandewalle et al. [19] obtained the following relationship for the post-peak response of SFRC beams:

δ = 0.85CMOD +0.035 ...(10)

From the tests performed here, the experimental data of deflection versus CMOD in the post-peak phase, for all the dosages of steel, polypropylene, glass fibres and plain concrete are as seen in Figures 16 to 19.

A fit of the experimental data in the post peak phase gives the relations between CMOD and deflection (δ) as in Eqns. 11 to 14, for SFRC, PFRC, GFRC and plain concrete, respectively. Note that the dotted line shown in each plot of Figures 16 to 19 is a linear least square fit of the data points and R2 is the coefficient of determination of regression.

Figure 15. Rigid body model of the beam test

Figure 16. Post peak CMOD-deflection curve of SFRC

Figure 17. Post peak CMOD-deflection curve of PFRC

Figure 18. Post peak CMOD-deflection curve of GFRC


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deteRmInatIon oF FlexuRal touGhneSS paRameteRS From the experimentally-obtained load-CMOD and load-deflection curves, toughness parameters can be determined

Table 4. Flexural toughness parameters (mean and COV)

Fibre Type Concrete

Fibre dosage in kg/m3

fctf

(LOP)(MPa)

fR,1 (MPa)

CMOD = 0.5 mm

fR,2 (MPa)

CMOD = 1.5 mm

fR,3 (MPa)

CMOD = 2.5 mm

fR,4 (MPa)

CMOD = 3.5 mm

feq,2(MPa)

feq,3(MPa)

Steel

M40SF0 - 5.22 (±8%) - - - - - -

M40SF10 10 5.33(±8%)

2.19(±8%)

2.11(±10%)

2.13(±11%)

2.05(±14%)

1.99(±16%)

2.04(±13%)

M40SF15 15 5.44(±9%)

2.35(±6%)

2.48(±8%)

2.55(±11%)

2.48(±12%)

2.09(±9%)

2.40(±10%)

M40SF20 20 5.58(±4%)

3.40(±16%)

3.73(±15%)

3.80(±10%)

3.68(±8%)

3.23(±18%)

3.62(±13%)

M40SF30 30 5.38(±8%)

4.04(±15%)

5.21(±12%)

5.54(±14%)

5.48(±14%)

3.89(±18%)

5.03(±12%)

M40SF45 45 5.47(±8%)

5.30(±15%)

6.70(±18%)

6.63(±23%)

6.53(±21%)

5.29(±16%)

6.33(±19%)

Polypro-pylene

M40PF0 - 5.0(±4%) - - - - - -

M40PF2.5 2.5 5.13(±8%)

1.52(±9%)

1.30(±6%)

1.27(±4%)

1.16(±11%)

1.26(±11%)

1.33(±4%)

M40PF3.75 3.75 5.06(±9%)

2.00(±7%)

1.92(±7%) 2.01 (±6%) 1.90

(±7%)1.66

(±16%)2.05

(±11%)

M40PF5 5 5.51(±5%)

2.33(±12%)

2.55(±10%)

2.58(±13%)

2.29(±13%)

2.13(±10%)

2.49(±10%)

Glass

M40GF0 - 5.06(±8%)

0.82(±18%) - - - - -

M40GF5 5 5.13(±11%)

1.34(±10%)

0.58(±16%)

0.39(±13%)

0.23(±56%)

1.04(±17%)

0.54(±33%)

M40GF10 10 5.47(±4%)

1.80(±14%)

1.13±13%)

0.70(±39%)

0.57(±45%)

1.48(±26%)

0.95(±23%)

M40GF15 15 5.77(±6%)

2.42(±21%)

1.22(±32%)

0.75±41%)

0.52(±38%)

2.51(±22%)

1.49(±40%)

δ = 0.80CMOD + 0.017, for SFRC ...(11)

δ = 0.77CMOD + 0.038, for PFRC ...(12)

δ = 0.95CMOD - 0.002, for GFRC ...(13)

δ = CMOD - 0.0012, for Plain Concrete ...(14)

From Eqns. 11 to 14, it can be seen that in the post peak phase, the relation between CMOD and deflection (δ) varies depending on the type of fibre. As deflection increases, the increase in CMOD is highest for plain concrete followed by GFRC, SFRC and PFRC. Using the rigid body model (see Figure 15) proposed by Armelin and Banthia [18] it is obtained that, the location of the plastic hinge from the top of the beam c was found to be 0.6 mm for plain concrete, and for SFRC has the value of c of 0.5 mm, for PFRC, c =3.5 mm, and for GFRC it is much greater, c =21.4 mm. This implies that at the end of CMOD = 4 mm, the crack length for SFRC and plain concrete is almost the same, and slightly higher than PFRC, whereas the crack length for GFRC is significantly less, reflecting less crack bridging in the latter case.

Figure 19. Post peak CMOD-deflection curve of Plain Concrete


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according to the EN 14651:2005 and RILEM TC 162-TDF. The limit of proportionality (LOP) and residual flexural tensile strength fR,j at CMOD of 0.5, 1.5, 2.5 and 3.5 mm were obtained from Eqns. 1 and 2, respectively. The equivalent flexural strengths feq,2 and feq,3, were obtained as per Eqns. 7 and 8, respectively. The parameters for all mixes are given in Table 4, as mean values and coefficients of variation (COVs).

There is some marginal increase in the flexural strength (LOP) due to the incorporation of fibres but there is no clear trend or dependence on the fibre dosage or type, though the highest values were obtained with the GFRC.

The values of residual flexural strength (fR,j) at different crack openings of 0.5, 1.5, 2.5 and 3.5 mm reflect the ability of FRC to maintain the load-carrying capacity over different crack openings. It was observed in steel fibre reinforced concrete that the residual strengths increase after a crack mouth opening of 0.5 mm, for all dosages, reflecting the ‘hardening’ behavior. For PFRC, the residual flexural strength values are almost the same, for all crack openings beyond 0.5 mm, indicating constant residual load carrying capacity. However, in GFRC the residual flexural strength decreases significantly at larger crack openings, reflecting the softening response of the post peak curves. The equivalent flexural tensile strengths, i.e., feq.2 and feq.3, show the same trends as the residual values with respect to increase in dosage. The residual strengths obviously give the same trends as the load-CMOD curves and have an advantage over the equivalent strengths in that they can be used to find intermediate values.

The CoVs obtained for the toughness parameters (see Table 4) are generally higher than those of the flexural strength and the compressive strength. It can be observed that the variations in the toughness results are, in general, much lower for polypropylene fibres in comparison with the hooked ended steel and glass fibres, as reflected in the

CoVs. This lower scatter of polypropylene fibre specimens compared to steel fibre ones was also reported by Buratti et al. [8] which is due to homogenous distribution of fibres and higher number of polypropylene fibres across the fracture surface. It is also seen that the COVs of fR is found to be slightly higher than feq in the case of SFRC and GFRC but almost in the same range for PFRC.

RelatIonShIp between equIvalent and ReSIdual FlexuRal tenSIle StRenGth paRameteRS The relations between feq,2 and feq,3 and between fR,1 and fR,4

for SFRC, PFRC and GFRC are shown in Figures 20 to 25, wherein a linear trend was observed, similar to that reported by Barros et al. [14] for SFRC. On comparing the parameters of all the concrete mixes (see Table 4), feq,3 and fR,4 was found

to be having higher values than feq,2 and fR,1 for both SFRC and PFRC, however, the difference is greater in the case of SFRC, and for PFRC there is only a slight variation which is also

Figure 20. Relation between feq,2 and feq,3 of SFRC

Figure 21. Relation between feq,2 and feq,3 of PFRC

Figure 22. Relation between feq,2 and feq,3 of GFR


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visible in load-CMOD curves which is much flatter for PFRC as seen in Figures 7 and 8. Nevertheless, in the case of GFRC, feq,3 and fR,4 was found to be having lower values than feq,2

and fR,1 which is visible in the softening load-CMOD curves. This also indicates that energy absorption and post cracking capacity is maintained at large crack mouth openings in the case of PFRC and increases at larger crack mouth openings in the case of SFRC especially at higher dosages. From Figures 20 to 22, it can be observed that, good correlation exists between feq,2 and feq,3 for SFRC, and a reasonably good correlation for PFRC and GFRC. However, in the case of fR,1 and fR,4 (see Figures 23 to 25), although there is good correlation existed for SFRC and PFRC, no correlation exists for GFRC, which can be explained by its softening response after the peak and, consequently, its low load carrying capacity after a crack width of about 1.5 mm.

