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Alwyn Laubscher & Associates (Pty) Ltd Springfontein, Cape Town, South Africa

SPRINGFONTEIN SEA WALL CONCEPT DESIGN

Coastal Engineering – Specialist Study

REV.03

09 July 2018

Alwyn Laubscher & Associates (Pty) Ltd Springfontein, Cape Town, South Africa

PRDW 5th Floor, Nedbank Building, Clock Tower Precinct, Victoria & Alfred Waterfront Cape Town, South Africa | PO Box 50023, Waterfront 8002 T: +27 21 418 3830

www.prdw.com

Cape Town, South Africa

Santiago, Chile

Perth, Australia

Seattle, USA

Vitoria, Brazil



Concept Engineering

S2001-84-TN-CE-001-R2 Springfontein Sea Wall Concept Design

09 July 2018

REV. TYPE DATE EXECUTED CHECK APPROVED CLIENT DESCRIPTION / COMMENTS

0 A 17/10/2017 CMB/AHH AHH AHH Draft for comment

1 A 25/10/2017 CMB CDH AHH Draft for comment

2 A 07/11/2017 CMB AHH AHH Draft for comment

3 C 09/07/2018 CMB AHH For Approval

TYPE OF ISSUE: (A) Draft (B) To bid or proposal (C) For Approval (D) Approved (E) Void

Alwyn Laubscher & Associates (Pty) Ltd Coastal Engineering – Specialist Study Page I of II

SPRINGFONTEIN SEA WALL CONCEPT DESIGN S2001-84-TN-CE-001-R3 Springfontein Sea Wall Concept Design.docx

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TABLE OF CONTE NTS

TABLE OF CONTENTS Page N°

1. INTRODUCTION ....................................................................................................................................................... 1

1.1 Background .................................................................................................................................................. 1 1.2 Study Scope .................................................................................................................................................. 3

3. SITE CONDITIONS .................................................................................................................................................... 3 3.1 TOPOGRAPHY AND BATHYMETRY ................................................................................................................ 3 3.2 CLIMATE CHANGE ........................................................................................................................................ 4 3.3 WATER LEVELS ............................................................................................................................................. 5

3.3.1 Astronomical Tidal Level ................................................................................................................. 5 3.3.2 Storm Surge .................................................................................................................................... 5 3.3.3 Design Water Level ......................................................................................................................... 5

3.4 EXTREME WAVE ESTIMATION ...................................................................................................................... 6 4. OVERTOPPING ESTIMATE ........................................................................................................................................ 7

4.1 Observed wave overtopping at Springfontein during 7 June 2017 storm event. ........................................ 7 4.2 Empirical and Qualitative Overtopping Assessment .................................................................................... 9

5. SEAWALL DESIGN AND COST ESTIMATE ................................................................................................................ 12 5.1 Concept Design .......................................................................................................................................... 12 5.2 Cost Estimate ............................................................................................................................................. 13

6. CONCLUSIONS AND RECCOMENDATIONS ............................................................................................................. 15 7. REFERENCES .......................................................................................................................................................... 16

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TABLES Page N°

Table 3-1: Increase in design parameters due to climate change to the year 2068. .................................................... 4

Table 3-2: Tidal characteristics at the Port of Cape Town (SANHO, 2017). .................................................................. 5

Table 3-3: Storm Surge at the Port of Cape Town - excluding and including adjustments for climate change (to the

year 2068). ................................................................................................................................................... 5

Table 3-4: Design water levels - excluding and including climate change. ................................................................... 6

Table 3-5: Nearshore wave heights for different offshore heights (Peak period of 18 s, foreshore slope of 1:30,

water depth of 1.33 m). ............................................................................................................................... 6

Table 3-6: Nearshore wave heights for different seabed slopes (Peak period of 18 s, offshore significant wave

height of 10 m, water depth of 1.33 m). ...................................................................................................... 6

Table 4-1: Description of overtopping limits. ................................................................................................................ 9

Table 5-1: Seawall cost estimate summary. ................................................................................................................ 14

FIGURES Page N°

Figure 1-1: Location of Grootte Springfontein – Farm 1 (Toms & Badenhorst, 2015). ................................................. 1

Figure 1-2: Flood lines as well as the extent and location of proposed seawall for quarry area (Toms & Badenhorst,

