SCI P159- Structural Fire Design to EC 3 4 and Comparison With BS5950

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TECHNICAL REPORT SCI PUBLICATION 159

Structural Fire Design to EC3 & EC4, and comparison with BS 5950

R M Lawson BSc(Eng), PhD, ACGI, CEng, MICE, MlStructE G M Newman BSdEng), CEng, MIStructE, MlFS

Published by:

The Steel Construction Institute Silwood Park, Ascot Berkshire SL5 7QN Telephone: 01 344 23345 Fax: 0 1 3 4 4 2 2 9 4 4

P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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SCI Technical Reports

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Please forward your comments to Dr R M Lawson, The Steel Construction Institute, Silwood Park, Ascot, Berkshire, SL5 7QN.

0 1996 The Steel Construction Institute

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Publication Number: SCI-P- 159

ISBN 1 85942 036 2

British Library Cataloguing-in-Publication Data. A catalogue record for this book is available from the British Library

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FOREWORD

Eurocode 3: Design of steel structures, Part 1.2: Structuralflre design and Eurocode 4: Design of steel and composite structures, Part 1.2: Structuralflre design were both formally approved by CEN (ComitC EuropeCne de Normalisation) in the latter part of 1993. These Eurocodes will be published by the British Standards Institution and by the various national standards institutes in 1996 and will have the status of ENV (or pre-norm) standards. Their BSI identification numbers are BS ENV 1993 Part 1.2 and BS ENV 1994 Part 1.2 respectively. Both Eurocodes are supplemented by National Application Documents (NADs), which give partial safety factors and other local requirements for use in the country concerned. The UK NADs will be published together with the Eurocodes.

It is intended that all ENV Eurocodes will have temporary status in order to obtain feedback on their use and to incorporate improvements. Eventually (after 3 to 4 years), they will be formally issued as EN Eurocodes which will have an equivalent status to, or will act in place of national standards. It is therefore important that British engineers gain experience of using all Eurocodes during their ENV period.

This guide will assist in this familiarization r81e. It provides an introduction to the provisions of the Eurocodes dealing with the fire resistance aspects of steel and composite structures, and gives background and comparison to existing British Standards.

The authors of the guide were Dr R M Lawson and Mr G M Newman of The Steel Construction Institute with assistance from the following:

Dr B R Kirby British Steel (Technical) Mr J T Robinson British Steel (Sections, Plates & Dr K F Chung The Steel Construction Institute Mr J C Taylor The Steel Construction Institute Mr K Baltzer The Steel Construction Institute Mr M A Wadee The Steel Construction Institute

The work leading to this guide was funded by British Steels) and by the Department of the Environment initiative.

Commercial Steels)

Steel (Sections, Plates & Commercial through the Partners in Technology

Some of the information used in this guide is adapted from Background Papers presented to the Project Teams for these Eurocodes. The Convenors of the Project Teams have agreed to the use of this information in the guide and their permission is gratefully acknowledged. The Project Teams were as follows:

Eurocode 3: Part 1.2:

Mr M Law Convenor, UK Mr J C Taylor Editor, UK Dr J Kruppa France Dr L Twilt Netherlands Professor P J Dowling Chairman of Eurocode 3 : Part 1 .1

111 ...


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Eurocode 4: Part 1.2:

Dr J B Schleich Convenor, Luxembourg Dr J Kruppa France Dr R M Lawson UK Dr P Schaumann Germany Dr L Twilt Netherlands Professor R P Johnson Chairman of Eurocode 4: Part 1.1

Note: This Technical Report (dated June 1996) was prepared before the UK NADs for EC3 and EC4, Parts 1.2 had been completed. It is intended to revise the report later to take the NAD requirements more fully into account. The NADs will be prepared in late 1996.

The report is therefore not in its final form and is issued as a Technical Report in order that users are informed of the likely scope and content of these Eurocodes.

iv P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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CONTENTS

FOREWORD

SUMMARY

Page No. ... Ill

viii

INTRODUCTION 1.1 Introduction to Eurocodes 1.2 Introduction to fire safety 1.3 Scope of Eurocodes 3 and 4: Parts 1.2 'structural fire design' 1.4 Definitions 1.5 Symbols 1.6 References t o Eurocodes and BS 5950 1.7 Materials grades in Eurocodes

FIRE LIMIT STATES 2.1 Fire resistance test 2.2 Partial safety factors 2.3 Design values of actions at fire limit state 2.4 Material factors 2.5 Load level

PERFORMANCE OF MATERIALS AT ELEVATED TEMPERATURES 3.1 Physical properties of structural steel 3.2 Physical properties of other forms of steel 3.3 Physical properties of concrete

CRITICAL TEMPERATURES OF STEEL MEMBERS 4.1 Introduction to critical temperatures 4.2 Members in tension 4.3 Members in bending 4.4 Members in compression 4.5 Critical temperatures for simple design

THERMAL RESPONSE OF PROTECTED AND UNPROTECTED STEEL MEMBERS 5.1 5.2 5.3 5.4 5.5 5.6

5.7 5.8 5.9 5.10 5.1 1

Definition of section factors Traditional and modern fire protection materials Theoretical behaviour of unprotected steel sections in fire Temperatures in unprotected columns and beams Theoretical behaviour of protected steel sections in fire Traditional method of appraising the performance of fire protection materials Assessment of material properties Comparison of protection thicknesses Proposed simplified approach Partial protection to bare steel beams and columns Computer methods for predicting thermal and structural response

1 1 2 4 5 7 7 8

9 9

10 12 13 15

17 17 24 27

32 32 33 34 43 51

53 53 56 58 59 60

63 69 73 74 75 76

V P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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6

7

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1 1

12

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14

PARTIALLY AND FULLY ENCASED COLUMNS 6.1 Introduction 6.2 Partially encased columns 6.3 Fully encased I sections 6.4 Columns with concrete block infills

PARTIALLY AND FULLY ENCASED BEAMS 7.1 Partially encased beams 7.2 Partially encased composite beams 7.3 Fully encased beams 7.4 Slim floor beams 7.5 Shelf angle floor beams

COMPOSITE BEAMS 8.1 Introduction 8.2 Composite beams with board or spray protection 8.3 Composite beams with partial encasement

COMPOSITE DECK SLABS 9.1 Minimum slab depths 9.2 Bending resistance

CONCRETE FILLED HOLLOW SECTION COLUMNS 10.1 Introduction 10.2 Normal design 10.3 Fire design

DETAILING REQUIREMENTS 1 1 .l Partially encased sections 1 1.2 Fully encased sections 1 1.3 Concrete filled sections 11.4 Beam to column connections

EXTERNAL STEELWORK 12.1 Influence of location of steel members 12.2 Heat transfer to a steel member

CASES NOT COVERED BY EUROCODES 13.1 Portal frames in fire 13.2 Castellated beams 13.3 Walls and roofs 13.4 Ceilings 13.5 Bracing 13.6 Escape stairways

NATURAL FIRES 14.1 Important parameters in determining fire temperatures 14.2 Concept of time-equivalent 14.3 Fire loads in buildings 14.4 Temperatures in steel sections in natural fires 14.5 Influence of active protection measures

77 77 78 88 90

91 92 95 95 96 97

99 99 100 106

109 109 112

1 20 1 20 121 1 23

138 138 138 139 139

143 143 144

146 146 146 147 148 148 148

149 149 153 154 155 155

vi P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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15 RE-USE OF STEEL AFTER A FIRE 15.1 Mechanical properties 15.2 Inspection and appraisal 15.3 Re-use of unprotected sections 15.4 Re-use of protected sections 15.5 Re-use of composite deck slabs

157 157 158 158 159 159

REFERENCES 160

APPENDIX A: WORKED EXAMPLES 165

APPENDIX B: SECTION FACTORS OF STEEL MEMBERS 185

vii P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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SUMMARY

This guide provides background information, design tables and useful guidance on Part 1.2 ‘Structural Fire Design’ of Eurocodes 3 and 4, dealing with structural design in steel and composite construction respectively. This report is presented in the form of a guide on structural fire design, and makes cross-reference to the relevant clauses of the Eurocodes and to their forthcoming UK National Application Documents. Detailed comparison is made with the relevant British Standard, BS 5950: Part 8, in order to assist designers in interpreting the Eurocodes for UK practice.

The forms of construction that are covered by these Eurocodes include protected and unprotected steel beams and columns, composite beams and slabs, concrete encased and concrete-filled hollow sections, and special structures such as external steelwork. The publication reviews the scope and use of these Eurocodes, and extends the number of simplified tables on some of these forms of construction that are considered to satisfy the requirements of the Eurocodes. Modifications to some tables in EC4: Part 1.2 are proposed. Worked examples on the fire resistant design of certain of these cases are presented in detail. References to background information are presented. Additional information on aspects not covered by EC3 and EC4 are included to provide a comprehensive review of fire resistant design of steel and composite structures.

v111 ...


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l INTRODUCTION

1. l Introduction to Eurocodes

Eurocodes are intended to provide methods of design of buildings and other structures that are accepted throughout the European Union and associated countries. Compliance with Eurocodes should be of equivalent status to national standards, and indeed in many countries, Eurocodes are used in the absence of relevant standards.

Work began on the preparation of Eurocodes in the early 1980’s and drafts for comment were released on steel, concrete and composite construction by the European Commission in 1985. These drafts were subsequently improved and followed to the CEN procedures for preparation and approval. Finally, in 1992, Part 1.1 of these major Eurocodes were approved for publishing by the national standards organisations throughout Europe.

Each ENV Eurocode is supplemented by a National Application Document (NAD) for the country concerned. The NADs provide numerical values of partial safety factors for materials, and local requirements, but are not permitted to modify the basic principles and application rules of the Eurocodes. The NADs are intended to be withdrawn when the EN versions of the Eurocodes are issued after a period for comment and revision (see Foreword).

The Eurocodes that will be primarily referred to in this guide are:

BS ENV 1993: Eurocode 3: Design of steel structures(’) BS ENV 1994: Eurocode 4: Design of composite steel and concrete structured2)

Part 1. l of these Eurocodes (known in shorthand as EC3 and EC4) refer to ‘General rules and rules for building’, which may be considered to be appropriate for ‘normal’ design.

Parts 1.2 of these Eurocodes refer only to ‘Structural fire design’ and are published separately. It is assumed that the normal design of the structures is in compliance with Part 1.1 of the relevant Eurocode or with the appropriate British Standard, and therefore compliance with the supplementary requirements of Part 1.2 achieves the required fire resistance of the structure.

The equivalent British Standard for steel and composite construction is BS 5950: for which the relevant parts in the context of these Eurocodes are:

Part 1:

Part 3:

Part 4: Part 8:

Code of practice for design in simple and continuous construction: hot rolled sections Section 3.1 : Code of practice for design of simple and continuous composite beams Code of practice for design of composite slabs with profiled steel sheeting Code of practice for fire resistant design.

1 P159: Structural fire design to EC3 and EC4 and comparison with BS 5950 (1996 Edition)

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The drafting of EC3 and EC4 Parts 1.2 began in 1990, although European Recommendations had existed since 1985(3). In 1994, EC3 and EC4, Parts 1.2 were approved by CEN and since then effort has gone into the editorial stage and preparation of their NADs. The published version of these Eurocodes will be available in their ENV status with their NADs shortly. All ENV standards will have the status of ‘pre-norms’ but can be used for design within the scope of the UK Building Regulations.

However, some further codification, particularly on the method of evaluating fire protective materials, is still in progress under the auspices of CEN, and therefore the guidance on fire resistant design is not yet complete.

All Eurocodes are presented in a limit state format; i.e. partial safety factors are used to increase or ‘factor’ the specified loads, and also to reduce the material strengths to obtain the design member resistances. These partial safety factors are separately termed; partial factors for loads, and partial factors for materials. In principle, the numerical values of the partial factors for materials are established to account for the inherent variability of the strength of the materials, and therefore to obtain an equivalent level of safety in the structural use of all materials (refer to Section 2.4). The global safety factor is not a single value, but for most steel or composite structures, it lies in the range of 1.5 to 1.7 for ‘normal’ design.

Reduced partial factors of safety are adopted in fire conditions in order to take account of the relatively low probability of occurrence of fire and extreme loading, the acceptance of greater level of damage in fire in comparison to normal conditions, and the concept that the regulatory requirements for fire resistance are sufficiently severe to provide the necessary margin of safety.

Part 1.2 of EC3 and EC4 provide some new concepts with which structural designers in the UK may not be fully familiar, although most of the principles are already incorporated in BS 5950: Part 8, which was published in 1990. Importantly, there are advantages of harmonization of design procedures throughout Europe, particularly for design in the wider European context.

Furthermore, the terminology and use of symbols may be new to designers in the UK, but familiarity with the language of Eurocodes should not discourage their use. By way of an example, the word ‘actions’ in Eurocodes includes both applied loads and other thermally induced forces.

1.2 Introduction to fire safety

The Interpretative Document of the Construction Products Directive issued by the European Commission and titled Essential requirement: safety in case ufflre, sets out the main objectives of fire safety, in simple terms, as:

the load bearing capacity of the construction can be assumed to be adequate for a specified period of time. the generation and spread of fire within the works are limited.


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the spread of fire to neighbouring construction works is limited. occupants can leave the works, or be rescued by other means. the safety of rescue teams is taken into account.

In principle, these provisions fall within the general requirements for building design and choice of materials, but only the first is entirely the responsibility of the structural designer.

All fire regulations are primarily concerned with safety of occupants and fire fighters in buildings, and to prevent the probability of major fire spread (conflagration). However, in practice, regulations also endeavour to provide for a measure of property protection, particularly with regard to disproportionate damage in small fires. It is extremely difficult to separate these requirements and also to take into account special fire protective measures such as automatic detection, use of sprinklers, smoke barriers, etc., which have a major effect on both life safety and damage limitation. These measures are known as ‘active’ measures because they seek to reduce the severity of a fire.

In most cases the structural designer is concerned with so called ‘passive’ protection, which is used in order to achieve adequate load resistance by preventing the steel or other elements from heating up excessively, and hence losing a major proportion of their strength in fire. Methods of analysis are well established to determine the thermal and structural response of the primary elements of construction (e.g. beams or columns) when protected in this way. For many common forms of construction, simplified design tables have been prepared to offer quick and safe design with limited design effort.

Nevertheless, there are many special and more complex structural forms which demand more accurate treatment, and design codes often include provisions for these methods in order not to stifle innovative design.

British Standards have been prepared that offer acceptable methods of design within the context of the UK Building Regulations. The Eurocodes will also operate within the same fire resistance requirement of the UK Building Regulations, although harmonization of both design loads (actions) and fire resistance requirements throughout Europe is a longer term intention.

‘Fire resistance’ is a notion that is applicable only to a standard fire resistance test using a standard time-temperature curve, as specified by I S 0 834(4) and BS 476: Part 20(5) (see Figure 1). Therefore, fire resistance in itself is only a scalar representation of the relative performance of elements of construction (beams, columns, walls), and is used as a means of calibrating the likely performance of members in fire conditions. For example, a beam with one hour fire resistance will not fail at exactly one hour in a real fire, and so on. Indeed, there is evidence to show that real fires do not lead to conditions as severe as that of a standard test.


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‘Fire engineering’ using the concept of so-called ‘real’ or ‘natural’ fires is a useful tool to justify the use of bare steelwork in structures with low potential fire severity. Examples of these types of structure that benefit from the fire engineering approach are car parks, sports halls, railway and airport terminals. These methods are outlined in Section 12 of this publication.

Fire resistance can be established by two distinct methods: from the results of fire tests by calculation.

Fire test results themselves are conclusive for the particular configuration tested, but further calculations are often required to extend the test result to cover other practical cases. Indeed, the test data can be used to validate a particular design method, and hence to provide tabular data for general design within the limits of the test parameters.

Calculation procedures are normally adopted for forms of construction not amenable to test, or for well established design methods that use physical parameters for the materials employed. An example of this second approach is the required protection thickness to steel sections, which is a function of the stress state in the member, the geometry of the section, and the thermal characteristics of the protection material.

1.3 Scope of Eurocodes 3 and 4: Parts 1.2 ’structural fire design’

Both Eurocodes 3 and 4, Parts 1.2(’9 2, and BS 5950: Part 8 @)are only concerned with calculation methods to achieve the required degree of fire resistance of the steel or composite section. By their nature, these calculation methods are conservative and take account of the uncertainties in the materials and analysis methods used.

Three levels of calculation method are given in Parts 1.2 of the Eurocodes: tabular methods simple calculation models advanced calculation models.

The tabular methods are used for direct design knowing certain parameters related to loading, geometry, reinforcement, etc. Design tables are often relatively restrictive, but cover common cases of design.

The simple calculation methods are generally ‘hand’ calculation methods which can be made more amenable by computer programming. They are not necessarily ‘simple’ to use, but are based on well established principles, such as plastic analysis of sections. These calculation models are used for general design and will lead to more economic design than the tabular methods.


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Code Clause

The advanced calculation models are used only in understanding frame behaviour or in developing design aids in the future. The models are only appropriate for computer analysis and not for general design.

Eurocode 4: Part 1.2 (EC4 Part 1.2) refers extensively to Eurocodes 2 and 3 Parts 1.2 for material properties, although there is some duplication of information to assist in understanding these new codes. EC4: Part 1.2 also deals with certain forms of construction that may

only be designed to act compositely at the fire limit state, and are otherwise designed as steel members to Eurocode 3: Part 1 . 1 .

In common with all Eurocodes, EC3 and EC4 Parts 1.2 have clauses which are principles and others that are application rules. Principles must be followed in design, whereas other application rules may be adopted if they can be shown to satisfy these principles.

Similarly, nonnative annexes are those that contain additional information to supplement the application rules. Informafive annexes are included only for background information. They may duplicate other Eurocode clauses, and are intended to be removed in the final EN versions.

1.4 Definitions

The following key definitions are important in interpreting Eurocodes 3 and 4 Parts 1.2. The corresponding term in BS 5950: Part 8 is also identified.

EC3/EC4 cl 1.3

Action: Force applied to a member due to external loading. In fire conditions, this may also include thermal actions.

Adaptation factor: A factor used to multiply the applied loads in fire conditions (or to reduce the resistance) in order that the critical temperature of the member is correctly determined.

Critical steel temperature For a given load level (see below), this is the temperature at which failure is expected to occur in a steel element exposed to a uniform temperature distribution. The critical temperature is approximately the same as the limiting temperature in BS 5950: Part 8.

BS 5950 cl. 1.2.7


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Code Clause

Fire load: Combustible contents of the enclosure, usually expressed as MJ/m2 over the floor area of the enclosure.

Fire compartment: A space or enclosure within a building surrounded by compartment floors and walls that offer a separating function in order to prevent a possible fire spreading to other parts of the building.

Integrity:

Load level:

Net heat flux:

A criterion relating to prevention of the passage of smoke or flame across a compartment boundary.

This is the load acting on a member as a proportion of its resistance at room temperature. The equivalent term in BS 5950: Part 8 is the 'load ratio'.

The energy per unit time (W/m2) that is absorbed by a heated surface. This parameter may be obtained from the operation of standard furnaces, and used in more general analytical models.

Resultant emissivity : This is the ratio between the actual radiative heat flux to the member and the net heat flux that would occur if the member and its radiative environment are considered as black bodies.

Specific heat of material: Quantity of heat stored (joules) in a unit mass of a material (kg) for 1 "C temperature rise.

Standard fire : A term used to define a standard IS0 834 time-temperature curve in a fire resistance test.

BS 5950 cl 4.2.2.2

Thermal conductivity: Quantity of heat in unit time (W) which passes through a unit cross-sectional area of a material for a unit temperature gradient (i.e. 1°C temperature change per unit length).


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Code Clause

1.5 Symbols EC3/EC4

The important symbols and subscripts in Eurocodes 3 and 4 Parts 1.2 cl 1.4

are as follows:

Ed design effect of actions for normal design Efi d design effect of actions for fire design fay,20 characteristic strength of structural steel at 20°C fay,e effective yield strength of structural steel at temperature, 8 R, design resistance of a member fc,20 characteristic cylinder strength of concrete at 20°C fc,e characteristic strength of concrete at temperature, 8 E , resultant emissivity yM,fi partial material factor in fire conditions h thermal conductivity of a material q load level

In general, the following subscripts to the main symbols are used:

a structural steel (acier) c concrete cr critical d design value of action, or resistance fi at fire limit state k characteristic value S reinforcing bar 8 parameter at elevated temperature

1.6 References to Eurocodes and BS 5950

In this publication a column at the right hand side of the text contains appropriate references to EC3 and EC4 Parts 1.2 and BS 5950: Part 8 with the following prefixes:

EC3 = EC3: Part 1.2 EC4 = EC4: Part 1.2 BS 5950 = BS 5950: Part 8

Other standards are referred to in full, for example EC3: Part 1.1.

The relevant clause, table or figure number is given to assist in cross-referencing. If equally relevant throughout the section, the reference clause is only given at the top of the section in this


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Code Clause

publication. Tables directly taken from the Eurocodes or BS 5950: Part 8 are acknowledged and are included by permission of the British Standards Institution.

1.7 Materials grades in Eurocodes

Structural steel grades are defined by yield strength. EC3: Part 1.1 refers to three general grades in BS EN 10025(7). Their comparable common grades to BS 5950 are as follows:

BS EN 10025 BS 5950 S235 (formerly Fe 360) -

S275 (formerly Fe 430) Grade 43 S355 (formerly Fe 500) Grade 50.

The numerical value corresponds to the yield strength of the steel (hence 275 N/mm2 for S275 steel). S235 steel is not used in the UK, but is widely used in Continental Europe. It is the reference grade in EC3: Parts 1.1 and l .2 for development of some design formulae. Grade designations 43 and 50 refer to the approximate ultimate tensile strength values (divided by lo), which are now replaced by the new designations in terms of yield strength.

In all cases, the steel strength is the guaranteed minimum value, which is lower than the characteristic strength adopted by limit state codes. The difference may be significant when considering appropriate partial safety factors for the ultimate and fire limit states.

Concrete is defined by its cylinder or cube strength, the cylinder strength being 20% smaller than the cube strength on average. Therefore, C25/30 grade corresponds to a characteristic compressive strength of 25 N / m 2 as a cylinder, and 30 N/& as a cube (the latter being the normal standard in the UK). Typical structural concrete grades are C25/30 to C40/50 in reinforced concrete or composite construction.

Reinforcement grade is defined by its proof strength rather than its ill-defined yield point. S500 is the standard grade (500 N/mm2 proof strength). This designation has recently replaced the existing British Standard, which is based on 460 N/mm2 proof strength. The ratio of the ultimate tensile strength to proof strength is relatively low (approximately 1.1 to 1.2) for most types of reinforcing bar, which leads to use of a higher partial factor of safety than for structural steel. Furthermore, the ductility of this new grade of reinforcement is relatively low. S500 B has a guaranteed elongation of 5 % at failure.


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Code Clause

2 FIRE LIMIT STATES

2.1 Fire resistance test

Three criteria are imposed by the standard fire resistance test to BS 476: Part 20(5) (formerly Part 8), and in IS0 834(4):

(1) Insulation: A fire on one side of a wall or underside of a floor acting as a compartment boundary should not cause combustion of objects on the unexposed side. Limits of temperature rise of 140°C (average) or 180°C (peak) above ambient temperature are specified in the standard fire resistance test.

(2) Integrity: A wall or floor acting as a compartment boundary should not allow passage of smoke or flame from one compartment to another as a result of breaks or cracks in the wall on floor. Both the insulation and integrity criteria also apply to members embedded in walls or floors.

(3) Load-carrying capacity: The members in a structural assembly should resist the applied loads in a fire. Failure criteria for beams and columns are defined as:

Beams: (a) A limiting deflection of spad20 is reached or, (b) For deflections greater than span/30, a rate of deflection

of span2/(9000 X member depth) is exceeded. The units of rate of deflection are mm/min when the dimensions are in mm.

Columns : Failure to support the applied load. In practical terms, this corresponds to a rapid rate of increase of vertical and lateral deflection (limit undefined). Columns fail rapidly in buckling after a relatively small increase in lateral deflection.

The fire resistance test follows a temperature-time curve (defined in BS 476: Part 2O@) and IS0 834) (4) . This curve is presented in Figure 1. It is described by the formula:

BS 476 Part 20

where:

T = furnace temperature ("C) t = elapsed time (mins).


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Code Clause

The fire resistance test is not intended to reflect the temperatures, and hence structural behaviour, that would be experienced in real fires. It is used as a means of obtaining a measure of the relative performance of structures and materials within the capabilities of standard gas-fuelled furnaces.

I I I

0 5 101520 30 60 90 W

Time (min)

Figure 1 Time-temperature curve of /SO 834 and BS 476: Part 20

Other temperature-time curves are defined for the more severe hydrocarbon fires.

2.2 Partial safety factors

Fire in buildings is a rare occurrence and for calculation purposes is treated as a form of ‘accidental’ loading. Safety factors reflect the fact that the structure is required to ‘survive’ extreme events such as fire but without need for further reserve of strength. This is because the probability of overload and inaccuracies in the method of calculation (or testing) are considered to be small and of less significance than those under normal loading situations.

EC3lEC4 cl 2.4.2 &

2.4.3

In limit state terms, structural safety is satisfied provided:

Ed 5 R, where Ed = X:yf E and R, = __ R

Y r n where:


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Code Clause

E = effect of applied action (loads in general) R = corresponding resistance of the member Ed = design value of the effect of the actions R, = design resistance of the member yf = partial safety factor on actions (loads) ym = partial safety factor on materials.

The multiple yf ym broadly represents the minimum overall factor of safety that is required in normal conditions.

The actions (or loads) often act in combination and their total effect EC1 Part 1

on the design value of the actions at the ultimate limit state is taken into account as follows:

where :

yG and y Q are the individual partial factors for permanent and variable actions.

$lo is a factor due to the probability of one form of loading acting in combination with another. Qk is the individual variable load considered in combination, and Gd is the permanent load (normally dead loads).

At the fire limit state, which is treated as an accidental situation, the ECI Part 1 design effect of the actions is expressed differently as:

where *l and & are factors due to the probability of loads acting individually or in combination. Implicit in this formula is that yG and yQ are set to unity.

The partial safety factors on loads in Eurocodes 3 and 4 are presented in Table 1 for design at both the ultimate and fire limit states, in comparison to those of BS 5950: Part 1 and 8@). These factors include the relevant load combination factors, q0 at the ultimate limit state, and and % at the fire limit state, which are defined in Table 2. In principle, design to the Eurocodes requires more potential load combinations than to BS 5950.

The serviceability limit state corresponds to control of deflections, cracking, etc. and is checked using unity partial safety factors. There is no equivalent serviceability condition in fire design. Therefore, structures are only assessed for their fire resistance and not for limitation of damage in modest fires.


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Code Clause

2.3 Design values of actions at fire limit state

Loads or actions at the fire limit state take into account the low probability of extreme loading co-incident with a fire of the severity necessary to cause structural collapse. These loads are defined in EC3/EC4

Eurocode 1 Part 2.2(8). Loads may be classified as those that are cl 2.4.3 &

likely to occur over a long period, (i.e ~ permanent loads) and loads Fig.2. l

that are inherently variable in magnitude. The partial factor on permanent loads is taken as unity, but the partial factor on variable EC1 Part 2.2 imposed loads at the fire limit state is reduced to as low as 0.5 for domestic and office buildings. This is the factor $, in Table 2. The corresponding value in BS 5950: Part 8 is 0.8 for variable loads.

BS 5950 cl 3.1 & Table 2

A fire in a compartment may be assumed to ‘consume’ a major part of the loading in it. However, the floor above a compartment is both subject to load and to the effects of fire from beneath. The factor therefore reflects only the probabilistic effects of loading in accidental situations and not to any loss of loading in a fire.

Wind loads may be considered to act alone, or in addition to vertical loads using the load combination factors in Table 2. As defined, there are two possible load combinations depending on whether or not wind is the dominant action. If wind actions are small, then their effect is ignored (q2 = 0). (The equivalent value is 0.33 in BS 5950: Part 8). If wind actions are dominant, as may be the case in sway frames, then a partial factor of 0.5 for wind actions is used ($, = O S ) , but the effect of the other actions is reduced (q2 = 0.3). No relaxation is given for the minimum height of building below which wind loads are not considered. (The equivalent height is 8 min BS 5950: Part 8).

No account is taken of forces or actions induced due to thermal or second order effects in fire conditions. This is because of the indeterminate nature of these actions and because these forces may also reverse in direction and act as restraints at large deflections. In general, neglecting thermally induced actions will not lead to premature failure in ‘robust’ construction.

Snow loads would not normally be considered in combination with other loads in design of frames, although the Eurocode does make provision for the case where snow loads are the predominant actions, such as on roofs. This case corresponds to $ = 0.2 on snow loads in Table 2.

BS 5950 Cl 3.1

In general, the design values of the actions and hence forces and moments at the fire limit state will be lower according to the


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Eurocodes than in conventional UK practice. Regulators throughout Europe may well adopt a different view on these partial factors, although the Eurocodes values reflect increasing confidence on the actual performance of buildings comprising all structural materials in real fires.

2.4 Material factors

Partial safety factors on materials take into account the inherent variability of the material strength, the member geometry and possible workmanship errors. These factors vary with the limit state considered and are discussed below for the relevant materials. The values are summarised in Table 3 .

2.4.1 Steel

The partial safety factor for steel at the ultimate limit state is taken as 1.05 in the UK NAD to Eurocode 3 : Part 1.1. This factor was largely adopted in order to obtain an equivalent overall factor of safety to that given by design to BS 5950: Part 1 . Many countries have adopted a partial factor for steel of 1 .O, whereas others have retained the ‘boxed’ values of 1.1 in the original draft Eurocode. It is expected that with further justification in terms of actual material strengths and tighter control of member geometry, a value of l .O could be adopted universally. This factor is set to unity at the serviceability limit state.

EC3lEC4 cl 2.3

BS 5950 cl 3.2

The partial factor for steel at the fire limit state is also taken as 1 .O, which implies that the design strength of steel in fire is its minimum guaranteed value (see Section 1.7). Logically, this value should be lower than that of concrete, but a decision was made by the CEN Co-ordinating Group that all materials should have a unity partial factor in fire conditions.

2.4.2 Concrete

The partial safety factor for concrete at the ultimate limit state is EC2: Part 1.2 taken as 1.5 in the UK NAD to Eurocode 2(’). The partial factor for concrete at the fire limit state is reduced to 1 .O, which is below the equivalent value of 1 .3 used in conventional UK practice. (Refer to the Institution of Structural Engineers publication(”)).

2.4.3 Reinforcement EC3: Part 1.2

The partial safety factor for steel reinforcement at the ultimate limit state is taken as l . 15 in the UK NAD to Eurocode 2(’). This factor is reduced to 1 .O at the fire limit state.


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Table 1 Partial safety factors for loads at the ultimate and fire limit states

ACTION FIRE L.S. ULTIMATE L. S.

EC 1 /EC3 BS 5950 EC 1 /EC3 BS 5950

Self weight and dead

1 .o 1 .o 1.6 1.5 Imposed loads -

loads 1 .o 1 .o 1.4 1.35

permanent

Imposed loads - variable 1.5 1.6 0.5 to 0.9

0.33 0 or 0.5 1.2 1.05 Wind loads*

0.8

* acting in combination with imposed loads (see Table 2)

Table 2 Load combination factors at ultimate and fire limit states in Eurocode 7 Part 2.2i81

Domestic Office Congregational Shopping Storage

Snow Wind

Load Combination Factors

90 9 2 91 0.7

0.8 0.9 1 .o 0.6 0.7 0.7 0.6 0.7 0.7 0.3 0.5 0.7 0.3 0.5

0.6 0.2 0 0.6 0.5 0

Table 3 Comparison of partial safety factors for materials

MATERIAL I ULTIMATE L. S. FIRE L.S. 1 EC3 BS 5950 EC3 BS 5950

Structural steel I 1.05 I 1.0 1 .o 1.0 I Concrete 1.3 1 .o 1.5 1.5

Reinforcement

1 .o 1 .o 1.25 1.25 Shear connectors

1 .o 1 .o 1.15 1.15

-


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Code Clause

Load level

The concept of ‘load ratio’ or ‘load level’ is very important in BS 5950 understanding the behaviour of structures in fire. The term load ratio cl 4.4.2 was first introduced in BS 5950: Part 8, and is defined as the proportion of the member resistance that is utilised at the fire limit state or explicitly as:

Load ratio = Forces or moments at the fire limit state Member resistance to these forces or moments

( 5 )

The forces or moments are determined using the relevant partial factors at the fire limit state (see Section 2.3). The member resistance is that corresponding to the ‘normal’ or ‘cold’ condition, using the partial factors for materials corresponding to the ultimate limit state. Combination of actions (such as moments and axial forces) may be treated as in Section 4.4.2.

The term ‘load level’ is used in Eurocodes 3 and 4 Parts 1.2, which has the same broad definition at the load ratio, and may be expressed as a mathematical ratio. The various terms in the Eurocodes which represent the actions and resistances of a member, are defined as follows:

Ed = action effect at the ultimate limit state Ef,,d,t = action effect at the fire limit state

= member resistance at the ultimate limit state Rti,d,t = reduced member resistance at exposure time, t , at the

fire limit state.

The load level at exposure time, t , at the fire limit state may therefore be defined as a proportion of the resistance of the member:

%,t - - ‘%i,d,t

&l

As noted above, R, is determined using the partial factors at the ultimate limit state. In design, it is necessary to demonstrate that at the relevant fire resistance period:

EC3lEC4 cl 2.4.3

Because of the different treatment of load and material factors in EC3 and EC4 Parts 1.2 relative to BS 5950: Part 8, the maximum permitted load level given by Equation (7) will be slightly greater than the load ratio in the BS 5950: Part 8 by the multiple of the


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Code Clause

partial factor for steel (i.e. 1.05). It should also be noted that direct comparisons between these codes are not straightforward because of the different partial factors applied to variable imposed loads (see Table 2).

The concept of load level is confused in EC3: Part 1.2 by the EC3 Cl 2.4.3 8~ introduction of a further term, qfi which is the reduction factor for the design load level at the fire limit state. Effectively, this term is the proportionate loading at the fire limit state (qfi = Efi,d,t/Ed).

Taking a typical case where half of the total applied load is variable, it follows from Tables 1 to 3 that the corresponding reduction factor for the design load level, qfi, for a member at the fire limit state would be 0.53 according to EC3 : Part 1.2, and the equivalent load ratio would be 0.6 according to BS 5950: Part 8. This assumes that the member reaches its design resistance at the ultimate limit state.

EC3 Fig 2.1

Many members are designed to satisfy various serviceability criteria and consequently have some reserve of strength. In practice, it could be expected that the actual load levels or load ratios of typical members would be in the range of 0.45 to 0.55, according to Equations ( 5 ) or (6).