The relations between feq,2 and fR,1 and between feq,3 and fR,4,

for SFRC, PFRC and GFRC are shown in Figures 26 to 31, and it was found that there is good correlation that existed between the equivalent and residual flexural tensile strength parameters for SFRC and PFRC, with better correlation for SFRC than PFRC. A similar study conducted by

Figure 23. Relation between fR1 and fR4 of SFRC

Figure 24. Relation between fR1 and fR4 of PFRC

Figure 25. Relation between fR1 and fR4 of GFRC

Figure 26. Relationship between feq.2 and fR1 of SFRC

Figure 27. Relationship between feq.2 and fR1 of PFRC

Figure 28. Relationship between feq.2 and fR1 of GFRC


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Barros et al. [14] for SFRC, also reported a good correlation between feq,2 and fR,1 and between feq,3 and fR,4 for SFRC and the relation between them was almost the same as shown in Figures 26 to 31. In the case of GFRC there is poor correlation between feq,2 and fR,1 and between feq,3 and fR,4.

claSSIFIcatIon oF FRc baSed on FIb model code 2010 FRC has been included in the fib Model Code 2010 for structural elements, at both serviceability and ultimate limit states. A material classification has been included for FRC based on the characteristic flexural toughness parameters fR,1k and fR,3k, which characterize the material behavior at the serviceability limit state (represented by CMOD = 0.5 mm) and at the ultimate limit state (represented by CMOD = 2.5 mm), respectively. The characteristic strength is defined as :

σchar = σm - ks ...(15)

where σchar and σm are the characteristic and mean values, respectively; s is the standard deviation and k is the risk factor corresponding to a failure rate of 5%, obtained as a function of the number of specimens tested N, as shown in Table 5 (RILEM TC 162-TDF [21] ). Here, kxunknown is considered, since the standard deviation of the population is unknown, and consequently for N = 6, k is taken as 2.18.

The class of the FRC is denoted with a number followed by a letter: the number is based on the value of fR,1k rounded off to

the nearest 0.5 MPa and the letter depends on the ratio fR,3k/ fR,1k ratio, as follows:

a, if 0.5 <fR,3k / fR,1k ≤ 0.7

b, if 0.7 <fR,3k / fR,1k ≤ 0.9

c, if 0.9 <fR,3k / fR,1k ≤ 1.1

d, if 1.1 <fR,3k / fR,1k ≤ 1.3

e, if fR,3k / fR,1k >1.3

Table 5. Risk factor (k) as a function of number of specimens in a sample (from RILEM TC 162-TDF [21] )N 1 2 3 4 5 6 8 10 20 30 ∞

kxknown 2.31 2.01 1.89 1.83 1.80 1.77 1.74 1.72 1.68 1.67 1.64

kxunknown - - 3.37 2.63 2.33 2.18 2 1.92 1.76 1.73 1.64

Figure 29. Relationship between feq.3 and fR4 of SFRCFigure 30. Relationship between feq.3 and fR4 of PFRC

Figure 31. Relationship between feq.3 and fR4 of GFRC


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Further, the Model Code 2010 specifies that the FRC can only be used at the ultimate limit state (ULS) if the relationships given in Eqns. (16) and (17) are fulfilled, where fLk is the characteristic value of the LOP [22].

fR,1k/ fLk ≥ 0.4 ...(16)

fR,3k/ fR,1k ≥ 0.5 ...(17)

Based on the data obtained, the FRCs tested here can be classified according to fib Model Code 2010 [12], using in Eqns. (16) and (17), as in Table 6.

From Table 6, it can be observed that, as the dosages of steel and polypropylene fibres in concrete increase, the class reflects the improvement in flexural toughness performance. For example, the class in which 10,15,20,30 and 45 kg/m3 dosage of SFRC belongs are 2b, 2c, 2.5e, 3e and 4b, respectively. However, none of the dosages of GFRCs considered here, could be classified as they fail to satisfy the restrictions on the minimum residual strength (fR,3k / fR,1k ratio is found to be less than 0.5). Among all the mixes considered here, this condition for ULS was satisfied only for the concretes with steel fibres, and for 3.75 kg/m3 dosage of polypropylene fibres.

Since the classification is based on ranges of values, different dosages of fibres can result in the same class of concrete and the same volume fractions can result in different classes.

For example, concretes with 15 kg/m3 dosage of steel fibres, 3.75 and 5 kg/m3 dosages of polypropylene fibres belong to the same class 2c. For the same volume fraction of 0.4%, the classification according to fib Model Code 2010 gives the ‘classes’ of M40SF30 and M40PF3.75 as ‘3e’ and ‘2c’.

In accordance with this classification, the designer has to specify the grade of concrete and fibre class for a particular application, for example if M40-4b is specified, it is required to provide a concrete of M40 grade with fR,1k of 4-5MPa and fR,3k / fR,1k ratio of 0.7-0.9.

concluSIonSThe post-cracking behaviour of steel, polypropylene and glass fibre reinforced concrete was assessed using three-point bending tests on notched specimens for M40 grade of concrete. From the studies of concretes, fibre types and dosages as those considered here and from the results obtained, the following conclusions can be drawn:

• The residual flexural tensile strength (fR,j) and equivalent flexural strengths (feq) values of notched beam testing show an appreciable increase in toughness when compared to plain concrete and the performance is better as the dosage increases irrespective of the type of fibre, as expected.

• From the load-CMOD and load-deflection responses in the notched beam test, it can be concluded that at

Table 6. FRC Classification based on fib Model Code 2010

Fibre Type Concrete

Fibre dosage in kg/m3;Fibre Volume fraction

Vf in %

fL,k(MPa)

fR,1k(MPa)

fR,3k(MPa) fR,3k / fR,1k Class fR,1k / fLk

Steel

M40SF0 - 4.30 - -

M40SF10 10;0.13 4.48 1.80 1.61 0.9 2b 0.40

M40SF15 15;0.19 4.46 1.85 1.92 1.0 2c 0.42

M40SF20 20;0.26 4.66 2.35 3.06 1.3 2.5e 0.50

M40SF30 30;0.38 4.46 2.71 3.91 1.4 3e 0.61

M40SF45 45;0.57 4.58 3.62 3.25 0.9 4b 0.80

Polypro-pylene

M40PF0 - 4.66 - -

M40PF2.5 2.5;0.27 4.44 1.30 1.14 0.9 1.5b 0.30

M40PF3.75 3.75;0.40 4.28 1.70 1.77 1.0 2c 0.40

M40PF5 5;0.54 5.03 1.85 1.82 1.0 2c 0.37

Glass

M40GF0 - 4.23 0.49 - 0.12

M40GF5 5;0.19 4.14 1.06 0.28 0.3 - 0.26

M40GF10 10;0.37 5.10 1.26 0.11 0.1 - 0.25

M40GF15 15;0.56 5.11 1.33 0.07 0.1 - 0.26


TECHNICAL PAPER

large crack openings ,the hooked-ended steel fibres are more efficient in retaining the post-crack load-carrying capacity in comparison with polyproplene and glass fibres for the same volume fraction. SFRC had the highest residual and equivalent flexural strengths followed by PFRC, and lastly GFRC for a volume fraction of 0.4%, at all crack openings.

