Steel-Concrete Composite Coupling Beams, Behavior and Design - Binginan Gong & Bahram M. Shahrooz

Engineering Structures 23 (2001) 1480–1490www.elsevier.com/locate/engstruct

Steel-concrete composite coupling beams — behavior and design

Binginan Gonga, Bahram M. Shahroozb,*

a S&B Infrastructure, Ltd., 3535 Sage Road, Houston, TX 77056-7011, USAb University of Cincinnati, Department of Civil and Environmental Engineering, 741 Baldwin Hall, PO Box 210071, Cincinnati,

OH 45221-00171, USA

Received 12 September 2000; received in revised form 16 January 2001; accepted 10 April 2001

Abstract

Structural steel/composite beams provide a viable alternative for coupling individual reinforced concrete wall piers. Well-estab-lished guidelines for shear links in eccentrically braced steel frames form the basis of current design guidelines. However, theseprovisions ignore the effects of nominally reinforced concrete encasement which typically surrounds the coupling beam, and arebased on overly conservative assumed deformation demands. A coordinated analytical and experimental research program at theUniversity of Cincinnati has focused on cyclic response of steel/composite coupling beams, their connections to reinforced concretewalls, and overall behavior of composite coupled wall systems. Using the results from this study, guidelines for proper design anddetailing of steel/composite coupling beams and beam-wall connections have been developed. This paper summarizes the researchprogram, and highlights the basic concepts, important findings, and recommendations. 2001 Elsevier Science Ltd. All rightsreserved.

Keywords: Composite construction; Coupling beams; Coupled walls; Cyclic testing; Mixed construction; Seismic design

1. Introduction

An efficient structural system can be achieved if theopenings in structural walls are arranged in a regularpattern. In this manner, a number of individual wall pierscan be coupled together to produce a system with largelateral stiffness and strength. The structural performanceat or near ultimate state can also be optimized by properdetailing of the coupling beams (i.e., the beams that con-nect the individual walls). Coupling beams should beproportioned to avoid over coupling (i.e., a system thatacts as a single pierced wall) and light coupling (i.e.,a system that performs as a number of isolated walls).Extensive past research [1–8] has led to well establishedseismic resistant design guidelines for reinforced con-crete coupling beams. Current design provisions [9] typi-cally result into diagonally reinforced deep beams inorder to satisfy the stiffness, strength, and energy dissi-pation demands. The diagonal reinforcement consists of

* Corresponding author. Tel.:+1-513-556-3677; fax:+1-513-556-2599.

E-mail address: [email protected] (B.M. Shahrooz).

0141-0296/01/$ - see front matter 2001 Elsevier Science Ltd. All rights reserved.PII: S0141-0296 (01)00042-6

relatively large diameter bars which have to beadequately confined to avoid buckling at advanced limitstates. Anchorage of the reinforcing bars in wall pierscan pose difficulties.

Structural steel or steel–concrete composite beamsprovide a viable alternative, particularly for cases withfloor height restrictions. In contrast to conventionallyreinforced concrete members, steel/composite couplingbeams can be designed as a flexural-yielding or shear-yielding member. Therefore, a desirable mode of energydissipation is achieved depending on the particular case.The main design issues are (a) proportioning anddetailing of steel/composite coupling beams, and (b)beam–wall connections. Depending on whether the wallboundary element consists of structural steel columns orreinforced concrete elements, the coupling beam–wallconnection is different. In the former case, the connec-tion is similar to steel beam–column connections. Theconnection in the latter case, which is the focus of thispaper, is achieved by embedding the coupling beaminside the wall piers and interfacing it with the wallboundary element. The embedment length will clearlyhave a major influence on the performance ofsteel/composite coupling beams.

1481B. Gong, B.M. Shahrooz / Engineering Structures 23 (2001) 1480–1490

A number of recent studies at the University of Cin-cinnati have been focused on examining seismicresponse of such composite systems. An overview of theresearch program is provided herein. Current designguidelines [10] are evaluated, and a number of changesare recommended.