Therefore, a design load level of approximately 0.5 would correspond to an appropriate value for fire resistant design of most structural members in buildings.


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Code Clause

3 PERFORMANCE OF MATERIALS AT ELEVATED TEMPERATURES

All materials lose strength at elevated temperatures and it is the purpose of fire resistant design that the reduced resistance of structural members exceeds the applied forces at the fire limit state. Other important properties of materials in fire are the coefficient of thermal expansion, the specific heat and thermal conductivity.

3.1 Physical properties of structural steel

3.1.1 Introduction

Steel begins to lose strength at temperatures above 300°C and the EC3 cl 3.2

decrease in strength is approximately at steady rate with increasing EC4 cl 3.2.1

EC3lEC4 temperature until about 800°C. The residual strength of steel, Annex A although small, then reduces more gradually until reaching its melting temperature at approximately 1500°C. In comparison, temperatures in a standard fire resistance test do not exceed 1200°C (after 4 hours).

Conventionally, steel is assumed to retain half its strength at 550"C, which is often considered to be its 'failure temperature'. In practice, structural steel displays a higher strength retention at this temperature, although the above assumption is valid for reinforcing bars.

The behaviour of steel in fire is affected by the rate of heating, as BS 5950 cl 2.1 there is a component of deformation arising from creep at temperatures above 450°C. For this reason, research has concentrated on the effect of heating rate. Assuming that the rise in temperature of a section is linear and that a temperature of 600°C is reached after 30 and 120 minutes respectively, then practical rates of heating are S"C/min for well insulated sections, to 20"C/min for unprotected or lightly insulated sections. Small-scale tensile tests using rates of heating in this range would therefore be representative of the performance of structural members in fire.

cl 2.2

Data on the physical properties of steel are given in Eurocodes 3 and EC4 Annex A 4 Parts 1.2 . Eurocodes 2, 3 and 4 are consistent in terms of material properties. However the Annexes on steel properties in EC4: Part 1.2 are informative in order that the same properties are presented in EC3: Part 1.2 as 'normative'. Informative Annexes may be deleted later in the EN version. For comparison, the property data given in BS 5950: Part 8 follows the Eurocodes closely, although the specific heat and thermal conductivity of steel are given as simplified single values in the BS.

EC3 Annex A


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Code Clause

3.1.2 Methods of test

The strength retention of steel is a function of the heating rate, the strain limit used for determination of the strength, and the method of test itself. There are two basic methods of test:

Isothermul or steady state tests have been traditionally used for mechanical engineering applications, where the specimen is subject to constant temperature and further strain is applied at a constant rate. The stress-strain curve is therefore appropriate for a given temperature.

Anisothermal or transient test are ones where the specimen is subject to a constant load and the rate of heating is set at a pre-determined amount. The resulting strains are measured. The effect of thermal strains are deducted by using 'dummy' unloaded specimens subject to the same temperature conditions. Stress-strain curves for a particular temperature are obtained by interpolation from a family of curves at different stresses. The reference heating rate is generally taken as 10Wminute (i.e. 600°C rise in 60 minutes).

Anisothermal tests result in lower strengths than isothermal tests, but may be considered to be more realistic for structural applications. Stress-strain curves for steel at different temperatures are shown in Figure 2. It is apparent that the strength retention increases gradually with strain, which is a form of strain hardening at elevated temperature.

The choice of strain limit is therefore very important. The former ECCS Re~ommendations(~) used an effective 'yield' criterion of 0.5 % strain for temperatures above 400°C. However, BS 5950: Part 8(6) increased this strain limit to 1.5% for beams because fire tests on these members had shown that strains well in excess of 3 % could be reached at the limiting deflection of the tests (see Section 4.3). As a result of a considerable amount of research and testing undertaken by British Steel(") and Arbed Recherches, an effective strain limit of 2% is now adopted for determining the strength of steel at elevated temperatures in EC3 and 4 Parts 1.2.

However, it should be recognised that the faster the rate of heating, the higher the temperature at which a particular steel strength is attained at a certain strain. Therefore, unprotected members often fail at slightly higher temperatures than protected members for the same load level. Despite this effect, it is generally considered that the effective strain limit is appropriate for most steel members failing in a flexural mode.


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Code Clause

3.1.3 Strength retention of steel

The values of strength retention of structural steel of grades S235, EC4 Table 3.1 S275 (Grade 43) and S355 (Grade 50) are very close when ‘normalised’ with respect to their room temperature yield strengths. These normalised data are presented in Table 4 as a function of temperature and strain. The normalised strengths corresponding to 2% strain are used in the design of all types of structural members designed to EC3 and 4 Parts 1.2. Empirical formulae are given in EC3 Fig 3.1 EC3: Part 1.2 which define the form of the stress-strain curve when used in computer analyses. No further strain hardening is included at elevated temperatures.

In EC3: Part 1.2, the strength retention of steel is presented as a strength reduction factor.

The strength reduction factors for these steel grades according to the former ECCS recommendation^,(^) BS 5950: Part 8 and EC3: Part 1.2 (‘1 are presented in Figure 3 , The three strain limits in BS 5950: Part 8 apply to different types of members. A strain limit of 1.5 % is used for beams whose protection materials remain intact in a fire. A higher strain limit of 2 % is used for composite beams. As noted earlier, a strain limit of 2% is used throughout in EC3: Part 1.2 and modification are made for different types of members.

Strain

BS 5950 Table 1

Figure 2 Stress-strain curves for steel at elevated tempera tures


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Code Clause

Table 4 Stress-strain relationship at elevated temperatures for grade S 275 steel

I I Effective yield strength at elevated temperature, relative to yield strength at 20 "C.

I Steel temperature 0, ["C]

100 800 700 600 500 400 300 200

0 . m o.Oo0 0.000 o.Oo0 0.000 0.000 0.000 o.Oo0 o.Oo0 0,0005

0,058 0,091 0,212 0,407 0,482 0,611 0,687 0,764 0.0010 0,034 0,050 0,118 0,229 0,267 0,305 0,344 0,382


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Code Clause

In BS 5950: Part 8, lower strain limit of 0.5% is used for columns and reflects the influence of instability of these members. The difference in the strength retention of steel between 1.5 % and 2 % strain is small, and therefore the design values in BS 5950: Part 8 and EC3: Part 1.2 are very close (when applied to flexural members).

These comparisons of strength reduction factors are presented in 1x3 Table3-1 Table 5. The relevant strength reduction factor used in design to EC3: Part 1.2 is ky.e. (This is known as km,, in EC4: Part 1.2.) Table 1

Values are also presented for the proportional limit, kp,e, although this parameter is not used in structural fire design. Yielding is not discernable from the stress-strain curves at high temperature and, therefore, the proportional limit is taken as at a strain of approximately 0.1 % . A further parameter kx,e is defined which corresponds to a strain of approximately 0.5 % , and is used where control of deformations is required in fire conditions.

EC4 Table 3.2 BS 5950

;; 1.0 c 0 cu - 0.9 c 0 .- c

U 2 0.8

E r, 0.7

v) E 0.6 c 0)

0.5

0.4

0 . 3

0.2

0.1

0

Variation in strength : - - Eurocode 3: Part 1.2

BS5950: Part 8 - - 2% strain

ECCS Recs.

elastic modulus 'l- (EC3 Part 1.2)

0 100 200 300 400 500 600 700 800 900 1000

Temperature ("C)

Figure 3 Variation of the strength retention of steel at elevated temperature according to various codes


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Code Clause

Table 5 Strength reduction and elastic modulus factors for structural steel according to EC3 Part 1.2 and BS 5950 Part 8

ky,e corresponds to the steel strength at 2% strain relative to its normal yield strength (this parameter is presented as kmax,e in EC4: Part 1.2)

kx,e corresponds to approximately 0.5 % strain kp,e corresponds to an effective yield value (= 0.1 % strain) kE,e is the initial tangent modulus divided by 200 kN/mm2

Stress-strain data is also given in EC4: Part 1.2 for higher strength S460 steel which is currently not widely used in the UK, and for which no general design rules are given in EC3 and EC4 Parts 1.1.

3.1.4 Other properties of steel

The coefficient of thermal expansion of steel, aT, increases slightly with temperature. At room temperature aT is taken as 12 X 10-6/oC, but at temperatures of 200 to 600°C, aT may be taken as 14 X 10-6/"C. At around 73OoC, steel undergoes a phase change and there is a marked change in the expansion characteristics as energy is absorbed and the material adopts a denser structure. This behaviour is illustrated in Figure 3.3 of EC4: Part 1 .2(2) and in reference (12).

The specific heat of steel is the heat stored (in joules) in a unit mass of steel for a 1 "C temperature rise. For simple calculations an average specific heat of steel C, of 600 J/kg"C may be adopted. A precise relationship between C, and temperature 8 is given in EC3 and EC4 Parts 1.2, and is illustrated in Figure 4.

EC3 cl 3.3.1.1 EC4 cl 3.3.1

EC3lEC4 Fig 3.3

BS 5950 cl 2.1

EC3 cl 3.3.1.2 EC4 cl 3.3.1

EC3lEC4 Fig 3.4


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Code Clause

The specific heat rises dramatically at about 700°C due to the phase change noted above, and there may be a slight delay in the temperature rise of a steel member at this point in a standard fire test. Above 750"C, the specific heat falls to a lower level. The precise formula for the specific heat of steel is recommended for use in computer analysis as it can have an important effect at these high temperatures.

The thermal conductivity of steel is the amount of heat per second 1x3 cl 3.3.1.3 (Watts) that passes through a unit cross-sectional area of material for EC4 cl 3.3.1

a unit temperature gradient (i.e. 1 "C temperature change per unit length). This parameter is not as important for steel as for fire protective materials, because the thermal conductivity of steel is 50 times greater than that of concrete, and 500 times greater than that of common fire protective materials. The thermal conductivity of steel is approximately 54 W/m"C at room temperature, reducing to an average value of 45 W/m"C at higher temperatures. A precise formula for the thermal conductivity of steel is given in EC3: Part 1.2. A single value of 37.5 W/m"C is given in BS 5950: Part 8.

Poisson's ratio for steel may be taken as 0.3 and the density of steel may be taken as 7850 kg/m3. Both values are temperature independent.

Specific heat [ J / kg K ]

1500

1000 ~-__-

/ l -

\ 500-- -- 0 7--- 0 200 400 600 800 IO00 1:

Temperature

EC3lEC4 Fig 3.5

BS 5950 cl. 2.1

00

c ."C 1

Figure 4 Specific heat of steel at elevated temperatures


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3.2 3.2.1

Code Clause

Physical properties of other forms of steel

Reinforcing steel

Reinforcement should conform to BS EN 10080('3) (formerly BS 4449 was operational in the UK until early 1996). Important characteristics are the yield and ultimate strength, and minimum elongation at failure. Reinforcing bars used in a primarily structural r61e (as opposed to links and ties) are available in standard diameters of 12, 16, 20, 25, and 32 mm. The standard grade of steel for reinforcing bars and mesh is S500 (i.e. 500 N/mm2 yield strength). This grade corresponds to the former grade of 460 N/mm2 in BS 4449 and BS 8110(14). Mesh (or fabric) reinforcement is also covered by BS EN 10080. It is usually specified in terms of its cross-sectional area of steel per unit width (e.g. A142, 193, 252 etc.).

The ductility (elevation at failure) of S500 steel is very low and even the more ductile designation S500 B has a guaranteed value of only 5 % , unlike the 12% in BS 4449. This relatively low ductility has implications for structural fire design, where significant redistributions of moment are required.

The strength of cold worked reinforcing bars at elevated temperatures is presented in EC2 and EC4 Parts 1.2, and in Table 6. The design values in the Eurocodes differ slightly from the publication 'Design and detailing of concrete structures for fire resistance' published by the Institution of Structural Engineers(*') in 1978, and in BS 8110: Part 2(14). This is due to the higher strain limit now adopted in the Eurocodes.

Table 6 Strength reduction factors for cold worked reinforcing bars

I I Strength Reduction Factors I Temperature

100 200 300 400 500 600 700

Eurocode 4

l .o 1 .o l .o

0.94 0.67 0.40 0.12

BS 8110

1 .o 1 .o 1 .o

0.81 0.62 0.42 0.20

Elastic Modulus Factor

1 .o 0.88 0.72 0.56 0.40 0.24 0.08

Cold worked reinforcing bars lose strength more rapidly than hot rolled bars above 600"C, due to the loss of the cold working effect. However, it is the purpose of the insulation provided by the concrete cover to keep the temperatures of the reinforcing bars below this


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Code Clause

temperature. EC4: Part 1.2 states that the strength of hot rolled bars at elevated temperatures may be taken as for structural steel.

The strength reduction of prestressing wires at elevated temperatures is much greater than of reinforcing bars, and therefore greater concrete covers are usually required for prestressed elements in order to insulate the prestressing wires in fire.

3 .2 .2 Cold formed steel

EC3: Part 1.2 does not cover the use of cold formed steel sections and the designer should refer to BS 5950: Part 8 for guidance.

BS 5950 Appendix B

The anisothermal high temperature properties of galvanized cold formed steel to BS EN 10147(15) (formerly BS 2989) have been determined by British Steel for temperatures up to 600°C. Strength reductions of cold formed steel are greater than for structural steel in the important temperature range of 400 to 600°C. However, this data is based on a 95 % confidence limit (i.e. 5 % chance that the steel strength will fall below the specified value), and is therefore lower than the mean value used to assess the design values for structural steel.

Guidance on the fire protection of cold formed steel members used as primary structural members is given in a recent SCI Publication(16). Where these members perform a secondary structural r61e as purlins or side rails, they are generally not fire protected. An important exception, for example, is in a compartment wall in a warehouse.

3 .2 .3 Stainless steel

EC3: Part 1.2 does not cover the use of stainless steel sections and the designer should refer to other published guidance(12).

Stainless and other high alloy steels display different properties at elevated temperatures. Modern austenitic stainless steels (3 16 S31 and 304 S16 grades) lose strength relatively rapidly at about 1OO"C, and a 50 % strength reduction is reached at about 400 "C. However, at higher temperatures, the strength retention stabilizes, and at 800"C, the strength of stainless steel is still about 40% of its original value. This suggests that it might be possible to design stainless steel members as unprotected if they are subject to low stresses.

The coefficient of thermal expansion of stainless steel is higher, and the thermal conductivity is much lower, than for carbon steels.


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3.2.4

3.2.5

Code Clause

Bolts

The two main types of bolts used in structural applications are grade 4.6 and 8.8 bolts (formerly to BS 3672). Grade 4.6 bolts are forged from mild steel, whereas grade 8.8 bolts are manufactured from micro-alloy steel which is quenched and tempered to obtain their high strength. The margin between the strength at 0.2% proof strain and ultimate tensile strength is much lower in grade 8.8 bolts than in grade 4.6 bolts.

There is not any data or guidance on the strength of bolts at elevated temperature in the Eurocodes. However, recent research by British Steel(17) has quantified the performance of these types of bolts in tension and shear. In principle, the strength retention of these bolts is at least equivalent to the strength of the parent steel at elevated temperatures, but at 600 to 700°C there is a more marked loss of strength. However, actual bolts strengths are always considerably higher than the specified minimum, and the normal partial safety factor of 1.25 for bolts is set to 1.0 in fire. Therefore, the actual strength retention of bolts is proportionately higher than the adjacent steel members in fire conditions.

The evidence of fire damaged buildings substantiates this conclusion, because generally bolted connections remain intact even though the frames or adjacent members may have distorted or failed. A series of fire tests have been carried out on common forms of steel and composite connections to assess their behaviour('*). In all cases the connections failed at higher temperatures than predicted, and the moment-rotation curves were 'ductile'. As a simple rule, the connections should be fire protected to the same level as the adjacent members. However, special measures may be required if the bolts in an otherwise protected member are directly exposed.

Welds

The strength of welds decreases markedly in the temperature range of 200 to 400"C, but then stabilizes close to the reduced strength of the parent steel. British Steel have recently carried out a comprehensive series of tests on welds at elevated temperatures. In BS 5950: Part 8, the strength reduction factor for welds was taken as 80% of the value corresponding to the strength of steel at 0.5% strain. The British Steel tests showed that this approach is conservative, especially when the partial safety factors for welds are set to unity in fire conditions.

No data or guidance on welds is presented in EC3 : Part 1.2. No special measures need generally be taken for welds, provided the members and their connections are protected to the same amount.


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Code Clause

Exceptions are welded shear blocks supporting beams that may necessitate additional protection.

3.2.6 Cast and wrought iron

EC3 : Part 1.2 does not cover the use of cast and wrought iron and the designer should refer to other published guidance(19).

Cast iron is a brittle and variable quality material and is not able to undergo significant distortion in fire because of its weakness in tension. Failure of cast iron columns in old buildings in fire often occurred because of secondary expansion of floors, or differential heating and cooling after the fire rather than direct loss of strength. In principle, the elevated temperature properties of cast iron relative to its room temperature properties are similar to those of mild steel.

Wrought iron may also be variable in quality, but is relatively ductile. Wrought iron loses strength more rapidly than mild steel at temperatures above 450°C but the strength retention values may be taken from Table 5 up to this temperature.

3.3 Physical properties of concrete

3.3.1 Normal weight concrete

Normal weight concrete can be composed of either siliceous aggregate (e.g. gravel or granite) or calcareous aggregate (e.g. limestone). The elevated temperature properties of concrete depend to some extent on the aggregate type. Concrete loses strength less rapidly than steel at temperatures above 300°C, but can suffer from the phenomenon of spalling, that is breaking away of the concrete cover to the reinforcement. For bending members, it is therefore the rise in temperature and the consequent loss of strength of the reinforcing bars that determines the fire resistance of the member.

EC4 cl 3.2.2

Calcareous aggregate concretes possess lower thermal conductivity and have less tendency to spa11 than siliceous aggregate concretes. The strength retention data in Eurocodes 2 and 4 Parts 1.2 are based on the CEB Recommendations(20), which are significantly different to the data in BS 81 10: Part 2. No distinction is made between these 1x4 Table 3.3

two concrete types in EC4: Part 1.2 and the lower strength values corresponding to siliceous aggregate concrete are tabulated, although reference is made to data in EC2 Part 1.2 for other concrete types.

EC2 Part 1.2

The strength reduction factors for these two concrete types at elevated temperature are presented in Table 7(a). As noted above, the siliceous concrete values are used in EC4 Part 1.2. The strains corresponding to these maximum strengths range between 0.25 and


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Code Clause

l .5 % over the temperature range of 100 to 800°C. For comparison, the strength reduction factors for concrete in BS 8110: Part 2 are presented in Table 7(b). It is apparent that BS 81 10 takes no strength reduction for temperatures up to 350°C and the strength values are generally higher than for siliceous aggregate concrete in EC4: Part 1.2. However, the higher partial safety factor for concrete in BS 81 10 at the fire limit state than in EC4: Part 1.2 is more significant than slight variations in strength.

The stress-strain data for concrete in compression is presented in the 1x4 Cl 3 . 2 ~ ~ Eurocodes as a function of temperature, and an empirical formula is given for this relationship in both EC2 and EC4: Part 1.2. The EC4 Annex B declining portion of the stress-strain curves are defined in Annex B of EC4: Part 1.2 (which is informative) for use in advanced computer analyses. Strains of up to 5 % can be experienced before the concrete loses all its strength. In simple calculation models, the strength retention corresponding to the average temperature of the concrete is used in 'plastic stress' block analysis.

Figure 3.2

The elastic modulus of concrete is used in the design of composite columns, and the initial tangent modulus is the relevant parameter for determining the stiffness of the concrete. Values are also presented in Table 7 . The secant modulus corresponds to the point of maximum strength and is taken as two-thirds of the initial tangent value. It is apparent that the elastic modulus of concrete reduces more rapidly than that of steel at elevated temperatures. Over the important range of temperature of 300 to 500"C, the ratio of the elastic moduli of steel to concrete increases to approximately 30 from its initial value of 7 at 20°C. This means that the stiffness of composite members reduces proportionately more than that of a steel section, due to the greater loss of stiffness of the concrete.

In principle, concrete in tension is ignored in fire design; the concrete simply provides insulation to the steel in the tensile zone. The concrete cover to the main reinforcing bars is assumed to maintain the steel temperature below 550"C, so that at least half its strength is retained in fire conditions. The minimum concrete covers that are specified in EC2 Part 1.2 (and BS 81 10: Parts 1 and 2) increase with fire resistance period. Supplementary mesh or links are often required where the concrete cover exceeds 40 mm in order to prevent spalling of the concrete. It should be noted that the 'axis distance' to the centre of the reinforcing bar is used to define the bar placement in EC2 and EC4 Parts 1.2, which is greater than the cover to the edge of the bar (by half the bar diameter).

The physical properties of normal weight concrete are presented in EC2 Part 1.2. Its dry density is generally taken as 2350 kg/m3 (increasing to approximately 2450 kg/m3 when reinforcement is

EC4 cl 3.3.2


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included). The specific heat of concrete depends on the type of aggregate, but may be taken as an average value 1000 J/kg"C over the normal temperature range in fire. The thermal conductivity of concrete is much lower than that of steel and may be taken as an average value of 1.6 W/m"C. The actual value reduces gradually with temperature from 2 W/m"C at room temperature to 1 W/m"C at 700°C. The coefficient of thermal expansion of concrete is slightly higher than that of steel, but the differences are not significant to the structural behaviour of reinforced or composite sections.

3.3.2 Lightweight concrete

Lightweight concrete used in structural applications consists of pelletized aggregate made from either sintered fuel ash (known as Lytug) or expanded clay. The dry densities of structural lightweight concretes are in the range 1600 to 2000 kg/m3, although for C20 to C40 strength grades, densities exceeding 1800 kg/m2 are usually required. In the UK, Lytug aggregate with sand fines is often specified as this is readily pumpable with workability additives. Its dry density is in the range of 1850 to 1950 kg/m3. Other types of lightweight aggregate are used elsewhere in Europe.

Lightweight concrete possesses much better fire resistant properties, as it has a lower thermal conductivity and coefficient of thermal expansion, and higher strength retention than normal weight concrete at elevated temperatures. However, its specific heat is lower, which can influence the response of some structural members in fire. This EC2 Part 1.4 data is presented in EC2: Part l .4(21), a special Part dealing with lightweight concrete design both in normal and fire conditions, and also in EC4: Part 1.2.

Typical physical properties of lightweight concrete that may be used EC4 cl 3.3 .3 in structural fire design are presented in Tables 7 and 8. These values are appropriate for all lightweight concretes in the broad density range of 1600 to 2000 kg/m3. The properties in Table 8 are averaged over the range of 20 to 600°C. In BS 81 10 Part 2, it is assumed that Lytug concrete is used, which gives no strength reductions up to 500°C. At higher temperatures, strength retention values are significantly better than in EC4: Part 1.2.

A general requirement of EC4: Part 1.2 is that lightweight concrete is only considered for composite floors or composite beams in conjunction with composite floors. This restriction is made because of the limited use of lightweight concrete for encased columns and beams in many parts of Europe, rather than lack of design or test data.


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The better insulating properties of lightweight concrete will lead to higher fire resistance of composite slabs than with normal weight concrete. For simple calculation models, the temperature rise in lightweight concrete may be taken as 90% of the values in normal weight concrete. Futhermore, this means that thinner slabs are possible with lightweight concrete in order to satisfy the necessary insulation requirement in fire (see Section 8.1).

No stress-strain data is given for lightweight concrete for use in advanced computer analysis. The values of initial elastic modulus given in Table 7(a) are appropriate to the properties of Lytug aggregate concrete.

Table 7(a) Strength reduction factors and elastic modulus for concrete at elevated temperatures according to EC4 Part 7.2

Temperature (c) Strength Reduction Factors I Elastic Modulus

I NWC(S)

100

0.08 900 0.15 800 0.30 700 0.45 600 0.60 500 0.75 400 0.85 300 0.90 200 0.95

NWC(C) 0.97 0.94 0.91 0.85 0.74 0.60 0.43 0.27 0.15

(kN/mmZ)

LWC LWC NWC

1 .o 12.5 9.0 1 .o

7.5

0.3 0.3 0.28 0.5 0.5 0.40 l .o 1 .o 0.52 1.4 1.7 0.64 2.2 3 .O 0.76 3.2 4.5 0.88 4.5 6.5 1 .o 6.0

Notes: NWC(S) means siliceous aggregate concrete (as used in EC4 Part 1.2) NWC(C) means calcareous aggregate concrete LWC means lightweight aggregate concrete Elastic modulus is the initial tangent modulus

Table 7(b) Strength reduction factors for concrete at elevated temperatures according to BS 8 l 10 Part 2

Temperature (c)

300 400 500 600 700 800 900

Strength Reduction Factors U

NWC

1 .o 0.91 0.73 0.56 0.38 0.20 0.03

LWC

1 .o 1 .o 1 .o 0.8 0.6 0.4 0.2

EC4 Table 3.3


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Table 8 Typical properties of normal and lightweight concrete at ele va ted tempera tures

I ProDertv I NWC I LWC I [Density (kg/m3)

~

l 2350 1850 approx. I SDecific heat (J/kg"C) I lo00 I 700 I Thermal conductivity (W/m"C) 1.6 0.8 1


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Code Clause

4 CRITICAL TEMPERATURES OF STEEL MEMBERS

4.1 Introduction to critical temperatures

In Section 3, the way in which steel loses strength with increasing temperature was described. All steel members lose strength in a similar manner, but because of aspects such as temperature gradients and instability effects at elevated temperatures, it is not necessarily appropriate to apply the same strength reductions to structural members as to the material itself.

Consider a simple design model for a member in pure tension. The reduction in tensile resistance of the member would exactly follow the reduction in strength of the material. In this case the temperature at which the member would fail in fire is determined from the load level applied to the member, which corresponds exactly to the strength reduction factor for the material. Mathematically, this is given by the simple equation at the limit of the resistance of the member:

Efi,d,t is the load level applied to the member as a proportion of its Rd

design resistance (= qfi d in Equation (6) ) . , >

ky ,O is the strength reduction factor for steel at temperature, 8. Values, which are appropriate for different steel grades are given in Table 5 in Section 3.

A further parameter arises because of the different partial safety factors used to determine the load level at the fire limit state. Equation (8) is correct for unity partial factors at both the ultimate and fire limit states. However, R, is calculated taking into account a partial factor for steel ym of greater than 1 .O (1.05 in the UK NAD) at the ultimate limit state. It follows that the effective value of ky,O may be increased to compensate for the increased value of effective load level, taking into account the ratio of partial factors. Therefore:

EC3 cl 4.2


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The term k’y,e is not used in the Eurocodes and is used here for simplicity to avoid repeating the ratio of partial factors.

When using the UK NAD, it may be assumed that kty,O = 1.05 ky,e in all calculations. This factor of 1.05 applies to all steel and composite members.

The temperature e,, at which the member would fail under load is known as the ‘critical temperature’ in EC3: Part 1.2 and the ‘limiting temperature’ in BS 5950: Part 8. The concept is the same in both codes.

The critical temperature of the member is determined by iteration 1 x 3 Cl4.2.4(2) knowing all the parameters in Equation (9). An alternative empirical equation for e,, is given in Equation (4.1.8) of EC3 : Part 1.2 which fits the solution to this equation reasonably well, and may be used for computer analyses. It is only appropriate to uniformly heated members.

& Table 4.1

In principle, the critical temperature is only a function of the load level and the type of member, and not its shape or size, or the form of fire protection. The critical temperature can, therefore, be considered to be time and section independent.

The designer then has to calculate how much fire protection is required in order that the critical temperature of the member is reached at a time which is equal to, or greater than the specified fire resistance period. The higher the critical temperature of the member, the smaller the amount of fire protection that is required to achieve a given fire resistance. Therefore, low load levels and hence higher critical temperatures of members can be used to achieve some reduction in protection thicknesses, leading to economies in construction.

4.2 Members in tension

Tension members are typically in the form of bottom chords or E C ~ ,I 4.2.3.1

bracing members in trusses, and ties or wind bracing members in frames. In principle, their critical temperatures may be determined from the strength reduction factors for the material, as noted in Equation (9) above.

However, the strength reduction factors for steel in EC3 and 4 Part 1.2 are based on a 2% strain limit. At this strain, significant extensions of bracing members may occur in fire (200 mm for a 10 m long member), leading to possible second order effects in heavily loaded columns and frames.


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BS 5950: Part 8 recognises this effect by specifying a lower strain BS 5950 limit of 0.5 % for members in tension, which leads to lower critical temperatures because of the lower strength retention of steel at this strain (see Section 3.1 .3) . In EC 3 : Part 1.2, no limitation is made on the application of the rules to tension members.

cl 2.3

Where a designer is concerned about the effect of extension of the bracing members on the stability of the whole structure in fire conditions, it is prudent to adopt a lower strain limit or critical temperature for these members, as in BS 5950: Part 8. This may be achieved by adopting the same critical temperature for tension members as for stocky columns (see Section 4.4).

4.3 Members in bending

4.3.1 Classification of cross-sections in bending

Members in bending may be designed in accordance with various analysis principles, as determined by the classification of the cross-section, which is expressed in terms of the proportions of the compression flange and web. From the point of view of fire design, it is the section classification as influenced by elevated temperatures that is important. This is because of the high strains that are experienced at the fire limit state. The section classes for normal design of members in bending are as follows:

Class 1: The member resistance may be established on plastic analysis principles, and the frame may be designed by plastic hinge analysis.

Class 2: The member resistance may be established on plastic analysis principles, and the frame should be designed by elastic analysis without formation of plastic hinges.

Class 3: The member resistance should be established on elastic analysis principles, and the frame designed by elastic analysis.

Class 4: The member resistance should be established using reduced elastic properties taking into account local buckling, and the frame designed by elastic analysis.

The limiting proportions of the elements in compression for these Class 1 and 2 sections may be modified in fire conditions according to the ratio:

EC3 cl 4.2.2

EC3 cl 4.2.2(4)


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Code Clause

where:

b/t = actual width/thickness ratio of element (b/tjeff = effective ratio at elevated temperatures

fY %,e = yield strength reduction factor for steel at temperature,

= yield strength of the steel

8 (see Table 5)

temperature, 8 (see Table 5). ‘E,O = elastic modulus reduction factor for steel at

In EC3: Part 1.2, this equation is presented in terms of a EC3c1 4.3.2(4) modification to the term E. The effective widthhhickness ratio is compared to the classification limits in EC3: Part l . 1. However, the ratio (ky,e/kE,e) is greater than unity, and therefore it is possible that the class of the member in bending changes from ultimate limit state to fire limit state analysis. The value of this ratio at 550°C is approximately 1.35, and therefore many isolated beams may move from Class 1 to 2 and 2 to 3, in fire conditions. This aspect is not satisfactory for the fire resistant design of members in bending, as it requires an iterative procedure to determine the section class, as a function of 8. The limiting proportions of compression flanges of beams are presented in Table 9, for a critical temperature of approximately 550 O C.

EC3 Part 1 . 1 cl 5.3.1

However, there are important cases where this approach is not required. For example, when a beam is connected to an insulating element, such as a concrete slab, the section classification in fire may be treated as the same as for normal design. Therefore, no re-classification is required. Furthermore, no change in classification is required for Class 3 or 4 sections, because elastic properties are used in determining their bending resistance.

It is hoped that the effect of the change in of section classification BS 5950 will be rationalised during the ENV period of EC3: Part 1.2. The alternative approach, as adopted in BS 5950: Part 8, is to use a lower EC3 Table 3 .1 strain limit of 0.5 % for steel elements subject to local buckling. The strength reduction factor in EC3 in this case is k, 8 (see Table 5). It is apparent that k, 8 = kE,+ and so no re-classification is required. However, the use of k, 8 rather than ky,e means that the bending resistance of the member is reduced by about 15 % in the important temperature range of 500 to 650°C.

cl 2.3

> ,

An alternative approach is to treat all beams (other than those connected to slabs) in the same manner as columns by introducing an appropriate adaptation factor (see Section 4.4).


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Code Clause

Table 9 Limiting proportions of compression flanges of members EC3 cl 4.2.3.3(1) in bending at 20" C, and approximately 550" C, when using

the strength reduction factor k,,

4.3.2

Compression flange of members in bending

I sections:

S275 steel S355 steel

SHS sections:

S275 steel S355 steel

Class of steel section

Ultimate limit state I Fire limit state I ~~

1

9.2 8.1

30.5 26.8

Isolated class l or 2 members

Individual beams not connected to floors experience approximately uniform temperature throughout their cross-section. Members of I or C cross-section that do not suffer local or lateral torsional buckling may be expected to reach strains well in excess of 2 % in their flanges at the large deflections that occur in fire conditions. Strains vary considerably as a result of the variation of bending moment along the member. Hence it is not easy to calculate deflections in fire conditions by hand, nor is it required by codes. In practice, maximum flange strains of 2 % would lead to beam deflections of the order of spad40 for sensible span: depth ratios, which is consistent with the failure criterion of standard fire tests (see Section 2. l).

The method of analysis at the fire limit state is determined by the section classification, which is determined according to the limiting proportions of the compression flanges, as given in Table 9. For Class 1 or 2 sections, it is assumed that the plastic resistance of the cross-section can be attained and that deflections are indeterminate.

The reduction in plastic moment resistance of a uniformly heated EC3 Table 3.1 beam would follow the basic strength reduction of the material, as given by Equation (8). Therefore, the critical temperatures of isolated restrained beams would be the same as for members in tension, corresponding to the strength retention factor ky,e in Table 5, multiplied by 1.05 (the ratio of partial factors), as in Equation (9). Members failing by lateral torsional buckling would require more detailed analysis (see Section 4.3.4).

There may be situations in which control of deflection of beams is important in fire conditions. These beams may support walls at


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Code Clause

compartment boundaries, or shaft walls. In such cases, the use of a lower strain limit corresponding to the strength reduction factor, kx,o is appropriate (see Section 4.3.5).

4.3.3 Beams supporting concrete slabs

A steel beam supporting a concrete or composite floor will be subject to greater heating to its bottom flange than its top flange which is influenced by the 'heat sink' effect of the concrete floor. This results in a significant temperature variation across the depth of the section and the top flange and upper part of the web will be cooler than the rest of the section. Nevertheless, the important element for calculation of the critical temperature of the beam is its bottom flange.

EC3 cl 4.2.3.3(3)

The colder upper parts of the section resist greater force than the lower parts, which means that the plastic neutral axis of the section will rise with increasing temperature to a level where tension and compression forces due to bending are balanced. Typically, the temperature variation is of the form of Figure 5 , such that the top flange temperature may be only two-thirds of the bottom flange temperature, corresponding to a temperature difference between the flanges of approximately 150 to 200°C. The web temperatures are less affected by the connection of the top flange to the slab and may be treated conservatively as having the same temperature as the bottom flange.