• In notched beam test, a linear relationship was found to be existing between displacement and CMOD in the post-crack stage, for all the fibre dosages considered here (up to 0.6%Vf) for SFRC, PFRC and GFRC.

• For steel fibres and polypropylene fibres, good correlation exists between the equivalent and residual flexural tensile strengths.

• Classification of concretes with different dosages of hooked-ended steel, polypropylene and glass fibre reinforced concretes has been illustrated in accordance with the performance based approach of fib Model Code 2010. For the design of FRC structural elements, at both serviceability and ultimate limit states, concretes with all the dosages of steel fibres, and with 3.75 kg/m3 dosage of polypropylene fibres were found to be suitable.

Acknowledgements

The authors acknowledge the support of Bekaert Industries, Grace Fibres, and Owens Corning for having provided the fibres used in this study. The help of Ms. Navami Sunil, Mrs. Malarrvizhi, and the other staff of the Construction Materials Laboratory of IIT Madras is gratefully appreciated.

References

1. Mindess, S. and Banthia, N., Toughness Characterization of Fibre-Reinforced Concrete: Standard to Use? Journal of Testing and Evaluation, 2004, Vol. 32, pp 1-5.

2. Rossi, P., Coussy, O., Boulay, C., Acker, P. and Malier, Y., Comparison between plain concrete toughness and steel fibre reinforced concrete toughness, Cement and Concrete Research, 1986, Vol. 16, pp 303-313.

3. Balaguru, P. and Shah, S. P., Fibre Reinforced Cement Composites, McGraw-Hill, New York, 1992.

4. Gopalaratnam, V.S. and Gettu, R., On the characterization of flexural toughness in fibre reinforced concretes, Cement and Concrete Composites, 1995, Vol. 17, pp. 239-254.

5. Granju, J.L. and Balouch, S.U., Corrosion of steel fibre reinforced concrete from the cracks, Cement and Concrete Research, 2005, Vol. 35, pp 572 – 577.

6. Zerbino, R. L., Giaccio, G. and Gettu, R., Pseudo-ductile behaviour of steel fibre reinforced high-strength concretes, The Indian Concrete Journal, February 2006, Vol. 80, No. 2, pp 37-43.

7. Nayar, S. K., Gettu, R. and Krishnan, S., Characterisation of the Toughness of Fibre Reinforced Concrete - Revisited in the

Indian Context, The Indian Concrete Journal, February 2014, Vol. 88, No. 2, pp. 8–23.

8. Buratti, N., Mazzotti, C. and Savoia, M., Post-cracking behaviour of steel and macro-synthetic fibre-reinforced concretes, Construction and Building Materials, 2011, Vol. 25, pp 2713–2722.

9. Gettu, R., Mobasher, B., Carmona, S. and Jansen, C. D., Testing of concrete under closed-loop control, Advanced Cement Based Material Journal, 1996, Vol. 3, pp 54-71.

10. Test Method for Metallic Fibre Concrete – Measuring the Flexural Tensile Strength (Limit of Proportionality (LOP), Residual), EN 14651, CEN, Brussels, 2005.

11. RILEM TC 162-TDF, Test and Design Methods for Steel Fibre Reinforced Concrete – Final Recommendations, Materials and Structures, November 2002, Vol. 35, pp 579-582.

12. fib Model Code for Concrete Structures 2010, Earnest & Sohn, Germany, 2013.

13. Saldívar, H., Flexural Toughness Characterization of Steel Fibre Reinforced Concrete - Study of Experimental Methods and Size Effects, Doctoral thesis, Universitat Politecnica de Catalunya, Barcelona, Spain, 1999.

14. Barros, J.A.O, Cunha, V.M.C.F, Ribeiro, A.F. and Antunes, J.A.B., Post-cracking behaviour of steel fibre-reinforced concrete, Materials and Structures, 2005, Vol. 38, pp 47-56.

15. Nayar, S.K., Design of Fibre Reinforced Concrete - Slabs on grade and Pavements, Doctoral thesis, Indian Institute of Technology Madras, Chennai, India, 2015.

16. Jamet, D., Gettu, R., Gopalaratnam, V. S. and Aguado, A. Toughness of high strength concrete from notched beam tests, Testing of Fibre Reinforced Concrete, ACI: SP155, Eds. D. J. Stevens et al., American Concrete Institute, Detroit, USA, 1995, pp 23-39.

17. Oh, B.H., Mazzotti, C., Kim, J.C., and Choi. Y.C., Fracture behaviour of concrete members reinforced with structural macro-synthetic fibres, Engineering Fracture Mechanics, 2007, Vol. 74, pp 243–257.

18. Armelin, H. S. and Banthia, N., Predicting the flexural post cracking performance of steel fibre reinforced concrete from the pull out of single fibres. ACI Materials Journal, Vol. 94, 1997, pp 18-31.

19. Vandewalle, L., and Dupont, D., Bending Test and Interpretation, Proceedings of RILEM TC 162-TDF Workshop, Test and Design Methods for Steel Fibre Reinforced Concrete – Background and Experiences, PRO 31, RILEM Publication, 2003, pp 1-13.

20. Barr, B.I.G. and Lee, M.K., Round Robin analysis of the RILEM TC-162-TDF beam bending test: Part 2-Approximation of δ from the CMOD response, Materials and Structures, 2003, Vol. 36, pp 621-630.

21. RILEM TC 162-TDF, Test and Design Methods for Steel Fibre Reinforced Concrete, σ-e method, Materials and Structures, 2003, Vol. 36, pp 560-567.

22. di Prisco, M., Colombo, M. and Dozio, D., Fibre-reinforced concrete in fib Model Code 2010: Principles, models and test validation, Structural Concrete, 2013, Vol. 14, pp 342-361.


TECHNICAL PAPER

Sujatha Jose holds a Masters degree in Construction Engineering and Management from College of Engineering, Guindy, Anna University; pursuing her PhD at Building Technology and Construction Management Division, IIT Madras. Her research interests include special concretes, sustainability and durability in concrete.

Dr. Ravindra Gettu is a Professor of Civil Engineering at the Indian Institute of Technology Madras (IITM), Chennai. His current research interests are in fibre reinforced concrete, the effective use of chemical admixtures, self-compacting concrete, sustainability, housing and the mechanical characterization of construction materials. He is currently the Vice-President of RILEM, the International Union of Laboratories and Experts in Construction Materials, Systems and Structures.

Shabari Indhuja holds a Masters degree in Structural Engineering from Periya Maniammai University, Thanjavur. Her research interests include fibre reinforced concrete and studies on fresh properties of concrete.


POINT OF VIEW

Suggestions to improve method of installation of cast-in-situ driven and bored concrete piles

Kartik Chandra Ta

Construction of pile foundations requires a careful choice of piling system depending upon the sub-soil conditions, the load characteristics of a structure and the limitations of total settlement, differential settlement and any other special requirement of a project. The installation of piles demands careful control on position, alignment and depth and involves specialized skill and experience [1& 2] In this Paper suggestions on improvements in method of installation of only bored and driven cast-in-situ piles have been included. However, stipulations of relevant IS Codes i.e. IS 2911 (Part 1/ Sec 1)- 2010[1] and IS 2911 (Part 1/Sec 2)-2010[2] should be followed for ‘Design and construction of cast-in-situ driven piles’ and for ‘Design and construction of cast-in-situ bored piles’ respectively.

A) GENERAL

1. Sub-soil Investigations

Detailed sub-soil investigations should be conducted with trial borings, standard penetration tests(SPT), location of ground water table, chloride and sulphate content in soil and ground water etc for a depth at least 10m below proposed founding level, particularly in a major project site to check variations in strata, existence of any cavern, old well etc.