2. Summary of current design guidelines

Steel coupling beams are designed according to theprovisions outlined in the 1997 NEHRP RecommendedProvisions for the Development of Seismic Regulationsfor New Buildings [10]. These provisions are based onseismic detailing of steel links in eccentrically bracedframes. The coupling beam rotation angle is arbitrarilyset equal to 0.09 rad. Note that for link beams this angleis computed. The assumed coupling beam rotation israther conservative. For example, if the beam length isless than 1.6 Mp/Vp (Mp=plastic moment capacity;Vp=plastic shear capacity, i.e., 0.6 Fy (d�2tf)tw in whichFy=yield strength, d=beam depth, tf=flange thickness,tw=web thickness), web stiffeners at (30 tw�d/5) willhave to be provided. Considering that most couplingbeams are typically short, current design leads to closelyspaced web stiffeners. Steel coupling beams are oftenencased inside nominally reinforced concrete elements,e.g., in door lintels. However, due to lack of informationabout the effects of concrete encasement, current designguidelines are tacit about cases in which the couplingbeam is encased. Accordingly, most engineers ignore theinfluence of encasement apparently on the basis that (a)the encasement is nominally reinforced and hence notexpected to significantly contribute towards strength andstiffness, and (b) the design will be more conservativeby not including the contribution of the encasing elementaround the steel coupling beam. As a result, details forpreventing flange and web buckling are identical to thoseused for unencased coupling beams, and the embedmentlength is calculated to develop the design capacity of thesteel beam only.

No specific guidelines are provided for computing therequired embedment length, but references are made toprevious studies [11–14] for further information. Thesestudies examined the applicability of two models pro-posed by Marcakis and Mitchell [15] and Mattock andGaafar [16]. In Fig. 1, Mattock and Gaafar’s model isillustrated. The applied shear (Vu) is resisted by mobiliz-ing an internal moment arm between bearing forces Cf

and Cb. A parabolic distribution of bearing stresses isassumed for Cb, and Cf is computed by using a uniformstress equal to 0.85 f�c where f�c=concrete compressivestrength in MPa. The bearing stresses are assumed to bedistributed uniformly over the beam flange width. Fol-lowing these assumptions and calibration against experi-mental data for steel corbels subjected to monotonic

Fig. 1. Mattock and Gaafar’s model for computing embedmentlength.

loading, this model calculates the required embedmentlength (Le) according to Eq. (1) in which twall=thicknessof wall pier, bf=beam flange width, and b1=ratio of theaverage concrete compressive strength to themaximum stress.

Vu�4.05�f�c�twall

bf�0.66

b1bfLe�0.58−0.22b1

0.88+aLe

� (1)

In this equation, the inflection point is assumed to beat the midspan; hence, the value of “a” is taken as onehalf of the coupling beam span. The model proposed byMarcakis and Mitchel generally results in a slightlylonger embedment length; however, the differencebetween the two models is negligible [13,14]. AlthoughMarcakis and Mitchell [15] and Mattock and Gaafar [16]originally developed their respective equations fordesign of steel brackets attached to reinforced concretecolumns, previous studies at the University of Cincinnatiand elsewhere [11–14] have shown that these modelsresult in acceptable performance for steel couplingbeams which are subjected to cyclic shear. Moreover,the calculated embedment length from either model isadequate to ensure a desirable mode of energy dissi-pation for steel, unencased coupling beams by formingthe plastic hinges in the beam rather in the connectionregion. Note that the value of Vu in Eq. (1) should betaken as the plastic shear capacity of the steel beam (i.e.,Vp=0.6 Fy (d�2tf)tw) to ensure adequate performance.

1482 B. Gong, B.M. Shahrooz / Engineering Structures 23 (2001) 1480–1490

Fig. 2. Plan view of selected prototype structure.

3. Response of steel–concrete composite couplingbeams

A recent study [17,18] has examined the effects ofconcrete encasement. The test specimens were extractedfrom a 20-story prototype building shown in Fig. 2. Forthe chosen span length of the steel/composite couplingbeams, current design guidelines [10] require stiffenerplates to be placed at (30tw�d/5) on one side of the web.As part of the experimental phase of this study, whichwas conducted in two phases, seven specimens weretested. The main test variables, summarized in Table 1,were (a) presence or lack of encasement, (b) the amountof web stiffener in the steel beam, (c) presence or lackof face bearing plates at the wall–beam interface, (d) thelevel of shear force for which the beam–wall connectionis designed, and (e) floor slab around the coupling beam.The encasement was nominally reinforced with four 4.9mm longitudinal bars and 4.9 mm transverse reinforce-ment placed at one-half depth of the encasing element,