It is often found that the plastic neutral axis of a beam is close to the top flange when the critical temperature of the member is reached. This observation is relatively insensitive to the form or thickness of fire protection as a temperature difference of 150°C is sufficient for the compressive resistance of the top flange to exceed the combined tensile resistances of the bottom flange and web. The effect of the cooler top flange and lateral restraint by the concrete slab also avoids consideration of local or lateral buckling effects in a member in positive (sagging) bending. Therefore, no reclassification of the member is required.

The bottom flange and the majority of the web, therefore, act in tension to resist the bending moment applied to the section, which leads to a proportionate increase in the contribution of the web. Consider the case of a simply supported beam with a uniform temperature distribution in the web and bottom flange:


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Code Clause

, ^ , "

I . - , h . 3 " _

. . . " . " . " . ' ~ T-150 (3T-150

I I I Compression

I I T

Plastic neutral

7-- y axis

Tension

1 (a) Section through (b) Temperature ( c ) Stress

beam and slab variation variation

Figure 5 Temperature and stress variation in a beam connected to a concrete slab

Taking moments about the centre of the top flange means that the moment resistance of the heated section relative to its 'cold' plastic resistance varies according to the approximate formula:

where A , and A , are the cross-sectional areas of the flange and web EC3 respectively. cl 4.2.3.3(3)

Typically, the first geometric term in brackets is between 1.2 and 1.3 for most UB sections, which means that the member resistances and critical temperatures of beams supporting slabs are greater than given by the basic strength reduction of the material, ky,e .

In EC3 : Part 1.2, the performance of beams connected to concrete slabs is expressed in terms of an 'adaptation factor', K. This factor has the effect of reducing the magnitude of the loads, or alternatively increasing the effective resistance of the member, as was done above. The critical temperature of the member is given by modifying the load level or resistance in Equation (9), according to the expression:

or alternatively, A E f i d t - - kys' Rd K


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Code Clause

The critical temperature of the beam, OC, is determined by solving this equation by iteration, knowing the variation of steel strength, k with temperature 8 , as given in Table 5. As noted previously, ky , e = 1.05 ky,e in the UK NAD. p

The adaptation factor, K , is taken as 0.7 for I beams supporting EC3 concrete or composite floors. Currently, this is a ‘boxed’ value in EC3: Part 1.2 and is independent of the critical temperature, or member type, or form of protection. The K factor was determined empirically, based on fire test data. It is approximately the inverse of 1.3 which takes account of the shift in neutral axis position for non-uniformly heated members, as is given by Equation (1 1) for typical I sections. This K factor means that the load level is effectively reduced by 30% when determining the basic strength reduction of the steel and, hence, leads to an increase in the critical temperature, OCT, of the beam. The values of critical temperature of beams calculated using this approach are presented in Table 10. These values are appropriate for the ratio of partial factors of 1.05.

cl 4.2.3.3(8)

The limiting temperatures of BS 5950: Part 8 are determined by observations from fire resistance tests on unprotected beams. Implicit in these tests is that very high strains are experienced at failure so that actual material strengths would be higher than given by the 2% strain limit.

BS 5950 cl 2.3

The series of over 20 fire tests on unprotected beams was carried out by British Steel during the last 15 years as reported in the ‘Compendium of UK Standard Fire Test Data(22)’. The tests were carried out on beams of different sizes and with different load ratios. A lower bound through the test data leads to the variation of limiting temperature with respect to load ratio, as given in BS 5950: Part 8, and as shown in Figure 6. The EC3: Part 1.2 approach is compared in Table 9 which shows great consistency when K = 0.7. It is EC3 concluded that the differences between the Eurocode and BS 5950: Part 8 are relatively small and can be ignored.

cl 4.2.3.3(8)


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c 0 R

Code Clause

Test data (British Steel)

EC3: Part 1.2 and BS 5950: Part 8

5000 I I I 1 I I I I I I

1.0 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0

Load level or load ratio

Figure 6 Test results and critical temperatures of beams connected to concrete slabs

As noted earlier, the critical temperature of the beam is time independent, and is used to determine the amount of fire protection that is required for a given fire resistance. The critical temperatures of beams connected to concrete slabs may be 50 to 70°C higher than those of isolated beams. At a load level of 0.6, the critical temperature of a beam connected to a concrete slab is 620"C, which is significantly higher than the temperature of 550°C normally used in assessing fire protection materials (see Section 5). Possible reductions in protection thickness at this higher critical temperature may be of the order of 10%.

4.3.4 Beams failing by lateral torsional buckling

Isolated beams may fail by lateral torsional buckling if the restraints to the compression flange of the beam are spaced widely apart. EC3 : Part 1.2 recognises the possible influence of the high strains on the temperature dependent slenderness of the beam. This phenomenon is described in more detail when interpreting the behaviour of columns in fire (see Section 4.4.1).

EC3 cl 4.2.3.3(5)

In order to avoid further consideration of the influence of lateral torsional buckling at the fire limit state, EC3: Part 1.2 requires the use of an adaptation factor, K , of 1.2 for isolated beams whose slenderness ratio x, exceeds 0.4. For restrained isolated beams, or


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Code Clause

for beams with x, less than 0.4, K is taken as unity. The effect of the adaptation factor of 1.2 is to increase the loads or, alternatively, to reduce the bending resistance. Therefore, the critical temperatures of beams failing by lateral torsional buckling are up to 50°C lower than isolated restrained beams. Critical temperatures for these cases are presented in Table 10.

A further complication in this approach is the possible re- classification of the section in fire (see Section 4.3.1), and the modification in the temperature dependent slenderness ratio in the same manner as for columns (see Section 4.4.1). However, the EC3 adaptation factor of 1.2 partly takes into account any potential decrease in bending resistance due to buckling effects. In addition, the end connections may be assumed to provide partial fixity to the beam in fire conditions, so that for practical purposes, X L T , ~ = x,, for beams in all cases. Although not currently recognised in EC3: Part 1.2, it is intended that this interpretation is adopted in the UK NAD.

cl 4.2.3.3(6)

4.3.5 Beams requiring control of deformations

There may be situations where control of deformations is required at EC3 Table 3.1 the fire limit state. A good example is for beams supporting compartment walls or shaft walls, where excessive deformations may permit passage of smoke or flame. EC3: Part 1.2 recognises this possibility by use of a strength retention factor, IC,,,, corresponding to approximately 0.5 % strain in the steel member. In this case, deflections are unlikely to exceed spardl00 at the fire limit state and the walls are not subject to excessive deformation. Values of k, B are presented in Table 5.

An additional benefit of the use of this parameter in analysis terms is that kX,@ kE,o and therefore the section classification and effective slenderness are unchanged at the critical temperature of the member. However, the critical temperatures of beams requiring control of deformation are 50 to 80°C lower than other beams in general design, and therefore more fire protection is required for the same fire resistance period.

4.3 .6 Statically indeterminate beams

Continuous beams are assumed to benefit from redistribution of moment along the beam, so that a plastic mechanism can develop in c l 4.2.3.3(9) fire conditions. It is also considered that the zones towards the supports are slightly cooler than the rest of the beam. EC3: Part 1.2 uses the adaptation factor method by further reducing the K factor for the support zones of continuous beams by a multiple of 0.85. The relevant K factors are therefore 0.6 for continuous beams supporting concrete slabs, and 0.85 for isolated continuous beams.

EC3


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Code Clause

The effect of this reduction in K factor is to increase the critical temperatures of continuous beams by about 30°C. This difference is sufficiently small to be neglected in the design of continuous beams. However, a greater benefit in terms of critical temperature is to take account of continuity in fire conditions which is m utilised in normal conditions. In this case the effective load level on the beam can be reduced considerably (by up to half). This approach is reviewed in SCI publication 086('*).

4.3.7 Class 3 or 4 sections in bending

The preceding Sections apply to beams whose plastic resistance can EC3 cl 4.2.3.4

be developed at both the ultimate and fire limit states. Class 3 and 4 sections are those sections for which only elastic properties may be used, as they are susceptible to the effects of local buckling. Cross-sections may be classified according to the proportions of either the flanges or the web.

For beams supporting concrete slabs, it could be argued that section classification is irrelevant in fire conditions as the shift in the neutral axis puts most of the section into tension, and the cooler top flange is not crucial to the stability of the member. Therefore, these members may reach the same critical temperatures as Class 1 or 2 sections.

For isolated beams, it is necessary to consider the effects of local EC3 buckling (and lateral torsional buckling). In principle, the boundary between Class 2 and 3 sections, and Class 3 and 4 sections changes in fire conditions as discussed in Section 4.3.1. EC3: Part 1.2 recognises this by not requiring a change in the classification of Class 3 or 4 sections. An alternative approach is to use lower strain limit which has the effect of reducing the critical temperatures of Class 3 and 4 sections, so that no further consideration of local buckling need be made at the fire limit state.

cl 4.2.2(4)

In EC3: Part 1.2, a single critical temperature of 350°C is specified for Class 4 sections in all cases. This critical temperature is unnecessarily conservative. It may be argued that the strength retention factor kx,e corresponding to 0.5 % strain is appropriate for Class 3 or 4 sections, as k, 8 = kE,e (see Table 5). It follows that the effects of local buckling, a i determined by Equation (lo), would not increase disproportionately in fire resistant design. The corresponding critical temperatures of Class 4 beams using this strain limit are presented in Section 4.4.5. It is proposed that this approach may be used within the context of EC3 : Part 1.2.


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4.4

4.4.1

Members in compression

Members in compression are of two basic forms: either columns in frames, or struts in trusses, which are treated differently, as follows:

Columns in braced frames

Columns in braced frames are subject to compression and relatively small end moments. Because of the uncertainties about the magnitude of these moments in fire conditions, it is reasonable to assume that moments can be ignored in ‘simple’ construction in order that the results of isolated column tests can be used without modification.

Additionally, account may be taken of the reduced effective length of columns where they are connected to colder members. This might be the case in a multi-storey building where a fire is confined to one compartment between floors. It follows that the columns above and below the compartment are much stiffer than the heat affected members. This behaviour is illustrated in Figure 7.

’I Compartment

kW3 kW4 Fire compartment

Code Clause

EC3 cl 4.2.3.2

EC3 Fig 4.1

EC3 Fig 4.1

‘Failed beam Effective length of’ Failed column column

Y ,/Heated floor

Initial conditions Fire limit state

Figure 7 Effective length of columns in multi-storey buildings in fire

In multi-storey buildings, EC3 and 4 Parts 1.2 permit the effective EC3 length factor to be taken as 0.5 for internal columns in braced frames at the fire limit state. In the UK NAD, it is proposed that this cl 4.3.6.1(9)

approach is modified by reducing the slenderness factor in fire to 0.7 of the equivalent value used for the columns in normal design. This parameter assumes that the lowest effective length factor used in normal design is 0.7, and therefore the ratio of effective length factors is 0.5/0.7 = 0.7. UK NAD to

cl 4.2.3.2(4) EC4

EC3: Part 1.2


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Code Clause

This approach is also appropriate for single storey height columns, and for columns in the upper storey of a building. It is assumed that partial fixity is achieved in fire conditions, either by connection to the foundations or to the adjacent members. Therefore, effective length factor in fire is consistently taken as 0.7 of the normal design value (of 1 .O) for these columns. EC3 Part 1.2 uses an effective length of EC3 Fig 4.1 0.7 for this case, which is consistent with this approach.

This reduction in effective length is crucial to developing simple design methods for columns in fire. It is desirable for critical temperature methods to be largely independent of slenderness, or alternatively, for any beneficial effects of slenderness to be ignored in simple design.

The fire test data on columns is reviewed in Section 4.4.6. The behaviour of columns in fire is complicated by the facts that:

columns fail at lower strains than beams instability effects become more critical at higher strains column imperfections are variable end restraints are important to column buckling end moments occur in sway frames.

These effects are partially recognised in BS 5950: Part 8 by using a lower strain limit of 0.5 % for steel in appraising the column tests. However, in BS 5950: Part 8 a further factor of 1.15 is used to enhance the member strength due to the indeterminate effect of end restraints etc. in columns of modest slenderness. This factor is ignored in slender columns (A > 70), which therefore fail at slightly lower limiting temperatures than stocky columns.

BS 5950 Table 5

In EC3: Part 1.2, the strength retention factor for steel for column design is taken as corresponding to 2% strain. This means that both local and lateral buckling effects might become more critical for compression members, because of the proportionate decrease in the elastic modulus relative to the strength of steel at higher strains. This reduction in stiffness relative to resistance is important to understanding the behaviour of columns in fire.

This effect is taken into account by modifying the slenderness ratio EC3 of the member, x, to determine the temperature dependent slenderness ratio, x,, according to the expression:

cl 4.2.3.2(2)


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Code Clause

Values of ky,e and kE 0 are presented in Table 5, as a function of 8. The ratio of these two parameters is always greater than unity.

At 8 = 500"C, the effective increase in slenderness ratio due to the term J(ky ,e /kE,e ) is 1.14, and at 8 = 6OO0C, the increase is 1.23. A typical increase in slenderness ratio is 1.2 over the important temperature range of 500 to 600°C.

However, as noted earlier, this increase in slenderness may be UK NAD to conservatively multiplied by a end fixity factor of 0.7 for columns in fire conditions relative to normal conditions. The multiple of 1.2 X 0.7 leads to a ratio of effective slenderness ratios of 0.85 for internal columns in braced frames in fire conditions relative to normal conditions.

EC3 Part 1.2

In fire tests on isolated columns, the influence of end fixity on the temperature dependent slenderness, AB, is ignored, and it was found that the test results were over-predicted using a strength reduction for steel corresponding to 2% strain limit. Whilst it might have been EC3 C* 4 . 2 3 2 more logical to use a lower strain limit for members in compression, EC3: Part 1.2 uses the concept of the adaptation factor to obtain better agreement between tests and the basic material properties.

The basic relationship between load level and material strength, as defined in Equation (9), may be modified to represent the behaviour of columns, as follows:

or alternatively,

where the adaptation factor, K = 1.2 for columns, and

x (x,) is the buckling resistance factor for a column with a temperature dependent slenderness ratio, X,.

x (x) is the buckling resistance factor for a column with a normal design slenderness ratio, x.


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Code Clause

It is apparent that the critical temperature of columns is dependent both on the load level and partly on the slenderness ratio, x. Accepting that x, = 0.85 x for the general design of columns, as noted above, it follows that the ratio of x factors will increase with increasing slenderness, x. Therefore, the lowest critical temperature e,, will occur for a stocky column, or for a column restrained against buckling in a braced frame (i.e. with x < 0.2).

The critical temperatures of columns are presented in Table 1 l , as a function of the load level applied to the column in fire conditions and the slenderness ratio under normal conditions. In this table, x is taken as not more than 1.5, as the strains experienced at failure of very slender columns are less predictable than in stocky columns. Even so, the increase in critical temperatures over the slenderness range is only 20 to 40°C. Therefore, it is reasonably accurate to adopt the critical temperatures for short or stocky columns for use in general design.

The critical temperatures of columns in braced frames according to EC3 : Part 1.2 approach and to BS 5950: Part 8 are summarised in Table 10. In this Table, x is assumed to correspond to a stocky column (x < 0.2). The differences between the two codes are relatively small ( C 20'C) for load levels up to 0.6.

4.4.2 Columns subject to combined moment and axial force

Members subject to bending and compression may be treated by determining the load level from the combination of the proportions of axial load and moment considered independently. This approach is explained clearly in BS 5950: Part 8, but is not treated explicitly in EC3 Part 1.2. The following approach is proposed here:

The effective load level for member subject to combined actions may be determined from the expression:

where:

BS 5950 Table 5

EC3 cl 4.2.3.5

BS 5950 cl 4.4.2.3

Ffi = axial force applied in fire conditions Mfi = maximum moment applied in fire conditions FRd = axial buckling resistance moment at the ultimate limit

MRd = buckling resistance moment at the ultimate limit state. state


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Code Clause

If axial force and moment each vary proportionately with applied load, the load level may be determined from the strength utilization of the member at the ultimate limit state times the proportionate loading at the fire limit state, qfi. This load level may then be used to determine the critical temperature of the member, in the same manner as for a member in compression.

In both cases of resistance to compression and moment, account should be taken of the slenderness of the member. As noted earlier for columns in braced frames, the effective slenderness in fire may be taken as the same as for normal conditions. For internal columns, in ‘simple’ construction, end moments may be ignored at the fire limit state.

It may be assumed that compression is the dominant action, as is the case for braced frames and trusses, and that the adaptation factor K is 1.2. Therefore, the critical temperatures of members subject to combined actions, may be assumed to be equivalent to the values for stocky columns, as in Table 11.

4.4.3 Columns in sway frames

Eurocode 3: Part 1.2 does not treat sway frames explicitly, as all columns are assumed to be subject to small, or negligible end moments. However, in principle, the same approach may be used as for columns in braced frames by considering the combination of axial force and moment due to horizontal and vertical loads on the frame, as may develop in fire conditions. In this case, the additional end fixity of columns in sway frames in fire is ignored. Therefore, the effective slenderness ratio should be increased to take account of instability effects at elevated temperatures, as given by Equation (14). The effective slenderness ratio for columns in sway frames in fire conditions may conservatively be taken as x, = 1.25 x, as derived from Equation (1 3).

The further use of an adaptation factor, K, of 1.2 for columns subject to combined moment and compression, leads to the critical temperatures as in Table 12. It is apparent that the critical temperatures for columns in sway frames are 30 to 100°C lower than those for columns in braced frames. The critical temperatures reduce with increasing column slenderness, unlike the case for braced frames. Reducing the K factor to 1 .O in cases such as portal frames where end moments are dominant, will increase the critical temperatures by about 40°C.


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Code Clause

Table 10 Critical temperatures of members to Eurocode 3 Part 1.2 and BS 5950 Part 8

Type of Member

Members in compression

Members in bending - supporting concrete floors

- not supporting concrete floors

- failure by buckling

Members in tension

Figures in brackets refer to

Load Level at Fire Lin

0.7 I 0.6 I 0.5 I 0.4

485 I 1;: 1 1 13): (510) (540) (580) (615)

600 (590) (620) (650) (680)

535 625 595 565 (520) (555)

525 485

(620) (585)

635 595 565 535

605 565

(460) (510) (545) (590) S 5950: Part 8 values as a functic

t State I I

0.3

(655) 700 650

0.2

(780) (725) 780 725

(710)

675 735 (660) (745)

650

735 675

700

(635) (690) L of load ratio

Table 11 Critical temperatures for columns in braced frames

Table 12 Critical temperatures for columns in sway frames and for struts

Slenderness Ratio, X Load Level at Fire Limit State of Member

0.6

605 550 500 430 2 1.5

615 565 520 460 1 .o 630 575 540 500 0.7

645 595 560 520 0.4

0.3 0.4 0.5

BS 5950: Part 8 also treats columns in sway frames more conservatively by adopting a single limiting temperature of 520°C for all column slenderness. The corresponding critical temperature of a column with a slenderness ratio of 1.0 and subject to a load level of


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Code Clause

0.5 would also be 520 "C according to EC3 : Part 1.2. It is proposed that a critical temperature of 500°C may be adopted conservatively for columns in sway frames to EC3: Part 1.2.

4.4.4 Struts in trusses

Struts and other compression members are relatively slender and would not be expected to benefit from end fixity in fire conditions where they are connected to other exposed members. In principle, the same approach may be adopted as for columns in sway frames, in which case the critical temperature is a function of the effective slenderness ratio of the member. Alternatively, a lower strain limit may be used, in which case the dependence on slenderness ratio is smaller. The 0.5% strain limit on steel strength as given by the parameter kx,e corresponds broadly to point where both the reduction yield strength and elastic modulus are equal. The critical temperatures corresponding to this strain limit are a safe lower bound to the strut behaviour and give values close to those of sway frames. Therefore, the critical temperatures of struts are also represented by the values in Table 12.

4.4.5 Class 4 members in compression

The design rules described previously for beams and columns apply to members whose elements are not subject to the effects of local buckling (i.e. Class 1 or 2 sections). For uniformly heated members, the strength of the compression elements will be the dominant factor. In order to avoid the disproportionate effect of local buckling on lateral buckling of columns, it is necessary to use a lower strain limit when determining the strength reduction of these elements. No guidance is given in EC3 Part 1.2.

For Class 4 members in compression, it is proposed to adopt a strain limit of 0.1 %, corresponding to the strength reduction parameter, kp,e in Table 5. This leads to the critical temperatures given in Table 13. For Class 4 members in bending, a strain limit of 0.5 % is appropriate (see Section 4.3.7).

EC3: Part 1.2 proposes also a single critical temperature of 350°C for Class 4 sections in the absence of further calculations. Clearly, this value is very conservative and could be improved. The SCI publication Building design using cold formed steel sections: Fire protection(16) adopts critical temperatures of 450°C for columns and 500°C for beams using these Class 4 sections.

EC3 cl 4.2.4

EC3 cl 4.2.4


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Code Clause

Table 13 Critical temperatures of Class 4 sections

Member Load Level at Fire Limit State

0.6

590 545 500 410 Columns

605 565 530 500 Beams

0.3 0.4 0.5

based on steel strain of 0.5% for beams and 0.1% for columns.

4.4.6 Fire tests on columns

The test data on bare steel columns in fire comes from British Steel (3 tests)(22) and from ECSC (18 tests), as reported by Janss and

, and as summarized recently by F r a n ~ s e n ( ~ ~ ) (133 tests).

In all test series, the columns were subject to axial load with pinned, or clamped ends. Column sizes ranged from HEA100 to HEA300 (continental sizes). The important parameters are the failure temperatures and the load level applied to the columns, rather than the failure times. Although these tests were performed on unprotected sections, the failure (or critical) temperatures are equally applicable to protected sections.

The member resistance was predicted using the steel strength ky ,e , at the relevant failure temperature and the temperature-dependent slenderness ratio (calculated as given in Equation (14)). Janss and

carried out a statistical analysis of their tests. They found that the required adaptation factor, K, was on average 1.23, in order to obtain equivalence between the compression resistance with the applied load at the failure temperature. The standard deviation of the data was 0.17.

It is argued by Janss and Minne that is appropriate to use the mean K factor of 1.2 because of the lower level of safety that is adopted in fire design, and because of the greater influence of end fixity on the stability of the columns in real buildings, and also the ability for some redistribution of load in steel frames. These data have been extended recently by F r a n ~ s e n ( ~ ~ ) who summarised 133 tests on isolated columns, which also included the Janss tests.

The comparison between these tests on columns, and the critical temperatures obtained according to EC3 : Part 1.2, is shown in Figure 8. Clearly, the family of test results on columns is inherently more variable than that on beams due largely to the influence of the initial out of straightness of columns on buckling.


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Code Clause

0' 1 I I 1 ; I

0 0.2 0.4 0.6 0.8 1 Load level

Based on temperature dependent slenderness

Figure 8 Test results and critical temperatures for columns

As noted earlier, it is more logical to use a lower strain limit when assessing the steel strength in columns. A strain limit of 0.5 % leads to broadly the same strength reduction as the variation in elastic modulus. Using this strain limit to re-analyse the same column tests leads to an adaptation factor, K , of 0.92 with a standard deviation of 0.1. Therefore, K may be conservatively taken as 1 .O if the lower strain limit of 0.5 % is used (corresponding to the parameter kx,e in Table 5 ) .

Nevertheless, examination of the data in Figure 8 shows that relatively few tests are significantly outside the EC3: Part 1.2 rules using the strength parameter ky,e divided by a K factor of 1.2. It is concluded that EC3 Part 1.2 rules based on isolated (pin-ended) columns are appropriate for structural fire design of columns, taking account of the beneficial restraint effects that occur in real buildings.

4.5 Critical temperatures for simple design

The critical temperatures of members are used for evaluation of the EC3 cl 4.2.4 thickness of fire protection. Members with lower critical (or limiting) temperatures require more protection than those with higher critical temperatures. Therefore, reducing the load on a member also leads to an increase in its critical temperature.

For simple design, it is sensible to adopt a load level corresponding to the use of most structural members in practice. For beams and columns in general application, the load level may be taken as 0.55 at the fire limit state, assuming that at least half of the total load is


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Code Clause

‘variable’ or short term in duration. For applications such as warehouses, where more of the applied load is likely to be ‘permanent’, a load level of 0.65 is appropriate.

The critical temperatures of certain classes of structural members reviewed in this Section , and of two other forms of construction reviewed in later sections, are presented in Table 14 which correspond to a load level of 0.55. These critical temperatures have been ‘rounded’ by 20°C or so to simplify the classification procedure. It is apparent that members with protection Class A may employ an assessment temperature significantly above the traditional value of 550”C(25), whereas Classes C and D lead to assessment temperatures 50 to 100°C below 550°C. The corresponding ‘boxed’ values quoted in EC3 : Part 1.2 are also given in Table 14.

A further reduction of 50°C is proposed for members subject to permanent loads, but these cases are relatively rare in normal building applications.

It should be noted that the critical temperatures in Table 14 are proposals at present, but are considered to be appropriate for evaluation of the required thickness of fire protection in ‘simple’ design.

Table 14 Critical temperatures for simple evaluation of protection thickness for certain classes of members

Protection Class

A

B

C

D

Type of Member

EC3 : Proposal

Critical Temperatures

Part 1.2

- Beams supporting concrete or 620 C composite slabs

- Restrained beams (Class 1 or 2 ) - Composite beams - Columns in braced frames

550°C

- Beams failing by buckling - Columns in unbraced frames 500°C - Struts or ties in trusses - Class 3 beams - general

550°C

510°C

- Class 4 sections - general - Composite deck slabs - Concrete filled tubular sections

350°C 450°C

Note: For applications where the applied load is permanent, reduce these critical temperatures by 50” C.


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Code Clause

5 THERMAL RESPONSE OF PROTECTED AND UNPROTECTED STEEL MEMBERS

EC3 cl 4.2.5

The fire resistance of a steel member is determined by the time at which its temperature reaches the critical temperature, as defined by its load level and application (see Section 4). The actual rise in temperature of the steel member with time is evaluated by the methods given in EC3: Part 1.2, which are reviewed in this Section. The same methods are also presented in EC4: Part 1.2.

Steel sections may be unprotected, in which case the temperature rise is relatively rapid, or they may be protected by materials of low thermal conductivity, so that the temperature rise in the steel is less rapid. Traditional forms of protection that are applied on site include board and sprayed materials, and intumescent coatings. The term ‘insulation’ is used in EC3 and EC4 Parts 1.2 to denote protection using these materials.

There are also many examples of partial protection of steel sections, so that part of the section is exposed to the fire and the remaining part is protected. Partial protection of a steel member by concrete is covered in Section 6 .

5.1 Definition of section factors

The controlling parameter that determines the temperature rise in a E C ~ 4.2.5.1 steel member is the ratio of the heated perimeter, Hp, to the EC4 Cl 4.3.3.2 cross-sectional area, A , of the member. This ratio of Hp/A in UK practice is normally presented in units of m-’. It is known as the ‘section factor’. The corresponding term in EC3 and 4 Parts 1.2 is A N , where A now represents the heated surface area per unit length, and V is the volume of the member per unit length. The magnitude of the terms Hp/A and A/V is the same.

Typical values of section factor are between 100 and 250m-’ for the size range of hot rolled I sections. Sections with low A/V factors respond more slowly to heat and, therefore, achieve higher periods of fire resistance than sections with high A/V factors. Some unprotected sections with very low A N factor heat up so slowly that they can achieve 15 to 30 minutes fire resistance.

The definition of the heated surface area of an unprotected member, or a member with protection that follows the shape of the section, is relatively straightforward.


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For a fully exposed I section, the heated surface area per unit length is :

A = (4b + 2d - 2t)

where b and d are the overall breadth and depth of the section and t is the web thickness.

Where a beam supports a floor, it is assumed that the floor material is of such a low conductivity that heat does not pass through the floor into the upper surface of the flange. The heated surface area, A , reduces to (3b + 2d - 2t). A suitable insulating material offering partial protection would be concrete (for a floor or a wall), or brickwork (for a wall). In the case of timber floors, the heated surface area corresponding to a fully exposed member should be used conservatively.

Formulae for heated surface areas of various protected members are 1x3 Table 4.3 given in Table 4.2 of EC3: Part 1.2 (and in Table 3 of BS 5950: Part 8). The basic formulae for sections with 3 and 4 sided exposure are given in Figure 9. Published data also takes into account the root fillets between the web and flange in rolled sections.

Values of Hp/A or A N for standard UB and UC sections and hollow sections are presented in Tables B1 to B6 in Appendix B (taken from reference (25)).

There are three main types of fire protection to steel members that should be considered (see Figure 9):

Projiile Protection is where the fire protection follows the surface of EC3 the member. Therefore, the section factor is defined by the proportions of the exposed steel member and relates to the use of sprayed and intumescent coatings.

Tables 4.2 & 4.3

Box Protection is where there is an outer casing of boards around the member. The heated perimeter is defined as the sum of the inside dimensions of the smallest possible rectangle around the section, neglecting air gaps, etc. The cross-sectional area is that of the steel section. The thermal conductivity of the protection material is assumed to be much lower than that of steel and, therefore, the temperature conditions within the area bounded by the box protection are assumed to be uniform.

Solid Protection is where the member is encased (typically by concrete). This is a more complex case because of the non-uniform thermal profile through the encasement. If only part of the member is exposed (for example the lower flange), then the heated surface


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Code Clause

area may be taken around the portion of the section that is exposed. This assumes that the passage of heat through the concrete to the steel is much smaller than through the exposed steel surface.

There are various other practical cases that should be considered. Sections with openings, such as castellated beams, may be considered as having the same section factor as the parent steel section. EC3 cl 3.3.2 Rectangular and circular hollow sections are exposed to heat from only one side and, therefore, have a relatively low section factor for the same use of steel as an I section. Consequently, hollow sections are more fire resistant, which can also be enhanced by concrete or water-filling (refer to Section 9.3).

EC4 cl 3.3.4

Profile protection Box protection

L

2h+3b+2t A

3-sided protection

-- A - 2h+4b+2t v A

4-sided protection

-- ~

A - 2h+2b v A

, e

e

A - 2h+2b v A -- ___

Box with air gaps

Figure 9 Section factor definitions


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Code Clause

5.2 Traditional and modern fire protection materials

The most important parameter defining the performance of an appropriate fire protection material is its thermal conductivity. There are a number of readily available materials of relatively low thermal conductivity, and hence high insulating value. For most good insulating materials, it is the presence of voids distributed within the structure of the material that largely contributes to its insulating performance.

Examples of traditional fire protection materials are concrete (both normal and lightweight), brick- and blockwork and plasterboard. These materials are not necessarily the best insulants but combine robustness and practicality, and have stood the test of good performance in service. This is important because the protection must not break away from the steel, nor allow large gaps to form, thereby increasing the heating rate of the steel section. Materials must possess good ‘stickability’ in fire (see Section 5.6.2).

In the UK, ‘deemed to satisfy’ requirements for traditional protection materials, such as concrete and bricks, are given in the BRE publication by Morris, Read and Cooke( 26), which is referred to in BS 5950: Part 8. Some of these rules have found their way into EC4 : Part 1 .2 , and are discussed in Section 6.3.3. Both concrete and blocks rely on their ‘heat sink’ effect, due to their thermal capacity rather than their insulating properties.

Examples of modern fire protection materials are those based on mineral fibre, vermiculite, calcium silicate and lightweight cementitious materials, in spray or board form. Expanded vermiculite, in particular, is one of the best insulants, but can be soft and easily broken away from the steelwork. Mixtures of cementitious materials and vermiculite are more robust. Other examples of modern fire protection boards include fibre-silicate, and mineral fibre mats.

Intumescent coatings are a separate category of materials which although originally applied in thin layers (as little as 1 to 2 mm thick), expand on heating or ‘intumesce’ in a fire. The foam-like expansion products offer the appropriate insulation to the steel. Some spray applied intumescent coatings are thicker and can provide over 90 minutes fire resistance. Thin film intumescent coatings are generally used for architectural reasons where the steelwork is left fully exposed. Good examples of the use of intumescent coatings are in the many cast and wrought iron structures which have been renovated in recent years.

BS 5950 cl 4 .3 .3


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Code Clause

Gypsum and plasterboard are also considered to be generic and traditional protection materials, despite the fact that they are products manufactured by a small number of suppliers. As materials, plaster and gypsum, and gypsum plasterboard are much better insulants than normal weight concrete or brickwork. Their moisture contents can be as high as 20%, which suggests that the heat (and hence time) required to drive-off excessive moisture can be significant and can enhance the fire resistance offered.

Manufacturers British Gypsum and Lafarge provide comprehensive information on the use of their products and for selection of appropriate fixing details. Two layers of board are often required when encasing steel members. The boards are held in place by through fixings onto light steel ‘furring’ pieces, or in some systems by direct fixing between boards.

Sprayed fire protection is currently popular in commercial steel buildings where the floor soffit is hidden, and where additional casings are provided around the steel columns. Box or board systems are more popular where the protection to the beams and columns is left exposed.

Sprayed systems are usually applied in a number of layers. A priming coat applied to the steel section may be recommended by the manufacturer. The main advantage of sprayed systems is that they can easily protect complicated beam-column junctions, trusses and secondary elements. Their main disadvantage is the mess and dust created during spraying.

Board systems often use additional noggins and filler pieces between the flanges of the beam to which the boards are attached. Their method of jointing is important in order to prevent gaps opening up. Pre-formed box systems are also used.

There are many proprietary materials of the above types that are marketed, each with different properties and methods of installation. Information on these products and their suppliers is given in the ASFPCMKCI publication(25).


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Code Clause

5.3 Theoretical behaviour of unprotected steel sections in fire

The incremental rise in temperature of a uniformly heated bare steel section in time interval 6t due to one direction of heat as:

where:

C, and p a are the specific heat and density of steel Values are given in Section 3.1.4 (see also Figure 4).

flow is given

respectively.

temperature ("C) of the furnace (or temperature in a natural fire) at a particular time t (secs)

temperature of the steel section ("C), which is assumed to be uniform, at time t

section factor (m-l) of the exposed steel member per unit length

coefficient of heat transfer by convection. (This is normally taken as a constant of 25 W/m2 "C although a precise figure is given in EC 1 Part 2.2)

coefficient of heat transfer by radiation, where

The parameter E,,, is the resultant emissivity and represents the radiation transmitted between the fire and the metal surface. Its magnitude depends on the degree of direct exposure of the element to the fire. Elements which are partially shielded from the radiant effects of the heat of the fire would have a lower value of E,,,. It is also dependent on furnace characteristics that are appropriate to each test laboratory, and therefore values differ slightly among the countries who have carried out major testing of steel structures.