In order to protect green pile concrete with reinforcements from action of high chloride and sulphate content in soils and adjoining ground water, stipulations of Table 4 of IS 456:2000 [ 3 ] should be strictly followed. If chloride content in ground water is too high (i.e above 25, 000 ppm) then expert advice shall be taken for remedial measures and /or use of alternative solution.

The author encountered at a project site chloride content 130, 000 ppm in ground water in a creek land filled with dredged material from Arabian sea where on the basis of expert advice precast concrete piles were driven under a multi-storeyed building. By use of precast piles concrete strength was fully attained before driving and the effect of chloride content on such concrete would be inconsequent.

It has been established in a recent research at CSIR-SERC , Chennai [4] that the chloride ion penetrability (tested through RCPT) of Portland Pozzolona cement was nearly one-third of corresponding OPC concrete and the RCPT value of concrete with Calcium Nitrite Inhibitor (CNI) was found to be significantly more than that of other concretes without CNI. Thus before using Corrosion Inhibitor, properties of the particular Inhibitor shall be thoroughly checked through relevant tests.

2. Survey

a) Survey of pile positions should be done using precision survey instruments handled by experienced Surveyor and not in a casual manner. Additional and special care shall be taken in survey particularly for location of single and two-pile groups as large static eccentricities of load due to pile being out of position would induce moments in piles for which they are not normally designed.

b) Pile Nos should be indicated on piling plan and the same should invariably be mentioned in all records maintained.

3. Sequence of installation of piles

a) In an isolated building installation of piles should be started from centre fanning outwards. Again in a large


POINT OF VIEW

pile group similar trend should be followed i.e start from centre and proceed towards outer edges.

b) Disturbance of freshly laid concrete in a driven/bored pile shall be avoided by not installing the adjacent pile in the same group. Rather, principle of staggering should be followed and piling rig should be rolled/shifted by a suitable distance to start next pile in a large pile group before return again to first pile location. By this heaving of soil between piles in the same group also can be avoided. If pile groups are small , installation of pile(s) in two such groups at a distance should be taken up simultaneously.

4. Location of bore well at piling site

No bore well for withdrawal of ground water for various uses at site shall be installed under or near the edges of the building with pile foundation within influence zone of drawdown curve of the bore well to avoid settlement of building foundation.

The Author had a bitter experience of working in construction of a six storeyed warehouse building of flat slab construction resting on cast-in-situ driven piles in dock area in Kolkata where a few bore wells were installed in restricted building area itself for drawing ground water for gravel wash, mixing concrete etc . The level of bottom of piles 20 m long and of the strainer of borewells was more or less the same. By the time the roof of the structure (approx 150 m x 55m) with two expansion joints was cast with gradual progress in brickwork, flooring etc, at lower floors cracks were noticed at various locations in the structure alongwith uneven settlement at expansion joints. Also along length and width of the building settlement was noticed in a saucer shape after checking levels of top of pile caps which were initially maintained at datum. Maximum settlement at middle of length and width of the building was 18 cm in spite of the building resting on piles installed by a reputed Co.

As a remedial measure to save the building and for its future use , drawing water from the bore wells had to be immediately stopped and after detailed investigations a short dummy basement floor was introduced connected to the pile caps and tie beams.

5. Monitoring of settlement of pile founda-tion

It is advisable to record settlement of pile foundation particularly under high rise buildings and major structures during construction and for some time thereafter till full

live load is applied to ascertain that total/differential settlement is within the allowable settlement for the particular structure. Normally such levels are measured on top of pile caps or at a fixed height on basement/ground floor columns/piers.

Group effect with long term settlement in tall buildings should be considered in deciding pile capacity particularly when resting on silty/clayey soil with high water table. If the entire foundation area of a building is occupied by piles in mixed type of soil, due to bulb effect the whole foundation as a unit may produce settlement in the structure more than permissible limit for the particular type of cladding and finishes used in the building even before load capacity of individual pile based on pile load test is reached.

The Author has the knowledge of reducing height of Tata Centre building in Kolkata , one of the tallest buildings in those days(1965-66), by three floors when due to bulb effect settlement monitored for the building resting on maximum No of cast-in-situ driven piles at minimum spacing for the whole building area was about to exceed the permissible limit.

Again, it is reported that Urbana Towers in Kolkata (45 storeyed) on pile foundation (55 m deep ) recorded a settlement of 1 mm during Nepal earthquake (7.8 in Richter scale)in April, 2015 while still under construction.

6. Pile Concrete

a) Grade of concrete in piles should not be less than M25 and cement content not less than 400 kg/cum of concrete when ground water table is met.

b) Slump of concrete at the time of placing shall be between 150 and 180 mm whether in dry bore or below ground water table.

7. Record of installation of piles

Record of installation of piles shall be maintained as per format in IS code for cast-in-situ driven piles [1]and as per format in IS code for cast-in-situ bored piles [2].

8. Pile Integrity Test (PIT)

i) In major projects now-a-days 100 p.c piles are subject to low intensity PIT and such test is included as a schedule item in contract for which cost is not very high and many agencies in India can conduct this test.


POINT OF VIEW

ii) Even if not included as a regular item in Bill of Quantities in contract, in case of abnormality and/or doubt in quality of pile noticed from installation data in record i.e driving/boring record, quantity of concrete consumed, disruption in pouring of concrete, dislocation of pile shoe, honeycombing, cross-sectional changes, dis-continuity etc, low intensity pile integrity test should be conducted on all such piles before proceeding with construction of pile caps. If result of such test is not satisfactory, load test may have to be conducted on such pile or else pile may have to be rejected as per decision of the Engineer.

B) DRIVEN CAST-IN-SITU CONCRETE PILES

1. Pile driving formula

Any established dynamic pile driving formula may be adopted for driving of piles at site giving due consideration to limitations of various formulae. Minimum grade of concrete to be used in pile foundation is M25 [1].

2. Choice of pile type

a) This type of piles is widely used where at a particular site variations in soils strata is too much and length of pile cannot be easily ascertained in advance from sub-soil investigation results. At such sites wide variation of length of piles is normally noticed.

b) If piles are to be driven close to adjoining buildings not supported on pile foundations, proper precautions shall be taken to avoid any damage to the building structure. In such situations installation of bored piles would be a better option.

c) Due to limitation in capacity of pile driving rigs available in India, maximum dia of cast-in-situ driven piles is limited to 600 mm. Thus there is limitation of use of driven piles where load per pile is very high.

3. Pile Shoe

Cast iron conical pile shoes were mostly in use for driven cast-in-situ concrete piles to start with. But due to inconsistency in quality of casting of such shoes a large percentage of shoes started breaking and got damaged during driving. Remedial measures involved additional work by way of extraction of the tube, refilling of pile hole by soil /sand and re-installation of pile causing wastage of time and money apart from embarrassment to the Co.

As a solution now-a-days mild steel pile shoes of flat type are commonly used as shown in Figure 1 (with hexagonal and round base plate). However fabrication of shoes should be done carefully to avoid too much play in space between inner and outer ring. This may cause slanting of pile shoe during driving which may allow entry of adjoining soil. Also design of pile shoe should be done properly with adequate factor of safety to avoid damage during hard driving, if any.

4. Dolly

Timber dolly had been in use in India since inception of driven piles for placing on top of drive cap for transmission of load from hammer to pile tube. Depending on quality of timber used in dolly, hammer weight and height of fall, one timber dolly, normally 20 cm thick may be used in installation of one or more piles. Replacement of worn-out dolly during pile driving is time consuming.

For economy, alternative material e.g masonite board 10cm thick is in use in India now-a-days with which 20-25 piles may be installed even using pneumatic/ hydraulic hammer of higher capacity.