2015). ........................................................................................................................................................... 2

Figure 1-3: Sketch of proposed development of Springfontein quarry area (CNDV, 2017). ......................................... 2

Figure 3-1: Potential pool area showing extents (dotted hatching) of in-fill level required for pool facility. ............... 3

Figure 3-2: Historical satellite image showing low tidal level used a basis for an estimation of the Mean Sea Level

(MSL) contour............................................................................................................................................... 4

Figure 3-3: Typical cross section through the cliff. ....................................................................................................... 4

Figure 4-1: Summary of observed wave overtopping at Springfontein quarry area during 7 June 2017 storm event

by Andre Beukes. ......................................................................................................................................... 7

Figure 4-2: Measured wave data at Slangkop Waverider: June 2017 (CSIR, 2017). ..................................................... 8

Figure 4-3: Predicted tide on 7 June 2017. ................................................................................................................... 8

Figure 4-4: Cross section of the Three Anchor Bay promenade relative to the rugged rock cliff at Springfontein. ... 10

Figure 4-5: Aerial photograph of Three Anchor Bay during calm conditions. ............................................................. 10

Figure 4-6: Progression of extreme wave overtopping event at Three Anchor Bay during June 2017 storm event. . 11

Figure 4-7: Characterisation of an overtopping event with mean overtopping rate of 75 l/s/m and an individual

wave overtopping rate of 3000 l/s/m. ....................................................................................................... 12

Figure 5-1: Cross section of concept seawall. ............................................................................................................. 13

Figure 5-2: Extent of seawall considered for cost estimate. ....................................................................................... 13

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Alwyn Laubscher & Associates (Pty) Ltd



Concept Engineering

1. INTRODUCTION

1.1 Background

PRDW has been requested to provide high-level specialist coastal engineering inputs for the proposed

Springfontein development, located 50 km north of Cape Town as indicated in Figure 1-1.

Figure 1-1: Location of Grootte Springfontein – Farm 1 (Toms & Badenhorst, 2015).

Specifically, PRDW’s scope of work relates to a historic quarry area in the South Node which was previously

excavated for roadway stone. The depressed quarry area is to be developed into a pool facility with associated

amenities as part of the greater Springfontein development.

A coastal setback study was undertaken to identify suitable locations for development within the quarry area

(Toms & Badenhorst, 2015). Resultant flood lines as well as the location and extent of the proposed seawall

required to protect the quarry area as proposed from this study are shown in Figure 1-2.




Figure 1-2: Flood lines as well as the extent and location of proposed seawall for quarry area (Toms & Badenhorst, 2015).

Subsequent to the coastal setback study, CNDV Landscape Architects developed a concept layout of the

development planned for the quarry area as shown in Figure 1-3.

Figure 1-3: Sketch of proposed development of Springfontein quarry area (CNDV, 2017).




1.2 Study Scope

The scope of high-level specialist coastal engineering inputs is as follows:

▪ Characterise wave overtopping and associated implications for property development in this area;

▪ Develop a concept seawall cross-section; and

▪ Provide a high-level direct cost estimate.

2. FUNCTIONAL REQUIREMENTS

2.1 Design Life

A design life of 50 years is considered for the seawall. Assuming that construction of the seawall is completed

by 2018, the end of service life will be reached by 2068.

2.2 Design Storm Event and Encounter Probability

A 1 in 100 year return period is considered for the design storm event. This translates to an annual exceedance

probability of 1.0% and an encounter probability of the design wave event of 40% during the life of the

structure.

3. SITE CONDITIONS

3.1 TOPOGRAPHY AND BATHYMETRY

An overview of available topography is presented in Figure 3-1. Available survey information does not extend

below the +1 m MSL contour. To estimate the MSL contour, historic Google Earth satellite images were used

to identify low tidal levels. An example of a low tide water level at the site is shown in Figure 3-2.

Figure 3-1: Potential pool area showing extents (dotted hatching) of in-fill level required for pool facility.




Figure 3-2: Historical satellite image showing low tidal level used a basis for an estimation of the Mean Sea Level (MSL) contour.

The coast and foreshore clearly consist of bedrock. The natural tidal pool to the west of the quarry area is

virtually dry under low tidal levels. An assumed typical section through the cliff is presented in Figure 3-3. The

location of the section is indicated in Figure 3-2.