The approach of EC3: Part 1.2 is different in presentation, but follows the same basic method except that the heat flux acting on the surface of the steel is defined by the parameter hnet,d in EC1: Part 2.2, so that Equation (16) becomes:


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Effectively, hnet,d includes two parameters due to convection and radiation, and the radiation term is given by Equation (17). The resultant emissivity is defined as a multiple of two parameters EC3 Cl 4.2.5.1 Ef (= 0.8) and ern(= 0.625), giving E 0.5. In the proposed UK EC1: Part 2.2 NAD for EC3, E, is given as a 'boxed' value in order to obtain good correlation with existing fire test data in UK furnaces. The proposed value of E,, (i.e. e , Ef) is 0.625 in the UK NAD for EC3: Part 1.2.

EC4 cl 4.3.3.2

This differential equation for the temperature rise may be used to determine the steel temperatures by incremental integration, if the variation of the fire (or furnace) temperature with time is known. The method may be used for any type of fire, provided the appropriate value of E,,, is used. Recommended time steps for analysis of unprotected sections are not more than 5 seconds.

An approximate formula for the time to (now expressed in minutes), at which a limiting steel temperature of 0, is achieved in an unprotected section is given in the former ECCS Re~ommendations(~) as :

te 1 0.54 (0, - 50) (Am/V)-o.6

The linear variation of 0, with time is valid in the range of steel temperatures from 400" to 600°C.

5.4 Temperatures in unprotected columns and beams

The thermal response of unprotected I section columns and beams has BS 5950 been measured in test furnaces and have been correlated with Table 4 analytical methods. For reference purposes, these temperatures may be presented in terms of the section factor or, alternatively, flange thickness tf. For some heavy sections with low section factors, 30 minute fire resistance can be achieved (see Table 15).

No guidance is given on the use of unprotected steel sections in EC3 Part 1.2, and the designer may refer to BS 5950: Part 8.

In BS 5950: Part 8, design temperatures are defined as those that the critical element (lower flange of beams, for example) of an unprotected steel section would reach at the appropriate period in a standard fire resistance test. Design temperatures may be presented as a function of flange thickness as for beams, in particular, this gives a slightly better measure of performance than the 'average'


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section factor of the member. This data is presented in BS 5950: Part 8 for unprotected columns and for unprotected beams supporting concrete floors.

Although, equivalent data for design temperatures, nor minimum BS 5950 section factors of unprotected members, are given in EC3: Part 1.2, but the values of BS 5950: Part 8 may be used to comply with the principles of the Eurocode.

Tables 6 & 7

Table 15 Minimum section factors to achieve 30 minutes fire resistance for unprotected steel members according to BS 5950: Part 8

I I Minimum section factor (m-') I Beams supporting concrete I slabs

Fully exposed columns I I Partially exposed columns I 70 (based on exposed portion) I

5.5 Theoretical behaviour of protected steel sections in fire

The one-dimensional passage of heat through a thin fire protective material into a steel section can be expressed by the following equation:

These parameters are defined in Section 5.2, except as follows:

h, is the thermal conductivity of the protection (or insulation) material (W/m"C) (see definitions in Section 1.4). (The corresponding term in BS 5950: Part 8 is ki) .

d, is the thickness of the protection material (m).

A,/V is the section factor of the protected section (m-'), which may be either in the form of profile or box protection.

Time dependent properties for h, and can be introduced in an incremental integration for e,, knowing the variation of the fire or furnace temperature, Of, with a time, t . Recommended time steps for


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Code Clause

analysis of protected sections are not more than 30 seconds. The temperature rise of the steel section is of the same basic form, but at a slower rate compared to an unprotected section.

However, Equation (20) ignores certain beneficial factors. Firstly, there is a certain component of heat transfer by convection and radiation between the fire and the outer surface of the protection. Secondly, thicker heavier insulation materials have some thermal capacity (they store heat) thereby reducing the heat flux to the steel section. Thirdly, most protective materials have some natural moisture content and a certain amount of heat is required to vapourise this moisture. This causes a 'dwell' in the rise of temperature of the protection materials at approximately l00 "C.

The generalised differential equation for the temperature rise in a EC3 C* 4.2.5.2 protected steel section, taking these factors into account, is given in EC3 Part 1.2 and is:

EC4 cl 4.3.3.2(5)

where :

Cp = specific heat and pp = density of the fire protection material.

This one dimensional heat flow behaviour is illustrated diagrammatically in Figure 10.

The term 4 is determined from an assumed variation of temperature in the fire protective material. It defines the relative amount of heat stored in the protective material. The final term in 68, in this equation is relatively small, but should be included in an 'exact' analysis. Moisture content effects are considered in Section 5.6.4.

Equation (21) ignores the surface radiation and convection effects. This term is important in unprotected sections, but it is small in comparison to the term representing the insulating capacity of the fire protection material, and hence is neglected in EC3 and EC4: Parts 1.2.


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Temperature profile

Protection l.___

material

Figure 10 One dimensional heat flow through a thick fire protection material

The standard time-temperature curve may be input into Equation (21) knowing all the temperature-dependent material properties. An example of the temperature rise in a protected steel member is shown in Figure 11. The 'dwell' in the temperature rise due to moisture is apparent. The temperature rise follows the same basic form as the standard fire but at a slower rate. The temperature corresponding to a natural fire may also be input (see Section 1 l), although this should be used with caution where material properties are strongly temperature-dependent.

Critical temperature

l l l , in a protected section

Fire resistance 'Dwell'

Time

Figure 11 Typical heating curve for a protected steel section


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5.6

5.6.1

Code Clause

The use of Equation (21) in practical applications is considered in Section 5.7. The corresponding method of BS 5950: Part 8 is described in the following Section.

Traditional method of appraising the performance of fire protection materials

Regression formula used in the UK

Traditionally, the performance of fire protective materials was evaluated using the method in the ASFPCM/SCI publication (‘Yellow Book’)(25) coupled with an independent testing procedure. Two forms of test are required: insulation tests and ‘stickability’ tests. The former is concerned with determining the thickness of fire protection needed to keep the average steel temperature at or below 550°C at a given fire resistance period. The latter is needed to ensure that the fire protection remains intact over the fire resistance period.

As noted in Section 5.1, the important parameter in determining the thermal performance of the steel section is the section factor, H p / A (or A,IV in EC4: Part 1.2). In interpreting the thermal performance, the reference temperature of a beam section in a fire test is taken as the average of the lower flange and the mid-point of the web.

A number of small-scale unloaded fire tests are carried out on different sections and with a range of thicknesses of protection. From these tests, a ‘regression formula’ can be developed of the form:

- - a. -+ al dp A/Hp + a2 dp (23)

where:

dp = thickness of protection (mm) t550 = time for the steel temperature to reach 550°C ao, al and a are regression coefficients for the protection material.

ao, al and are determined for the material under test by multiple linear regression over the range of the test data. This equation can then be used to determine the thickness of protection required for other sections and fire resistance periods. Physically, this formula has little meaning other than to broadly represent the influence of the two main parameters of section factor and protection thickness. For a given section size, the protection thickness varies linearly with fire resistance period.


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From Equation (23), the required thickness of fire protection can be determined as a function of Hp/A (or %/V) and the fire resistance period required. Examples of data in the ‘Yellow Book’(25) for typical sprayed and box protection are presented in Tables 16 and 17. Any thickness of sprayed protection can be specified, but box protection is only available in certain thicknesses (typically 5m m increments).

In principle, it is also possible to carry out this regression analysis at other limiting temperatures (such as 650°C for beams). This analysis will lead to different regression coefficients (usually larger) in comparison to those calculated at 550°C.

More terms may be introduced into the regression analysis to improve its accuracy, but conversely, more tests are required to carry out the statistical analysis. The proposed CEN 127 requirements use a 7-term regression analysis, as presented in Section 5.7.5.

Table 16 Minimum thickness of a typical spray protection to an l section (taken from reference 1.25))

250

51 33 23 14 10 310

50 32 23 14 10 290

49 31 23 14 10 270

48 31 22 14 10

Linear interpolation is permissible between values of H p / A


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Table 17 Minimum thickness of a typical box protection l section (taken from reference (24))

A or B

l/’?

A or B

I

E

- 210 -

D

- 113 - C

- 90 - B

- 71 -

A or B

I ”2

I

H 110 - K 91 - J

69 75 -

F I

62 - D

- 58

A or B I I

2 4 3

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to an

Flre rc\iatance perlod (hours)

5.6.2 Stickability of fire protection materials

The ‘stickability’ of the fire protection material is assessed using loaded beams and columns in a large test furnace. The recommended size of beam in UK practice is 305 X 127 X 42 kg/m UB spanning 4.5 m (4.0 m heated length). However, according to the proposed CEN Committee 127 requirements, this section size will increase to a 400 mm deep beam. Two beams are tested with the minimum and maximum thicknesses of protection that are likely to be used. Typically, tests would be terminated at a deflection exceeding span/40. At this deflection, strains of at least 3% are developed in the steel flanges.

Legend (mm) A - 16mm B -188 c -20 D -25 E -30 F -3s G --40 H -45 I -50 J -55 K -60 L -65 M -70 N -15 0 -80

The purpose of both the beam and column tests is to identify if the protection cracks, or if gaps open up, so permitting heat to reach the steel section prematurely. The ‘stickability’ limit therefore determines the permitted maximum fire resistance period, irrespective of the insulation criteria. The test is also important as it gives a measure of the strains that can be developed, and hence determines


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the strength retention factor that may be used in analysis of the steel section (see Section 3.1).

Stickability tests on columns are carried out using a recommended 203 x 203 x 52 kg/m UC of 3 m height. However, column tests are usually less critical from a stickability point of view because of the lower strains at failure in comparison to beams.

The testing regime proposed by CEN Committee 127 is significantly different in detail, but not in principle from that described above for UK practice.

5.6.3 Generalized approach in BS 5950: Part 8

The approach of BS 5950: Part 8 is presented in order to assist designers in making a comparison with the approach of EC3: Part 1.2.

The passage of heat through a fire protection material into a steel section is controlled by the differential Equation (21) in Section 5.4. Considering this general equation for 'heavy' protection materials and ignoring the minor terms, the temperature rise in the steel section is:

where:

A,/V = section factor of the member (= Hp/A in BS 5950:

hp = thermal conductivity of the fire protective material

dp = thickness of material (= di) 8, and 8, are the steel and fire temperatures respectively at

Part 8)

(= ki)

time t . (e, = 8, in BS 5950: Part 8).

Most 'heavy' insulants have a specific heat of 1100 to 1200 J/kg"C. The specific heat of steel is taken as 600 J/kg"C, which is approximately half of the above value. Therefore the value of 5, defined by +/3 in Equation (22), may be approximated to:

5 = -*- dP PP PS V

BS 5950 Appendix D

where pp and ps are densities of the insulation and steel respectively.


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BS 5950: Part 8 then draws on the ECCS recommendation^(^) which the steel temperature in a standard fire (derived from Equation (24)) is expressed in terms of an empirical formula. This is valid in the range of 8, between 400 and 600°C. The time at which a steel temperature of 8, is achieved is given by:

where:

te - - in minutes

)LP - A p = in m-l

e, = in "C.

- in Wlm "C

- V

Rearranging the above equation in terms of the required thickness of protection, dp, when the steel temperature, 0,' reaches its critical temperature, Ocr, gives:

where :

If = an insulation factor in terms of the fire resistance period and the critical temperature of the steel.

In BS 5950: Part 8 Appendix D, Equation (27) may be extended to cover 'heavier' forms of fire protection by introducing a reduction factor, F,. The protection thickness is evaluated from the following equations (using the terminology as used above):


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Practically, F,, has a value of between 0.6 and 1.0. Numerical data for If and F, are presented in Tables 18 and 19.

The limitation to this method is the lack, at present, of representative thermal properties of specific rather than generic protection materials. The BS 5950: Part 8 method is much simpler than the EC3 Part 1.2 method, but is unlikely to be used in the long term after the fire protection manufacturers have re-assessed their products to the CEN requirements.

Table 18 Insulation factor (lf} in BS 5950 Part 8

e resistance period of

2 hour

3000

4 hour 3 hour

2200 1530 900

2450 1690 1000

2750 1890 1120

3100 2150 1260

3550 2450 1440

4100 2850 1670

4900 3350 1980

5900 4100 2400

7400 5 100

Table 19 Values of F, as a function of effective density, p

Fire protection material weight factor 1 I I I 1 I I

CL

0.50 0.55 0.62 0.73 0.92 0.95 1 .o F,

2.0 1.5 1 .o 0.5 0.1 0.05 0 .o

5.6.4 Effect of moisture content

Most fire protection materials contain moisture. When temperatures within the material approach ZOO'C, further heat input does not increase the temperature of the material but vaporizes any free moisture (latent heat of vaporization). This causes a delay or 'dwell'


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5.7 5.7.1

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in the temperature-time response of the protected steel section (Figure 11). This is a well-known phenomenon, and can be significant in materials where the stored moisture exceeds 3 % by weight.

The ‘dwell time’ t, can be determined by the approximate formula:

t V - - c PP d p 2 minutes (3 1)

where c is the percentage moisture content in the material.

The other units are as defined in Section 5.6.3.

Hence the total time taken to reach a limiting steel temperature e,, is (re + t,) (minutes). However, if it is required to determine $ in terms of te, then it is necessary to modify Equation (28). This is done by increasing the effective density of the insulation, according to the approximate relationship given in BS 5950: Part 8:

pp’ = pp (l + 0.03 c) (32)

This formula has been shown to be accurate for protection materials achieving a fire resistance of up to 120 minutes. pp‘ now replaces pp in Equation (28).

The proposed CEN method takes account of the influence of moisture content by determining the ‘dwell time’ and calculating the effective thermal conductivity using the modified time-temperature curve (see Section 5.7.3)

Assessment of material properties

Approximate properties

The important parameters in determining the required thickness of fire protection are the thermal conductivity h,, the density and moisture content of the fire protection material. Data exists on the performance of generic materials, such as vermiculite spray or gypsum board, and these are given in Table 20. However, for many materials there is a change in thermal conductivity with temperature and therefore the values given are averaged over the normal temperature range. The thermal response is less sensitive to changes in the density of the fire protection material.

The effective thermal properties of a particular product can be determined by ‘back-analysis’ of existing fire tests. However, the


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value of thermal conductivity that is obtained from this analysis is not constant because of factors such as location of the specimen in the furnace, variations in protection thickness, and its moisture content, and also any inaccuracy of the calculation model.

It is therefore necessary to carry out a statistical assessment of the results of these tests. This is the basis of the european approach adopted by CEN Committee TC 127.

Table 20 Examples of the properties of some common fire protective materials

Protective Material

Normal Weight Concrete Lightweight Concrete Cellular Blocks Brickwork

Gypsum Boards Fibre Silicate Sheets Perlite or Vermiculite Sheets Mineral Wool Slabs

Vermiculite or Perlite Cement Sprayed Mineral Fibre

~ ~~~

Density

(by weight)

% Moisture (kglm') Content

2350

4-5 1800

3

2.5 1000 2000

800 15 -20 600-800 3 500 - 800 10- 15

200 2

550 20

300 I1 I

Specific Heat Cp (J/kg"C)

Thermal

AD (W/m"C) Conductivity

1300

1200

0.6-0.8 1200

1.3-1.7

0.4 1200 1.2

1700 0.2 1200 0.15 1100 0.15

1100 0.2

1100 0.12

1200 0.10

The calculated value of h, is also dependent on the method of analysis (and hence back analysis) that is used. In principle, it is not possible to transfer values of h, between methods, although differences may in fact be small. The value of h, is also slightly dependent on the critical temperature used in the evaluation.

5.7.2 Approach adopted by CEN TC 127

In parallel with the activities of the Project Teams for EC3 and EC4 Parts 1.2, CEN Technical Committee 127 have been endeavoured to prepare rules for testing procedures and evaluation of fire test results, which will inevitably affect design procedures in the Eurocodes.

EC3 and 4 Parts 1.2 both specify a one-dimensional heat flow equation, as presented in Section 5.5. They refer to methods of EN yyy 5 Part 4 which is being developed by CEN Committee 127, the European committee responsible for fire testing standards.

CEN TC127 have accepted a hierarchy of three methods, as follows:


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(1) Methods based on the differential equation approach, using input thermal parameters.

(2) Methods based on a statistical interpretation of the test results (regression analysis).

(3) Graphical or approximate methods (appropriate for intumescent coatings).

A larger number of indicative test specimens may be required than in conventional UK methods. There are also various ways in which the results of these tests may be assessed within the scope of these methods.

In the differential equation method, two approaches may be envisaged. They both use the generalized differential equation, as presented in Equations (21) and (22). The approaches are:

Variable thermal conductivity method. Constant thermal conductivity method.

These two approaches within the context of the EC3 Part 1.2 and the CEN TC127 requirements, are described in the following sections.

5.7.3 Variable thermal conductivity method

This method is based on the assumption that the thermal conductivity of a protection material will vary with its temperature. As the temperature of the protection material varies between the temperature on the outside, which is close to the fire temperature, and the temperature of the steel on the inside, an average temperature is assumed to be simply the linear average of these temperatures. It has been found that, provided this assumption is used in analysing tests and also in predicting performance, errors in the non-linear temperature variations tend to cancel themselves out.

For small increments in time and rise in steel temperature obtained from the test records, the instantaneous value thermal conductivity from each indicative test is calculated by rearranging Equation (21). This value is then expressed in terms of the average protection temperature. The process is repeated for all test samples and the results of this analysis are then grouped into average temperature bands of 50°C in the protection material. For each such band, the average and standard deviation of the set of test results is calculated. Using values of the so-called “characteristic thermal conductivity” taken from these bands, the actual test results are then predicted by back-analysis.


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This back-prediction process is repeated until the acceptance limits are met. This acceptance criterion involves increasing the values of the 'characteristic thermal conductivity' by adding a fraction of the standard deviation from the family of tests to the average value within each 50°C band, as follows:

where:

hchar = characteristic thermal conductivity among the tests Aaverage = average thermal conductivity among the tests 'Ai = standard deviation of the results Kst = modification factor to obtain the appropriate acceptance

limit.

Acceptance criteria have been established. These are based on 80% confidence limit in the time taken to reach specified temperatures band from back analysis using the characteristic thermal conductivity. Typically, K,, will be approximately unity.

An important property of most sprayed protection materials is their ability to retain moisture which is held in the form of either free water, or water of crystallisation. This "dwell" time of a temperature of approximately 100°C has been found to vary as the cube of the material thickness, as follows:

dwell time = dp3 kdwell (34)

For dwell times measured in minutes, and d,, in millimetres, "kdwell" is approximately 0.0004. In solving the differential Equation (21), a time step equal to this dwell time, is included in the temperature time response at 100°C.

5.7.4 Constant thermal conductivity method

In this method it is assumed that for the duration of the heating of one test sample to a specific critical steel temperature, the thermal conductivity is constant. Each sample is analysed and a single average value of conductivity is found for each steel temperature. For example, it might be found that the test time to finally reach a steel temperature of 550°C leads to a thermal conductivity of 0.12 W/m"C. However, to reach 650"C, the value might be 0.13 W/m"C. The thermal conductivity is evaluated without taking any account of the moisture content of the material.


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From all the test results, a relationship between thermal conductivity, steel temperature and protection thickness is found using a simple linear regression formula, as follows:

The regression coefficients a. and q are then determined, which leads to the appropriate value of effective thermal conductivity for use in subsequent analysis, as a function of the protection thickness and critical temperature of the steel.

5.7.5 European regression method

An alternative approach may be used within the content of the CEN TC127 requirement, which is based on a regression method. The form of regression proposed in ENyyy5: Part 4 is more complex than that commonly used in UK (see below). It is similar to a regression method already used in Australia. It differs from the UK method in that it applies over a range of steel temperatures and is presented as follows:

l0 - -

a. +al d , +a2 - d p +a3Qa+a4dpQa A I V

+a5 d , - Q a Q , a7 A I V f a 6 - f - A I V A I V

where :

43 = time (in minutes) to reach the critical steel temperature,

dP

0, = thickness of fire protection (mm)

A N = section factor (m-') of test specimen a. etc = regression coefficients determined by analysis of the

test data.

The equation is linear in thickness and steel temperature for a given section factor. The regression equation is solved statistically over the set of data from the unloaded thermal tests. Again, the regression coefficients are established based on certain acceptance criteria, as noted in Section 5.7.3.

5.8 Comparison of protection thicknesses Although there appear to be many methods of assessing the required thickness of fire protection, all the methods described in Section 5.7 tend to give similar results. If tests were carried out for


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a particular combination of thickness and section size, the variations in fire resistance would not be significant. However, in a competitive market these small differences are important to the manufacturers, The differences between the various European methods and BS 5950: Part 8 are summarised in Table 21. The thicknesses are shown for typical board and spray materials for a section factor of 200 m-', and for 60 minutes fire resistance.

Table 21 Summary o f protection thickness for 60 minutes fire resistance according to four methods

Material Thickness (mm) for critical Method temperature o f

550°C I 650°C I 750°C

5.9 Proposed simplified approach

The methods in BS 5950: Part 8 and EC3: Part 1.2 do not lend themselves to rapid calculation of the required protection thickness. Based on analyses carried out by SCI, it is possible to use an alternative simplified approach to calculate the thickness of protection as a function of the critical (or limiting) steel temperature. If the protection thicknesses for 550°C, as specified in reference (25), are assumed to be the 'base level', it is possible to specify reduction factors for the required protection thickness appropriate for higher critical temperatures. For typical protection materials, these reduction factors are given in Table 22. The reduction factors apply to all periods of fire resistance and protection thicknesses. However, these factors must only be taken as a guide to possible thickness reductions, and the relevant manufacturer must be consulted for final design.


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Table 22 Approximate thickness reduction factors for protection materials (relative to thickness at 550" C)

l Critical temnerature PC) I Thickness reduction factor I I 550 I 1 .oo I

600

0.73 700 0.81 650 0.90

750 I 0.68 1 In almost all cases, the actual protection thicknesses determined by a rigorous calculation procedure will be significantly smaller than given by the reduction factors in Table 22 particularly, for shorter fire resistance periods (compare with Table 21). Both tables assume that the materials possess the necessary stickability at higher temperatures.

5.1 0 Partial protection to bare steel beams and columns

Partial protection to steel members is often provided by the other elements to which the members are attached. Examples are concrete floor slabs and masonry walls. In simple terms, their effect is to insulate a certain portion of the steel member from the direct effects of the fire, and therefore to reduce its exposed perimeter.

EC3 : Part 1.2 gives guidance on partial protection using concrete encasement, but does not cover partial protection by masonry. Nevertheless, the following description is appropriate to understanding the behaviour of these members.

Consider a steel column that is embedded in a cavity wall so that half of its perimeter is exposed to the fire. This member may be conservatively treated as having a section factor (heated perimeter/cross-sectional area) of half its original value. It is assumed that the heat gained by the member is spread uniformly throughout the section. However, the temperature distribution through a partially protected member is far from uniform. The steel temperature in the insulated portion may be less than half of that in the exposed portion, despite the relatively high conductivity of the steel.

The cooler portions of the cross-section have higher strength retention, and help to 'support' the hotter weaker portions. These effects contribute to a significant enhancement in the fire resistance of steel columns located within the inner leaf of masonry walls so that the flange and part of the web are exposed. Tests suggest that 30 minutes fire resistance can be achieved for such partially protected


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steel members that alone would have a fire resistance of barely 15 minutes.

Another practical example of partial protection is that of columns with ‘blocked-in’ webs using concrete blocks bonded between the flanges of the section. The effective reduction in section factor is by some 70% relative to that of the exposed member. Tests have shown that these ‘blocked-in’ column can achieve 30 minutes fire resistance. Guidance is given in BRE Report 317(27).

The required thickness of fire protection in a partially protected member may be evaluated using its reduced section factor according to the methods described in Section 5.5 and 5.7.

5.1 1 Computer methods for predicting thermal and structural response

The analysis method presented in Sections 5.3 and 5.5 dealt with EC3 cl 4.3.2

one-dimensional heat flow through uniform protection materials. There are many protected sections where heat flow occurs in two dimensions because of the non-uniform thickness of protection, or because of the shape of the section. An example of this is concrete encasement to an I section.

Various computer programs have been developed to analyse generalized heat flow through sections comprising different materials. In principle, the cross-section is modelled as a series of rectangular or trapezoidal elements and the basic heat flow equations are solved between the elements for each time interval. The surfaces exposed to the fire are subject to the precise time-temperature relationship. Other surface emission properties take into account the local heat transfer by radiation. These methods can incorporate time dependent material properties. Some more complex methods can take account of the moisture content of the materials and internal voids.

The reduced resistance of the elements of the cross-section may be determined using the material laws presented in Section 3.1 and a function of the temperature distribution calculated as above, or as obtained from tests. Equilibrium at critical cross-sections is then established with respect to the applied forces assuming plane sections as in conventional theory. Failure occurs when the member resistance is insufficient to support the applied loads.

These methods of all termed ‘advanced calculation models’ according EC4 Annexes

to EC3 and EC4: Part 1.2. All advanced methods should be calibrated against results of temperatures and structural performance obtained from representative tests.


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6 PARTIALLY AND FULLY ENCASED COLUMNS

6.1 Introduction

The principle of using other materials, such as concrete or masonry, to offer partial protection to steel sections in fire is well established. Concrete encasement provides a high degree of fire resistance by insulating the steel section and also by providing some composite action. However, concrete encasement is a relatively labour intensive and slow method of construction.

The attention of the steel industry and researchers has therefore turned towards methods of achieving 60 or 90 minutes fire resistance by partial protection in which some of the steel section is exposed to the fire. Various examples have been codified and are reviewed in the following sections. In other well established and common design cases, ‘deemed to satisfy’ requirements are appropriate (as in reference 26)

The analysis of partially or fully encased members is covered in Eurocode 4 Part 1.2 rather than Eurocode 3, as these sections depend on some degree of composite action, in fire even if this effect is discounted in normal design. Nevertheless, EC4: Part 1.2 assumes that the member resistance is calculated taking account of the composite action in fire conditions. Ignoring composite action in normal design has the effect of reducing the effective load level on the member in fire conditions by over-designing the member in normal design.

The various forms of partially protected or ‘composite’ columns that may be encountered are:

columns with concrete infilling between the flanges columns with full concrete encasement columns built into masonry walls columns with masonry infilling between the flanges concrete filled hollow sections

EC4: Part 1.2 only considers the use of concrete with partial or full encasement. Composite columns using concrete-filled hollow sections are covered in Section 10.

The different forms of concrete encasement to columns are illustrated in Figure 12. Partially encased columns may be designed as composite or non-composite at the ultimate limit state.


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Code Clause

V

4 Axis distance

Shear links welded to web

' Reinforcing bar

Partially encased column

Reinforcing bar

-. - - - Shear link

Cover to bars

4 Cover to steel section

Concrete encased column

Figure 12 Different forms of concrete encasement to columns

Tabular guidance is appropriate for simple and quick design, accepting that this approach will be necessarily more conservative than calculation methods. In most cases, both tabular and calculation approaches are covered in different Sections of EC4: Part 1.2. More comprehensive design tables are presented in an earlier publication by ECCS(28). The following sections review the design methods and present the simplified tables (with appropriate modifications).

6.2 Partially encased columns

Concrete infilling between the flanges of an I section column can E C ~ 4.2.3.3 readily increase the fire resistance of the column from less than 15 minutes as a bare steel section to over 60 minutes as a partially protected section. Longer periods of fire resistance can only be achieved by using reinforcing bars embedded within the concrete, so that the loss in strength of the steel section is balanced by the strength retention of the reinforcement.


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Code Clause

It is possible to include the compressive resistance of the concrete infill in the normal design of the column, provided it is continuous vertically and can resist any loads transferred from the beams. In practice, this is normally achieved by using welded fin plate connections to the major and minor axes of the column so that the beam to column connections may be made outside the line of the column. The concrete may either be cast in-situ using formwork attached to the tips of the flanges, or pre-cast and the partially encased columns are delivered to site. In the second case, end bearing of the individual columns may be achieved by an end plate detail to the column.

The benefit of composite action on the ‘normal’ design resistance of the partially encased column is typically 20 to 50%, depending on the size of the section. For simple design, this composite effect is often neglected and the normal design is based on the steel section alone. The reinforcing bars contribute little to the normal design resistance of the member and are neglected except under fire conditions.

6.2.1 Test results or partially encased columns

As a background to understanding the behaviour of partially encased columns, four fire tests on concrete infilled steel columns carried out by SCI in 1992. The concrete provided partial encasement except for the upper part of the steel section. This might be the case where conventional beam-column connections are used and the concrete is not continuous. The tests are reported in SCI Publication 124(29) and are summarised in Table 23.

Table 23 Results of SCI tests on web infiled steel columnsf291

Test Notes Fire Resistance Load Shear Steel Column Size Grade Ratio Connectors

1

72 0.55 500 mm centres S275 254 UC X 73 ke/m 4

Additional welded stiffener 69 0.45 500 mm centres S355 203 UC X 60 kg/m 3

71 0.35 500 mm centres S275 254 UC X 73 kglm 2

No additional measures 58 minutes 0.50 300 mm centres S275 254 UC x 73 kglm

Notes: 1 . The load ratio was based on the steel section designed to BS 5950: Part 1.

2. The shear connectors were HILT1 HVB80 ‘shot fired’ shear connectors were attached to the web.

3. In tests 3 and 4, a welded horizontal stiffener was attached to the top of the steel column to transfer force directly to the concrete.


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Code Clause

On the basis of these results, the steel columns can achieve 60 minutes fire resistance with no additional reinforcement and with nominal shear connectors (in this case HILT1 HVB shot-fired shear connectors). In two tests, a stiffener was welded between the flanges at the top of the column in order to ‘contain’ the concrete and to provide a more effective local compression transfer into the concrete. This detail provided a significant enhancement in the fire performance of the column (compare tests 1 and 4). The stiffener might be required for structural reasons and would therefore perform a dual function.

Other tests have been performed on partially encased columns in Germany, but the concrete was often heavily reinforced and welded shear connectors were used.

6.2.2 Squash resistance

The approach of EC4: Part 1.2 is based on ECCS TN 55(28) and is EC4 ,, 4.3.6,1

more general than the SCI publication(29). However, EC4: Part 1.2 EC4 F often requires the use of reinforcing bars within the concrete encasement. Both the concrete and reinforcement are assumed to be continuous so that there is effective compression transfer. Table 4.6 in EC4: Part 1.2 gives the minimum column dimensions for use of the method. In EC4: Part 1.2 the load level corresponds to analysis of the column as a composite section under both normal and fire conditions.

The tabular approach does not lead directly to the required amount of reinforcement, which has to be determined by calculation, based on the data given in Annex F of EC4: Part 1.2. Reinforcement is required in all cases, and additional shear links are either welded to the web, or passed through holes punched in the web. These links are required to prevent bursting of the longitudinal bars in compression.

Analysis at the fire limit state takes into account composite action of all the elements of the cross-section of the column. In principle, reduction factors are applied to the strength of all the elements, as a function of the temperatures that they reach after a given fire exposure time. A typical case is illustrated in Figure 13. The reduced strength of all the materials is obtained using the material laws given in Section 3.

The axial (or squash) resistance of a partially encased column at the EC4 fire limit state is therefore given by the following formula, which is Annex F.6(1)

a simplification of that in EC4: Part 1.2:


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Code Clause

where:

A , = cross-sectional area of one steel flange

A , = cross-sectional area of the steel web

AS = total cross-sectional area of the reinforcing bars in the concrete

4 = cross-sectional area of the concrete (minus the steel section and reinforcement)

f,,, f y s and& are the characteristic strengths of the steel, reinforcement and concrete respectively

= strength reduction factor due to the temperature of the EC4 Annex F exposed flange (see Table 24). Table F.4

= strength reduction factor due to the average temperature 1x4 Annex F in the web. The internal portion is fully effective. Table F.2

= strength reduction factor due to the temperature in the EC4 Annex F reinforcement at axis distance, U (defined in Figure 14). Tables F.5 &

F.6

= factor corresponding to the average temperature of the concrete. The outer portion of the concrete encasement is ignored. The 0.85 factor due to the influence of sustained loading on the concrete in normal column design is set to unity in fire.

Factors r1 to r4 are not presented explicily in EC4: Part 1.2. The partial safety factors for all the materials are all set to unity in fire, and are not stated explicitly.

Table 24 Average temperature of the flanges in a partially encased column

Fire Section width (mm) Resistance l I I 1

I 30 I 807 I 743 I 704 1 679 1 660 I c I I I 60 I 935 I 871 I 833 I 807 1 789 I I I I 1 I 90 I 969 I 928 I 903 1 887 I 875 I

EC4 Annex F Table F. 1

120 - I 993 I 974 I 962 I 953


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Code Clause

X ’ Y

4

r 4 = 0.9

Figure 13

r 3 = 0.6

Outer layer neglected

90 30 120 40

Reduced section of a partially encased column in fire (60 minute fire resistance)

Figure 14 Definition of axis distances to reinforcing bars

The corresponding ‘cold’ or normal resistance, Npg,Rd of the column is obtained by setting the factors rl to to unity, divided by the appropriate partial safety factors for these materials at the ultimate limit state. In this case, the concrete strength is taken as 0. Sf,. The reinforcement term, A,, may be ignored in determining the normal resistance of the cross-section. It is therefore treated as ‘fire reinforcement’. Design values of the compressive resistance of partially encased columns, Npl,Rd, are given in reference (30). The maximum load level that can be attained in a partially encased column restrained against buckling may then be presented as Nfi,pl.Rd/NpI,Rd. This ratio has the same definition as fiv in Equation (6 ) .

The approach to calculating the temperatures and reduction factors in Annex F of EC4: Part 1.2 is semi-empirical. In simple terms, a reduced effective cross-section may be defined at each fire resistance


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Code Clause

period. The reduced section is of the form of Figure 13, showing also the zone of concrete that is ignored. Typical values of the reduced strength of the various elements for the R60 case are presented in this figure.

6.2.3 Buckling resistance

The buckling resistance of a partially encased column is also adversely affected because the flanges, which largely provide the stability of the section, are hotter than the remaining part. However, minor axis buckling of a column is normally the dominant effect, in which case the section does not lose its buckling resistance disproportionately. In a full analysis, account may also be taken of EC4 c, 4.3.6.1 the reduced effective length of columns in fire conditions (see Section 4.4). A design study has been carried out of the influence of the effective length on the buckling resistance of these columns in fire conditions, and the results are presented in Figure 15 for a typical partially encased column.