Table 1Test specimens and test variables

Specimen I.D. Encasement Spacing of web stiffeners Force for connection design Face bearing plate Floor slab

Phase I1 No L Vsteel section No No2 Yes L Vsteel section No No3 Yes 2L Vsteel section No No4 Yes N/Aa Vsteel section No NoPhase II5 Yes N/Aa Vcomposite section No No6 Yes N/Aa Vcomposite section Yes No7 Yes N/Aa Vcomposite section Yes Yes

a No web stiffeners were provided.

see Fig. 3. A low-strength concrete (f�c=12 MPa) wasused for the encasement in order to accentuate nominalconfinement around the steel coupling beam. In all thespecimens, auxiliary transfer bars had been attached(through the use of mechanical half couplers) to the topand bottom flanges at two locations (Fig. 3) in an effortto aid in the transfer of bearing forces to the surroundingconcrete [13,14]. Face bearing plates in specimens 6 and7 consisted of a pair of 4.75 mm thick stiffeners weldedon both sides of the web. The face bearing plates werelocated inside the wall boundary element transversereinforcement. Another pair of stiffeners were also addedat 125 mm from the face bearing plates under the auxili-ary bars. The resulting concrete struts between theseplates (shown schematically in Fig. 3) are expected toenhance the performance by reducing the contributionof bearing stresses against the top and bottom flanges.The benefits of face bearing plates have been demon-strated in past studies on steel beam-reinforced concretecolumn connections [19], and as part of testing of speci-men No. 6. Additional details regarding the test speci-mens and other aspects of the experimental program areprovided elsewhere [18]. The focus of this paper is onspecimens No. 1, 4, 5, and 7.

3.1. (a) Strength characteristics

The load-deflection responses of the unencased speci-men No. 1 and encased specimen No. 4 are plotted inFig. 4. The hysteresis loops are stable, and exhibit a sig-nificant level of energy dissipation. Sudden drops duringthe last cycles are primarily due to fracture of wall trans-verse reinforcement passing through the steel beam web,and fracture of the weld around the mechanical halfcouplers which were used to attach the auxiliary bars tothe beam flanges [18]. Both specimens could developand exceed the capacity computed based on the meas-ured material properties [18]. For specimen No. 1, thecapacity is equal to the plastic shear capacity, i.e., 0.6Fy(d�2tf)tw. The shear capacity of specimen No. 4 was


Fig. 3. Specimen details.

computed by (a) superposition of the shear capacities ofthe steel section and concrete beam, or (b) a fiber crosssection analysis incorporating flexure as well as shear.The average value from these two analyses is plotted inFig. 4.

The nominally reinforced encasement around the steelcoupling beam is apparently adequate to prevent weband flange buckling at advanced yielding. Note thatspecimen No. 4 could develop 107 kN at a shear angleof 0.057 rad. At this shear deformation, current NEHRPprovisions [10] require web stiffener plates at 1.5 timesthe spacing used for specimen No. 2. Despite having nostiffeners, specimen No. 4 could develop shears corre-sponding to 1.10 Vp. Hence, nominally reinforcedencasement around steel coupling beams is sufficient toprevent web and flange buckling; and web stiffeners arenot needed.

3.2. (b) Energy dissipation characteristics

The dissipated energies are compared in Fig. 5. In aneffort to examine the performance of the specimens, thedissipated energy was separated into two parts: (a) theenergy dissipated by plastic hinges in the beam(“Beam” ), and (b) the energy dissipated by inelasticdeformations in the connection region (“Connection” )[18]. For specimen No. 1, the participation of beamtowards energy dissipation was more significant thanthat for specimen No. 4. Although the beam capacitiescould be developed, the performance of the encasedspecimen was not as desirable because inelastic action inthe connection region contributed more than the plastic

hinges in the beam. This behavior is attributed to theprovided embedment length.

3.3. (c) Revised embedment length

As indicated in Table 1, the required embedmentlength for specimen No. 1 and No. 4 was, accordingto current design guidelines and practice, computed todevelop the shear capacity of the steel coupling beamonly, i.e., 0.6 Fy(d�2tf)tw. The experimental data shownin Fig. 4 suggest that nominally reinforced encasementcan increase the capacity by as much as 23%. Clearly,if the provided embedment length is calculated todevelop a smaller shear, plastic hinges form in the con-nection region before fully mobilizing the full capacityof the composite beam. This performance is not desir-able. A capacity design methodology, in which theembedment length is computed to develop the capacityof the composite beam and not the steel beam alone, isproposed. This method requires that the contribution ofconcrete encasement towards shear capacity is takeninto account.