Figure 3-3: Typical cross section through the cliff.

3.2 CLIMATE CHANGE

The effect of climate change to the end of the structure service life is considered in accordance with PRDW’s

Position Paper for Climate Change (PRDW, 2010) as well as the IPCC’s Synthesis Report on Climate Change

(IPCC, 2014). Increases to design parameters to the year 2068 (end of structure service life) are presented in

Figure 3-1.

Table 3-1: Increase in design parameters due to climate change to the year 2068.

Storm surge increase [%] Wave height increase [%] Sea level rise [m]

14 12 0.55




3.3 WATER LEVELS

3.3.1 Astronomical Tidal Level

The predicted astronomical tidal water levels at the Port of Cape Town are presented in Table 3-2. At the Port

of Cape Town, Land Levelling Datum (LLD), commonly referred to as Mean Sea Level (MSL) is 0.825 m above

Chart Datum (CD).

Table 3-2: Tidal characteristics at the Port of Cape Town (SANHO, 2017).

Description Water Level [+m MSL]

Highest Astronomical Tide (HAT) 1.195

Mean High Water Springs (MHWS) 0.915

Mean High Water Neaps (MHWN) 0.435

Mean Level (ML) 0.155

Land Levelling Datum (LLD) (1 January 2003 onwards) 0.000

Mean Low Water Neaps (MLWN) -0.125

Mean Low Water Springs (MLWS) -0.575

Lowest Astronomical Tide (LAT) -0.825

3.3.2 Storm Surge

Actual water levels at a site will vary (both positively and negatively) from the predicted astronomical tidal

level due to changes in atmospheric pressure and wind effects. This phenomenon is referred to as storm

surge. Calculated extreme storm surges representative for Cape Town are presented in Table 3-3.

Table 3-3: Storm Surge at the Port of Cape Town - excluding and including adjustments for climate change (to the year 2068).

Return period [years]

Storm Surge (best estimate) [m]

Excluding climate change

Including climate change

10 0.60 0.68

20 0.65 0.74

50 0.70 0.80

100 0.75 0.86

3.3.3 Design Water Level

Design water level were calculated in accordance with Equation (3-1):

𝐷𝑒𝑠𝑖𝑔𝑛 𝑊𝑎𝑡𝑒𝑟 𝐿𝑒𝑣𝑒𝑙 = 𝑇𝑖𝑑𝑒 + 𝑆𝑡𝑜𝑟𝑚 𝑆𝑢𝑟𝑔𝑒 + 𝑆𝑒𝑎 𝐿𝑒𝑣𝑒𝑙 𝑅𝑖𝑠𝑒 (3-1)

Calculated design water levels are presented in Table 3-4.




Table 3-4: Design water levels - excluding and including climate change.

Parameter Design Water Level [m MSL]

Excluding climate change Including climate change

Tide (MHWS) [m MSL] 0.915 0.915

Storm Surge [m] 0.75 0.86

Sea Level Rise [m] - 0.55

Design water level [m MSL] 1.67 2.33

3.4 EXTREME WAVE ESTIMATION

Extreme waves at the site will be influenced by deep sea wave heights and the water depth at the toe of the

cliff in front of the sea wall. Deep sea waves will break as they move into shallow water where the height of

the broken waves is limited by the local water depth. This is referred to as a depth limited wave. As deep sea

wave heights increase they cause an increase in shallow water depths which is referred to as wave set-up. A

site specific wave refraction study with accurate seabed bathymetry will be required to quantify wave

transformation to the site. This is not justified for the present level of study and therefore typical extreme

deep sea heights were used to check the sensitivity of local depth limited wave heights to this. As shown in

Table 3-5, increasing the offshore significant wave height (Hs) from 9 m to 11 m does not have a significant

effect on the depth limited significant wave height. In the absence of a refraction study it is considered

reasonable to use an offshore design wave height of 10 m to estimate nearshore depth limited waves.

Table 3-5: Nearshore wave heights for different offshore heights (Peak period of 18 s, foreshore slope of 1:30, water depth of 1.33 m).