The Author used imported cushion 12 cm thick with 7 t hydraulic hammer in installation of driven piles in a project at Bhavnagar(in 1997) and it was possible to drive upto 400 Nos piles (40cm dia x 20 m long)in marine clay layers with one cushion.

5. Pile Tube

a) No of joints in pile tube of required length should be as minimum as possible. Seamless pipes, each of large length, say 8 m and above should preferably be used to minimize No of joints. Welding in joints where used

Figure 1. Mild steel pile shoe with hexagonal and round base plate.


POINT OF VIEW

shall be done with V-cuts by engaging skilled welders to maintain a straight line avoiding any protrusion of weld inside the tube. Again, cooling of joints after welding shall be done without development of any crack. Shifting and /or lifting of pile tube after welding should be done only after the joints are properly cooled to ambient temperature. Conventional A-frame driven piling rig is shown in Figure 2.

b)At a site where the length of pile tube falls short of length of pile to be installed, piling contractor tends to use ‘Follower Tube’ to make up the deficit length retaining the height of piling rig already mobilized at site. However, such proposal should not be encouraged as some amount of energy gets dissipated at junction between main pile tube and follower tube resulting in total length of pile installed with follower falling short of length required to sustain design load.

If on the basis of sub-soil investigation the depth of pile is decided and piling rig and pile tube are mobilized at site and subsequently during initial load test it is ascertained that length of pile required would be more, in that case either the length of rig should be extended at site , if possible , or else a taller rig should be mobilized along with lengthening of pile tube.

6. Problems normally faced during installa-tion and their solutions

i) Where ground water table is high or artesian condition persists, during driving of pile tube the bottom of tube may get filled with ground water for some depth. After final set of pile tube, inside of it should be checked for presence of accumulation of ground water. If depth of such water is more than 30 cm the tube should be extracted, pile bore refilled with selected soil and re-driving of pile tube should be done.

However, if such water accumulation is frequent and of routine nature, caulking/sealing of groove in pile shoe where the tube rests during driving, should be done with raw jute and bitumen to prevent entry of ground water during driving.

The Author has the knowledge of two piling projects in Kolkata in which due to not checking such water accumulation of water at bottom of pile tube on completion of driving , the piles failed during load test as the bottom concrete was of sub-standard quality. Since it was not possible to extract such piles, load carrying

capacity had to be reduced, redesign of pile groups done and additional piles installed by the contractor at his own cost resulting in cost and time overrun.

ii) a)Particularly when driving is very hard, due to excessive pull during extraction of pile tube with concrete and pile cage inside, sometimes snapping of wire rope or breakage of pulley block/sheave etc may take place. In such a situation it is a common practice to reduce friction by punching one or more empty bores and scoop out soil around the pile tube for its easy extraction . The earlier it is done, the better. Subsequently extracted tube is cut into pieces of suitable length preferably at welded joints where feasible, break and remove concrete and reinforcements and then re-weld carefully maintaining straight length for the tube and leaving no protrusions inside the tube.

To avoid such a situation, regular maintenance of piling engine and winch should be done alongwith increasing No of sheaves in pulley block and replacement of wire rope whenever first sign of wear is noticed.

b)The pile bore and punched holes after removal of jammed pile tube shall be filled with soil/sand as specified and left for sometime before re-driving pile at the old location.

iii) If on a rare occasion pile tube during extraction is lifted with pile cage and concrete inside since green concrete is still to set the tube should be removed outside pile location and by hammering on top of the tube it should be emptied. The empty pile bore should then be filled up and further procedure followed as in (ii) (b) above.

Figure 2. Conventional A-frame rig moving on roller over timbers.


POINT OF VIEW

iv) If pile cage alone is extracted during extraction of pile tube leaving behind concrete inside the bore, again on a rare occasion, due to protrusion of weld inside tube/ breakage of cover blocks resulting in touching of cage on side of pile tube etc , such pile shall be rejected, pile group to be redesigned and additional pile(s) as required would be installed.

v) Sometimes stiff clay or dense sand layer of relatively small thickness 1 to 1.5 m exists followed by soft layers before reaching founding stratum . It may cause very hard driving with available equipment mobilised at site being unable to puncture this stiff layer. In such cases attempt shall be made to extend the pile to required depth by mobilizing hammer of heavier weight , increasing height of fall of hammer and using heavier pulley block for extraction etc.

The Author has the knowledge of foundation of multi-storeyed Terminal building at Kolkata airport where cast-in-situ driven piles were installed by one piling contractor (in 1963-64) using steam hammer when with the system and equipment available with the Co, a stiff clay layer about 1.5 m thick at approx 18m depth could not be punctured and piles were terminated somewhere in that layer. During load test, however, the piles failed to carry the design load. Subsequently, additional piles had to be installed on the basis of re-design of pile groups and pile caps at contractor’s own cost.

During execution of second phase of the project, with detailed sub-soil information available, the selected contractor installed driven piles to required length and capacity using higher capacity of drop hammer and increased height of fall of hammer successfully puncturing the stiff clay layer and pile lengths were between 22 to 24 m.

7. Concrete compaction and consumption in piles

a) Tamping of pile tube during extraction after concreting is a standard practice which should be rigorously followed for artificial compaction of pile concrete in the bore apart from gravity compaction under weight of concrete mass with high slump.

b)In normal mixed soil condition volume of concrete consumed in driven cast-in-situ pile does not exceed theoretical volume of pile for the driven length.

However, the Author came across a piling site at Bhavnagar near sea coast where thick marine clay deposits existed. During installation of driven piles when pile tube was extracted, marine clay of about 5 cm thick used to get stuck to the outer periphery of the tube. As a result, actual volume of concrete consumed in 40 cm dia pile was between 20-25 p.c. more than theoretical volume of pile. After extraction of tube it used to take some time to get the clay layer dried up in prevalent windy condition at site when cracks appeared in clay deposit on tube surface which was then removed fast by hammering before shifting the tube to new pile position.

8. Sophistication in pile driving and quality control of pile concrete

i)The Author used crawler mounted Pile Driver(imported), Hydraulic Hammer (7 t and 10 t), Vibrator cum Extractor, concrete batching plant and transit mixer in installation of driven piles 40 cm and 45 cm dia upto 22m length in marine clay deposit for a plant at Bhavnagar, Gujarat . refer Figures 3 and 4. Standard practice followed by piling contractors in India is transporting concrete from 10/7 or 14/10 mixer by chute cart and placing into pile tube with help of winch and wire rope. However, at Bhavnagar site concrete from automatic batching plant was transported by transit mixer and directly poured into pile tube.

By this record No of piles could be installed in a day by one Pile Driver i.e 40 Nos 40 cm dia 20-22 m long, achieving 36 minutes cycle time.

Figure 3. Pile Driver with Hydraulic Hammer and Vibrator-cum-Extractor on crane.


POINT OF VIEW

ii) For fast completion of installation of large No of piles in the project, a few conventional driving pile rigs were mobilized. Even for pouring concrete directly into the pile tube at an elevated position after part extraction of the tube, special arrangement was made as shown in Figure 5.

iii) Because of existence of soft soil strata at top, adjacent piles could not be installed in sequence in the same group and the pile driver had to be shifted to next group (s) and later return to the first group after allowing enough time for concrete to set. Since the Pile Driver was crawler mounted its movement was very fast as against conventional piling rigs moving on rollers over timbers .

iv) At bottom level of pile cap pile concrete should be of sound quality ( ref Figure 6). If not, dismantling shall be done upto the level of sound concrete and then pile is rebuilt with design mix concrete.