In lieu of detailed techniques such as fiber based mod-els [18], a simple method based on combining the shearcapacity of the steel beam (Vsteel) and encasement (VRC)appears to be a reasonable alternative, i.e., Vn=Vsteel+VRC

in which Vsteel=0.6 Fy(d�2tf)tw and VRC is computedbased on standard methods for reinforced concrete mem-

bers (e.g., [20:318–99]), i.e., VRC=0.166√f�cbd+Avfyd

swhere b=width of the encasement, d=effective depth ofthe encasement, Av=total area of transverse reinforce-


Fig. 4. Hysteresis responses of encased and unencased couplingbeams.

ment, and s=spacing of the transverse steel. In this sim-plified approach, the constitutive models for concreteand steel are idealized, and factors such as strain harden-ing of the longitudinal and transverse steel are ignored;the shear capacity of the reinforced concrete encasementis based on the ACI method in which the concrete contri-butions towards shear capacity from aggregate interlock,dowel action of the longitudinal bars, and the uncrackedconcrete are lumped together [21]; and the contributionof the concrete beyond cracking is not included [22].Therefore, this simple model needs to be calibrated sothat the computed capacity would be comparable to thatobtained from a more detailed analysis such as fibermodeling. A “correction factor” was determined by com-paring the capacity computed from detailed fiber sectionanalysis based on the modified compression field theory[22] and the capacity from the superposition method, i.e.,Vn=Vsteel+VRC. The parametric study included 24 casesin which the concrete compressive and the steel yield

Fig. 5. Distribution of dissipated energy.

as well as the ultimate strength, dimensions of the steelsections, and the overall encasement dimensions werechanged systematically. The ratio between the two com-puted capacities ranges between 1.31 to 1.85 with anaverage of 1.61 and a standard deviation of 0.075 [18].Within the limitations of this parametric study, a “cor-rection factor” of 1.6 is recommended and hence theshear capacity of composite coupling beams is taken asVn=1.6 (Vsteel+VRC). Additional studies covering a widerrange of parameters is necessary to further refine the rec-ommended correction factor.

3.4. (d) Evaluation of revised embedment length

In an effort to evaluate the performance of couplingbeams for which the embedment length is calculatedbased on the proposed capacity design method, strength


and energy dissipation characteristics of specimens No.4 and 5 are compared. Specimen No. 5 was generallysimilar to those tested in the first phase, except for alonger embedment length that the revised procedurewould require, and the testing method which includedwall overturning moment in contrast to the first fourspecimens for which the wall overturning was not simu-lated.

Normalized load-shear angle envelope curves ofspecimens No. 4 and 5 are compared in Fig. 6. The verti-cal axis is normalized with respect to the shear at theonset of web yielding (Vy). Shear angle at this stage (gy),was used to normalize the measured shear angles. Priorto yielding, the two specimens exhibit almost identicalload-deformation relationships. The response of speci-men No. 5 shows a remarkable improvement in termsof achieving higher ductilities and reduced strengthdeterioration beyond the maximum load. For example,at shear angle of 0.0885 radians, the load had droppedonly to 96% of the peak value. Specimen No. 5 alsodeveloped a larger capacity in the positive directionwhen the wall overturning moment produced compress-ive stresses around the connection. Under negative bend-ing, the two specimens developed rather similar loads,although specimen No. 5 reached a slightly lower loadthan specimen No. 4. This difference is attributed to thepresence or lack of wall overturning moment. The walloverturning moment in specimen No. 5 resulted in ten-sile stresses that reduced the level of bearing stress trans-fer between the beam flanges and the surrounding con-crete in the connection region. Despite these stresses, theload carrying capacity did not drop significantly belowthat for specimen No. 4 in which the connection regionwas kept under a constant gravity compressive stress.The longer embedment length in specimen No. 5, whichis the result of using the revised design methodology,

Fig. 6. Normalized load-deflection curves.