Location Significant wave height (Hs) [m]

Deep sea (offshore) 9 10 11

Nearshore (1.33 m water depth) 2.30 2.38 2.50

Depth limited wave heights increase with increasing peak wave periods (Tp). A peak wave period of 18 s is

considered representative of an extreme design condition. The foreshore slope is uncertain as no detailed

bathymetry is available. Goda’s method of irregular wave transformation (Goda, 2000) was used to

characterise wave conditions at the toe of the cliff. It is noted that Goda’s method assumes a constant

foreshore slope which will not be the case, especially in the presence of submerged reefs. As a sensitivity

check, the effect of foreshore slope on nearshore wave height is illustrated in Table 3-6.

Table 3-6: Nearshore wave heights for different seabed slopes (Peak period of 18 s, offshore significant wave height of 10 m, water depth of 1.33 m).

Description Unit Value

Significant deep sea wave height m 10 10 10

Seabed slope (horizontal distance for 1 m vertical drop) - 30 20 15

Nearshore significant wave height (1.33 m water depth) m 2.38 2.54 3.22

The nearshore wave height is sensitive to foreshore slope and in the absence of detail bathymetry, a

conservative approach should be followed for the design. It is recommended that a bathymetric survey be

done prior to finalising the design.




A check was done on Goda’s wave transformation results with (Delft Hydraulics, 1990), indicating that Goda’s

model was slightly more conservative. Wave set-up was also calculated for an offshore wave height of 10 m,

foreshore slope of 1:30, peak wave period of 18 s and water depth of 1.32 m. Set-up of 1.1 m was calculated

indicating that the water depth would increase from 1.33 m to 2.42 m under the assumed design condition

(which explains why a significant wave height of 2.38 m can occur in a “depth” of 1.33 m).

It should be noted that the actual bathymetry differs significantly from a uniform slope. The only reliable

method of determining extreme waves is with detailed modelling. This would best be achieved in a physical

model that would enable checking overtopping, optimising wall dimensions and quantifying pool

hydrodynamics.

4. OVERTOPPING ESTIMATE

4.1 Observed wave overtopping at Springfontein during 7 June 2017 storm event.

Valuable insights into wave overtopping at Springfontein were made possible due to observations made by

Andre Beukes during a recent extreme storm event on 7 June 2017. A summary of Andre’s observations is

presented in Figure 4-1.

Figure 4-1: Summary of observed wave overtopping at Springfontein quarry area during 7 June 2017 storm event by Andre Beukes.

A detailed characterisation of the June 2017 storm event is beyond the scope of work. It is however clear by

comparing available wave height measurements from the Slangkop Waverider buoy during the storm to

historical records, that the June 2017 storm was one of the largest storms on record. As shown in Figure 4-2,

a maximum significant wave height (Hm0) of approximately 11.5 m and peak wave period (Tp) of 18 s was

measured.

A high-level estimate of the return period associated with the June 2017 storm indicates a return period of

approximately 50 years or more (taking account of the westerly direction of the storm event).




Figure 4-2: Measured wave data at Slangkop Waverider: June 2017 (CSIR, 2017).

It is noted that the peak of the storm coincided with mid-tide as shown in Figure 4-3. As wave conditions at

the site are depth limited during extreme wave conditions, wave overtopping observations made my Andre

Beukes (at 12:00 on 7 June 2017) must be understood in this context. A storm event with lower return period

but coinciding with a high tide could result in similar wave overtopping.

Figure 4-3: Predicted tide on 7 June 2017.

Based on Andre’s observations, the presence of seaweed and driftwood up to the dotted line “C” in Figure 4-1

indicates the extent of flooding due to wave-overtopping on site at a maximum tide level of 0.74 m MSL. No

information regarding increased water level due to storm surge is available. This tide level is 0.73 m below

the assumed design water level (including sea level rise). It is therefore not representative of the maximum




overtopping that can be expected over the lifetime of the structure, but provides good insight into

overtopping implications for major storm events without sea level rise.

4.2 Empirical and Qualitative Overtopping Assessment

To estimate the risk of damage to developments landward of the seawall, an attempt is made to quantify

overtopping for extreme storm events including the impact of climate change.

As a guide, Eurotop (EurOtop, 2016) and the Rock Manual (CIRIA; CUR; CETMEF, 2007) provide descriptive

guidelines equating mean overtopping rates to expected levels of damage behind the defence. In this study,

the primary defence is defined as the seawall of the pool/ in-fill area. Descriptions of example overtopping

limits are presented in Table 4-1.