EC4: Part 1.2 permits use of an effective length factor of 0.5 for vertically continuous columns in braced frames, and 0.7 for other columns, in fire conditions. For the case shown in Figure 15, it is apparent that even with an effective length factor of 0.7, there is only a slight reduction in buckling resistance of the column with increasing column length. For an effective length factor of 0.6, the buckling resistance is largely unaffected, and for a factor of 0.5, it increases with column length.

The proposed UK NAD for EC3 and EC4 Parts 1.2 recommends use UK NAD to

of a multiple of 0.7 on the column effective length, which for normal design is 0.85. It follows that the effective length factor in fire conditions is 0.7 X 0.85 = 0.6 rather than 0.5.

EC4: Part 1.2

Therefore, for simple analysis, it would be reasonable to define the maximum load level for a partially encased column in a braced frame by its reduction in squush resistance, and independent of its length or slenderness. The maximum load level may be expressed in the same form as Equation (6), as follows:

where qfi,t - __._

- Efi,d,t

Rd


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Code Clause

The maximum load level qfi,t is then used as a multiplication factor applied to the normal design resistance of the member, which would, in this case, take account of the influence of buckling through its slenderness. If composite action is ignored in normal design, then the maximum load that may be applied, as a proportion of the resistance of the steel section, will increase. A minimum load level of 0.5 would normally be sufficient to ensure that the column possesses adequate fire resistance in most practical load cases.

c 0.7 .- 0

0 a c

* 0.6

V c 0.4 0

.- c 0

0.3 U ld 0 J 0.2

254 uc 73 S275 steel

R60 L,=0.6L

Le =0.7L

Le=1.0L

R90 Le=0.7L

0 0 1 2 3 4 5 6

Column length (m)

Figure 15 Influence of column length and effective length on fire resistance of partially encased columns (Le = effective length in fire conditions)

6.2.4 Tabular design

For normal design based on the steel section alone, the SCI tests(29) demonstrated that 60 minutes fire resistance can be achieved on 254 UC columns without the use of additional reinforcing bars. The same observation is not so clear from EC4: Part 1.2.

Adopting the empirical approach of Annex F leads to a modified table for simplified design, as in Table 26. This table and EC4: Part 1.2 assume that the column acts compositely at both the ultimate and fire limit states. In general, four reinforcing bars are required for the R60 case.


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Code Clause

The load level that can be achieved depends also on the steel grade and thickness of the section, as the higher the steel strength, the smaller the benefit of composite action in fire. Because of the slight dependence of buckling resistance with slenderness, the load level is calculated for a typical column whose effective length is 15 times its width, including also a further effective length factor of 0.7 in fire conditions.

The axis distance to the reinforcing bars is defined in Figure 14. The reinforcement area is distributed equally to the four corners of the section. These bars are held in place by links.

This analysis has been carried out for Universal Column sections, rather than continental HE sections designed as composite columns, which was the basis of Table 4.6 in EC4: Part 1.2. Table 25 applies to S355 steel, and smaller reinforcement percentages may be used for S275 steel. A further requirement is that the maximum percentage of reinforcement, expressed in terms of the area of concrete, should not exceed 6% I

Table 25 Minimum cross-sectional dimensions and reinforcement areas of partially encased composite columns (modified from Table 4.6 of EC4: Part 1.2)

Fire Resistance (min)

R30 R120 R90 R60

EC4 Table 4.6

Load Level qfi = 0.3 Minimum dimensions

60 50 40 Minimum axis distance 0.8 0.5 0.3 0 Minimum reinforcement 300 250 (300) 200 (260) 150 (160)

Load Level qfi = 0.5 Minimum dimensions 200 250 (300) 300 Minimum reinforcement 0.2

50 40 30 Minimum axis distance 1 .o 0.5

Load Level qfi = 0.7 Minimum dimensions 250 300 Minimum reinforcement 0.4 1 .o Minimum axis distance 30 40

The minimum reinforcement area is expressed as a ratio of the cross-sectional area of one flange

EC4 values in brackets These values are appropriate for S355 steel The values of reinforcement area may be reduced in proportion to the steel design strength, f,,/355.


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6.2.5

Code Clause

Minimum dimensions have been modified slightly to conform to UK test data and UC section sizes. The new table now gives minimum percentages of reinforcement expressed as a ratio of the cross- sectional area of one flange, which were not given previously in Table 4.6 of EC4: Part 1.2. The ratio of web to flange thickness should be greater than 0.6, which is consistent with all UC sections. Changes in minimum dimensions are noted. It is proposed that this UK NAD to modified table is included in the UK NAD.

This approach described in Section 6.2.3 may be used to calculate 1x4 Cl 4.2.3.3 load levels for all UC sections. Using this analysis, maximum load levels for partially encased columns are presented in Table 26, corresponding to 60 minutes fire resistance. The difference between the load levels adopted for ‘steel’ and for ‘composite’ sections in Table 26 arises where the normal design of the column is based on the steel section alone. Greatest benefit is achieved for the lightest steel sections.

For 90 minutes fire resistance, the same trend is evident, although all the elements of the cross-section become progressively weaker. The maximum load levels are presented in Table 27. In this case, the axis distance to the reinforcing bars is increased by 10 mm in order to compensate for the loss of strength of the section. The slight dependence on column slenderness is the same as for the lower fire resistance period.

In practice, many of the lighter steel sections can achieve a load level in excess of 0.4 at 90 minutes fire resistance, provided normal design is based on the resistance of the steel section alone. Longer periods of fire resistance can realistically only be achieved by full concrete encasement (see Section 9.4.2).

In conclusion, the modified Table 4.6 in EC4: Part 1.2 (see Table 25) provides the necessary information in the format for general application. This table has been re-evaluated for UC sections and now gives the minimum area of reinforcement as a ratio of the cross-section area of the bottom flange of the steel section. It is conservative for the full range of UC or HE sections.

Combined moment and compression

At the fire limit state, the combination of moment and axial force on partially encased sections is more problematical because of the greater loss of strength of the flanges of the steel section which contribute proportionally more to the bending resistance. In principle, the tabular methods only apply to columns in ‘simple’ construction with small or modest moments (for example, due to eccentricity of axial load).

EC4: Part 1.2

& Table 4 .6


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No guidance on combined moment and compression is given in EC4 Part 1.2. In reference (28), it is proposed that a simple method of taking account of moment is to use a magnification factor of 1.5 on the major axis moments. The effective load level is given by the combination of axial force and moment, as follows:

Nti,d Mfi,d

Npe,Rd Mpe,Rd rlfi,t - >- + (39)

where Nfi,d and kf,,d are the axial force and major axis moment applied at the fire limit state respectively. A similar linear interaction term may be included for minor axis moments. No magnification factor is applied to minor axis moments because the section is less dependant on the flanges in this case. The load level is then compared to the maximum values given in Tables 26 and 27 for UC sections.

This approach to combined moment and compression is proposed for inclusion in the UK NAD.

Table 26 Maximum load levels of partially concrete encased columns for 60 minutes fire resistance

Section Size Steelf,, = Steelf, = 275 N/mm2

Steel Steel Composite 203 UC 46 0.68

0.40 0.37 0.45 86 0.44 0.40 0.51 71 0.50 0.44 0.57 60 0.55 0.47 0.62 52 0.60 0.50

254 UC 73 0.64 0.46 0.56 89 0.56 0.43 0.50

107

0.41 0.39 0.45 167 0.45 0.41 0.50 132 0.45 0.40 0.50

305 UC 97 0.77 0.56 0.69 118 0.68 0.50 0.61 137 0.67 0.50 0.60 158 0.61 0.48

0.41 0.40 0.46 283 0.44 0.41 0.48 240 0.49 0.44 0.53 198 0.55

555 N/mmz --l i ComDosite (no?:dia)

0.44 4 x 20 1 0.42 4 x 20

4 x 20 0.34 4 x 20 0.36 4 x 20 0.39

0.42 4 x 20 0.39 4 x 20 0.36 4 x 20 0.38 4 x 25 0.36 4 x 25

0.50 4 X 25 0.46 4 x 25 0.47 4 x 32 0.44 4 x 32 0.41 4 x 32 0.38 4 x 32 0.35 4 x 32

Code Clause

UK NAD to EC4: Part 1.2

~ ~ ~~~ _____~

Notes: Data for C25/30 concrete intill ‘Steel’ means normal design based on steel section. ‘Composite’ means normal design based on composite section. The reinforcing bars are placed at an axis distance of 40 mm. Load levels are calculated for a column effective length of 15 X column width.


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Code Clause

Table 27 Maximum load levels of partially concrete encased columns for 90 minutes fire resistance

Section Size

203 UC 46 52 60 71 86

254 UC 73 89

107 132 167

305 UC 97 118 137 158 198 240 283

Steel f, =

Steel

0.48 0.44 0.40 0.36 0.32

0.46 0.40 0.36 0.32 0.30

0.61 0.53 0.48 0.44 0.38 0.35 0.33

-

!75 N/mm2 1 Steel f.. = 355 N/mm2 I Bars ~ ~~~

Composite Steel Composite (no. X dia)

I I I

0.35 0.33 0.31 0.29 0.26

0.34 0.31 0.29 0.27 0.25

0.43 0.39 0.37 0.35 0.32 0.30

l 0.29

0.42

0.25 0.31 0.27 0.35 0.29 0.38 0.31

0.22 0.27

0.40 0.30 0.36 0.28 0.32 0.26 0.29 0.24 0.26 0.22

0.54 0.38 0.47 0.36 0.43 0.33 0.39 0.32 0.35 0.29 0.32 0.27 0.30 0.25

4 x 25 4 x 25 4 x 25 4 x 25 4 x 25

4 X 25 4 x 25 4 x 25 4 x 25 4 x 25

4 x 32 4 x 32 4 X 32 4 X 32 4 X 32 4 X 32 4 X 32

Notes: Data for C25/30 concrete infill ‘Steel’ means normal design based on steel section ‘Composite’ means normal design based on composite section The reinforcing bars are placed at an axis distance of 50 mm Load levels are calculated for a column effective length of 15 X column width

6.2.6 Detailing rules

Detailing rules have to be observed in terms of the minimum cover (axis distance minus half bar diameter) to the bar reinforcement, and the maximum spacing of the shear links around the bars. These links should be of a minimum diameter of 6 mm and spacing of 300 mm, and should be adequately connected to the web of the section. Their r6le is to prevent bursting of the main reinforcing bars and to achieve composite action with the concrete. The links are necessary only where additional bars are used, and are not needed where some other method of shear connection is used. General detailing rules are presented in Section 11.

Fully encased I sections

Fully concrete encased I sections are only covered by a simple E C ~ 4.2.3.2 tabular method in EC4: Part 1.2. These sections are illustrated in Figure 12. The concrete cover to the steel section permits the steel to retain a good proportion of its strength. Table 28 presents the minimum external dimensions of the columns, and the minimum cover to the reinforcement and to the steel section. Using these minimum dimensions, the composite column can achieve the required


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Code Clause

fire resistance, irrespective of the load level in fire conditions (in practice, the maximum load level is 0.6).

No guidance is given on the amount of longitudinal reinforcement, although the minimum bar diameter should be 12 mm. Otherwise, the amount of reinforcement should be consistent with the normal design requirements to EC4: Part 1.1. A further requirement of EC4: Part 1.2 is that the reinforcing links should be of minimum diameter of 6 mm and spaced at not less than 250 mm vertically. These links prevent bursting of the reinforcement and spalling of the concrete.

Where the concrete provides only an insulating function (i.e. the section is otherwise designed as non-composite), a separate simplified table may be used in the same form as for encased beams (see Table 3 1). In this case the minimum concrete cover to the section is much reduced. Fabric reinforcement (4 mm minimum bar diameter and 250 mm maximum spacing) is placed around the section to prevent spalling. In the R30 case, only partial encasement is required, which is a safe lower bound to the design of these sections (see Section 6.2.4).

Table 2% Minimum dimensions of fully encased composite columns

Fire Resistance (mins)

R60 R90

60 50 40 40 Cover to steel section

400 350 250 200 Column width (mm)

R180 R120

Cover to reinforcing bar 20 20 30 40

Minimum Dimensions

Column width = external concrete encasement

The methods of fire resistant design in EC4 Part 1.2 are broadly similar to those in BS 5950: Part 8, although comparisons also depend on the methods used for normal design. The ‘encased strut’ method of BS 5950: Part 1, for example, is a hybrid approach where some account is taken of the beneficial effects of the concrete encasement to reduce the column slenderness and hence to increase the column’ S buckling resistance.

EC4 Table 4.4


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Code Clause

6.4 Columns with concrete block infills

Partial protection of steel columns in a masonry wall can be achieved if only one flange and a modest proportion of the web is exposed. Two fire tests have been carried out on columns shielded in this way. A column with a section factor of 77m-’ (based on the exposed portion of the section) achieved a fire resistance of 30 minutes when subject to a load ratio of 0.6. Enhanced fire resistance can be achieved by fully infilling the web of a column built into a wall so that only one flange is exposed.

Concern has been made about thermally induced curvatures in the columns causing distress in the masonry walls. Provided the columns are part of a braced or sway frame (i.e. they are not isolated), this effect is relatively small and should not be critical to the integrity of the walls.

A further method of fire protecting steel columns is to use blockwork infills between the flanges of isolated columns. The concrete blocks offer a ‘heat sink’ so that the core of the steel section retains more of its strength.

Although they are not covered by the requirements of EC3: Part 1.2, columns with ‘blocked in’ webs are covered by BS 5950: Part 8 and BRE Digest 317(27). A fire resistance of at least 30 minutes can be achieved when the ratio of the exposed perimeter to the cross-sectional area of the steel column is less than 69m-’. The fire resistance of the column can be enhanced by using intumescent coatings on the exposed steel flanges.


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Code Clause

7 PARTIALLY AND FULLY ENCASED BEAMS

Concrete infilling or full encasement can increase the fire resistance of a steel beam, although the benefits are usually less than for columns protected in a similar way. This is because more of the concrete is in tension than compression for a member in bending, which means that it contributes less to the composite action in fire. The principal effect is the insulation provided by the concrete to the steel section and any additional reinforcement.

Three cases are considered to various extents in EC4: Part 1.2: beams with partial encasement composite beams with partial encasement steel or composite beams with full encasement.

These cases are considered in the following Sections. The forms of construction are illustrated in Figures 16 and 17.

Shear links welded to web or through holes in web

' Reinforcing bar

Non-composite partial ly encased beam

Axis distance ' - S k

Shear connector

Reinforcing bar

Composite partial ly encased beam

Figure 16 Different forms of partially encased l beams


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Code Clause

Concrete encased beam

Slim floor beam

Figure 17 Different forms of encased beams

7.1 Partially encased beams

Concrete infilling between the flanges of an I section beam can EC4 cl 4.2.2 increase its fire resistance to 60 minutes, although additional EC4 Tables embedded reinforcement is required for longer periods of fire resistance. Minimum cross-sectional dimensions are presented in EC4: Part 1.2 Table 4. l for composite beams. No cases are tabulated for partially encased non-composite beams, except for the simple case of 30 minutes fire resistance, which does not require reinforcement.

4.1 & 4.2

However, the general calculation methods given for partially encased composite beams may also be applied to non-composite beams by determining the temperatures and strength reductions of all the elements at the required fire resistance period. The bottom flange gets much hotter and loses more strength than the web and top flange, which is assumed to support a concrete slab. Hence the plastic neutral axis of the beam rises so that most of the web goes


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Code Clause

into tension. As a good approximation for lightly reinforced beams, the neutral axis lies close to the top flange in fire conditions.

The model for calculating the reduced bending resistance of the steel section as presented in EC4: Part 1.2 Annex E, is based on establishing reduced effective areas of the cross-section, as illustrated in Figure 18. It is similar to the reduction factor method described below, but is semi-empirical and is based on various tables and formulae.

In the extreme case, by taking moments around the top flange of the steel beam, it follows that the moment resistance of the partially encased beam may be given by the approximate formula:

where the terms are defined as in Section 6.2, except:

his the distance between the centre of the flanges of the section, U is EC4 Annex E the axis distance to the bar reinforcement from the top of the lower flange. This distance is a variable and is used in determining r3. No guidance is given on minimum axis distances, but the minimum concrete cover is 20 mm to the steel flange. In practice, the axis EC4 c, 5,1(6) distances should be sufficient to provide the necessary insulation to the bars and the valuesgiven in Table 25 are appropriate also for partially encasedbeams.

Table E.6

Therefore, the maximum load level that a partially encased beam can resist in fire conditions is defined by:

The embedded reinforcement A,, is ignored in the normal design of the steel beam, whose plastic bending resistance is therefore given by :

The ratio of bending resistances in qfi depends largely on the areas of bar reinforcement relative to the aiea of the bottom flange.


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Code Clause

Concrete in compression

Reinforcement in tension

Neglected zone

Plastic neutral a

Web in tension

Steel ignored

lxis

Figure 18 Reduced section of a partially encased beam in fire

Following the general calculation method of Annex E of EC4: Part 1.2 leads to the results in Table 29 for 60, 90 and 120 minutes fire resistance. These results are expressed as the minimum load level for a given section range. The dependence on the weight of steel section is relatively small, as it is the web of the section and the reinforcement that contributes the majority of the tensile resistance of the beam in fire.

Table 29 Maximum load levels for partially encased steel beams supporting a concrete slab (using the analysis of EC4: Part 1.2 Annex E)

Section Range

203 X 133 UB 254 X 146 305 x 127 305 x 165 356 x 127

406 x 140 406 x 178

356 X 171

457 X 152 457 X 191 533 x 200 610 X 229 610 X 305 686 X 254 762 x 267

T

I R60 0 bars

0.38 0.40 0.47 0.43 0.50 0.49 0.54 0.53 0.56 0.56 0.60 0.63 0.59 0.66 0.69

R90 2 X 20 mm dia.

0.32 0.34 0.36 0.39 0.41 0.42 0.46 0.46 0.46 0.48 0.55 0.52 0.55 0.50 0.57

R120 2 x 25 mm dia.

axis < 60 mm 0.27

axis < 60 mm 0.33

axis < 60 mm 0.36 0.38 0.40 0.39 0.43 0.47 0.50 0.47 0.53 0.55

EC4 Annex H

All load levels are minimum for the section range in S355 steel Axis distance to reinforcing bars = 50 mm for R90 and 60 mm for R120


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Code Clause

An important observation is that the unreinforced section is still reasonably effective for 60 minutes fire resistance provided the load level is not greater than 0.5 for larger beam sizes, for which this method of protection is generally more appropriate. This load level may be achieved for 90 minutes resistance by using two reinforcing bars of 20 or 25 mm diameter adjacent to the bottom flange. Some narrow sections cannot be used because they do not achieve the necessary minimum axis distance to these bars.

7.2

7.3

Partially encased composite beams

Partially encased composite beams are treated explicitly in EC4: EC4 cl 4.2.2 Part 1.2, and design tables are presented for direct design. Both the & Tables 4.1 minimum section width and reinforcement percentage are tabulated as a function of load levels of 0.3, 0.5 and 0.7. The minimum axis distance to the reinforcement is presented in EC4: Part 1.2 Table 4.2.

& 4.2

The approach is reviewed in Section 8.3, under the chapter heading of composite beams. A simplified table for UB sections in the same form as Table 29 is presented.

Fully encased beams

Complete encasement provides the necessary insulation to the steel section and does not rely on composite action. Table 30 presents the minimum depth of cover to the flanges of an encased steel section, dependent on the fire resistance required. No additional reinforcement is needed except for light fabric reinforcement around the section. The R30 case corresponds to a partially encased section EC4 cl 5.1(6)

with no reinforcement. The data is similar to that given in the ‘deemed to satisfy’ tables of the Building Regulations. The fabric reinforcement should have a maximum spacing of 250 mm between the wires.

The behaviour of fully encased composite and non-composite beams is similar, and the same design table (Table 30) may be used in both cases.

Table 30 Minimum concrete cover to fully encased beams and columns

Fire Resistance (min) 50 40 30 25 0 Concrete Cover to Steel Section (mm)

180 120 90 60 30


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Code Clause

7.4 Slim floor beams

Slim floor beams comprise a UC or HE steel section with a welded bottom flange plate that supports a concrete slab, so that all of the steel section, except the bottom plate, is encased in concrete. A typical beam is shown in Figure 17. The floor slab may be of the form of a precast concrete unit, or, alternatively, a deep composite slab. In both cases, a fire resistance of 60 minutes can be achieved without further fire protection. Design guidance exists in two SCI publications - P1 and P127(32).

This form of construction is not specifically covered by the requirements of EC4: Part 1.2, although the general principles apply. Normally, slim floor beams are designed as non-composite, unless provisions for composite action are made by the use of shear connectors. In fire conditions, however, there is sufficient test evidence to show that significant composite action occurs between the concrete encasement and the steel section without shear connectors to improve the load and fire resistance of the section. This benefit is partly responsible for the good performance of these sections in fire.

The temperature distribution through the steel and concrete components is presented in the two SCI p~blications(~ ‘ 3 32) . There is an important thermal interface resistance between the plate and section which reduces the bottom flange temperature by approximately 100°C relative to the exposed plate. At 100 mm above the bottom flange, the steel web and concrete temperatures are less than 300°C, at which temperature they may be assumed to be at full strength.

The maximum load level that slim floor beams can resist is given by the moment resistance of the heat affected ‘composite’ section divided by the moment resistance of the steel section (and plate). The results for various common design cases are presented in Table 31. The maximum load levels for slim floor beams that are designed as composite under normal conditions are lower than those that are designed as non-composite, because the ‘reserve’ of composite action that can be achieved in fire is smaller in the first case.

The maximum load levels for slim floor beams with deep composite slabs are higher than those with precast concrete slabs because of the greater composite action experienced in the first case. Fire tests carried out on the slim floor beams utilising the space between the ribs of the composite slabs for service openings through the beam showed that the maximum load ratios were reduced typically by a further 10 % .


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Code Clause

In all cases, a load level in excess of 0.5 indicates that fire resistance is not likely to be critical to the design of the slim floor beam, and therefore additional fire protection is not required. However, for longer periods of fire resistance, sprays, boards or intumescent coatings may be successfully applied to the exposed flange. The required protection thickness is based on the section factor of the exposed flange and plate.

l 'able 31 Maximum load levels for various forms of slim floor construction at 60 minutes fire resistance (revised values from SCI Publication 12 7(311)

Universal I Slim floor construction with: Column as a

beam Precast concrete slabs Deep deck slabs

60

0.68 86 0.67 71 0.65

254 UC X 73 -

89

0.74 167 0.71 132 0.68 107 0.67

305 UC X 97 -

118

0.71 158 0.69 137 0.67

C ~ ~~

Approximately- 10% lower than NC values.

Depends on depth of in-situ

slab

Not appropriate

NC H C

0.82 0.50

0.46 0.71 0.44 0.68 0.43 appropriate 0.63 0.41 Not 0.61 0.50 0.65 0.76 0.47 0.59 0.71 0.46 0.55 0.68 0.46 0.52 0.66 0.47 0.49 0.66

0.56 0.76 0.53 0.77

appropriate 0.51 0.79 Not

Lightweight concrete as in-situ concrete in all cases S355 steel section and plate NC = non-composite under normal conditions C = composite using welded shear connectors H = non-composite, with 160 mm diameter web openings

7.5 Shelf angle floor beams

Shelf angle floor beams comprise a UB steel section with welded or bolted angles attached to the web which support a concrete slab, so that the upper part of the steel section is encased in concrete. The floor slab is usually made of precast concrete units with or without an in-situ concrete topping. However, the steel section is infilled with concrete. The legs of the steel angles are placed upwards so that they are also encased. Fire resistances of 30 and 60 minutes can be achieved, although in the latter case, only a modest proportion of the section can be left exposed.

BS 5950 Appendix E

This form of construction is not specifically covered by the requirements of EC4: Part 1.2, although again the general principles apply. The temperature distribution through the steel and concrete


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Code Clause

components presented in BS 5950: Part 8@) may be used. The moment resistance of the section can be determined from first principles by determining the reduced strength of all the steel elements in fire. The fully exposed elements possess little tensile strength at 60 minutes fire exposure, but the cooler shelf angle can resist significant tension and helps to retain some of the moment resistance that is otherwise lost.

The maximum load level that the shelf angle floor beam can support is again given by the moment resistance of the heat affected section divided by the moment resistance of the steel section. The results are expressed as a function of the depth of section that is exposed. Design tables are presented in SCI publication 126(33). A beam with half of its section exposed can achieve a load level of approximately 0.4 at 60 minutes fire resistance. It is not normally cost-effective to consider using shelf angle floor construction if the section has to be fire protected.


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Code Clause

COMPOSITE BEAMS

8.1 Introduction

Composite beams comprise steel sections structurally attached to a concrete or composite slab and are designed to the requirements of Eurocode 4 Part 1.1 and also BS 5950: Part 3(34). Shear connectors, usually in the form of welded studs, provide the necessary longitudinal shear transfer between the steeI and concrete. The benefits of composite action under normal conditions are considerable and lead to an increase in moment resistance of 1.5 to 2.5 times that of the steel section, depending on the proportions of the composite section. The nature of composite action is such that most of the steel section is in tension and the plastic neutral axis of the composite section is close to, or in the concrete slab.

Two forms of composite beam construction are covered by EC4: Part 1.2. These are:

(a) Composite beams with traditional board or spray protection around the steel section.

(b) Composite beams with concrete infills between the flanges of the steel section.

Both forms of construction are analysed by ignoring the concrete encasement under normal conditions. Design tables for composite beams are presented in reference (36). The two forms of fire protection are treated separately in terms of the temperatures that are experienced in fire conditions and are covered in Sections 7.2 and 7.3. In the second case, additional reinforcement may be embedded in the concrete encasement to provide enhanced fire resistance.

In both cases, the temperature distribution in the cross-section is independent of the structural action. However, the structural performance of composite beams is subtly different to that of steel beams suporting a concrete slab in fire conditions. Therefore, it was necessary to prepare different rules for composite beams.

All the methods in EC4: Part 1.2 assume that the voids created by the deck shape over the beam are properly filled. This is done to ensure that the top flange of the steel section remains relatively cool. Guidance is given in the SCI publication (36) on the use of unfilled voids for 60 minutes fire resistance.


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Code Clause

8.2 Composite beams with board or spray protection

Two methods of calculation for composite beams with traditional fire protection are presented in EC4: Part 1.2:

(a) General analysis method: In this approach the temperatures of all the elements of the cross-section are calculated independently and the moment resistance determined from the first principles using the reduced strength of the elements.

(b) Critical temperature method: This approach is analogous to the adaptation factor for non-composite beams, and is more direct although more approximate than method (a).

The two methods will be discussed in the following sub-sections. Designers are nevertheless encouraged to use the critical temperatures presented in Section 8.2.4 for direct selection of the thickness of fire protection, when using the methods of Section 5.

8.2.1 General analysis method

This general analysis method in EC4: Part 1.2 assumes that the EC4 c, 4,3.3.2 temperature of the three steel elements, that is the top and is bottom flanges and the web, may be determined independently using the standard heat flow differential equation, as presented in Section 5.3. In using these equations, the member may be either unprotected or protected, although in practice an unprotected composite beam could not realistically achieve a fire resistance of 30 minutes. The moment resistance of a composite beam is then evaluated using the method in Annex D of EC4: Part 1.2.

The temperature and stress distributions in a composite beam are presented in Figure 19. For profile protection, temperatures are evaluated using the section factor of the lower flange, which is given by the term 2(b, + tf)/b, t,. The section factor of the upper flange, which is connected to the slab, is given by the term (b, + 2tf)/bftf. However, as a simplification to assist the analysis, the upper flange temperature may conservatively be taken as 70% of that of the lower flange temperature, and the web temperature may be taken as equal to that of the lower flange temperature.

For box protection, the temperature of the section is determined assuming a uniform temperature throughout all the elements, and therefore the section factor is based on a member with 3-sided exposure (as in Figure 9).


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Code Clause

The temperature in the concrete slab may be determined from EC4: EC4 Part 1.2 Table 4.10, but in practice it is reasonable to assume that the concrete strength is unaffected at the modest temperatures experienced. The temperature of the shear connectors is taken as 80% of that of the upper flange of the beam and their resistance is determined from Equation 6 . 1 3 of EC4: Part 1 .1 , taking account of the reduction in strength of the steel at this temperature. The concrete temperature is not critical to the strength of the shear connectors.

Table 4.10

As noted previously, partial factors for all materials are set to unity EC4 cl 4.3.3.5 in fire conditions, which means that initially there is an apparent rise in the strength of both the concrete and shear connectors relative to that of the steel section (due to their partial factors of 1 .5 and 1.25 respectively under normal conditions).

Cross-section

P 0.85 f

fy /Ya Ultimate

limit state

< f Y

Serviceability limit state

Actual Theoretical

Temperature distributions

fYkY,e

Fire limit state

Stress distributions in composite beams

Figure 19 Temperature and stress variation in composite beams in fire conditions

Later, at higher temperatures, the shear connectors lose strength less rapidly than the main part of the steel section, which causes an


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Code Clause

increase in the effective degree of shear connection in fire conditions. Therefore, composite beams designed for low degrees of shear connection under normal conditions retain more of their moment resistance less rapidly under fire conditions than fully composite beams.

It should be noted that the general analysis method is not appropriate for general design, but rather for preparation of design tables and software.

8.2.2 Adaptation factor for composite beams

The concept of an adaptation factor applied to either the load level or EC4 c, 4.3.3.3 resistance of steel beams and columns in fire conditions has been established in EC3: Part 1 .Z('), and is described in Section 4.3 .3 . In principle, the adaptation factor is chosen so that the performance of the member in fire is equivalent to the basic performance of steel at elevated temperatures. For a steel beam supporting a concrete slab, the top part of the section is cooler than the bottom part (see Figure 19). This leads to an increase in the moment resistance, and hence critical temperature, of the beam in relation to a uniformly heated member

The adaptation factor, K, is defined by solving Equation (12) and the same formula applies for composite beams, except for the value of K , as follows:

The material law for the strength of steel at elevated temperature, 8, is defined by the term ky,e in Table 5. The ratio of partial factors in brackets is 1.05 in the UK NAD. The maximum load level, qfi,t, is defined as the proportion of the normal bending resistance of the composite member that can be reached in fire conditions, as follows:

For a steel beam supporting a concrete slab, the critical temperature, e,,., at which the member fails is given by the value of 8 satisfying Equation (43). In Section 4.3.3, a unique value of adaptation factor, K was derived, which is taken conservatively as 0.7.

A similar approach may be adopted for composite beams, which also benefit from the top part of the section being relatively cool due to its attachment to the slab. In this case the load level r)fi,t is defined


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Code Clause

as for the load or moments acting on the composite section. However, because the neutral axis of the composite section does not rise significantly in fire, there is only a modest gain in the moment resistance of the composite beam relative to a uniformly heated composite beam. This action is illustrated in Figure 20. Therefore, the adaptation factor for a composite beam is closer to unity than for a steel beam supporting a slab.

However, the concept of degree of shear connection is important in the design of composite beams. In normal conditions, the moment resistance of a composite beam decreases with decreasing degree of shear connection, and a minimum of 40% shear connection is permitted in EC4: Part 1.1. In fire conditions, the shear connectors are partially insulated and retain more of their strength than the steel section. As noted earlier, the ‘effective’ degree of shear connection increases in fire conditions, leading to a less rapid decrease in moment resistance with temperature.

Degree of shear ,” connection = 0.7

,,’ Test result g’ (degree of shear

Degree of shear 4- connection = 0.4) connection = 1.0

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7

Load level on composite beam

Figure 20 Critical temperatures for composite beams in fire conditions

The result is that the degree of shear connection, N/Nf, of a composite beam under normal conditions affects the adaptation factor, K . The same definition as noted above applies, but the load level in Equation (44) is redefined in terms of the moment resistance


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Code Clause

of the composite section designed with an increased degree of shear connection in fire conditions.

The adaptation factor, K , for composite beams has been shown to be closely approximated by the relationship:

K = 0.7 + 0.2 (N/Nf) (45)

When N/N, = 0, the equation reduces to that of a steel beam (i.e. K = 0.7), as presented in Section 4.3.3 I

This approach is simplified in EC4: Part 1.2 by adopting a single EC4

value for the adaptation factor of 0.9 (corresponding to N/Nf = l ) , cl. 4.3.3.3(3)

for all composite beams, which is conservative for lower degrees of shear connection. The critical temperature of a composite beam is obtained by solving Equation (43) with K = 0.9.

8.2.3 Test data on composite beams

As a background to understanding the behaviour of composite beams, four full scale fire tests were carried out by SCI in 1991(36) in order to verify these approaches, and also to determine the influence of not filling the voids above the steel beam created by the shape of the composite slab. As noted earlier, this measure improves the economy of composite construction. All the tests used a 305 UB 33 steel section in S275 steel supporting a 120 mm deep composite slab. The load level in the tests was 0.66 based on a calculated degree of shear connection of 37%.

Test 1 consisted of 18 mm of board protection with filled voids above the top flange. The fire resistance in the test was 68 minutes and the temperature of the bottom flange at failure was 590°C. The corresponding upper flange temperature was 490 "C.

Test 2 was a repeat of test 1, but with unfilled voids above the top flange. The fire resistance was 61 minutes and the temperature of the bottom flange at failure was 580°C. The upper flange temperature was 63OoC, due to greater heat flow to the section through the unfilled voids.

Test 3 consisted of 23 mm of cementitious spray also with unfilled voids. The fire resistance was 74 minutes and the temperature of the bottom flange at failure was 550°C. The upper flange temperature was 720 "C, although at 61 minutes, the same temperature as in Test 2 was recorded.

Test 4 consisted of a thin film intumescent coating also with unfilled voids. The fire resistance was 48 minutes and the temperature of the


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Code Clause

8.2.4

bottom flange at failure was 570"C, which was approximately equal to the top flange temperature.

From these test results it was concluded that the critical temperature of a composite beam may be safely taken as 580°C for a load level of up to 0.6 when the voids above the top flange are filled, and 550°C for 60 minutes fire resistance when the voids above the top flange are unfilled. The difference in these critical temperatures leads to about 6% increase in protection thickness in the latter case, but importantly for economy, to the omission of the void fillers.

Critical temperatures of composite beams

The analysis of Section 2 . 2 , as confirmed by the test data reviewed EC4 c, 4.3.3.3 above, leads to the establishment of critical temperatures of composite beams. These critical temperatures may be used to determine the thickness of fire protection that is required for a given load ratio and fire resistance period.

The critical temperatures obtained by solving Equations (43) and (44) are presented in Table 32, as a function of the degree of shear connection implicit in normal design. However, EC4: Part 1.2 stted that the most conservative case of full shear connection (N/Nf = 1 .O, and K = 0.9) is used when obtaining the critical temperature of a composite beam.