Fig. 7. Distribution of dissipated energy for specimen No. 5.

delayed the connection failure until a higher capacitycould be developed, and hence the enhanced behavior.

A similar observation is made by evaluating theenergy dissipation characteristics of specimen No. 5shown in Fig. 7 in which the total dissipated energy(“Total” ) is divided as discussed previously. The inputenergy was predominately dissipated by inelastic defor-mations in the coupling beam. Beyond shear angle of0.05 rad., the participation of the connection was gradu-ally increased although the plastic hinges in the beamoutside the connection continued to dissipate a reason-able portion of the total energy. This trend is differentfrom that observed for specimen No. 4 (see Fig. 5). Thelonger embedment length in specimen No. 5 evidentlyenhanced the energy dissipation characteristics by reduc-ing the contribution of the connection region.

Note that the contribution of floor slab is not includedin the proposed design method because (a) as seen inFig. 8 the contribution of slab is relatively negligiblebecause the additional tensile forces from the slab bars

Fig. 8. Moment-curvature response of composite coupling beam inprototype structure.


Fig. 9. Distribution of dissipated energy for specimen No. 7.

are relatively small in comparison to the tensile force inthe flanges of a typical steel beam, and (b) the floor slabwraps the connection region and reduces its participationas illustrated in Fig. 9 [18]. The distribution of dissipatedenergy clearly indicates that the slab and beam dissipatedthe majority of the input energy, and the connectionessentially did not participate. Therefore, the contri-bution of floor slab is neglected, and the embedmentlength needs to be calculated to develop only 1.6(Vsteel+VRC).

3.5. (e) Stiffness of composite coupling beams

The variation of peak-to-peak stiffness against shearangle is plotted in Fig. 10. The initial stiffness of speci-men No. 4 is 25% larger than that for specimen No. 1.The initial stiffness for specimen No. 5 is less than thestiffness of specimen No. 4 because this specimen wasaccidentally loaded due to difficulties in the computer-based control of one of the actuators used for loading.Specimen No. 5 had apparently experienced some minorcracks before the testing program was started.

Fig. 10. Stiffness degradation of various composite coupling beamspecimens.

The floor slab clearly increases the initial stiffness ofspecimen No. 7. However, at shear angle of about 0.005rad., the significance of slab is effectively lost. Beyondthis small level of deformation the stiffness of specimenNo. 7 drops to a level comparable to the initial stiffnessof the specimens without slab. When specimens wereloaded up to a shear angle of about 0.06 rad., all thethree specimens had reached about the same stiffness.Therefore, although floor slabs increase the initial stiff-ness of coupling beams, the contribution of the slabdeteriorates under small deformations and may beignored in seismic design and analysis.

4. Evaluation of impact of encasement on overallstructural response

The effects of the additional stiffness due to nominallyreinforced encasement around steel coupling beams,which is ignored in current design guidelines, wereevaluated analytically by examining the overall responseof the prototype structure (Fig. 2). The larger stiffnessobviously results in smaller vibration periods. Such ashift could influence the dynamic behavior if the fre-quencies coincide with the frequency band of the designground motion with high input energy content.

The demands in the walls and coupling beams areaffected by the changes in the coupling beam stiffness.The concrete encasement increases the coupling beamstiffness which in turn increases the level of couplingaction between the individual wall piers. For instance,the wall axial load in the first floor of the prototype struc-ture increases by 40% when the influence of encasementis taken into account [17]. Such a large increase couldpose stability problems if the wall boundary elementsare designed for forces calculated based on ignoring theencasement. In addition, the foundation system needs tobe designed for the increased demands in the walls. Theincrease in wall shear force, which is about 10%, is notperhaps as critical. The increases in beam design forcesare offset by the corresponding additional capacity dueto encasement. Therefore, design of walls and foun-dation systems needs to incorporate the effects ofencasement around steel coupling beams. The numericalvalues stated above are particular to the prototype struc-ture, and are based on elastic analysis. For other struc-tures with different geometries and stiffness character-istics, the increase in the stiffness and hence the changesin the design forces may be more or less. Moreover,cracking of the encasement around the steel couplingbeam will reduce the stiffness of the composite couplingbeam, and hence the increased coupling action will beless than that predicted from a simple elastic analysis.Nevertheless, in view of the potential higher designforces, the engineer should consider the increased stiff-ness of composite coupling beams as one of the variablesin the design model.