Table 4-1: Description of overtopping limits.

Description Overtopping limit (q)

[m³/s/m] Reference

Damage to equipment set back 5-10 m ≤ 1 (EurOtop, 2016)

People at seawall (with clear view of sea) 1

Damage to paved or armoured promenade behind seawall

200 (CIRIA; CUR; CETMEF,

2007) Damage to grassed or lightly protected promenade or reclamation cover

50

During storm events it is recommended that access to the seawall be restricted. To prevent damage to

infrastructure behind the seawall, a mean overtopping limit of below 50 l/s/m is considered for this study

(this assumes that areas landwards of the seawall will be lightly armoured / protected).

The extreme shallow water conditions that exist on site are largely beyond the range of validity of available

empirical equations used to estimate overtopping. There is therefore a lot of uncertainty in estimating

overtopping with these equations. Initially, an investigation into the extensive Eurotop overtopping physical

model database was undertaken to find a ‘best fit’ physical model setup that best represented the site

conditions to estimate wave overtopping. Only vertical wall examples were found that would have lower

overtopping values compared to the rugged bedrock slope of approximately 1:2 (V:H). Unfortunately, no

physical model setup was found that sufficiently represented site conditions.

A pragmatic approach was subsequently followed to characterise overtopping. Extreme overtopping along

Three Anchor Bay promenade during the 7 June 2017 storm event (described in Section 4.1) was captured on

video. This information serves as a valuable basis to infer overtopping rates at Springfontein under similar

extreme conditions.

A cross section of the Three Anchor Bay promenade relative to the rugged rocky cliff at Springfontein is

presented in Figure 4-4.




Figure 4-4: Cross section of the Three Anchor Bay promenade relative to the rugged rock cliff at Springfontein.

An aerial photograph of the relevant section of promenade at Three Anchor Bay during calm conditions is

shown in Figure 4-5, while progressive snapshots of extreme wave overtopping event at Three Anchor Bay

during the June 2017 storm event are presented in Figure 4-6.

Figure 4-5: Aerial photograph of Three Anchor Bay during calm conditions.

Due to the shape of Three Anchor Bay, additional wave set-up due to the focusing of waves as they enter the

bay is likely (as compared to regular wave-setup associated with an open coastline). While no calculations

have been done to estimate this additional wave set-up, it can reasonably be assumed to be in the order of

0.5 m under extreme storm conditions. This is similar in magnitude to the sea level rise considered for

Springfontein. Overtopping at Springfontein including sea level rise is therefore expected to compare fairly

well to the existing conditions at Three Anchor Bay (Springfontein will not have any significant focussing

effects).




Figure 4-6: Progression of extreme wave overtopping event at Three Anchor Bay during June 2017 storm event.

A website hosted by HR Wallingford (www.overtopping-manual.com) provides videos of overtopping waves

that enables an appreciation of overtopping values. A screen shot of the maximum overtopping video on the

site, shown in Figure 4-7, shows an average overtopping of rate of 75 l/s/m with an individual wave

overtopping rate of 3000 l/s/m.

http://www.overtopping-manual.com/




Figure 4-7: Characterisation of an overtopping event with mean overtopping rate of 75 l/s/m and an individual wave overtopping rate of 3000 l/s/m.

Based on a comparison of the Three Anchor Bay video and that from Wallingford’s website, the overtopping

rate for Three Anchor Bay is estimated at between 100 l/s/m and 200 l/s/m. Above approximately 200 l/s/m

one can expect damage to paved promenades (see Table 4-1).

Eurotop (EurOtop, 2016) formulae were used to attempt calculating overtopping rates for the Three Anchor

Bay case. Reasonable estimates were obtained. However, when a vertical wall was introduced to estimate

reduced overtopping for acceptable design values, unrealistic results were obtained. This indicates that the

cliff and seawalll geometry probably lies outside the range of conditions tested for the empirical formulae.

The height of the seawall was therefore based on simply raising the 1:2 slope to a level where the overtopping

reduces to acceptable levels. Selecting an acceptable level should relate to avoiding damage to any

infrastructure landward of the seawall edge.

As the seawall serves as the only defence against wave attack, a crest elevation that results in an overtopping

rate that would avoid damage to grassed or lightly protected (i.e. paved) areas behind the wall during the

design storm event is therefore targeted. A mean overtopping rate of 35 l/s/m, achieved for a seawall with

crest elevation of +7 m MSL, is considered suitable.