9. Piles in rake (Raker piles)

In high seismic zones to cater to high horizontal forces, it may be economic to provide piles in rake under tall structures, bridges and heavy machineries etc. Such piles would be able to resist large horizontal forces apart from

vertical load. There are some special requirements for installing these piles as stated below:

i) Vertical conventional piling rig is normally converted into a raker pile rig attaching one trapezoidal piece of pile frame at bottom of a vertical rig. Precautions shall be taken in fabrication of the components for matching and accurate assembly so that the hammer moves through inclined guide frame smoothly without getting detached which may cause serious accident. Also close inspection of the hammer and guides shall be done to check wear and tear of hammer and guides.

ii) Checking of rake in pile tube shall be done frequently during driving, particularly when stiff resistance is encountered in driving through hard soil strata.

iii) Rake should not normally be flatter than 1H in 6V and preferably be limited to 1H in 7V as beyond this slope,

Figure 4. Vibrator- cum- Extractor in grip of pile tube for extraction

Figure 5. Set-up for pumping concrete from Transit Mixer to pile tube at a height.

Figure 6. Sound pile concrete at pile cut-off level.


POINT OF VIEW

compaction of concrete and the quality of pile may be doubtful.

C) BORED CAST-IN-SITU CONCRETE PILES

1. Dry bore with no ground water table

At some sites where water table is not encountered upto termination level of pile and the bore can stand on its own, there is a tendency to install the pile without the aid of bentonite mud. In such cases it is to be checked against collapse of soil, if any, till time of concreting including when pile reinforcement cage is lowered inside the bore. If any appreciable soil deposit is noticed at pile bottom due to above which can cause settlement of pile under load, flushing of bore with water or bentonite should be done for its removal.

2. Use of M.S. Liner

a) It is a common practice to use Mild Steel (M.S) liner for top 2.0- 2.5 m length of pile where sub-soil water table is high and top soil is soft. Thickness of liner plate should be such that round shape of the pile is retained. Depending on contract condition such liner may be retained or extracted after pile concreting.

b) If length of temporary liner used in any project exceeds 3 m such additional length should be provided through threaded joints with pieces 1.5 to 2 m in case of Hydraulic rigs (refer Figure 7) for easy and faster installation as well as extraction on completion of concreting in pile.

c) As per standard practice pushing of liner should precede boring and not that after boring, the liner should be pushed in the hole.

d) For special use of large diameter bored cast-in-situ piles in marine structures reference is drawn to Annex E of IS 2911( Part 1/Sec 2)[2] and refer Figure 8.

When liner is provided as sacrificial in case of piles in standing water or for marine structures for part or full length as applicable, thickness of liner plate should be sufficient as required for handling and driving/pushing the fabricated liner of required length, retaining its circular shape for the particular dia of pile. Again, all precautions shall be taken for safety of mobile gantry, piling rigs mounted on the same and stability of piles during high floods with velocity and current in water, while constructing river front structures. Suitable M.S bracings are to be provided to connect liners/piles under temporary condition for such stability till permanent connection to pile heads is provided, refer Figure 9. Attempt shall be made to extend piles to safe depth below maximum scour level.

Construction of pile muffs with steel plates is a standard practice adopted for support of precast beams where used for deck structure.

3. Checking of alignment

i) Alignment of piles, vertical or in rake, where provided, should be checked during installation of pile at depth intervals of 2 m, unless any closer interval is mentioned in contract document.

ii) During installation of raker piles through Hydraulic Rotary Rig inclination as required shall be provided in Figure 7. Bored piles under installation by Hydraulic Rig.

Figure 8. Marine bored piles in open sea condition


POINT OF VIEW

the mast and piling should proceed. Once in a month or earlier in case of doubt mast inclination shall be checked.

iii) If tripod rig (refer Figure 10 ) is used with bailer / chisel for installation of raker piles in marine condition in particular where liner of large length is exposed and standing above bed and if the rake is flatter than 1 Hor. on 6 Vert. then thickness of permanent liner shall be suitably increased to compensate for the loss of liner thickness due to friction developed during movement of bailer / chisel along inside face of liner in inclined position. If thin liner plate is used it may result in formation of pockets due to wear and tear of liner plate and loss of mortar / concrete through such pockets resulting in formation of poor quality of pile.

The Author has seen in one marine job formation of big openings in 6 mm. liner when pile was installed in rake 1 Hor : 2.5 Vert and lot of cement mortar and concrete leaked through the openings with compromise in quality of such permanent piles.

4. Maintenance of records during boring

i) Details of sub-soil strata as encountered during boring operation should be recorded with change of strata along with special information, if any as boring is advanced upto founding level. For any major and very important project with large No of piles, an experienced Geologist should be deployed for ascertaining correctly the types of strata as encountered.

ii) Standard Penetration Tests (SPT) should be conducted to determine founding level of pile suitable for carrying design load with required factor of safety.

5. Collapse during pile boring operations

i) During pile boring collapse of soil in borehole results in increase of dia of pile causing irregular increase of clear cover over reinforcement bars which would affect quality of pile apart from increasing cost due to increase in quantity of concrete consumed.

ii) On completion of pile boring after withdrawal of DMC / RMC pipes and lowering of reinforcement cage flushing of pile bottom is done generally using tremie pipes. After such flushing if concreting is not started soon and the bore is allowed to remain unattended for long, collapse of bore hole is very likely to take place affecting quality of pile adversely due to deposition of loose soil at pile bottom. Under load soft deposits would get compressed and settlement of pile would take place. In such conditions flushing of bore shall be repeated before commencing concreting in pile.

iii) In case of boring done with rotary hydraulic rig, cleaning bucket shall be attached to the Kelly when founding level is reached, for cleaning the base of each pile. Airlift technique may be used as an alternative method for cleaning of borehole.

6. Reasons for collapse of soil in pile bore

These mainly depend on type of soil i.e. whether cohesive or non-cohesive, its density, location of water table and pore water pressure, vibration effects, height of unsupported vertical face of wall and liquid inside borehole.

Presence of liquid inside a borehole increases the stability of the vertical cut earth surface of the borehole. When soil collapses total active pressure becomes more than cumulative passive pressure. Presence of liquid inside the borehole exerts counter pressure to resist the active pressure on the soil and facilitates the stability of the soil. Quantum of pressure exerted by the liquid on the Figure 9. Mobile piling gantry.

Figure 10. Bored piling using tripod rig


POINT OF VIEW

vertical cut surface of the hole depends on the density of the liquid.

7. Density requirement for Bentonite mud

a) Specification for Bentonite powder is covered in IS 6186 – 1986 (reaffirmed 1997) which is a characteristic type of fine grained clay being an alteration of volcanic ash containing not less than 85 p.c. mineral montmorillonite. It is used for various purposes other than stabilisation of pile bore. When used in bored piling work its property requirement satisfying IS 2911 (Part 1/ Sec. 2)[2] as stated in paragraph 7 (b) & (c) below shall be satisfied.

b) In cases other than mentioned in para B- 1 above borehole is generally stabilized by use of bentonite mud. For fresh bentonite suspension density should be between 1.03 and 1.10 g/ml. depending on pile diameter and type of soil encountered. Density of bentonite after contamination with drilled material in the bore may rise upto 1.25 g/ml. which shall be brought down to at least 1.12 g/ml. by flushing with fresh bentonite before concreting.

c) Also other recommended tests for bentonite mud in IS 2911 (Part 1) Sec. 2 – 2010 [2] are :

i) Liquid limit of bentonite shall be 400 p.c. or more when tested as per IS 2720 (Part 5).

ii) Marsh viscosity of bentonite suspension when tested by a standard Marsh Cone shall be between 30 to 60 sec. In special cases it may be allowed upto 90 sec.

iii) The pH value of bentonite suspension shall be between 9 and 11.5.