5. Evaluation of shear angle

Well established guidelines for shear links in eccentri-cally-braced frames [23] form the basis of current guide-lines [10] for design and detailing of steel/compositecoupling beams. The expected coupling beam rotationangle plays an important role in the required beamdetails such as the provision of stiffeners. Current designguidelines [10] arbitrarily set the coupling beam shearangle equal to 0.09 rad., and the beam is detailed accord-ingly. Such a large angle results in closely spaced webstiffeners. Note that the shear angle is computed for steellink beams in eccentrically braced frames.

In order to examine whether this level of shear angle(0.09 rad.) is reasonable and to understand the maximumexpected range of shear angle, the response of the proto-type structure was evaluated. A pseudo three-dimen-sional model of the prototype structure was constructed(Fig. 11(a)). The torsional and vertical springs used inthe model are intended to simulate the outrigger actionof the transverse members. The walls were modeled byan element (Fig. 11(b)) that incorporates axial, flexural,and shear deformations in the elastic and inelastic range[24]. Both dynamic and static push-over analyses wereconduced [18]. For static analyses, the lateral loads were

Fig. 11. Analytical modeling of prototype structure.

assumed to be distributed uniformly or triangularly, andthree ground motion records (1940 El Centro NS, 1989Loma Preita, and 1994 Northridge N60E) were used forthe dynamic analyses. Three different analyses wereconduced in which (a) the coupling beams were unen-cased steel members, (b) the influence of encasementaround the steel coupling beams was taken into account,and (c) flexibility at the foundation level was approxi-mately taken into account by placing vertical androtational elastic springs under the column and wallelements as shown in Fig. 11(c). The spring coefficientswere computed based on basic principles by assuming amodulus of sub-grade of 54,260 kN/m3.

The maximum computed coupling beam shear anglesfor various analyses are summarized in Table 2. Thelevel of shear angle for encased and unencased couplingbeams is considerably less than 0.09 rad. Only when thefoundation flexibility was taken into account and lateralloads were assumed to be distributed triangularly did thecoupling beam shear angle approach the value of 0.09rad. stipulated in NHERP provisions [10]. However, atthis shear angle the roof lateral drift exceeds 10% of thebuilding height, which is well above acceptable levels,and the base shear approaches 40% of the buildingweight. For other cases with reasonable drifts, the com-


Table 2Maximum coupling beam shear angle (rad.)a

Analysis Unencased fixed base Encased fixed base Encased “fl exible base”

Static push over: Rectangular lateral loads 0.0177 0.0198 0.0269(1.2%) (1.3%) (2.0%)

Static push over: Triangular lateral loads 0.0328 0.0378 0.0884(2.7%) (3.2%) (10.2%)

Dynamic:1940 El Centro NS 0.00391 0.00355 0.00388Dynamic:1989 Loma Prieta 0.00709 0.0075 0.00673Dynamic:1994 Northridge N60E 0.00208 0.0188 0.00209

a The values in the parentheses are calculated roof drifts at maximum reported shear angle.

puted shear angles are significantly less than 0.09 rad.It is deemed that similar observations are made for otherwell-proportioned buildings employing coupled walls.

Despite an effort to perform a reasonably completeanalysis of the prototype structure, major simplificationshad to be made, e.g., (a) the three-dimensional behaviorof the structure is modeled rather crudely, (b) simulationof the soil-structure action is very approximate, and (c)the parameters in the wall and beam hysteretic modelswere established based on subassembly tests with bound-ary conditions different from those in an actual building.The main reason behind the reported analytical studywas to explore the rationality of the assumed couplingbeam shear angle of 0.09 rad., in current design codes.Additional detailed studies that overcome the stated limi-tations of the current study are necessary.

Despite the simplicity of the analytical studies usedhere, the current assumed shear angle of 0.09 rad.,appears to be questionable and can lead to unnecessaryconservative detailing of steel coupling beams. Note thatthe reported test results (Fig. 4) show that stiffeners canbe eliminated for steel–concrete composite couplingbeams; therefore, the focus of this discussion is on steelcoupling beams. More rational techniques for computingthe value of shear angle are needed.