The only reliable method to determine overtopping and transmitted wave heights is by conducting a physical

model study. If the cost of such a study is not justified, some uncertainty will remain with regard to required

seawall heights, potential wave action in the pool and required protection landward of the seawall. While it

is expected that the proposed seawall are conservative, some maintenance after extreme storm events

cannot be ruled out.

5. SEAWALL DESIGN AND COST ESTIMATE

5.1 Concept Design

A typical cross section of the concept seawall design is presented in Figure 5-1. The seawall design assumes

that the wall is founded on suitable bedrock (i.e. consistent, hard and absent of any significant fissures or

cracks). The seawall is to be founded at the top of the cliff which varies in elevation between approximately




+5.0 m MSL and +5.5 m MSL. A crest elevation of +7.0 m MSL is considered in accordance with the

overtopping assessment presented in Section 4. A crest width of 1.5 m is considered as per the architect’s

requirements. The dimensions of the concept seawall design require verification as part of detailed design to

confirm stability when subjected to extreme wave loads.

Figure 5-1: Cross section of concept seawall.

It is noted a traditional gravity wall structure (i.e. blockwork wall) could also be considered as an alternative.

5.2 Cost Estimate

The scope of the cost estimate is limited to the direct construction cost of the reinforced concrete seawall. A

seawall section of approximately 70 m has been considered for this cost estimate as illustrated in Figure 5-2.

The remaining section of approximately 54 m has been excluded from cost estimate on the basis that a more

conventional pool/retaining wall is adequate along this section due to its protected location resulting in less

onerous design conditions.

Figure 5-2: Extent of seawall considered for cost estimate.




Together with the layout and concept design information presented in Figure 5-2 and Section 5.1, the

following was considered:

▪ The estimate is set at FEL 1 Concept level (Class 5) with a target level of accuracy of -50% to +50%;

▪ The estimate has been derived using estimated preliminary quantities and typical corresponding

current or escalated unit rates largely based upon PRDW’s internal rates database. Built-up rates and

prices have been used where no existing rates or prices were available and new rates or prices are

required;

▪ General assumptions:

▪ Base Date: October 2017.

▪ It is assumed that the wall is founded on suitable bedrock.

▪ Unrestricted access for construction purposes

▪ Exclusions:

▪ General site preparation;

▪ Mechanical, electrical and pipe work installation;

▪ Preliminary and general costs;

▪ Design development costs;

▪ Professional fees;

▪ Local or other authority approvals;

▪ Purchase / lease of land;

▪ Pre-tender and post contract escalation;

▪ Environmental, EIA and EMP costs;

▪ Geotechnical and additional site investigation fees and costs;

▪ Construction monitoring and Principal Agent fees;

▪ Market Adjustment and Post Contract Contingencies;

▪ Value Added Tax or other Taxes and import duties.

The direct construction cost estimates for the seawall, considering a total length of 70 m, is summarised in

Table 5-1.

Table 5-1: Seawall cost estimate summary.

Item Amount (ZAR)

In-situ concrete pool wall R 730 000

Supply, drill and grout Y25 (1.25 m long and 1.5m centers) R 310 000

Foundation preparation R 10 000

Wood floating (finishing) R 10 000

Total R 1 060 000

The direct construction cost estimate presented in Table 5-1 equates to approximately R 15 150 per metre

length of seawall.




6. CONCLUSIONS AND RECCOMENDATIONS

High-level specialist coastal engineering inputs have been provided for the development of the quarry area

(South Node) of Springfontein. The depressed quarry area is to be developed into a pool facility with

associated amenities as part of the greater Springfontein development.

Wave overtopping and associated implications thereof for the development of property within the quarry

area was characterised; a concept level cross-section of the required seawalls and associated cost estimate

were developed.

Available topography only extends down to +1 m MSL. The MSL contour was estimated based on aerial

photos. A foreshore slope of 1:30 (V:H) was assumed for the area seaward of MSL. The seawall is to be

founded at the top of the cliff which varies in elevation between approximately +5.0 m MSL and +5.5 m MSL.

Extreme waves were estimated at the toe of the cliff using Goda’s method of irregular wave transformation.