8. Bentonite types and how bentonite sus-pension helps in stabilization of bore

i) Normally two types of bentonite are available i.e. Sodium based and Calcium based. Bentonite is mostly available as calcium based in nature but as per requirement it is altered to sodium based to acquire requisite properties for specific purpose of stabilisation of vertical cut soil surface. It is not possible to distinguish between calcium based and sodium based bentonite only by mere visual inspection. For this pH value and swelling index of the two have to be tested.

ii) Sodium based bentonite has got higher liquid limit than other clay minerals and even much more than

calcium based bentonite. Due to thixotropic properties the sodium based bentonite solutions form a gel and in turn due to osmosis effect the gel is transferred to cake. Sodium based bentonite normally takes 12 hours to get the maturity.

iii) Action of bentonite in stabilizing the sides of boreholes is primarily due to thixotropic property of bentonite. This property permits the material to have the consistency of a fluid when introduced into a trench or hole. When left undisturbed it forms a jelly like membrane on the borehole wall and when agitated it becomes a fluid again.

iv) In case of granular soil bentonite suspension penetrates into sides under positive pressure and after sometime it forms a jelly. The suspension then gets deposited on the sides of the hole and makes the surface impervious and imparts a plastering effect.

In case of existence of impervious clay in the bore however the bentonite does not penetrate into the soil but deposits only as thin film on the surface of the borehole. Under such condition stability is derived from hydrostatic head of the suspension [5]. It is thus recommended to have top level of bentonite suspension in a borehole at least 1.5 m. above ground water table.

9. Concentration of bentonite suspension[5]

To achieve requisite density of the bentonite solution the concentration of bentonite (as p.c. of bentonite powder by wt. to be mixed with potable water ) is calculated as under

γs = 1 + 0.006 × Cs Where γs = Density and Cs is concentration of bentonite by wt.

Example : To achieve density 1.05 g/ml. of bentonite suspension,

1.05 = 1 + 0.006 × Cs or Cs = 0.05 ÷ 0.006 = 8.33 p.c.

i.e. in 1 m3 (1000 litres) of water 83.3 kg. bentonite powder will have to be added to get density of 1.05 g/ml. of bentonite suspension.

10. Flow of bentonite mud

For economy in use of bentonite mud the overflowed bentonite solution from the pile bore during boring is collected by gravity flow through a temporary channel normally made on ground by clayey material and stored


POINT OF VIEW

in a tank called ‘Contaminated Bentonite Tank’. The bored muck mixed with bentonite suspension is collected in this tank and after a lapse of time the bored muck is precipitated and the rejuvenated bentonite solution is calmly overflowed and collected in another adjacent tank called ‘Rejuvenated Bentonite Tank’. To maintain required density of this rejuvenated bentonite solution and to continue boring with this solution either matured bentonite jelly or fresh bentonite solution will be mixed time-to-time with this rejuvenated bentonite solution, agitated or recirculated for homogeneity.[5]

To eliminate calcium ions generated from cement slurry and accumulated in the rejuvenated tank, special additives e.g. Phosphates and ‘Sodium Hydroxide’ may be used. Again when sand content in contaminated bentonite is high particularly in case of piles in sandy strata bentonite solution is passed through De-sander where most of the sand particles are separated out and bentonite solution coming out of De-sander is circulated in the system after checking density requirement.

The Author has noticed at several bored piling sites use of only one bentonite tank for mixing and storage of used bentonite which is against standard practice.

11. Grouting of pile base/sides

a)In order to limit settlement of bored pile particularly where differential settlement between consecutive pile groups in continuous spans is critical, even in case of piles resting on rock cement grouting of pile base is sometimes adopted in important structures.

At Sirsi circle flyover in Bengaluru such cement grouting has been adopted to seal fissures in rock and improve bearing capacity.

c) Even cement grouting of sides and bottom of bored pile leaving MS Pipes at certain depth below cut-off level is done to increase side friction and ultimate skin friction resistance of pile in some projects in mixed soil condition.

At a gas based power plant at Kawas, Gujarat , this practice has been followed at a pressure 50 kg/sq.cm.

12. Bored Piles in rock

i) Mechanical Rotary rigs are suitable for installation of bored pile in normal soils and not at all suitable for rock strata where Hydraulic Rotary Rigs are to be deployed.

ii) Socketing of pile with low capacity Hydraulic Rotary rigs and ordinary rock cutting tools is difficult beyond 15 to 20 cm depth in hard rock strata of strength above 200 kg/sq cm. Specially designed rock cutting tools i.e rock auger with high capacity of rigs is essential to extend socketing between D and 1.25 D (where D = dia of pile) normally recommended in such rocks , ref Figure 11.

iii) Weight of chisel and capacity of operating winch shall be commensurate with diameter of pile, ultimate crushing strength(UCS) of rock, depth of socketing required in rock etc.

The Author has seen at a few project sites use of chisel of wt between 1.5 t and 2 t with winch capacity 3.5 t to install 1200 mm dia pile in hard rock with UCS 300 kg/sq cm and based on penetration of 2cm per hour the pile has been terminated even before achieving 1 D penetration in rock which is not a sound practice. One such chisel is shown in Figure 12. In such cases at least 3.5 t chisel, refer Figure 13 operated by 7.5 t winch should have been used for a satisfactory job.

iii) In hard rocks where Standard Penetration Test (SPT) cannot be conducted beyond N value greater than 200, the capacity of pile socketed in rock should be based on strength of rock as per design guidelines.

Figure 11. Rock Auger fitted to Hydraulic Rig


POINT OF VIEW

iv) When rock strata is reached and Hydraulic Rotary Rig is unable to drill through the same sometimes after one hour of drilling with negligible progress the pile is terminated at that level which is not a fair practice.

Such pile should be tested for load capacity and fulfillment of other criteria as per design requirement.

v) If rock socketing to a particular depth is mandatorily required and the Hydraulic Rig is not able to drill through the same with rock cutting tools available with the contractor it is a common practice to use tripod rig with heavy duty chisel of suitable weight. After drilling through overburden using Hydraulic rig the same is shifted and chiseling through rock is started with tripod rig.

However, risk with use of free fall type chisel in drilling in rock is that if rock strata is very hard and strength above 200 kg/sq cm control of alignment of chisel without a proper guide is extremely difficult. As a result a proposed vertical pile may be installed in rake.

The Author has experience in a marine project in basalt rock in Mumbai when some such vertical piles installed with tripod rig and chisel had to be exposed during adjoining excavation and rake measured was to the extent of 1 H in 5 V, ref Figure 14.

Figure 12. Rock cutting chisel of sub-standard quality in use.

Figure 13. Heavy duty rock cutting chisel

Figure 14. Vertical bored piles installed in hard rock as raker pile ( rake 1 H in 5 V )


POINT OF VIEW

To avoid such a situation, where Hydraulic rig is used, depending on crushing strength of rock , cutting tool i.e rock auger should be purchased on advice from machine suppliers or suitable equipment i.e Reverse Circulation Drill should be mobilized under expert advice.

13. In case of Hydraulic Rotary Rig length of total telescopic Kelly mobilised with the particular rig should be sufficient to install bored pile upto the specified founding level. Also condition of joint between Kelly pieces and between Kelly and bottom tool should be checked periodically so that during boring / drilling operation neither any Kelly piece nor the tool gets detached and is lost in the borehole.

It is extremely risky to send a diver in bentonite suspension in a pile borehole to salvage a Kelly bar or tool detached from the Kelly of the rig.

14. In case while boring in silty / sandy strata with Hydraulic rig using drilling bucket a hard strata of soft clay or similar is encountered it is always advisable in such a situation to withdraw the bucket and fit drilling Auger to drill through this stiff layer as otherwise there is a possibility of damage of the bucket and detachment from the Kelly end. The bucket may be lost in the bore which may have to be abandoned and redesign of pile group with installation of additional piles etc. may be involved which is a very costly and time consuming process.