The coupling beam shear angle is computed with ref-erence to the collapse mechanism shown in Fig. 12which corresponds to the expected behavior of coupledwall systems, i.e., plastic hinges at the base of walls andat the ends of coupling beams. The value of plasticinterstory drift angle (qp) is taken as Cdqe (Cd=deflectionamplification factor defined by NEHRP [10]), where theelastic interstory drift angle, qe, is computed under codelevel lateral loads (e.g., [9,10]). Knowing the value of

qp, shear angle, gp, is calculated as gP=qP

Lwall

Lin which

Lwall is the distance between center lines of the wall piersand L is the clear span of the coupling beam. Previousexperimental data suggest that the “effective fixed point”of steel or steel–concrete composite coupling beams isabout 1/3 of the embedment length from the face of thewall [13,14,18]. Therefore, for consistency with experi-mental observations it is recommended to take Lwall as

Fig. 12. Analytical model for computing shear angle of couplingbeams.

L+0.6 Le in which Le is the embedment length of thecoupling beam inside each wall pier. Note that with theexception of the assumed collapse mechanism and therelationship between the shear angle and drift angle, theproposed method is similar to that used for steel shearlinks in eccentrically braced frames.

6. Summary and conclusions

Seismic behavior of steel and steel–concrete com-posite coupling beams was evaluated through a coordi-nated experimental and analytical research study. One ofthe main objectives of the reported study was to scrutin-ize current design guidelines, and to recommend modi-fications if necessary. Based on the reported study, thefollowing conclusions are drawn. These conclusions areclearly based on a relatively limited number of tests andanalytical studies. Additional test data from more com-plete subassemblies and more detailed analytical studiesare recommended to supplement the results reported her-ein.


1. Nominal encasement around steel coupling beamsprovides an effective means for preventing web buck-ling. Hence, web stiffeners can be eliminated. Currentdesign codes need to be re-evaluated and relaxed forthe cases where the steel coupling beam is encased.Available provisions appear to be overly conserva-tive.

2. Although current design procedures result in a con-servative design and detailing of encased steel coup-ling beams, the increased strength and stiffness dueto the surrounding concrete encasement could havedetrimental effects on the overall performance if theyare not taken into account as part of the design.

3. Nominally reinforced encasement around steel coup-ling beams is expected to increase the stiffness. Theadditional stiffness increases the level of couplingbetween walls, which in turn affects the distributionof design forces. Most importantly, the wall axial loadat the base could substantially increase. The increasedstiffness of encased coupling beams has to be incor-porated in design of coupled walls as well as foun-dation systems.

4. Unless design calculations consider the contributionof encasement towards shear capacity of compositecoupling beams, a significant portion of the inputenergy will be dissipated by inelastic deformations inthe connection region, which is not desirable. There-fore, the embedment length has to be computed todevelop the expected shear capacity of the compositesection. In lieu of refined fiber analyses, the shearcapacity may be taken as 1.6 times the sum of theshear capacity of the steel beam and encasement. Theenergy dissipation characteristics, ductility, and load-carrying capacity of composite coupling beams aresubstantially improved by using the proposed meth-odology which leads into a longer embedment length.

5. The contribution of floor slab towards stiffness andstrength of composite coupling beams may beignored. The additional stiffness due to floor slab islost after rather small deformations. Contribution offloor slab is less than that expected for reinforced con-crete beams because the area of slab reinforcing barswithin the effective slab width is small in comparisonto the flange area of the coupling beam. Theadditional strength may be ignored when theembedment length is computed as the slab preventsthe formation of plastic hinges in the connectionregion.

6. Relatively detailed inelastic static and dynamic analy-ses suggest that the maximum expected couplingbeam shear angle in reasonably proportioned coupledwalls is probably less than the value assumed by cur-rent design provisions. Until the availability of furtherdata, a simple procedure, similar to a well establishedmethod for link beams in eccentrically braced frames,is proposed to compute a more reliable estimate of

the expected coupling beam shear angle demands, andhence the amount of stiffeners, if necessary, can beestablished more rationally.

Acknowledgements

The research presented herein is based on an investi-gation sponsored by the National Science Foundationunder grant no. BCS-9319838, with Dr. Shih Chi Liu asthe program director. Any opinions, findings, and con-clusions or recommendations expressed in this paper areof those of the writers and do not necessarily reflect theviews of the sponsors.

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Documents

Steel-Concrete Composite Coupling Beams, Behavior and Design - Binginan Gong & Bahram M. Shahrooz