It is apparent that the critical temperatures of composite beams in Table 32 are approximately 40°C below those of the equivalent non- composite beams in Table 10. In practice, this difference will lead to the requirement of about 7% increase in protection thickness for composite beams designed on the same basis. However, the thicknesses of traditional protection materials assessed in the UK for a critical temperature of 550°C will be unaffected, as the assessments are conservative at critical temperatures higher than 550"C(25).

Table 32 p 1.0+

Critical temperatures ("C) of composite beams

Load Level of Composite Beam at Fire Limit State

0.7

690 650 615 585 555 695 660 625 595 565 700 665 630 610 5 80 725 685 655 625 600

0.3 0.4 0.5 0.6

* steel beam + recommended value in EC4: Part 1.2 N/Nf = degree of shear connection at ultimate limit state


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Code Clause

8.3 Composite beams with partial encasement

8.3.1 General method

The use of concrete infills between the flanges of steel beams was E reviewed in Section 7.1. The same general principles apply when the beams are designed to act compositely with a concrete or composite slab by the provision of welded shear connectors. The full design method is presented in Annex E of EC4: Part 1.2.

The temperature distribution through the steel section, concrete encasement, and any reinforcement in the concrete is determined as for non-composite section. The concrete slab and shear connectors remain relatively cool. The strength reduction (or alternatively area reduction) of all the elements may be determined as a function of the temperatures that they experience at a given time.

In positive (or sagging) bending, no restriction is put on the effective breadth of slab acting with each beam, other than that in EC4: Part 1.1. In negative (or hogging) bending, the effective breadth is EC4 limited to 3 times the beam width, due to the influence of cracking of the slab in this zone. Deflection of the composite slab between the beams also may reduce the bending resistance of the composite beam slightly.

cl. 4.3.4.5(1)

The moment resistance of the composite section is obtained by first equating tension and compression vertically through the heat affected section, as in Section 7.1. The plastic neutral axis is normally found to lie within the slab, and therefore the effect of the concrete encasement in tension can be neglected.

As noted previously for non-composite sections, it is usually necessary to introduce two large diameter reinforcing bars within the concrete encasement in order to achieve more than 60 minutes fire resistance.

8.3 .2 Tabular methods

Design tables are presented in EC4: Part 1.2 to enable quick selection of beam sizes and reinforcement ratios. Designers are encouraged to EC4 Tab,e 4,

use these tables before turning to more precise methods. This table is reproduced in Table 33, showing also the modifications proposed in the UK NAD for UC and UB sections. The EC4 values are unconservative in some cases for higher load levels. UK NAD to

EC4 Part 1.2

The minimum axis distance to the bars for use of Table 33 is given in Table 34, which is taken directly from EC4: Part 1.2. Narrower sections may be used provided the required axis distance can be achieved.


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Code Clause

Various restrictions are placed on the use of the tabular method which are not unduly restrictive. Void fillers are required above the top flange if composite slabs are used. The reinforcement required may be reduced proportionately if S275 steel is used. In practice, reinforcement ratios greater than 0.6 indicate that this technique is not appropriate in relation to other methods of fire protection.

The model of EC4: Part 1.2 Annex E may be applied to the range of partially encased UB sections with standard reinforcement EC4 Table 4.2 arrangements. Maximum load levels for these composite beams are presented in Table 35. These values may be compared to the equivalent values for non-composite sections in Table 29. It is apparent that the maximum load levels in Table 35 are consistently higher. Table 35 shows that modest sizes of bar reinforcement achieve the required fire resistance and load levels.

Detailing rules for provision of shear links are presented in Section 9.3.

Table 33

Load Level

q 2 0.3

uc UB (hr 1.5b) UB (hr2b)

q 2 0.5 uc UB (hr 1 3 ) UB (hr2b)

q > 0.7 uc UB (hr 1 Sb) UB (hr2b)

~

Minimum cross-sectional dimensions and reinforcement in partially encased composite beams (modified from Table 4.1 of EC4: Part 1.2)

Fire Resistance (mins) I

The reinforcement is expressed as a ratio of the cross-sectional area of one flange and is placed adjacent to the bottom flange at the axis distance in Table 35.

The minimum beam width is given as min. b The minimum flange reinforcement is given as min. p EC4 values increased by 0.1 (*) or 0.2 (+) All values for S355 steel

EC4


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Code Clause

Table 34 Minimum axis distance to reinforcing bars in partially Table 4.2 encased composite beams

Minimum Fire Resistance (mins) Minimum section width

(mm) (mm) axis distance

R60 R180 R120 R90

170 120 100 U1

U 2

60 55 40 U 2

- 12 100 80 U1 200 60 45

U 2 35 50 60

60 60 45 25 U 2

80 70 50 40 U1 > 300

60

- -

250 120 90 75 60 U1

u1 = vertical distance above top of bottom flange u2 = horizontal distance from tip of flange

Table 35 Maximum load levels for partially encased composite

i beams (computer analysis based on EC4 Annex E)

I

Section Range

R60 0 bars

203 X 133 UB 0.55

305 X 127 0.59 305 X 165

0.60 356 X 177 0.60 356 X 127 0.56

406 X 178 0.62 457 X 152 0.64 457 X 191 0.64 533 x 200 0.67 610 x 229 0.68 610 X 305

0.69 686 X 254 0.64

0.71 762 X 267

254 x 146 0.55

406 x 140 0.61

Slab depth = 120 mm

Fire Resistance (min)

R90 2 X 20 mm dia

0.46 0.48 0.52 0.52 0.54 0.55 0.58 0.58 0.59 0.59 0.62 0.62 0.55 0.63 0.64

R120 2 X 25 mm dia axis < 60 mm

0.40 axis < 60 mm

0.45 axis < 60 mm

0.49 0.50 0.53 0.53 0.55 0.58 0.59 0.52 0.60 0.61

7

EC4 cl 4.3.1

All load levels are minimum for the section range in S355 steel Axis distance to reinforcing bars = 50 mm for R90 and 60 mm for R120


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9

Code Clause

COMPOSITE DECK SLABS

Composite deck slabs comprise steel decking and an in-situ concrete slab. The steel decking supports the concrete and other loads during construction and acts compositely with the concrete due to embossments in its profile. Design of composite slabs is covered by Eurocode 4 Part l . 1 which follows the same approach as BS 5950: Part 4 (1993 revision)(37). Both codes broadly adopt the same empirical rules based on tests.

The range of decking depths available in the UK is 45 to 80 mm, leading to slab depths of 100 to 170 mm, depending on the type of concrete used and the fire resistance requirements. Spans of composite deck slabs range from 2.5 to 4 m. Deep deck composite slabs are used in slim floor construction. In this case, the depth of the decking is 210 mm and the slab depth is typically 270 to 320 mm.

9.1 Minimum slab depths

The fire resistance criteria of insulation, integrity and strength (or 1x4 4.3.1.1

stability), as noted in Section 2.1, should be satisfied. The ‘insulation’ criterion determines the minimum depth of a composite slab. ‘Integrity’ is achieved largely through the continuous steel decking, and ‘strength’ is dependent on the provision of reinforcement (although R30 can be achieved in unreinforced slabs).

The minimum slab depth limits in EC4 Part 1.2 have evolved in recent years as a result of the greater amount of test data for a range of modern deck shapes and concrete types. The limits are compatible with concrete practice, as defined in BS 8110 Part 2(14) and Eurocode 2 Part 15’).

In EC4: Part 1.2, the minimum slab depth is expressed in terms of 1 x 4 cl 4.3.1.2 the ‘average’ slab depth as defined by two empirical formulae (see Figure 22(a)). These average depths are consistent with the values for solid concrete slabs in EC2 Part 1.2.

EC4 Table 4.8

The minimum depths may be reduced to 90% of the tabulated values EC4 in EC4: Part 1.2 when lightweight concrete is used. Lightweight concrete is defined as having a dry density of between 1600 to 2000 kg/m3, which is consistent with the use of Lytag structural concrete in the UK.

cl 4.3.1.2 (6)

In BS 5950: Part 8, the minimum slab depth is expressed in terms of BS 5950

the total slab depth for decks with a re-entrant shape, and the slab cl 4.9.2.2 EC4 Fig 4.1

depth over the deck for decks with a trapezoidal shape. A differential of approximately 10 mm is made between the minimum depths of


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Code Clause

slabs comprising light and normal weight concrete, reflecting the greater insulating effect of lightweight concrete. These two forms of decking are illustrated in Figure 21.

,,Mesh reinforcement

Minimum depth^\ . . . . . . . . . . . d' . . . . . . .

(to BS 5950: ' 1 Average depth tor Part 8) V . '

. . . . . . .

. . insulation [-."L, . : ' . .f--"--, ' . (to EC4: Part 1.2)

.'Bar (not needed for most cases)

(a) Trapezoidal decks in composite slabs

. . . .

Minimum depth ' . , . . ,

(to BS 5950: Part E)

' Average depth for

(to EC4: Part 1.2)

. . . . . . . . . . . . . . . . . . . . . ~ insulation . . . . .

. . . . . . . . .

V . . .

(b) Re-entrant decks in composite slabs

Figure 21 Composite deck slabs using two deck shapes

The insulation criterion is important because it is often found that it controls the design of composite slabs, and determines the weight of concrete to be supported. Lightweight concrete is beneficial in this respect.

In order to compare EC4: Part 1.2 and BS 5950: Part 8 directly, it is appropriate to use a typical deck depth of 50 mm, which for the trapezoidal profile is assumed to have a symmetric shape. The results are presented in Table 36. For adequate structural performance as a composite slab, the minimum depth of concrete over the deck is 40 mm, which will control the slab depths in most of the R30 cases.

It is apparent for Figure 22 and Table 36 that there are some BS 5950

significant differences (10 to 15 mm) in minimum slab depths, the Tables 13 & 14

Eurocode values generally being smaller than those in BS 5950: Part 8, which is known to be conservative. However, in most practical cases, the minimum slab depth is more influenced by moment resistance than insulation in fire conditions, which leads to greater slab depths in practice than given by Table 36.


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Code Clause

L N

+ 150

BS 5950 Part 8

EC4 Part 1.2

Data for t2=C3

50 ' I l I I I I I I I I 0 20 40 60 80 100

Deck height, h, (mm)

Figure 22 Comparison of BS 5950: Part 8 and EC4: Part 1.2 in terms of insulation depth of composite slabs (R90 case)


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Code Clause

9.2 9.2.1

112

Additional screeds may also be included in the minimum slab depth EC4 for insulation purposes, but the depth of screed considered to contribute to the insulation depth should not be taken as more than 20 mm.

cl 4.3.1.2(5)

Table 36 Comparison of minimum overall depths (in mm) Of

composite slabs to satisfy the insulation criterion in fire

120

180 165 200 180 170 150 185 170 240

160 150 175 165 145 135 160 150 180

130 130 145 145 115 115 125 125

Notes: Data for 50 mm deck profile NWC = Normal weight concrete LWC = Light weight concrete

The influence of deck profile height on the minimum slab depth for insulation purposes is presented in Figure 22(b) for the case of a slab with 90 minutes fire resistance. It is apparent that the different definition of slab depth in the two codes, leads to overall depths that differ by up to 25 mm over the range of trapezoidal deck heights and shapes used in practice. The difference is much less for slabs with re-entrant decks.

However, the EC4 definition of average slab depth is more logical 1 x 4 Table 4.8 for composite slabs with closely spaced ribs (< 300 mm apart) as tests and thermal analyses have shown that the ribs absorb heat from the adjacent thinner slab. It is, therefore, considered that EC4: Part 1.2 may be used with confidence for standard composite slabs. The exception to this rule is for deep decks with widely spaced ribs, in which case it is the depth over the deck profile that is relevant for insulation purposes.

Bending resistance

Plastic hinge analysis

The bending resistance of a composite slab in fire conditions depends on a combination of the sagging (positive) and hogging (negative) moment resistances of the section. Fire tests on continuous


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Code Clause

composite slabs have shown(37) that the plastic resistances of the section can be developed, and that the failure load at the fire limit state, Wfi, may be determined from the expressions:

9.2.2

or approximately :

internal span

where:

M , = sagging moment resistance of slab at fire limit state

MH = hogging moment resistance of slab at fire limit state

L = clear span between supports.

(= Mfi,Rd+ in EC4: Part 1.2)

( = Mfi, ~ d - )

Partial factors at the fire limit state are set to unity and are not presented in the above equations.

Clearly, the end span conditions will determine the design of the slab. The coefficient of 0.45 in Equation (47) is approximate, but assumes that MH = M,, and avoids iteration to determine Wfi.

The loads at the fire limit state include self weight, dead and imposed loads with the appropriate partial safety factors (see Section 2.2).

Sagging moment resistance

The approach of EC4: Part 1.2 to calculating the sagging moment EC4 cl 4 .3 .1 .4

resistance of a composite slab is empirical, and assumes that additional reinforcing bars are placed within the troughs of the deck. The effective heating rate of the bars is determined from the expression from the parameter, z, as follows:

where u l , u2 and u3 are the axis distances (in mm) of the reinforcing bars to the web and bottom of the steel deck, defined as in Figure 23 (EC4: Part 1.2, Figure 4.2).


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Code Clause

,,,,,/ Reinforcing bar '\,

\

sheet

Figure 23 Definition of dimensions to reinforcing bars in composite slabs

The temperature of the reinforcing bars are then determined from Table 37 as a function of z and the relevant fire resistance period. In using this equation, u3 2 35 mm and u1 or u2 2 50 mm.

Comparisons with the temperatures obtained from BS 5950: Part 8 are presented in Table 38. Generally, the results are similar for practical design cases. Equation (49) is also reasonably accurace for values of u1 to 15 below the EC4: Part 1.2 limits. No guidance is given for the use of lightweight concrete although it would be reasonable to take a 10% reduction in temperature in this case to account for its greater insulating effect, and to be consistent with BS 5950: Part 8 and with other clauses in EC4: Part 1.2.

Table 37 Temperature of reinforcement in EC4: Part 1.2 as a function of z

Fire Resistance Temperature of reinforcing bar ("C) (mm)

60

90

e, = 1175 - 350 c 8 1 0 ~ for ( z S 3,3)

e, = 1285 - 350 c 8 8 0 ~ for ( z S 3,6)

120 8, = 1370 - 350 z S 930°C for ( z 5 3,8)

180

240

e, = 1490 - 350 1000oc for ( z S 4,O)

e, = 1575 - 350 c 10500c for ( z S 4,2)

BS 5950 cl 4.9.2.1

EC4 Table 4.9


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Code Clause

Table 38 Comparison of reinforcement temperatures in composite slabs according to EC4: Part l .2 and BS 5950: Part 8

Geometrical EC4: Part 1.2 temperatures BS 5950: Pt 8 temperatures Darameters for fire resistances of: for fire resistances of:

Table 38 Comparison of reinforcement temperatures in composite slabs according to EC4: Part l .2 and BS 5950: Part 8

35 76 76 298 409 493 375 485 560

Note: BS 5950: Part 8 temperatures are based on flat slab temperatures *These values are below the limit of Equation (49)

The moment resistance of the composite slab is then determined taking account of the strength reduction of the reinforcement as a function of its temperature using the general material properties presented in Section 3. The strength reduction of the concrete in compression can be ignored in this case.

In EC4: Part 1.2, no account is taken of any strength retention of the steel decking, although tests have shown that some modest tensile strength remains, particularly for deck profiles which have a re-entrant portion. Table 39 presents some typical values of the strength retention of the steel decking, which may have a significant effect on the sagging resistance of lightly reinforced slabs. These values have been used in calibrating fire tests on composite slabs, which perform much better than given by calculation to the EC4 method.

Table 39 Percentage strength retention of steel decking in fire conditions I% of steel strength)

Fire resistance (mins) Deck type

60 > 120 120 90

Re-entrant decks ’

0 2.5 5 .O 7.5 Trapezoidal decks with

0 1 .o 3 .O 5.0 Plain trapezoidal decks

0 3.5 7.5 12.5

re-entrant portions


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9.2.3 Hogging moment resistance

The approach to calculating the hogging moment resistance of a composite slab is based on tabulated temperatures as a function of the depth into the concrete from the exposed surface. The approaches of EC4: Part 1.2 and BS 5950: Part 8 are the same in this respect. The temperature distribution is based on EC2: Part 1.2 and is presented in Table 40 for different fire resistance periods. In principle, the values in BS 5950: Part 8 and in BS 81 10 Part 2 are similar except that there are differences of up to 50°C for longer fire resistance periods. Significant differences are noted in Table 40.

Table 40 Temperatures in a flat slab at depth y from the exposed surface as given in EC4: Part 1.2

Depth mm

I

I I

L 100 I 60

60

705 642 (650)

581 525 (530)

469 42 1

374

327 , ;;; Data for normal weight concrete

Values in brackets are those in BS 5950: Part 8 that differ significantly from EC4: Part 1.2

The reduced compressive strength of the layers of concrete in the ribs may be determined incrementally (every 10 mm or so). The neutral axis depth is found by equating the tensile resistance of the reinforcing bars or mesh in the hogging moment region to the compressive resistance of the concrete. The hogging moment resistance is then determined directly by taking moments about the neutral axis position. Often the hogging moment resistance exceeds the sagging moment resistance when only mesh reinforcement is used.

Code Clause

EC4 cl 4.3.1.5 BS 5950

cl 4.9.2.1

EC4 Table 4.10 BS 5950 Table 12


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Code Clause

The temperatures in lightweight concrete slabs may be taken as 90% of the tabulated values. This approach is consistent with BS 5950: Part 8.

An additional requirement of BS 5950: Part 8 is that the reinforcement should have adequate ductility (elongation of failure) in order that it does not fracture prematurely at the large strains experienced in fire. No comparable requirement exists in EC2 and 4 Parts 1.2 and hence it is proposed that all reinforcing bars or mesh used in these fire engineering calculations should be specified as having a 5 % minimum elongation at failure (corresponding to grade S500 B in EN 10080(13)).

9.2.4 Fire test data

A considerable number of fire tests have been carried out on composite slabs in the UK since 1985. These tests are summarised in Table 41. Initially the concern was to develop general design recommendations and simplified load tables for up to 90 minutes fire resistance. Subsequently, fire tests have been carried out to justify extending these simplified load tables up to 120 minutes fire resistance. Although not specifically stated in EC4: Part l .2 , these tables effectively have a ‘deemed to satisfy’ status and can continue to be used in UK designs. This design guidance is represented in Tables 42 and 43 (taken from reference (38)). Span and depth limitations are noted in these tables.

Because of differences in furnace characteristics in continental Europe, it is not possible to use these simplified tables outside the UK. Nevertheless in Germany, the use of composite slabs comprising the re-entrant deck shape is widely accepted with light mesh reinforcement for up to 120 minutes fire resistance.

9.2.5 Protected composite slabs

It is possible to enhance the fire resistance of composite slabs by applying sprayed fire protection or protective ceilings to the underside. A simple method of ensuring that adequate fire resistance is provided is by limiting the temperature of the steel deck. A requirement of EC4: Part 1.2 is that the maximum critical temperature of 350°C should be used in determining the thickness of protection. Given the extensive amount of test data on unprotected steel decks, this limit is unnecessarily restrictive and a higher limit of 450°C would seem appropriate.

EC4 cl 4.3.2

This observation is confirmed by two recent tests on composite slabs with 18 mm of sprayed vermiculite cement on the soffit. These tests achieved over 4 hours fire resistance, The deck temperatures reached over 500°C at the end of the tests.


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L Table 41 Summary of recent fire tests in the UK on unprotected composite floors

DECK PROFILE

Robertson QL59 Robertson QL59 Robertson QL59 Holorib PMF CF46 Holorib PMF CF46 Robertson QL59 Metecno A55 Holorib (UK) Ribdeck 60 Ribdeck 60 Alphalok SMD R51 Quikspan Q5 1 Quikspan Q60 Multideck 60 Multideck 80 Holorib (protected) Ribdeck 60 (protected)

CONCRETE TYPE

~~ ~

LWC LWC LWC LWC LWC LWC NWC NWC NWC LWC LWC LWC LWC NWC NWC NWC NWC NWC NWC NWC

SLAB DEPTH

(mm>

130 130 130 120 110 100 135 140 140 150 1 40 140 130 140 140 150 150 150 120 130

SPAN (m)

3 .Os 3 .Oc* 3 .Oc 3.0c* 3 .Oc 3 .Oc 3 .Oc 3.6c* 3.6c* 3.0c* 3.0c* 3.0c* 3.6c* 3.0c* 3.0c* 3.0c* 3.6c* 4.0c* 3.0c* 3.0c*

IMPOSED LOAD (kN/m2)

6.7 6.7 6.7 6.7 5.25 5.75 6.75 6.7 6.7

10.0 5.6 8.5 6.7 6.7 5.0 5.0 6.7 6.7 6.7 6.7

~

REINFORCEMENT

A142 mesh A142 mesh A142 mesh A142 mesh Y5 @ 225 as mesh Y5 @ 150 as mesh Y5 @ 225 as mesh A193 mesh A193 mesh A193 mesh A193 mesh A252 mesh A252 mesh A 193 mesh A142 mesh A142 mesh A252 mesh A252 mesh A142 mesh A142 mesh

T SURFACE TEMP. ~~

I (

AFTER lh

73 70 95 60 110 90 85 66 65 45 64 56 92 96 52 79 74 69 35 40

'1 AFTER

1% h

100 110 100 135 120 95 98 95 61 93 77 110 102 78 97 89 87 65 70

The tests are in chronological sequence from July 1983 until January 1993 Surface temperatures are the average values on the unexposed surface failed prematurely because of the loss of protection to beams

S = simply supported c = continuous slab test * tests on long spanhhort span configuration

TEST PERIOD

(min)