During extreme storm conditions, waves are depth limited at the site. As such, the total water depth at the

point of interest will have a significant effect on the magnitude of the nearshore wave height. Extreme water

levels were calculated using a mean high water spring (MHWS) tide and storm surge as a basis. Additional

contributions to extreme water levels due climate change (such as sea level rise and an increase in storm

surge) were also considered. A final extreme water level of 2.33 m MSL was selected for design.

The sensitivity of offshore (deep water) wave heights and foreshore slopes on the calculated nearshore wave

heights were checked. While the nearshore wave height is not sensitive to offshore wave heights, the

foreshore slope does have a significant effect. Using the assumed 1:30 foreshore slope a significant nearshore

wave height (Hs) of 2.4 m was calculated at the toe of the cliff (+1 m MSL) considering an extreme water level

of 2.33 m MSL and a peak wave period (Tp) of 18 s.

Available design tools for estimating wave-overtopping on site were found to be out of range (validity). A

pragmatic approach was therefore followed to estimate overtopping. This was done by assessing the

overtopping observed at Three Anchor Bay during an extreme storm on 7 June 2017. A comparison of Three

Anchor Bay and Springfontein geometry allows an assessment of overtopping over the proposed pool wall

under extreme conditions. Observations of debris in the quarry after the 7 June storm also provide an

indication of the effects of overtopping under major storms in the absence of sea level rise.

As the seawall serves as the only defence against wave attack., a crest elevation that results in an overtopping

rate that would avoid damage to grassed or lightly protected (i.e. paved) areas behind the wall during the

design storm event is therefore targeted. A mean overtopping rate of 35 l/s/m, achieved for a seawall with

crest elevation of +7 m MSL, is considered suitable.

A number of assumptions were required in developing the seawall concept. The only way to ensure more

certainty and allow for optimisation is by carrying out a bathymetric survey and physical model study. If the

cost of these investigations is not justified it may be necessary to add additional conservatism during detail

design or to accept more damage than anticipated during extreme storm events.

A direct construction cost for approximately 70 m of reinforced concrete seawall of R 1 060 000 (or

R 15 150/m) was estimated. The cost estimate targets a concept level of accuracy (-50% to +50%).




7. REFERENCES

CIRIA; CUR; CETMEF, 2007. The Rock Manual. The use of rock in hydraulic engineering (2nd edition), London:

C683, CIRIA.

CNDV, 2017. Sketch of proposed Springfontein sea wall and pond.. Cape Town: CNDV Landscape Architects.

CSIR, 2017. Cape Point - Wave Data. [Online]

Available at: http://wavenet.csir.co.za/OnlineData/CapeTown/CapeTownwaveD.htm

Delft Hydraulics, 1990. Extreme shallow water wave conditions: Design curves for uniform sloping beaches,

s.l.: Delft Hydraulics.

EurOtop, 2016. Manual on wave overtopping of sea defences and related structures. An overtopping

manual largely based on European research, but for worldwide application. , www.overtopping-

manual.com.: Van der Meer, J.W.,Allsop, N.W.H., Bruce, T., De Rouck, J., Kortenhaus, A., Pullen, T.,

Schüttrump H., Troch, P. and Zanuttigh, B..

Goda, Y., 2000. Random Seas and Design of Maritime Structures, 3rd Edition, Singapore: World Scientific

Publishing.

IPCC, 2014. Climate Change 2014: Synthesis Report. Contribution of Working Groups I, II and III to the Fifth

Assessment Report of the Intergovernmental Panel on Climate Change [Core Writing Team, R.K. Pachauri

and L.A. Meyer (eds.)], Geneva, Switzerland: IPCC.

PRDW, 2010. Global Climate Change: Consequences for Coastal Engineering Design - Position Paper. Report

No. 939/1/001, Cape Town: PRDW.

SANHO, 2017. South African Tide Tables. 2017 ed. Tokai: South African Navy Hydrographic Office.

Toms, G. & Badenhorst, P., 2015. Grootte Springfontein - Farm 1: Coastal Setback Study, Cape Town: s.n.

Documents

SPRINGFONTEIN SEA WALL CONCEPT DESIGN › pdfdocs › Spring Pre-App Scoping...Coastal Engineering – Specialist Study Concept Engineering S2001-84-TN-CE-001-R2 Springfontein Sea