15. i) Bentonite pump mobilised for circulation and flushing of borehole in order to clean pile base to satisfaction shall be of adequate capacity for the dia and depth of piles under installation. As a guide, for installing 1200 mm dia pile of 20 m length, minimum capacity of bentonite pump for flushing and cleaning of pile base is 40 H.P.

The Author has seen use of only 15 H.P pump in such cases which is grossly inadequate for cleaning of pile base.

ii) Bentonite pump shall be located as near the pile bore as possible so that frictional loss in transport pipeline for bentonite upto the borehole location is minimum for maximum efficiency of the pump in cleaning of pile base.

16. Pile reinforcements

a) Where length of bored pile is large and dia is more than 600 mm it becomes difficult to handle, lift the pile

reinforcement cage in one length and place the same inside bore. In such case stiffener reinforcement in cage should be closely spaced and suitable spreader beam(s) may be used. If after all these it is not possible to place the cage in single length inside pile bore the same would be divided in two or more pieces and joints welded after hanging the lower piece of cage inside pile bore by steel bar of suitable capacity placed on top of temporary liner.

b) Clear cover to pile reinforcements, laterals as links or spirals and stiffener rings shall be provided as per Para 6.11.4 of IS 2911 (Part 1/Sec 2) [2]. Cover blocks shall be strong and provided in enough Nos so that the cage retains verticality and does not get tilted.

17. Actual quantity of concrete consumed in a bored pile of certain nominal dia and length should be compared with its theoretical volume and average for about 20 Nos. piles should be calculated to ascertain increase in quantity over theoretical quantity as ‘over break’ p.c. This should normally be between 5 and 8 p.c. If increase is beyond 8 p.c. reason should be investigated. Also if in any bored pile actual volume of concrete consumed is even less than theoretical volume it is considered unusual and reasons should be ascertained through Integrity Test if no other apparent reason is established.

18. Recent advances in technology for instal-lation of bored piles in India as noted by the Author

i)As an alternative to well foundation which has been considered as a tricky and time consuming process for construction of bridges in wide and deep waterway, large diameter bored piles are being installed using hydraulic rigs mounted on mobile piling gantry or jack-up platform/ spudded pontoon by contractors in India. One such arrangement has been shown in Figure 15.

ii) Bentonite being available in abundance in India and also being cheap is mostly used for stabilization of bore during installation of bored piles. However, bentonite having potential of polluting water is not allowed to be discharged in river or aquifer/water source from where drinking water is supplied. In some states in Europe use of bentonite is banned for stabilization of boreholes for piles, diaphragm walls etc.

In India, alternative material is available. It is highly concentrated dry synthetic polymer designed specifically for chemical interactions with diverse types of soil


POINT OF VIEW

profiles. It has been successfully used in pile foundation and diaphragm wall construction in a number of private high rise building projects and metro construction where specified in contract. This polymer is bio-degradable and degrades in contact with concrete, non-toxic during all stages of use and after use as well. It is costlier than bentonite but where pollution of water source is involved, use of this polymer gets a preference in any case.

iii) In Sea-Link project in Mumbai, commissioned in Mumbai a few years back, bored piles upto 2 metres dia have been installed under pylon structures in cabled stayed spans in rock strata using Reverse Circulation Drills mounted on Jack-up barges. Socketing in rock has been extended upto 15 m depth in hard rock strata of strength upto 50 MPa. Also, special techniques have been adopted by erecting a steel tubular pile cofferdam for casting of a 6 metres thick pile cap resting on 52 Nos piles.

iv) In a Container Terminal at Kochi 1200 mm dia 75 metres long bored piles have been installed using Hydraulic Rig and 50 mm dia tor bars in 6m lengths using Dextra Bartec couplers.

v) In a No of high rise residential/ commercial building projects upto 63 storeys height on-going or recently completed, bored piles have been installed upto 1200mm dia and depth upto 62 metres below ground level.

D. LOAD TEST ON PILES (VERTICAL)1. Static load test on piles shall be conducted as per IS 2911, Part 4 [6]

2. Initial load test on pile should be conducted in a fair way to finalise length of pile required for the particular dia to carry the design load with required factor of safety. On the basis of performance of test pile any revision in design should be done and also any improvement in construction practice, if necessary, may be adopted.

3. Following alternatives are normally followed for vertical load test on piles;

a) By erection of loading platform with concrete blocks, refer Figure 16 or with bags normally filled with stone chips or sand where sufficient space is available.

b) Where enough space is not available around test pile due to adjoining running road or obstructions etc normally 4 Nos Anchor/Tension piles with in-situ concrete are installed and suitable truss/girder arrangement is made to provide reaction. In such cases precaution shall be taken to provide length of anchor pile considerably less than test pile so that upward force mobilised through friction on sides of anchor piles does not have influence on vertical load capacity of test pile. Also spacing of anchor piles should not be too close to prevent free settlement of test pile and in any case, not less than 3 times dia of pile under test..

c) Soil anchors or rock anchors in rake may be used to provide reaction load for static load test where rock exists at or some depth below founding level of pile. Normally 4 or 8 Nos H.T.steel anchors of suitable capacity are used for the purpose. Figure 17 shows one such load testing arrangement using 8 Nos anchors for load test on pile in a river bridge project.

Figure 15. Hydraulic Rig mounted on mobile gantry Figure 16. Pile load test using concrete blocks on loading platform


POINT OF VIEW

d) By taking help of adjacent piles at specified distance where available in a pile group with suitable arrangement for transfer of load in case of routine load test.

e) Where due to space constraints load test cannot be conducted by any of the above methods High strain type dynamic testing of piles for load capacity are in use in Indian projects now although these are conducted as per ASTM method and there is no BIS Code for such

tests. There are some controversies in this regard and it is reported that static load test results on a pile in the same group where all piles were installed under identical condition varied widely being on the lower side compared to pile load capacity recommended on the basis of dynamic testing.

In view of above it is suggested that at least one static load test and one dynamic test be done on two piles preferably in the same group installed under identical conditions to compare the pile capacity determined by each method for comparison and to establish design load for pile before proceeding with further dynamic tests.

f) Before start of pile load test, finished level of top of test pile shall be recorded using precision leveling instrument and net settlement on completion of test as recorded through dial gauges used in load test shall be compared with level of top of pile measured through leveling instrument after rebound.

References

1. ______ Driven cast-in-situ concrete piles, IS 2911 (Part 1/Sec 1) : 2010, Bureau of Indian Standards, New Delhi.

2. ______Bored cast-in-situ concrete piles, IS 2911 (Part 1/Sec 2) :2010, Bureau of Indian Standards, New Delhi.

3. ______ Code of practice for Plain and Reinforced concrete, IS 456 : 2000, Bureau of Indian Standards, New Delhi.

4. Bhaskar Sangoju, Ravindra Gettu and B.H. Bharatkumar, Study of the parameters governing the chloride induced corrosion of reinforcement steel in cracked concrete, The Indian Concrete Journal, March, 2017, Vol. 91, No. 3, pp. 37-48.

5. ‘Collapse of soil in pile borehole and its stabilization using bentonite suspension’ by Mr D. Bhattacharyya (Indian Roads Congress- Vol 68-3, Oct, 2007 ).

6. ______ Load Test on piles, IS 2911 (Part 4) : 2013, Bureau of Indian Standards, New Delhi.

Figure 17. Pile load test using rock anchors

Kartik Chandra Ta, BE(Civil), FICE(UK) retd, FIE(I),MIGS, Certified Lead Auditor on QMS to ISO 9001:2008 has total work experience of 52 years in design and construction of deep foundations, flyovers and bridges, marine structures, buildings with high quality finishes, Quality Control and Quality Assurance (QC/QA). He worked with Consulting Engineering Services in India and abroad for 19 years and with Simplex Infrastructures Ltd for 28 years in various positions including the post of Technical Director (QC/QA) at corporate level. At present he is working as Short Term Consultant to World Bank, Delhi office.


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