60 105 90 90 101 87 120 90 90 120 136 149 128 135 126 122 135 92

240 240

~~~

TEST REF.

CIRIA 1 CIRIA 2 CIRIA 3 CIRIA 4 FRS-BS 1 FRS-BS2 FRS-BS3 CIRIA 5 CIRIA 6

R.LEES 1 R.LEES 2 R.LEES 3 ALPHA 1

SMD 1 QUIK 1 QUIK 2

WARD 1 WARD 2 PROT 1 PROT 2


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Code Clause

Table 42 Simplified design for composite slabs with trapezoidal decks

Maximum

(min Resista Span (m)

Fire

2.7 60

3.0 60 3.0

120 3.0 90

3.6 60 3.6 90 3.6 120

Minimum dimensions nce

t (mm)

NWC LWC

Slab depth (mm)

0.8 130 120

0.9 130 120 0.9

155 140 0.9 140 130

1 .o 130 120 1.2 140 130 1.2 155 140

Minimum Mesh Size

A 142

A142 A 142 A193

A193 A193 A252

For deck profiles > 60 mm, add increased deck depth to minimum slab depths in this table.

NWC Normal weight concrete LWC Lightweight concrete t = Minimum steel thickness

Table 43

Maximum Span (m)

2.5 2.5

3 .O 3.0 3 .O

3.6 3.6 3.6

Simplified design for composite slabs with re-entrant decks

I I Fire I Minimum dimensions I M ' . mmum

Resistance Mesh Size (min) t (mm) Slab depth (mm)

I I NWC LWC I 60 90

60 90 120

60 90 120

0.8 0.8

0.9 0.9 0.9

1 .o l .2 1.2

100 100 110 105

A142 A 142

120 110 A 142 130 120 A142 140 130 A193

125 120 A193 135 125 A193 145 130 A252

Re-entrant decks of depth S 50 mm.


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Code Clause

10 CONCRETE FILLED HOLLOW SECTION COLUMNS

IO. 1 Introduction Concrete-filled hollow sections are the most practical form of composite column. These sections may be of circular, square or rectangular form. The concrete may be plain (i.e. unreinforced), bar reinforced, or fibre reinforced (using chopped steel fibres). The merits of these different forms of infill depend largely on the fire resistance required. The use of fibre reinforcement is not covered by EC4: Part 1.2.

Although composite columns are intended to be designed to the combined requirements of EC4: Parts 1.1 and 1.2 (i.e. they are designed as composite under normal and fire conditions), it is also acceptable to design the columns solely as steel members (i.e. ignoring composite action under normal conditions), and to take advantage of the composite action in fire conditions. This second approach effectively under-designs the column under normal conditions, leading to the use of heavier steel sections than for fully composite columns.

The different forms of concrete filled hollow sections are illustrated in Figure 24. Circular or square hollow sections are used as columns. Reinforcement may be installed as 4 or 6 individual bars in larger sections, or single bars in smaller sections.

Concrete filled hollow sections do not rely greatly on the strength of the outer steel section in fire conditions because of the high temperatures and loss of strength experienced by the steel shell. EC4 cl 4.2.3.4 Therefore, the concrete core provides most of the necessary BS 5950 c, 4.6 compression resistance. The ratio of the ‘cold’ resistances of the steel and concrete components is therefore important in determining the potential reduced compression resistance, and fire resistance of composite columns in fire conditions. Larger diameter columns with relatively thin steel shells may be expected to retain more of their compression resistance in fire than smaller columns with thicker shells.

The steel shell also provides an important function in confining the concrete so that higher compressive strengths can be achieved than in reinforced concrete design. This triaxial confinement is much reduced in fire conditions because of the differential expansion of the hotter steel shell relative to the concrete core. Steel fibres mixed into the concrete or bar reinforcement help to avoid internal crushing of


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Code Clause

the core, and lead to higher fire resistances than in columns with plain concrete.

The approach of EC4: Part 1.2 follows that of normal column design to Part 1 . 1 , which is subtly different to BS 5950: Part Therefore, it is first necessary to describe the methods used for normal design of these columns under predominantly axial load.

EC4: Part 1.2 Annex G uses alternative material laws for the elevated temperature properties of steel, concrete and reinforcement. During the code-drafting period, it was not possible to consider the influence of using the general material laws in Section 3 of EC4: Part 1.2.

Unreinforced concrete-filled sections

Reinforced concrete-filled sections

Figure 24 Different forms of concrete filled sections

10.2 Normal design

The compressive resistance of a concrete filled section depends on the EC4 Part 1.1 sum of the resistances of the individual components. The design of cl 4.8.3.3

concrete filled columns in the UK has, up to now, been only covered by BS 5400: Part 5(39 1, which is based on the same general principles as EC4: Part 1 . 1 .


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Code Clause

It follows that the compressive resistance of a short column (not failing by buckling) is given by its squash load, which according to the two methods (and using the terminology of the two codes), is:

EC4: Part 1.1 :

BS 5400: Part 5 :

N U = 0.45fcu A , + 0.87fy As + 0.91 p y A

where fck = 0.8 f,,

In EC4: Part 1 .1 , the concrete strength is enhanced by triaxial effects and the steel strength is reduced due to hoop tension effects for columns with an effective slenderness, 1 < 0.5.

Slender columns fail by lateral buckling. This effect will generally only be significant when the height/minimum width of the column exceeds approximately 10.

The effective radius of gyration of the composite section involves EC4: Part l .1 calculating the contribution of all the components. The concrete may be converted into an equivalent steel section by dividing the concrete inertia and area by an appropriate modular ratio, a,. The effective inertia for short term loads is given by:

cl 4.8.3.5

(EOeff = (la + 0.8Zc/ae + l,) Ea (52)

where the subscripts to EZ refers to the stiffnesses of the steel, concrete and reinforcement respectively. The factor of 0.8 is empirical, and takes into account the non-linear stress-strain curve for concrete.

The modular ratio is the ratio of the elastic moduli of steel to concrete and is only used in determining elastic properties. The elastic modulus of concrete depends on the duration of loading, which causes creep in the concrete. In EC4: Part 1 .1 , the modular ratios for concrete are determined from tables of E,. In UK methods, an empirical formula is given for the elastic modulus of concrete. A modular ratio of 10 may be used conservatively in composite column design, although this value should be increased when the structure is subjected to predominantly long term loads.

Having determined the effective radius of gyration of the composite section (in steel units), the slenderness of the column may be


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Code Clause

calculated conventionally. This slenderness can then be used to determine the reduction factor, x , on the column resistance using the appropriate column buckling curve in Eurocode 3 Part 1 . 1 . Curve a is generally used for hollow sections, and curve c for I sections in normal design.

The reduction factor, x , is then used to multiply the compressive resistance of the whole composite section calculated in Equation (51) so that the buckling resistance of the column is given by:

The approach of BS 5400: Part 5(39), is semi-empirical and is based on tests. The EC4: Part 1 . 1 method gives higher compressive resistances for very slender columns than the BS method. Therefore, in the UK NAD for EC4: Part 1 .1 , its use is limited to composite columns with a slenderness ratio of less than 1 .0 , in the absence of test data outside this range. In practice, most composite columns in buildings will be within this slenderness range.

The treatment of combined moment and axial force in EC4: Part 1 .1 is based on first establishing the axial and moment resistance of the composite section independently. For slender columns, it is reasonable to consider a linear interaction between the combination of the effects of axial force and moment, but this approach is increasingly conservative for stocky columns. It is not appropriate here to review these methods and reference should be made to the SCI publication (30).

Design tables for composite columns to EC4: Part 1 .1 are also given in this SCI publication(30), which includes concrete filled and encased sections. The compressive resistance is presented as a function of the effective length of the column.

10.3 Fire design

10.3.1 Squash resistance

The compressive (or squash) resistance of a concrete filled section in z4 fire conditions follows the same basic approach as described in Section 10 .2 , except that reduction factors should be applied to all the components to take account of their reduced strength at elevated temperatures.

The compressive resistance of the section at elevated temperatures (using the same parameters as in Equation (50)) is given by:


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10.3.2

EC4 Table 4.10

Code Clause

PLf1 ' = f c k r1 + f y s A s r2 + f y A r3 (54)

The partial safety factors for concrete, steel, and reinforcement are EC4 Cl 4.3.6.1

set to unity at the fire limit state and are not presented in these equations. The reduction factors rl and r2 take account of the loss of strength due to heating of the concrete infill and bar reinforcement respecitively . The reduction factor r3 refers to the steel shell, whose temperature will be close to that of the fire. Nevertheless, the contribution of the steel shell to the resistance of the cross-section is important for thicker steel sections. The factors r1 to y3 are not presented explicitly in EC4: Part 1.2, but are used here to explain the principles I

Temperature distribution in the concrete core

For circular concrete-filled sections, analysis of the temperature distribution through the cross-section shows that the temperatures are greater than given by one-dimensional heat flow (such as through a wall). This is because the internal annular layers become progressively smaller and the heat flow into each layer increases. The same phenomenon is also observed for square sections, but in this case the corners of the section will be hotter than at the mid-sides of the section. Nevertheless, the average temperature of a layer closely approximates to that of an equivalent circular section.

The analysis of a range of section sizes show that simple multiplication factors may be applied to the temperatures obtained from basic one dimensional heat flow. The resulting temperatures then define the temperatures of the annular layers of the cross-section of a circular or square column. These multiplication factors are a function of the size of the section, and of the distance of the centre of the layer from the nearest exposed surface.

This simplified method is presented as an alternative to the general approach in EC4: Part 1.2 Annex G, and complies with the principles of EC4: Part l .2. This method uses various terms and values that are not given in EC4: Part 1.2, and is presented in the following sub-sections.

Influence of section size

The multiplication factors, C, for each annular layer in the concrete core are presented in Table 44 as a function of the section size. The width of these layers is taken as 20 mm for analysis purposes. The information in Table 44 is appropriate to both circular and square sections, as a reasonable approximation.

The multiplication factors are used to multiply the basic temperatures at a given depth in an infinitely wide section, which are presented in


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Code Clause

Table 4.10 of EC4: Part 1.2, and are reproduced in Table 45. These factors are shown to be relatively insensitive to the fire resistance period and are presented as independent of this parameter.

The reinforcing bars are located in the second or third layers at an axis distance of 30 to 50 mm depending on the fire resistance requirement. The temperature of the reinforcement may be calculated using the same multiplication factors as for the concrete. The results are presented in Table 45 for the relevant axis distance specified in Table 4.7 of EC4: Part 1.2.

For square sections, the main reinforcing bars are located in the corners of the section and will experience greater heating than given by their distance from the nearest surface. In this case the temperature of the bars should be calculated from an equivalent depth of half the axis distance, reflecting the fact that the bars are heated equally from two sides. It is found that the temperature of the bars at the appropriate axis distance specified in EC4: Part 1.2 is approximately 600 "C for square sections.

Influence of the steel shell

It is also found from thermal analysis of concrete filled sections that the steel shell influences the concrete temperatures, partly by shielding the concrete core, and partly by the heat capacity of the steel, which is greatly enhanced at temperatures of 700 to 750°C (see Figure 4). A further multiplication factor may be introduced that may be applied to all the previously calculated concrete and reinforcement temperatures to take account of the effects of the steel shell. This factor is dependent on the thickness of the steel shell, but its contribution reduces for longer fire exposure.

This further multiplication factor, C,, is also used to determine the temperature of the steel shell itself by multiplying by the fire temperature, determined from Equation (l), and given in Table 45. This factor is expressed as:

c, = (1 - 0.01 t) for I 60 minutes fire resistance (55)

c, = (1 - 0.005 t ) for 90 minutes fire resistance (56)

c, = 1 .o for 2 120 minutes fire resistance.

where t is the thickness of the steel shell (in mm).

The concrete core temperatures are then determined from the factor C, multiplied by the basic temperature in Table 45. The reinforcement temperatures may be determined by the same


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Code Clause

approach, using in this case the appropriate axis distance to the centre of the bars. Intermediate values may be interpolated to find the relevant value of C , .

The temperatures obtained from the ‘exact’ thermal analysis of a circular concrete filled section and the simplified approach based on these multiplication factors are presented in Figure 25 for one diameter of section at two fire resistance periods. The influence of steel thickness is apparent. It is demonstrated in this case that the temperature distribution is closely modelled by these C, and C , factors.

Strength reduction factors

The compressive (or squash) resistance of a concrete filled section can be determined from the relevant strength reduction factors of the various layers or elements at their respective temperatures. For the concrete core, the strength reduction factors for each progressive layer from the surface are multiplied by the cross-sectional area of the layer. The sum of the resistances of all the layers leads to the factor rl representing the average strength retention of the whole of the concrete cross-section relative to its compressive resistance ignoring partial factors.

The same analysis may be repeated for the reinforcing bars, knowing the temperature that they reach. The axis distance to the bars is selected so that they retain a significant proportion (-0.4) of their strength at the relevant fire resistance period (see Table 46). This is reflected in the factor, r,, which is determined from Table 6 for cold worked reinforcing bars.

The steel shell contributes only a modest amount of the compressive resistance of the column, as its strength reduction factor, r3, is often less than 0.1. However, for thicker steel sections at 30 or 60 minutes fire exposure, the effect of the steel shell can be significant, as it benefits also from the lower temperatures that it reaches in fire (see the factor C2). The strength reduction factor is determined from Table 5 for the relevant temperature of the steel shell.


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Code Clause

1,100

1,000

900

800 h

0, 700 2 2 600

500

400

300

200

l 0 0

0

c 3

al Q

l-

1,100

1,000

900

800

700 2 2 600

g 500 l-

400

300

200

100

0

h

3 c

al Q

Furnace temperature +

i \\ \

60 minutes fire resistance 244.5 diameter section

Thickness of steel shell

L -1 0 0 10 20 30 40 50 60 70 80 90 100

Distance from inside of steel shell (mm)

Furnace temperature

+

\

90 minutes fire resistance 244.5 diameter section

Thickness of steel shell

-10 0 10 20 30 40 50 60 70 80 90 100 Distance from inside of steel shell (mm)

Figure 25 Temperature variation in a circular concrete-filled section (60 and 90 minutes fire resistance)


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Code Clause

The maximum load level that a short column can resist in fire conditions is obtained by dividing the result from Equation (54) at the relevant fire resistance period by the normal squash resistance of the column obtained from Equation (50) using the appropriate partial factors. Alternatively, the normal compressive resistance of a concrete filled column may be obtained from the design tables in SCI publication(30).

Table 44 Multiplication factors C, to obtain temperatures of layers in concrete filled sections as a function of their size

Factors multiplied by the basic temperatures in Table 45

Table 45 Basic temperatures of infinitely large concrete sections exposed to fire on outer surface

Table 46 Approximate reinforcement temperatures and strength reductions in concrete filled sections

Fire Resistance (mins) Section Type Reinforcement Parameter 60 90 120 180

Circular Axis distance 5 10 550 580 600 Temperature 30 40 50 60

Strength reduction factor 0.64 0.53 0.46 0.40 Square

Take as o.4 for all caSeS Strength reduction factor 590 600 580 600 Temperature 30 40 50 60 Axis distance


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Code Clause

10.3.3 Fire design - slenderness effect

The effective slenderness of a concrete filled section also depends on EC4 cl 4.3.6.1

& Fig 4 .6 the stiffness of the various components, as influenced by their temperature. The elastic moduli of steel, reinforcement, and concrete may be determined from Tables 5, 6 and 7 based on the same temperatures used to calculate the squash resistance. The steel shell contributes to the stiffness of the cross-section and may be included, despite the high temperature that it experiences. The concrete contributes proportionally less to the stiffness of the cross-section than to its strength because of the high temperatures in the outer layers and the rapid decline of its elastic modulus with temperature. The same temperatures of the elements as previously, are used in these calculations.

The effective stiffness of the concrete filled section is given by:

where Ea,fi, Ec,fi and are the heat affected elastic moduli of the EC4 various materials. cl 4.3.6.1(5)

It is generally appropriate to use the initial tangent moduli of all materials for buckling calculations. However, EC4: Part 1.2 uses the secant modulus for concrete but no further modification factors, 4, are made to the stiffness of each material as was done for concrete encased sections. Equation (57) is appropriate to the use of the material laws in Annex G.

The effective slenderness ratio of a column in fire conditions is determined from:

where:

Np,,fi = the squash resistance of the cross-section (see Section 10.3.1).

Ncr,fi = the critical buckling load of the column given by:

The effective slenderness xfi is then used to determine the buckling EC4 reduction factor, x , for the column, considered to be an equivalent cl 4.3.6.1(2)


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Code Clause

steel section to EC3: Part 1 . 1 . EC4: Part 1.2 requires the use of buckling curve ‘c’ for all composite columns in fire conditions. It should be noted that buckling curve ‘a’ is generally used in the normal design of concrete filled columns. This difference in these buckling curves has a marked effect on the fire resistance of concrete-filled sections.

The reduction factor, x , is then applied as a multiple of the squash resistance of the whole cross-section at elevated temperatures (as in Equation (54)) to determine the design resistance of the column in fire.

The slenderness of a composite column in a multi-storey building is EC4 also influenced by its end fixity in fire conditions, due to the restraint of the parts of the column outside the compartment. In a rigorous analysis, EC4: Part 1.2, permits the use of an end fixity factor of 0.5 on the column effective length in fire design. The proposed UK NAD does not accept such a major reduction in the column slenderness, and rationalizes the design approach by taking the UK NAD to effective slenderness ratio of columns in braced frames as 0.7 of the value used for the same column in normal design.

cl 4.3.6.1(9)

EC4: Part 1.2

Unlike concrete encased sections, slenderness effects in fire have a disproportionate effect on concrete-filled sections. This is because the steel shell has a much greater effect on the stiffness of the cross-section than its compressive resistance. The maximum load level that a concrete-filled section can support is given by the ratio of its compressive resistance in fire conditions relative to its compressive resistance in normal design. The load level will be lower than the ratio of the squash resistances of the column in fire and normal design. The variation of load level with slenderness for a typical concrete-filled column is shown in Figure 26.

In order to take account of slenderness effects, it is necessary to define two geometrical properties of the column so that simple rules can be developed.

These are: L I 15d d 2 25t

where: L = column effective length in normal design d = minimum width (or diameter) of the column t = thickness of the steel shell.

It is found from these analyses that the maximum load level a column of these proportions can sustain may be determined by a factor of 0.8 times the ratio of the squush resistances, such that:


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Code Clause

S235 - C30137 - S400 2.5 'x Reinforcing, 8 bars, 25.0 mm axis distance

I 1600 -

-

l500 - -

1400 - -

1300 - -

g 1200 -

Y

0, -

a 3 1100

s B

- U a -

l- 1000

a

-

-

$ 9 0 0 - Q Q

-

5 800 -

M C

- .- B 700

B 500

z 6 0 0 -

-

-

a m - .- Q

C W - .- 01

-

400 -

-

300 - -

200 -

-

1 0 0 -

-

R60

R90

R120

0 I 2 3 4 Buckling Length at Elevated Temperature [m]

5

Figure 26 Variation of load level with slenderness of a square concrete-filled column (250 x 250 x 6.3 mm)


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Code Clause

o.8 Npl,fi,Rd

Npl , R d %,t 5

where the factor of 0.8 takes account of the effect of the increased effective slenderness of the column in fire conditions. The corresponding compressive resistance of a concrete-filled column in normal design may be obtained from the design tables in SCI Publication(30).

Alternatively, the column may be designed as non-composite in normal conditions, in which case its permitted load level in fire conditions is greatly increased, due to the reduction in N p l , R d .

10.3.4 BS 5950: Part 8 approval

The BS 5950: Part 8 method is subtly different from EC4: Part 1.2 in that an alternative definition of the load ratio is presented, as follows:

where K is a concrete core buckling factor, which is a function of the effective slenderness of the column (see Table 10 of BS 5950: Part 8).

The maximum values of q which can be sustained in fire conditions are tabulated in BS 5950: Part 8. The UK approach permits the use of unreinforced concrete-filled sections for 30 or 60 minutes fire resistance. Higher load ratios can be achieved if steel fibre mesh is used in the concrete.

10.3.5 Tabular approach

The simple tabular approach in EC4: Part 1.2 is relatively restrictive, even taking account of the different definitions of load ratio and load level. The design table (Table 4.7 of EC4 Part 1.2) presents the minimum diameter or width of the column required to achieve up to 180 minutes fire resistance. The minimum percentage of reinforcement in the design table ranges from 1.5 to 6%, the higher amount being required for cases exceeding 60 minutes fire resistance. In practice, use of 6% reinforcement and its additional shear links may cause difficulty in placing and proper compaction of the concrete in the column.

An alternative approach presented in the following tables uses the previously described analysis method based on first principles and using the material laws given in EC4: Part 1.2.

Bs 5950 cl 4.6.2.2 &

Table 10

EC4 Table 4.1

EC4 cl 3.2.1 cl 3.2.2 cl 3.2.3


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Code Clause

Using this analysis, the maximum load levels that may be applied to composite concrete filled columns at the fire limit state are presented in Table 47 for unreinforced sections and Table 48 for reinforced sections (with 2% reinforcement). The load level is expressed as a proportion of the ‘cold’ or normal design resistance of the composite column which includes also the influence of slenderness (see reference (30)). In these tables the reinforcement is ignored in calculating the design resistance of the composite column.

It is apparent that unreinforced concrete-filled columns can be designed efficiently for 60 minutes fire resistance, and larger columns with 2% reinforcement can be designed for 90 minutes fire resistance. Large, heavily reinforced columns can achieve 120 minutes fire resistance, but it is apparent that this form of construction is less economic for long fire resistance periods.

Table 47 Maximum load levels for unreinforced concrete-filled hollow sections designed as composite columns

Fire Resistance Section diameter (mm) (mins)

200 500 400 300

30 0.34

0.25 0.23 0.20 0.16 90 0.34 0.31 0.28 0.24 60 0.44 0.42 0.38

Notes: 1. Steel shell is assumed to be S275 steel and with d/t 2 25 2. Column effective length in normal design, L S 15 d

Table 48 Maximum load levels for reinforced concrete-filled hollow sections designed as composite columns

Fire Resistance Section diameter (mm) (mins)

200 500 400 300

30 0.47

0.26 0.23 0.19 0.14 120 0.35 0.32 0.27 0.22 90 0.47 0.43 0.38 0.32 60 0.59 0.55 0.51

Notes: 1 . Reinforcement is ignored in normal design 2. Reinforcement percentage 2 2 % of concrete area 3. Steel shell and column slenderness as in Table 48

It is expected that designs based on the general analysis method in EC4: Part 1.2 Annex G will lead to use of less reinforcement than the simple tabular method. Therefore, it is hoped to improve the guidance given in these tables during the ENV period of this

EC4 Annex G


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Code Clause

Eurocode in order not to adversely affect the economy of this form of construction.

It should also be noted that Annex G uses different material laws from those presented in Section 3.2 of EC4: Part 1.2. This was done to be consistent with previous CIDECT research and with ECCS TN 55(28). It is hoped to modify Annex G to be consistent with the general material laws during the ENV period.

The maximum load levels of concrete-filled sections using readily available circular sections in the UK is presented in Table 49. This table uses three standard bar diameters, increasing with section size and fire resistance. The table is based on the load level for the composite section in which the reinforcement is ignored in normal design.

The load levels in Table 49 may be increased proportionately if the column is designed as a steel (rather than composite) section at the fire limit state.

Special measures are required for square sections because of the greater heating of the bars located in the corners of the section. In order for the general design tables to be used for square sections, the UK NAD to EC4: Part 1.2 recommends that the axis distance to the bar reinforcement is increased by 10 mm for R60, and 20 mm for R90 or R120 classes in this case.

The existing Table 4.7 of EC4: Part 1.2 has been modified to conform better to the general analysis method for all shapes of concrete-filled columns. The new Table 50 proposed for the UK NAD gives lower percentages of reinforcement in a number of fire resistance classes, and to slightly modified minimum dimensions in some cases. Again, it is important to note in deriving this table that the reinforcement is ignored in normal design.

EC4 Table 4.7 UK Nad

The restrictions to the use of Table 50 include the use of S275 steel (unlike S235 in Table 4.7 of EC4: Part 1.2) and steel sections with a diameter/thickness ratio 2 25. Furthermore, EC4 Part 1.2 Table 4.7 states that reinforcement exceeding 3 % of the concrete area should not be included in the normal design of the column.


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Table 49 Maximum load levels of concrete-filled circular hollow sections designed as composite columns((with L/d S

15 and d/t 2 25, calculated using material laws in EC4: Part 1.21

Section Size

168 X 5.0 X 6.3

194 X 5.0 X 6.3 X 8.0

219 X 5.0 X 6.3 X 8.0

244 X 6.3 x 8.0 x 10.0

273 X 6.3 x 8.0 x 10.0

323 x 6.3 x 8.0 x 10.0 X 12.5

x 10.0 X 12.5

406 X 10.0 X 12.5 x 16.0

457 x 10.0 X 12.5 X 26.0

508 x 12.5 X 16.0 x 20.0

356 X 8.0

T Fire Resistance

Load Level

0.36 0.34 0.40 0.38 0.35 0.45 0.41 0.37 0.44 0.40 0.37 0.47 0.43 0.40 0.56 0.52 0.48 0.45 0.54 0.50 0.47 0.53 0.50 0.47 0.56 0.53 0.49 0.59 0.55 0.51

Bar Size

4 X 16 4 X 16 4 X 16 4 X 16 4 X 16

4 X 16 4 X 16 4 X 16 4 X 16 4 X 16 4 X 16 4 X 16 4 X 16 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25

4 x 16

4 x 25

t ~~ ____

90 mins

Load Level

0.21 0.20 0.26 0.25 0.23 0.32 0.30 0.27 0.34 0.30 0.27 0.37 0.34 0.30 0.48 0.44 0.40 0.36 0.46 0.41 0.37 0.45 0.40 0.36 0.54 0.49 0.44 0.51 0.45 0.41

Bar Size

4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 x 20 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 25 4 X 32 4 X 32 4 X 32 4 X 32 4 X 32 4 x 32

C30137 concrete and S275 steel Axis distances as in Table 46


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Code Clause

Table 50 Minimum cross-sectional, dimensions (in mm) and EC4 reinforcement of composite concrete filled sections (modified from Table 4.7 of EC4: Part l . 2)

Table 4.7 UK NAD

~ ~~ ~

Load Level qfi = 0.3

Minimum dimensions

Minimum reinforcement % Minimum axis distance


Minimum dimensions



Minimum dimensions


Increased axis distance to bars for sauare sections

T Fire

- O(1.5) l (200)

(260) (450)

+ l 0 + l 0

sistance (min)

R90 R180 R120

300 400 350 (220) (260)

40 60 50 6 6

400

3 (6)

500 450

60 50 40 6 6

450

50 40 6 6 (-) (550) 550

+20 +20 +30

1

Total reinforcement is expressed as a ratio of the cross-sectional area of the concrete core.

EC4 values in brackets (where different) These values are appropriate for S275 steel Tabulated values are for circular sections

10.3.6 Fire protected concrete filled sections

It is possible to enhance the fire resistance of concrete filled sections by applying fire protection to the outer surface of the column. This protection has the effect of reducing the temperature of the outer shell so that it retains more of its strength in fire. No direct guidance is given for determining the thickness of fire protection in this case in EC4: Part 1.2, although the general principles described for bare steel sections may be used. It may be more economic to apply fire protection to the surface of the section than to use a high percentage of bar reinforcement in the concrete core.


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Code Clause

BS 5950: Part 8 presents a method of determining the thickness of BS 5950 fire protection required for a concrete filled section as a reduction factor on the thickness required for an unfilled section of the same section factor. The reductions are relatively modest and range from 10% for a section factor of 100 r n - l to 45% for a section factor of 250 m-'. In the absence of further guidance, the BS 5950: Part 8 method(6) may be used for spray, board and intumescent protection.

cl 4.6.3


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Code Clause

11 DETAILING REQUIREMENTS EC4 Cl 5.3.2.5.4

Various recommendations are presented for detailing in EC4: Part 1.2. These requirements fall into the categories of

minimum cover to the steel section minimum axis distance to the main bar reinforcement minimum percentage and size of reinforcement minimum size and maximum spacing of links details at connections.

The minimum axis distance to the main reinforcement is generally presented in the simplified tables. The concrete cover (axis distance minus half bar diameter) should not be less than 20 mm, and not more than 50 mm, the latter being required to control spalling in fire. EC4 cl 5.1(5)

The detailing requirements for various forms of construction are considered below.

1 1.1 Partially encased sections

Where the concrete provides both a structural and insulating function, EC4 cl 5.3.1 additional shear links or stud shear connectors are required. The links should be of minimum diameter of 6 mm and spaced at not more than 250 mm along the member. The cover to the links should not exceed 35 mm. Three techniques for achieving the necessary shear transfer are permitted:

welded shear links, which a weld length not less than 4 X bar diameter. shear links and bars passed through holes in the beam web. shear links and 12 mm diameter welded stud shear connectors (or equivalent).

In the latter cases, the distance between the bars or studs should not exceed 400 mm along the member, or 200 mm vertically within the beam depth. The same recommendations apply for columns and beams. These detailing requirements are summarised in Figure 27.

l 1.2 Fully encased sections

Where the concrete provides a structural and insulating function, four additional longitudinal bars at the corners of the section confined by shear links are required. The bars should be of minimum diameter of 12 mm, and the links of minimum diameter of 6 mm and spaced


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Code Clause

at more than 250 mm along the member. No additional methods of shear connection are necessary as the shear links are assumed to prevent bursting of the bars and spalling of the concrete.

Where the concrete provides only an insulating function, it is permitted to use unreinforced concrete except for a fabric mesh around the section consisting of 4 mm bars spaced at not more than 250 mm along the section.

1 1.3 Concrete filled sections

Special requirements are made for concrete filled sections in order to ensure the integrity of the construction. Most importantly, it is necessary to form a 20 mm diameter hole at the top and bottom of each storey height of column (but not further apart than 5 m). These holes permit the steam from any entrapped moisture to be ejected, which avoids any tendency for damage to the outer steel shell. The corresponding minimum hole diameters are 12 mm in BS 5950: Part 8.

EC4 cl 5.3.7

Where additional bars are used in the core, the links containing the bars should be spaced at not more than 15 X diameter of the longitudinal bars. No additional form of shear connection is required in concrete filled sections.

11.4 Beam to column connections

Various forms of connections between beams and columns may be EC4 cl 5.4 used. The basic design requirement is that the fire protection afforded to the various elements of the connection (angles, plates, welds and bolts) should be at least equivalent to the minimum protection of the connected members. This ensures that the connections will not experience temperatures significantly different from the adjacent sections.

Although the high temperature properties of bolts and welds are no better than those of structural steel, the high factor of safety of these elements used in normal design and the thermal mass of the connections, ensures that they are not the ‘weak link’ in fire. This is borne out by the observations of fire damaged buildings.

Connections between partially encased concrete members require EC4 particular consideration, and typical design solutions are illustrated in Figure 28. Special detailing measures are made where part of the connection may be exposed, such as shear blocks used to support beams.

cl 5.4.1(6)


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For fire design, these blocks should be properly anchored to the concrete of a partially encased, or concrete filled section, by using welded shear connectors. If this is not done, the steel blocks should be fire protected conventionally.

Welded stirrups Welded shear connectors

I I

= stirrup diameter b

Maximum spacing (mm) Beam Column of stirrups along member 250 500

Partially encased columns (or beams)

& > 20 but Cover to bars

Maximum spacing of stirrups = 250mm

Maximum spacin of stirrups = 158,

Ful ly encased columns Concrete f i l led columns

Figure 27 Detailing requirements for partially and fully encased sections


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EC4: Part 1.2 also recognizes the fact that gaps between the connected parts will close up at large deformations in a fire. Therefore, nominally simple connections may behave as semi- continuous connections in fire. This effect is even more pronounced in composite beams where the reinforcement in the slab is relatively cool. Again, provided there is a measure of effective compression transfer at the bottom flange, hogging moments will be developed. A gap less than 15 mm between the beam flange and column is assumed to close up in fire.

End plate connections and some angle cleat connections may be 1 x 3 Cl4.2.5.4

assumed to develop hogging moments in fire. No guidance is given in EC4: Part 1.2 on how to account for this effect. However, it is assumed that a calculation of the moment resistance of the connection from first principles would lead to reduction in the effective load ratio on the beam. This analysis is covered further in reference (17).

& EC3 Annex C


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Welded bracket ,/

/ Extended end plate

,,' Bolt for location

Plate welded to primary beam

' Concrete encasement

I I

Typical beam-beam connection

Concrete encasement

Typical beam-column connection (also beam-beam connection)

Extended end plate

resisted by shear block. Welded shear block Bolts for location only . . . . .

Alternative beam-column connection to major axis of column

Pocket for bolt head (bolts project from column)

Concrete encasement

Figure 20 Typical connections in partially encased members


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Code Clause

l2 EXTERNAL STEELWORK

Eurocode 3 : Part 1.2 makes extensive reference to the temperatures experienced in fire by steelwork located outside the building envelope. This aspect is covered in Annex C of EC3 : Part 1.2. In principle, external steel columns and beams may be partially shielded from direct exposure to the fire plume discharging through windows and doors so that the heat flux is only due to radiant heat from the flame. Such a strategy relies on the use of robust cladding, which does not fail or radiate heat in fire, and carefully located openings in the facade, so that the exposed steel members are not directly heated.

Use of unprotected steel can usually be justified depending on the size and location of the steel sections. This technology has been used successfully on a number of steel framed buildings. The guidance in EC3 : Part 1.2 Annex C is based on the existing publication by Law and O’Brien(40), and uses the same basic physical equations, which are presented in summary as follows:

12.1 Influence of location of steel members

EC3: Part 1.2 Annex C considers two cases: Members not engulfed E C ~ Annex c 2 in flame, and members engulfed in flame. & c 3

The dimensions of the fire plume or flame emanating from an opening depends on whether or not there is a through draught. Flames tend to emerge from the upper two thirds of an opening, the cold air being drawn in below this level. The horizontal projection of the flame is approximately the same dimension, i.e. 2/3h increasing with the through draught. The flame width is approximately equal to the opening width, (see Figure 29).

The optimum location of the steel beams or columns is therefore strongly influenced by the dimensions of the flame. EC3 Annex C illustrates the relevant design cases. Clearly members engulfed in flame will experience higher temperatures than those that are partially shielded. Columns may be located between openings to reduce their heat flux, but beams are likely to be located directly above openings, and may be subject to greater heat flux. The projection of the flame can be altered by awnings above the openings, but these projections may be aesthetically undesirable. Alternatively, beams may be located some distance from the openings, but again, this is architecturally difficult to achieve.


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Code Clause

K ~ - % 2h13

M * 2h13

A

3

v

Plan h<1.25w, wall above No wall above or h z 1 . 2 5 ~

Figure 29 Projection of a flame through an opening

12.2 Heat transfer to a steel member

The heat transfer formulations are defined for the following four cases in EC3: Part 1.2:

Columns not engulfed in flame - Annex C2 Beams not engulfed in flame - Annex C3 Columns engulfed in flame - Annex C4

Beams engulfed in flame - Annex C5

The design formula for one case is presented as follows:

For members not engulfed in flame, the average temperature, T, of the steel member is determined from the following equation.

a p + a T = I , + I f + 293 a (62) EC3 Annex cl Cl .3

where :

a = Stefan Boltzmann constant = 56.7 X kW/m2 cc = convective heat transfer coefficient (kW/m2 K ) I , = radiative heat flux from the flame (kW/m2) I, = radiative heat flux from the opening (kW/m2).

Formulae are given for all these radiation effects. Reference is made to Eurocode 1 Part 2.2 (ENV 1991-2-2) for these various heat transfer coefficients. This analysis is not appropriate for hand calculation, although the publication by Law and O’Brien(40) does


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Code Clause

contain a worked example. It is considered that this publication is deemed to satisfy the requirements of EC3: Part 1.2 despite the subtle differences between the two methods.

Further description of EC3: Part 1.2 Annex C is not presented in this Guide as it is a specialist subject. However, there are many examples of where external steelwork has been used successfully.


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13

13.1

13.2

146

Code Clause

CASES NOT COVERED BY EUROCODES

Various common design cases may be encountered that are not covered by EC3 or EC4 Parts 1.2. Failing advice in the Eurocodes, the designer may refer to BS 5950: Part 8. Common cases are reviewed in the following sections.

Portal frames in fire

Building Regulations in the UK are concerned with the spread of fire from one building to another. Designers are sometimes required to demonstrate that the collapse of an unprotected portal frame rafter in fire will not cause collapse of a boundary wall. This check is only necessary where the boundary wall is required to possess fire resistance, for example where it is close to another building. BS 5950: Part states that the design is satisfactory if the bases and foundations of the boundary columns are designed to resist the forces and moments generated by rafter collapse and the columns are adequately fire protected. No requirement for this effect is given in EC3: Part 1.2.

In a severe fire, plastic hinges will form at the eaves and apex in a portal frame and the rafter will start to collapse. The rafter may buckle sideways in order to pass through the horizontal position before eventually hanging as some form of catenary. At this point the eaves will be pulled inwards. If prolonged heating occurs, the rafter will slowly subside until it touches the ground. The base overturning moment initially acts outwards and then it reverses to act inwards. To avoid collapse of the column, the maximum moment acting ‘inwards’ must be resisted.

Calculation methods are given in the Appendix to BS 5950: Part 8 and reference is made to the Steel Construction Institute’s publication Fire and Steel Construction: The behaviour of steel portal frames in boundary conditions(41). This publication is recognised by all the UK regulatory authorities. It should be noted that the requirement only applies to the use of unprotected rafters. The method given is not appropriate for frames with protected rafters as these frames will not undergo large deformations in a fire. This check is not necessary for sprinklered buildings.

Castellated beams

The behaviour of castellated beams in fire has been investigated by a series of six fire tests; three on castellated sections with sprayed fire protection and three with box protection(41). The conclusion from

BS 5950 Appendix F

BS 5950 cl 4.5.3.3


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these tests was that the greater heated perimeter of castellated beams leads to higher temperatures in the lower section of the web and flange. A simplified and conservative approach, as given in BS 5950: Part 8, is to increase the thickness of fire protection by 20% based on the section factor value of the uncut section. It is proposed to use this approach in design to the UK NAD for EC3: Part 1.2.

13.3 Walls and roofs

Internal walls are often designed as compartment boundaries and therefore should possess appropriate fire resistance. However, most cl 4.10 & 4. I 1 walls are not free standing and rely on lateral support by the structure.

BS 5950

The main concern is the deformation of the support structure in a fire and the effect it has on the integrity of compartment walls. BS 5950: Part 8 states that ‘where a fire resisting wall is liable to be subjected to significant additional vertical load due to the increased vertical deflection of a steel beam in a fire, either:

(a) provision should be made to accommodate the anticipated vertical movement of the beam or,

(b) the wall should be designed to resist the additional vertical load in fire conditions. ’

Walls constructed of masonry will usually be sufficiently robust that they can support any loads transferred from a member above. The more important case is that of a tall slender fire resisting wall such as in a warehouse where compartmentation is very important. Such walls are not normally able to resist significant vertical loads.

Guidance in method (a) is given in BS 5950: Part 8. In EC3: Part EC4 1.2 it is recommended that the critical temperature of beams cl 4.2.1(6)

supporting compartment walls is calculated using a strength retention factor kx,o corresponding to a 0.5% strain limit. This strain corresponds to a limiting deflection of about spadl00 for a fully exposed beam at the fire limit state. Most beams at compartment boundaries are only heated from one side, and it may be assumed that they deflect much less than this limit, and a deflection of spad200 is an appropriate design value.

The deflection of the support beam should be accommodated by sliding restraints in the plane of the wall which permit vertical but prevent horizontal movement. These details should not permit passage of smoke or flame are therefore crucial to maintaining the


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Code Clause

integrity of the wall. Manufacturers of purpose-made systems have made due allowance for this effect.

Generally, no special considerations are necessary for roofs as they are not normally required to possess fire resistance.

1 3.4 Ceilings

Suspended ceilings or ‘screens’ may, in some cases, be used to provide additional fire resistance to the floors or beams above. The main considerations in the effective use of fire protective suspended ceiling systems are that the tiles or boards do not become dislodged in a fire, and that the suspension hangars do not lose their strength. To overcome the possible disruptive effect of the thermal expansion, ‘cut-outs’ are introduced in the ceiling grid. Ventilation ducts and other openings should not impair the integrity of the ceiling. No EC4 cl 4.3.2

specific guidance is given in Parts 1.2 of EC3 or EC4, although in principle, ‘screens’ may be used to reduce the temperature of the steel members that are shielded.

1 3 . 5 Bracing

Vertical bracing provides the stability of simple frames and should be fire protected accordingly (see Section 4.2). If possible, bracing should be built into fire resistant walls so that the bracing needs no applied fire protection. Similarly, bracing within fire protected stairways or cores needs no additional protection (other than that provided around the stairway).

The lateral loading on low rise structures in fire is relatively small (see Section 2.3). In certain cases, it is possible to develop sway resistance of the structure in fire through the continuity of nominally simple connections. Other alternative load paths may be devised when considering the stability of the structure.

13.6 Escape stairways

BS 5950 cl 3.4

Escape stairways should be part of a fire-protected envelope as they are needed for safe evacuation of the building and subsequent fire-fighting. Steelwork contained completely within the fire-protected envelope does not need additional fire protection.


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Code Clause

14 NATURAL FIRES

‘Natural fires’ are those where a fire builds up and decays in accordance with the mass and energy balance within a compartment. They are significantly different in behaviour from ‘standard fires’ where gas is the fuel and where temperatures increase at a gradually diminishing rate. The prediction of the temperatures that are reached in natural fires has been the subject of considerable research. The CIB (Conseil International du Ba tk~~en t ) (~~) have developed predictive methods by which the severity of natural fires can be related to the standard fire.

Consideration of the effect of natural fires can result in considerable benefit, particularly in buildings where the amount of combustible contents are small. Examples are buildings of large volume such as sports halls, some retail premises, car parks and rail or air terminals. Some relaxation in the regulations for fire resistance is often possible, particularly where a ‘trade-off‘ between ‘active’ and ‘passive’ protection measures is envisaged.

Information on the fire load density, and equivalent severity of natural fires is not given in EC3: Part 1.2, but is contained in EC1 Part 2.2@). Although not expressed explicitly in EC3: Part 1.2, nor BS 5950: Part 8, the critical temperature method is applicable to any fire. Therefore, there may be circumstances where it is possible to compute the rise in temperature of a steel section in a natural fire, and then to follow the approach in Section 4 to establish its load carrying capacity. A review of fire engineering methods is given by Kirby“).

14.1 Important parameters in determining fire temperatures

14.1.1 Rate of burning

Temperatures in natural fires as compartments are dependent on: the fire load, the ventilation conditions and other properties such as the thermal conductivity of the compartment walls and roof. The ‘fire load’ should take into account the calorific value of all the combustible components and is usually expressed in terms of kg of wood-equivalent per unit area of the floor.

After ‘flash-over’, or rapid expansion of the fire from its source (after 2 to 5 minutes), the rate or burning of the fuel (i.e. wood) depends on whether the fire is ‘fuel’ or ‘ventilation’ controlled. If sufficient air is available, the rate of burning is influenced by the characteristics of the fuel. For ‘fuel-controlled’ fires, which in fire


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14.1.2

Code Clause

tests are characterised by wooden slats or cribs (normally 40 mm square at 40 mm spacing), the rate of burning is almost a linear function of the fire load. A typical rate of burning would be 50-60 kg/m2/hour for a fire load of 20 kg/m2. This suggests that a fire of this intensity would consume most of the fuel in roughly 30 minutes.

If there is insufficient air brought into the compartment through the openings to maintain the combustion process, the rate of burning will reduce. This is known as ‘ventilation controlled’ burning.

Wood produces an energy output of approximately 18 MJ/kg of material (i.e. a rate of approximately 4.6 k W k g ) . The equivalent energy content of synthetic materials, such as polystyrene, is up to 3 times greater than wood, indicating that these materials add significantly to the calorific value of the contents. Typically, 5 kg of air is required for the combustion of 1 kg of wood, whereas 13 kg of air is required for 1 kg of polystyrene.

For wood fires, the maximum rate of burning of the fuel in ‘ventilation-controlled’ fires is given by the empirical formula:

R = 0.09 A, 6 kg/sec (65)

where: A , = total ventilation area (m2) h = weighted mean height of the openings (m)

The opening height, h, is important because of considerations of gas flow into and out of the compartment. The hot gases are expended through the upper part of the opening, and the air necessary to maintain the fire is drawn in through the lower part of the opening.

In ventilation-controlled fires, the rate of burning of the combustible contents is approximately constant during a period in which the total weight of fuel reduces from 80% to 30% of its original value. The maximum temperature is experienced when around 50% of the fire load is consumed. For fuelcontrolled fires the behaviour is similar, but the maximum temperature occurs when 60 to 70% of the fire load is consumed.

Fires in small compartments

In a small compartment the temperature conditions at a given time are considered to be uniform. Temperatures are determined from the energy and mass balance equations of the form:

(a) Heat increase of air in the compartment = heat produced (by fuel burning) - heat lost (through walls, openings, etc.)


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(b) Mass of air leaving the compartment (through openings) = mass of air drawn in + mass of fuel consumed.

Peak temperatures in a compartment are usually experienced when there is just sufficient air for fuel controlled burning, but the size of the openings does not permit excessive heat loss. Openings in the roofs of compartments have a much greater effect than vertical openings (of the same size) in allowing the hot gases to escape. It is well known that roof-venting is a good method of reducing peak temperatures in fire. Various computer programs have been developed for predicting the temperature-time relationship in natural fires in small compartments.

A series of 21 fire tests in a modest sized compartment was carried out by British Steel and the Fire Research Station(45) in 1985. The brick and concrete compartment was 9.7 m long, 6.85 wide and 3.9 m high. A range of fire loads up to 20 kg/m2 and a variable width facade opening were the main parameters. An example of a typical temperature-time curve is shown in Figure 30. The case that most closely matches the early stages of a standard fire test is a fire load of 15 kg/m2 and a single opening of area equivalent to one quarter of one facade. This is shown as the curve identified by: 15(?4). The rates of burning were measured and agreed well with the theoretical values. All but one test was ‘fuel-controlled’.

1000 Standard fire

?! 3

l-

800

600

400

200

0

0 20 40 60 Time (mins)

Figure 30 Typical time-temperature curves for a natural fire


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Code Clause

14.1.3 Fires in large compartments

Fires in large compartments do not generally engulf the whole compartment. The same heat and mass balance equations apply as for a small compartment, but it is necessary to include the air and heat exchange between the fire volume (plume) and the remainder of the compartment. The fire plume (considered to be an inverted truncated cone in shape) over the area of burning fuel, expands in size according to an empirical formula. Air is drawn into the burning fuel to be replaced by air drawn in from the openings.

The above approach is a described as a ‘two zone’ model. More sophisticated zone models than this two zone model have been developed, but these are not yet practical methods. The single zone (small compartment) model would over-predict the temperatures experienced in a large compartment.

Fire tests on large compartments are few in number. British have recently carried out a series of 9 fire tests in a long compartment of 22.8 m X 5.6 m on plan and 2.75 m height. Ventilation was provided only at one end, and the fire was propagated at the other end of the compartment. The temperatures of the fire at various locations, and of different sizes of protected and unprotected steel members, were recorded.

A series of fire tests on a composite steel framed building was carried out in Germany(47). Various structural configurations including composite slabs and beams, and external steelwork were included. The building did not suffer excessive damage. Various tests on open(48, 49) and closed car park structures have also been carried out in the UK, USA and Australia to demonstrate that relatively low intensity fires are experienced in such buildings, leading to a relaxation in the regulation of requirements for fire protection of this form of structure.

An important series of fire tests was carried out by the Centre Technique Industriel de la Construction Metallique (CTICM) in a disused electricity generating hall in Paris (La Villette)(”). In this series of tests, fires were propagated in a small area (10% to 25%) of a 28 m X 39 m bay of the building. Various steel sections were suspended over the centre of the fire and their temperatures measured.

A serious fire also took place during the construction of Phase 8 of the Broadgate development in London in 1990. The assessment of the fire severity and structural response was reported(51). It was shown that the unprotected steel was subject to temperatures exceeding 600”C, and that the building continued to support its own


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Code Clause

weight without collapse. The damaged members were replaced or strengthened at less than 2% of the total repair bill.

Finally, a major series of fire tests is underway at the BRE test facility in Cardington, in the UK. British Steel and BRE are collaborating in the testing of an 8 storey composite frame. Fire compartments were created on certain floors and subject to various intensities of fire through gas, wood or furniture as fuels. This work will be reported in 1996, but the initial results have demonstrated the important influence of structural continuity in fire, leading to the good performance of unprotected composite beams.

14.2 Concept of time-equivalent

The concept of 'time-equivalent' has been widely used as means of measuring the intensity of a natural fire in relation to a standard fire. In principle, the greater the time equivalent, the greater the severity of the fire, and hence the greater the required fire resistance of the structure.

From an extensive series of fire tests in small compartments carried out worldwide, an empirical formula for the time equivalent te has been proposed. This was first established by considering the equivalent degradation of concrete elements in natural and standard fires. It includes the term A, J h which broadly defines the rate of burning in ventilation controlled fires. The formula for the time- equivalent fire severity(43) is:

where:

4f - - fire load expressed as MJoules/m2 of floor area (18 MJ = 1 kg wood)

A, = floor surface area A, = total surface area of the compartment.

In many references, qf is expressed in terms of the total surface area, rather than the floor area. This formula applies to enclosures of brick or concrete.

In EC1 Part 2.2, Equation (66) is replaced by a further simplified ECl Part 2.2 formula as follows:

te = yq kb qf minutes


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Code Clause

where:

k, = a factor accounting for the insulating properties of the walls.

yq = partial factor for risk and consequence of failure.

For large openings (greater than 10% of the floor area), the fire is fuel-controlled and te is independent of h. kb may be conservatively taken as 0.09 for highly insulating materials, reducing to 0.05 for less insulating materials.

In EC3: Part 2.2, the factor yq is a multiple of two factors, which take into account the risk of severe fires occurring, and also the consequences of failure. For offices up to 30 m high, yq is given by the multiple of 1.1 x 1.2. Account may also be taken of the influence of active protection measures, such as sprinklers, which will reduce the yq value by a factor multiple of 0.6. For parts of a building above 30 m height, the yq factor is doubled.

Using either of these equations, it is possible to evaluate the fire resistance required for a particular compartment within a building. Such an approach demands that the building should be unchanged during its life, particularly in terms of fire load and ventilation area.

14.3 Fire loads in buildings

The main problem relating to the use of the time-equivalent method is the selection of the fire load or fire load density that should be adopted. A number of surveys of fire loads have been conducted worldwide. Typically, the data have been presented in terms of mean ‘fire loads’ and characteristic values, i.e. 80% of rooms within the various occupancy groups have fire loads less than the values quoted.

All values are expressed in terms of MJoules/unit floor area. The fire load is determined by summing the mass of each object in the compartment times the calorific value of the materials. The fire loads typical of the following types of buildings are presented in EC1: Part 2.2, and characteristic values are given below:

Apartments, Houses, Offices, Institutional 500 MJ/m2 EC1 Part 2.2 Buildings, Car Parks. Table 7

Shops, Assembly Recreational Buildings

750 MJ/m2

Storage, Industrial Buildings 1000 MJ/m2


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If a fire load density of 500 MJ/m2 is inserted into Equation (67), the time-equivalent is approximately 60 minutes.

14.4 Temperatures in steel sections in natural fires

The temperatures in uniformly-heated steel sections e,, may be determined from Equation (21) in Section 5.5 for protected members, or from Equation (16) in Section 5.3 for unprotected members. These equations are based on one-dimensional heat flow into the section. More sophisticated finite element methods are required to determine the heat flow in more complex sections comprising different materials.

The natural fire temperatures 8, may be input into these equations. In a standard fire, the steel temperature rise follows the same form of curve as the fire temperature, the temperature difference between the furnace and the steel temperature being principally dependent on the section factor and the thermal resistivity (inverse of conductivity) of the protection material.

In a natural fire, the steel temperature rises and decays, but the maximum steel temperature does not occur at the maximum fire temperature. Rather it occurs on the declining portion of the fire temperature curve shortly after the point at which the fire and steel temperatures are equal. This is because when 8, exceeds e,, heat enters the section, and when 8, exceeds e,, heat leaves the section. The slight delay is because of the storage of heat in the fire protection material.

The effect of the fire protection to the steel member is both to slow down the rate of temperature rise and to ‘smooth-out’ any early peaks in the temperature-time history. For example, polystyrene in the fuel causes a very rapid rise of temperature early in a fire, but this also decays quickly. The maximum steel temperatures in a well protected member are not significantly different from the case where wood is the only fuel (for the same fire load density).

14.5 Influence of active protection measures

‘Passive’ protection refers to the fire resistance of members, whereas ‘active’ protection measures are those that contribute to a reduction in the severity of a fire. The most commonly-used active system is sprinklers. These are characterised in terms of their speed of reaction, rate of wetting, and total discharge capacity.


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There is a body of statistical evidence to indicate that sprinklered buildings have suffered less fire damage than non-sprinklered buildings. For steel buildings, such active systems are attractive because some trade-off between active and passive measures could result in the elimination of passive fire protection to the steelwork in many categories of buildings of low fire load. This aspect is recognised in EC1: Part 2.2. A secondary benefit is the increase in permissible compartment areas and volumes.

There are now many examples of buildings in the UK and Europe where such appraisal methods have been used. The Swiss were the first nation to introduce a risk evaluation matrix for buildings in fire (known as the Gretener system). This is administered by a monopoly-insurance bureau who offer building-structure insurance on a national basis. The method is based on a points system that heavily reflects the benefits of active measures. A similar calculation method is now used for industrial buildings in Germany. This is based on statistical evidence of fire severity in sprinklered buildings.

In principle, sprinklers, detection and other active measures are designed to prevent small fires from becoming large ones. However, there is some small risk (around 2% commonly quoted) of the systems malfunctioning or being unoperational (because of poor maintenance).


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Code Clause

l5 RE-USE OF STEEL AFTER A FIRE

Fire is a rare occurrence and it would normally not be economically viable to design for damage limitation and reparability after a severe fire. Indeed, regulations and fire tests are not concerned with re-use of the structure, provided it retains its load resistance for the appropriate period of fire resistance.

The potential for re-use of structural steel after a moderate to severe fire is a matter of engineering judgement. It may be expected that unprotected elements have distorted, and that the fire protection from some elements has become detached on cooling. Spalling of concrete encasement may have exposed the steel section.

Outside the zone of the building subject to the most severe part of the fire, there may be secondary damage because of thermal expansion of the floors or thermal bowing towards the fire. The criteria for re-use are the residual strength of the materials, and the distortion of the members. This second aspect is more important for members subject to compression, e.g. columns or struts.

No guidance on re-use or inspection of steel after a fire is given in EC3 or EC4: Part 1.2 and the reader has to refer to other published information(19).

15.1 Mechanical properties

Data on the mechanical properties of steel and other materials after being subject to elevated temperatures for up to 4 hours is presented by Kirby and Lapwood(”). The following sections summarise their observations and more recent research findings.

15.1.1 Structural steel

The mechanical strength properties of S275 steel are unaffected following a fire if the steel temperatures were less than 600°C.

BS 5950 Appendix

However, exposure of S275 steel to temperatures above 600°C may c.2.1

result in a reduction in strength of up to 10% below the minimum specified values for this grade. The mechanical strength properties of S355 steels are also unaffected for temperatures below 600°C. Exposure to higher temperatures may result in greater strength reductions than in S275 steel.


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Code Clause

15.1.2 Bolts and welds

The tensile strength reduction for grade 4.6 bolts follows that of S275 steel. For grade 8.8 bolts, the residual strength falls more markedly Appendix c.3 after exposure to temperature exceeding 450°C. The residual strength of bolts after exposure to higher temperatures reduces to about 80 % at temperatures of 600"C, and 60 % at 800°C. High strength friction grip bolts behave in a similar manner to grade 8.8 bolts.

BS 5950

The residual strength of welds is not significantly affected by exposure to high temperatures.

15.2 Inspection and appraisal

The initial inspection and assessment of any fire damaged structure BS 5950 is visual. This is followed by measurements of the Appendix 6.5 out-of-straightness of linear elements, particularly those that may be subject to compression. It may be necessary to carry out hardness tests on certain sections of steel to gain a measure of the residual strength of the steel. A portable Brinell Hardness Tester can be used to provide an indication of the ultimate strength of the steel and whether it is likely to meet the appropriate specification.

Comparisons may be made between fire damaged and only slightly affected or non-damaged members. Variations of 10% in Brinell Hardness Number can be regarded as significant. Where changes have occurred and the corresponding values of ultimate tensile strengths are below 5 % of the minimum requirements for S275 steel, or 10% for S355 steel, the mechanical strength properties of the steel may not meet the appropriate specification. In such cases, it may be necessary to seek advice from a metallurgist and to check the tensile properties of the steel by removing a coupon from a heat affected member and subjecting it to a standard tensile test.

15.3 Re-use of unprotected sections

No guidance on the re-use of steel after a fire is given in EC3: Part 1.2, but the following information is relevant to re-use.

Members that have not distorted beyond the normal BS 5950

out-of-straightness limits can be reused without further checking Appendix C . 1 provided the connections have not suffered damage. Members that have distorted slightly may be reused provide they are laterally restrained, or the stresses to which they are subjected are small. This is a matter of a further structural assessment.


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Code Clause

It may be necessary to replace bolts if they have been subject to temperatures above 600"C('7).

1 5.4 Re-use of protected sections BS 5950

Fire protection generally takes the form of spray or box protection or Appendix C.4

concrete encasement. In slightly damaged areas, spray protection can be resprayed having removed any damaged or friable material. However, some materials undergo chemical changes at 200-300°C and advice should be sought about their removal. Gaps may have developed between the individual panels of box protection as a result of the distortion of the structure, and it would be necessary to replace the affected panels.

In heavily damaged areas, the fire protection should be replaced. The steel members are unlikely to have been seriously affected by the fire, unless the protection had broken away locally.

Concrete is likely to have spalled from the steel and it will normally be necessary to replace the cover of the steel by a cement rich mortar or sprayed concrete (gunite) having broken back the damaged or cracked concrete.

1 5.5 Re-use of composite deck slabs

Some permanent deflection of all concrete or composite deck slabs would be experienced after a severe fire. In composite deck slabs it would normally be necessary to cut-away any sections of decking that have severely debonded from the slab, and replace it by reinforcing bars attached by shot-fired clips, between the ribs. The bars could then be 'gunited' in-situ to achieve the appropriate bond. In extreme cases, replacement of the complete section of slab between the steel beams should be considered. Load tests on fire damaged slabs have showed that moderately damaged sections will probably not require remedial work.


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REFERENCES

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BRITISH STANDARDS INSTITUTION BS ENV 1993: Eurocode 3: Design of steel structures Part 1.2 Structural fire design (including UK NAD) BSI, 1996

BRITISH STANDARDS INSTITUTION BS ENV 1994: Eurocode 4: Design of composite steel and concrete structures Part 1.2 Structural fire design (including UK NAD) BSI, 1996

EUROPEAN CONVENTION FOR CONSTRUCTIONAL STEELWORK European recommendations for the fire safety of steel structures: Design manual ECCS, 1985

INTERNATIONAL STANDARDS ORGANISATION IS0 834: Fire resistance tests - elements of construction ISO, 1985

BRITISH STANDARDS INSTITUTION BS 476: Fire tests on building materials and structures Part 2: Method of determination of the fire resistance of load bearing elements of construction BSI, 1987

BRITISH STANDARDS INSTITUTION BS 5950: Structural use of steelwork in building Part 8: Code of practice for fire resistant design BSI, 1990

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BRITISH STANDARDS INSTITUTION BS ENV 1992: Eurocode 2 : Design of concrete structures Part 1.2 Structural fire design (including UK NAD) BSI, 1995


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APPENDIX A: WORKED EXAMPLES


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Institute Subject Critical Temperature of Restrained Beam

Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01344) 22944

CALCULATION SHEET

Client Made by MA W June 1995

GMN May 1996

Date

Checked by Date

WORKED EXAMPLE l STEEL BEAM SUPPORTING A CONCRETE SLAB

Consider a beam, 406 X 140 X 46 UB supporting a concrete slab. It is assumed to be fully laterally restrained (i. e. no lateral-torsional buckling).

. . . . L .

\ Slab

406x140~46 UB section Yield stress : f, =275N/mrn2

UDL : W

1 A

L o a d i a

Dead loading: Gk = 3.0 kN/m2 Imposed loading: %,l = 3.5 k N / d

Maximum moment at Fire Limit State:

where: yGA = 1.00 qJ1,J = 0.50

3 x 62 8

:. M f i = [1.0x3.0+0.50x3.5]x- - - 64.1 kNm


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CALCULATION SHEET

Client Made by MA W June 1995

GMN Date May 1996

Date

'Checked by

Section classification: Class l

At ultimate limit state:

Wpl = 889 x Id mm3 (Plastic modulus) YMO - 1.05 - .- ~.

At fire limit state:

Load level:

where: uI = 0.7 because of non-uniform temperature across the section.

yfi - - = 0.28 232.83

It follows that:

ky,e = 0.2ax- O* = 0.184 (Effective yield strength reduction factor) 1.05

Therefore, from Table 3.1, this effective yield strength reduction factor is reached at a critical temperature of..

0, - - 700 -I- - O*04t5 x l o o = 738°C 0.120

The Steel Construction EC3: Part 1.2 Institute

BCF 558 Page 2 of 2

Subject Critical Temperature of Restrained Beam

-

-


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The Steel Construction EC3: Part l .2

BCF 558 Page l of 6

Institute // Subject Critical Temperature of Unrestrained Beam


(Client IMade by MA W lDate June 1995 1

CALCULATION SHEET I IChecked by GMN I Date May 1996

WORKED EXAMPLE 2 LATERALLY UNRESTRAINED BEAM

Point load arising from transverse beam

/ Transver 6x3m of

'se beam concrete

supporting floor

Region of -.

concrete floor,'

Critical temperature of point loaded beam AB to be determined to EC3: Part 1.2. Assume beam is 406 X 178 X 67 UB in S275 steel (grade 43 - UK).

Loading on Floor:

Dead loading: Gk = 3.0 kN/m2 Imposed loading: &,l = 3.5 kN/m2

Firstly, the beam has to be checked at the ultimate limit state against lateral-torsional buckling. Therefore, EC3: Part 1.1 should be used.

Beam Loading (Ultimate Limit State)

P = 3 X 6 X (1.35 X 3.0 + 1.50 X 3.5) = 167.4 kN

Lateral restraints are provided at points A, B and C, :. Buckling length: l = 3 m

To perform the check for lateral-torsional buckling, the following section properties are required, based on the section classification.


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The Steel Construction

lJob No: BCF 558 IPage 2 of 6 1 I I

Job Title EC3: Part l. 2

Telephone: (01 344) 23345 Fax: (01 344) 22944

Institute // Silwood Park, Ascot, Berks SL5 7QN I

-

lSubject Critical Temperature of Unrestrained Beam

1 Client (Made by MA W lDate June 1995 I

CALCULATION SHEET I IChecked by GMN IDate May 1996

Section: 406 X 178 X 67 UB tw = 8.8, t f = 14.3, d = 360.4, c = 178.8/2

Yield stress: fy = 275 N/mm2 :. E = [FT5 = 0.924

d/tW 72 E = 66.2 1 limits for Class l classification c/$ I 10 E = 9.2 1 (Table 5.3.1: Part 1.1).

For the section in question

dtw = 41.0 (2 66.2 :. OK) c/i) = 6.25 ( S 9.2 :. OK)

:. Section is C b s l

Lateral-torswnal buckling resistance (calculated using the simplified method in C-EC3- Concise Eurocode 3 for the Design of Steel Buildings in the United Kingdom. SCI fibliC&*OR P-l 16)

where:

P, = 1.0 for Class 1 or 2 sections

WAY = 1346 x I d rnd (major axis plastic modulus)

f b = bending strength to be detennined from the simplified method in C- EC3.

The beam has a single lateral restraint at its mid point.

Effective length factor, k = 1.0 (Table 5.21; C-EC3)

Therefore, it is subject to a linear variation of moment.

:. (k/C1)*.5 = 0.73 (Tdle 5.22; C-EC3)


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The Steel Construction Job Title EC3: Part 1.2

BCF 558 Page 3 of 6 Rev

-

-

Institute // Subject Critical Temperature of Unrestrained Beam

Silwood Park, Ascot, Berks SL5 7QN Telephone: 101 344) 23345 Fax: (01 344) 22944 Client Made by MAW I June 1995 Date

CALCULATION SHEET .Checked by GMN Date May 1996

To determine fb, Table 5.20 can be used, thus iLT and aLT should be determined.

~ L T - - _ l2 - minor axis radius of gyration 39.9 = 45.3 U Buckling parameter

- -- 0.88

~ L T - - iz X x = minor axis radius of gyration X torsional index

:. aLT = 39.9 X 30.5 = 1216.95

pWo- (k/Cl)o-sL/iLT = l . 0 X 0.73 X - 3O00 = 48.3 45.3

L/a,, = 3000 - - 2.47 1216.95

:. from Table 5.20 fb = 251.6 N/m&

Buckling resistance moment of beam:

Mb.Rd > 251.1 kNm (= Mmm)

406 X 178 X 67 UB can be used for normal temperature design.

FIRE LIMIT STATE

Calculating the critical temperature of this beam is an iterative process. One first has to guess a temperature, go through the procedure of determining an effective slenderness of the beam, and find the new critical temperature on the basis of this effective slenderness, If the final answer and the initial guess do not agree, then this new temperature becomes the initial guess and the process is repeated until the guess and the new temperature agree.

~~


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:I ::e

BCF 558

Subject Critical Temperature of Unrestrained Beam Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01 344) 22944

CALCULATION SHEET

Client Made by MAW June 1995

GMN May 1996

Date

Date Checked by

Beam loading (Fire limit state):

p = b1 x b2 x (YGA Gk @I , I %,l )

where: yGA = 1.0 @l,l = 0-5

:. P = 3 X 6 X (1.0 X 3.0 + 0.5 X 3.5) = 85.5kN

:. M,, - - -- - 85-5x6 = 128.3 kNm = Mfi 4 4

Critical temperature: initial guess: 8 = 550°C

High temperature lateral-torsional buckling slenderness: (EC3: Part l .2)

where:

.. Xm,e - -

XLT - -

0.625 1 from Table 3.1 of EC3: Part 1.2 0.455 1

45.7 (Cl. 5.5. 5(9) C-EC3)

93.9 E = 86.8

45.7/86.8 = 0.526

0.5 X [ l + 0.21 X (XLT.0 - 0.2) + XLT,e2J = 0.733

1 2 0.5

= 0.884 0.733 + (0.7332 - 0.616 )


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1:; ::e

BCF 558 Page 5 of 6 Rev

Subject Critical Temperature of Unrestrained Beam Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01344) 22944

Client Made by MAW June 1995

GMN May 1996

Date

Oate CALCULATION SHEET Checked by

Class l limits d/t, I 56.7 c/$< 7.9

Referring to Sheet 2, the ratios still satisfy the Class l limits.

x LT Wpl,y kYe f y = 0.884~ 1346 x lo3 x 0.625 x 275 -'* Mb,Rd,O =

? M I 1.2 x 1.0 x 106

The buckling resistance moment is greater than the required value of 128.3 kNm. Therefore the calculation is repeated assum'ng a higher critical temperature.

kY, 0 = 0.470 1 from Table 3.1 of EC3: Part 1.2 kE, 0 = 0.310 1

l = 0.76 + (0.76' - 0.6482)0.5

= 0.864


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The Steel Construction Institute Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01 344) 22944

CALCULATION SHEET

lJob No: BCF 558 (Page 6 of 6 lRev lJob EC3: Part 1.2 Subject Critical Temperature of Unrestrained Beam

Client Made by MAW lDate June 1995

l bhecked by GMN lDate May 1996 1

ALTERNATIVE METHOD

This is a simpler non-iterative method (kx,, method) where the temperature dependent slenderness is equal to the slenderness for normal design.

:. xLT,e = X L , = 0.526

The load level is given by:

1283 1.05 %=----

- Mb, Rd 322.5

which leads to:

kx,0 = 0.38

From Table 3.1, this effective strength is reached for a critical temperature 08

ea,,, = 590°C

This value is slightly lower than the value calculated by the rigorous method.


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The Steel Construction Job Title EC3: Part l. 2 Institute

~KD[~~ No: BCF 558 (Page l of 6 IRev 1 Subject Critical Temperature of Column

Silwood Park, Ascot, Berks SL5 7QN I Telephone: (01 344) 23345 Fax: (01 344) 22944

CALCULATION SHEET

Client Made by MA W July 1995

GMN Date May 1996

Date

Checked by

WORKED EXAMPLE 3 COLUMN IN SIMPLE CONSTRUCTION

Consider a 5 storey braced frame with a uniform column section withstanding loads from a 6 x 6mJloor area. Assume the loads are applied to the column centroid (i.e. axial loads only on the column) and that each storey is 4m high.

, 1 I I

WlkN I

Floor

adinp on Floors:

Dead loading: Gk = 3.0 k N / d Imposed loading: = 3.5 k N / d

Design a column section which can resist this loading, and then determine its critical temperature for fire resistance, using EC3: Parts 1.1 and 1.2 respectively.


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The Steel Construction Institute Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01344) 22944

CALCULATION SHEET

' I Job No: BCF 558 IPage 2 of 6 lRe" I lJob EC3: Part 1.2 I lSubject Critical Temperature of Column I

:. W = 6 X 6 X r1.35 X 3.0 + 1.50 X 3.51 = 334.8kN

Maximum load on column:

P = no. of storeys X W

:. P = 5 X 334.8 = 1674 kN (axial load)

No reduction in loading for the number of storeys is included in the analysis.

Normal desim of columrl

Yield stress: fy = 275 Nlmm2 E - - [?r = 0.924

TV: 254 X 254 X 73 UC \

A = 9290 m& (section area)

d/t, = 23.3 ] section fully in compression :. Class l section c/$ = 8.94 ] (Table 5.3.1 Part 1.1).

Axial load resistance:

Nb. Rd - - x PAAF,

Y m1

Where: x = reduction factor depending on slenderness and impe@ections PA = factor depending on the section classification Yml = partial factor dependent on the material

PA = 1.0 for Class 1-3 sections Yml = 1.05 for steel (UK NAD)

Where Q, = OS X (1 + a X (X - 0.2) + X2)


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Normalized slenderness: x - h

h1

- - (BJ.5

Slenderness h - - L Buckling length -= i radius of gyration

a1 = n: X (E/fy)O*s = 9 3 . 9 ~

Axis of buckling: minor axis (z-z)

:. Buckling curve c: CL = 0.49

x - - 61.9 X (1.0)o-5 = 0.714 93.9 x 0.924

4 = 0.5 X (l + 0.49 X (0.714 - 0.2) + 0.7142)

x - I - 0.881 +(0.8812 - 0.7142)0.5

0.716 x 1.0 x 9290 x 275 Nb.Rd -

- 1.05 x 103

- column section 254 X 254 X 73UC is ok

The Steel Construction EC3: Part l . 2 Institute

:l: ::e

BCF 558 lpage 3 of 6 lRev Subject Critical Temperature of Column


CALCULATION SHEET

Client Made by MAW July 1995

GMN May 1996

Date

Date Checked by

= 0.881

- - 0.71 6

= 1742.1 kN

To calculde the critical temperature of the column, the procedure is iterative. One first has to guess a temperature and then determine the reduced resistance of the column at this temperature, and then compare it to the loading at the fire limit state. If the reduced resistance of the column is equal to the loading at the fire limit state, then the critical temperature is correct. If they do not agree, then a new guess is made and the process is repeated. One iteration is usually sufficient.


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The Steel Construction

BCF 558

EC3: Part 1.2

Rev Page 4 of 6 Job Title

Institute \ \\ I

// lSubject Critical Temperature of Column


Client Made by MAW I July 1995 Date

I 1

CALCULATION SHEET bhecked by GMN I Date May 1996

Column loading (f ire limit state);

wfi = b1 X b2 X (YGA GK -+ ~ I , I &,l)

Where yGA - - 1.0

0.5

. .

*l,l - -

.. Wfi = 6 X 6 X (1.0 X 3.0 + 0.5 X 3.5) = 171 kN

Maximum load on column:

Pfi = 5 x wji = 855 kN (Wfi = axial load per storey)

Critical temperature: initial guess: 0 = 570°C

High temperature normalized slenderness of column:

New buckling length: Qfi = 0.7L

(UK NAD is expected to state that for the fire limit state the buckling length may be reduced by a maximum factor of 0.7)

:. Qfi - - 0.7 x 4000 = 2800 mm

=, afi - - 2800 64.6

= 43.3

xji - - 43.3 = 0.500

93.9 x 0.924

From Table 3.1: ky,e = 0.563 (Part l .2) kE,e = 0.397

0.595


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CALCULATION SHEET

Client Made by MAW July 1995

GMN Date May 1996

Date

'Checked by


BCF 558 Rev Page 5 of 6

Critical Temperature of Column -

- - 4 0 = 0.5 X (1 + 0.49 X (0.595 - 0.2) + 0. 59S2)

= 0.774

- l X J i - = 0.788

0.774 + (0.7742 - 0.595 ) 2 0.5

(Note: x = 0.72 in normal design)

where YM,B - - 1.0

t, Rd - - 0'788 x 9290 x 0.563 x - 275 = 944.5 kN

l .2 1.0

The buckling resistance is thus greater than the required value of 855 kN. must be repeated assuming a higher critical temperature.

Try 0 = 590°C ky,O = 0.501 kE,e = 0.339

40 = 0.5 X (l + 0.49 X (0.608 - 0.2) -I- 0. 60g2)

- l Xfi - 0.78 -I- (0.78' -. 6082)aS

= 0.79

0.79 9290 275 1.0

Nb,ji, t, Rd = X- X 0.501 X - = 842.6 kN 1.2 103

The calculation

= 0.608

= 0.78

Further iterations converge to approximately 588°C.


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The BCF 558 Steel Construction EC3: Part l. 2 Institute

lPage 6 of 6 lRev

Subject Critical Temperature of Column

- ALTERNATIVE METHOD

Using a simplified non-iterative method we obtain:

(UK NAD is expected to state that for the .fire limit state the buckling length may be reduced by a maximum factor of 0.7)

4 8 = 0.5 X (l + 0.49 X (0.500 - 0.2) -k 0.S2) = 0.699

xfi - l - 0.699 + (0.6992 - 0.500 ) 2 0.5

= 0.842

In this case, the adaptation factor is 1.0

- 0.842 9290 =* Nb,fi,Rd - X- x kx,e x 275 - - 855 kN

1.0 103

* Critical temperature ea,- = 583°C (from Table 3.1)

This value is slightly lower than the value calculated by the rigorous method.


Client Made by MA W lDate July 1995

I I

CALCULATION SHEET IChecked by GMN IDate Ma.v 1996


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Construction

Job No: BCF 558

EC4: Part 1.2

Rev Page l of 5 ~

Job Title

Institute // Subject Deep Deck Composite Slab


CALCULATION SHEET

Client Made by MAW I Seut 1995 Date

IChecked by I

GMN I May 1996 Date

MOMENT RESISTANCE OF DEEP DECK COMPOSITE SLAB TO EC4: PART 1.2

37.5>ci k-

Design D&

Normal weight concrete:

Decking:

Reinforcement:

Slab span, L

Fire resistance

Loadina

Dead loading, Gk

Imposed loading, Qk l

Ultimate Limit State

Cylinder strength fc,200c = 25 N/mm2

CF210 deck, rib spacing, b =

16 mm diameter bar fys,200c - -

= 5.5m

= 90 minutes

= 2.5 k N / d

= 3.5 kN/m2

600 mm centres

460 N/mm2

Factored moment at ultimate limit state, assuming a simply supported span and no composite action.


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The Steel Construction EC4: Part l .2

BCF 558 Rev Page 2 of S

- Institute // Subject Deep Deck Composite Slab

Silwood Park, Ascot, Berks SL5 7QN I Telephone: (01 344) 23345 Fax: (01 3441 22944 Jclient IMade by MAW lDate Sent 1995

I I

CALCULATION SHEET /Checked by GMN IDate May 1996

where: y~ = 1.35

= 1.50 yaj

:. M,, - - 0.4 x X (1.35 X 2.5 f 1.50 X 3.5) = 19.6 kNm per rib 8

Moment Resistance at Ultimate Limit State

y = 7.3 mm

I l

Resistances: Concrete in compression:

R C = 600 y X 0.85 fc,2pc = 12750 y

Reinforcing bar: (16 mm diameter)

For R, = R , it follows that the neutral axis depth in the concrete is:

y = 7.3 mm

Moment resistance of composite slab in positive (sagging) bending is:

MRd = 12750y2 -t 92489 x (230 - y ) ] x 1Q6

MRd = 20.94 kNm/rib

MRd ’ - Composite slab is adequate for design at the ultimate limit state.


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The Steel Construction Institute Silwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01 344) 22944

CALCULATION SHEET

Job Title EC4: Part 1.2 Subject Deep Deck Composite Slab

Client Made by MAW Sept 1995

GMN Ma.y 1996

Date

Date Checked by

FIRE LIMIT STATE

Applied moment (using appropriate partial factors)

M-- = x L2 x (YGA G k -l- @I,I &,r ) 8

where: YGA = 1.00

%,l = 0.50

:. Mfi - - 5*52 x (1.0 x 2.5 + 0.5 x 3.5) 8

Moment Resistance o f ComDosite Slab at Fire Limit State

From EC4: Part 1.2 clause 4.3.1.4.:

= 9.64 kNm/rib

By symmetry: ul = u2 = 50 mm

u3 = 70mm

Parameter z is determined from:

l 1 l 1 _ - - z & + K+ Ju1= 0*402 -

:. z = 2.48


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I Silwood Park, Ascot, Berks SL5 7QN I

l Telephone: (01 344) 23345 Fax: l01 344) 22944

I CALCULATION SHEET

Client Made by MAW I Seat 1995 Date

I IChecked by GMN lDate May 1996 I -

EC4: Part 1.2 Table 4.9: for 90 minutes fire resistance, the temperature of the reinforcing bar encased in the concrete rib is:

l = 1285 - 350 X 2.48 = 417°C

( If lightweight concrete was being used, the temperature would be reduced by 10%)

I From EC4: Part 1.2 Table 3.4, the reduced strength of the reinforcing bar is:

fsmax,e = 0.89 fsy,20" c

:. Strength of reinforcing bar = 0.89 X 460 = 409 N / m d

y = 6.

F-l


BCF 558 JPage 4 of 5 1''" Deep Deck Composite Slab

-

-

Resistances:

Concrete: R, = 600 y X 0.85 fc,2pc .a. R, - - 12750 y N

Reinforcing bar:

Rs = X 16 X fsm,e = 82234 N 2

4 when R, equals RP it follows that neutral axis depth in slab, y = 6.4 mm

I Moment resistance of composite slab in positive (sagging) bending:


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Construction lJob No: BCF 558 IPage 5 of 5 p e v - 1 I Job Title EC4: Part 1.2 I

Institute // Subject Deep Deck Composite Slab

Siiwood Park, Ascot, Berks SL5 7QN Telephone: (01 344) 23345 Fax: (01 344) 22944

CALCULATION SHEET

The Steel

Client Made by MAW

Date May 1996 GMN

Date Sept 1995 Checked by

* Mfi,Rd Mfi

:. Moment resistance of the slab is adequate for 90 minutes fire resistance.

As a second case, for 120 minutes fire resistance assume lighfweight concrete is used

Temperatures of reinforcing bar in LW concrete:

8, = (1370 - 3502) X 0.9 = 452°C

fsmax,B = fsy,2O"C

0.80

:. Strength of rebar = 0.8 X 460 = 368 N / d

Moment Resistance of Composite Slab at Fire Limit State

Concrete: R, = 6 0 9 X 0.85 X 25 = 1275Oy

for R, = R , neutral axis depth in slab, y = 5.8 mm

Moment resistance:

L J

- Mfi,Rd = 16.8 kNm/rib

*'- Mfi,Rd ' Mp

The moment resistance of the slab is adequate for 120 minutes fire resistance.


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APPENDIX B: SECTION FACTORS OF STEEL MEMBERS


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t

Table B . l Universal beams

Desigl

Serial size mm

Y14X419

Y14X305

838x292

762 x 267

686 x 254

610x305

610x229

533x210

457X l91

457x152

406X 178

406 X 140

356X 171

356X 127

305 x 165

305 x I27

305 X 102

254x 146

254x 102

203 x I33

2U3x 102 17xx 102 152XXY 127x76

Ion dass p metre kg

343 388

289 253

20 1 224

226 I94 176 I97 I73 147

152 170

140 125 238 I79 149 140 125 113 101 122 109 101 92 82

89 98

82 74 67 82

67 74

60 52 74 67 60 54 46 39 67 57

45 51

33 39

46 54

40 48 42 37

28 33

25 43

31 37

2x 25 22 30 25 23 I Y 16 13

- - -

Depth of

section D

mm 920.5 911.4 926.6 918.5 910.3 903

840.7 850.0

834.9 769.6 762 753.9 692.9 687.6 683.5 677.9 633 617.5 609.6 617 61 1.9 607.3 602.2 544.6

536.7 539.5

533. I 528.3

463.6 467.4

460.2 457.2 453.6 465.1

457.2 461.3

449. x 454.7

409.4 412.8

402.6 406.4

402.3 397.3

358.6 364

355.6 352 352.8 248.5 310.9 307.1 303.8 310.4 3M.6 303.8 312.7

304.8 259.6 256 251.5 260.4 257 254 206.X 203.2 203.2 177.X 152 4 127

- -

308.9

- Width

of section

B mm

418.5 420.5

307.8 305.5

303.4 3 0 4 . 1

293.8 292.4 291.6 268

265.3 266.7

255.8 254.5 253.7 253 311.5 307 304.8 230.1 229 228.2 227.6 211.9

210.1 210.7

209.3 208.7

192 192.8

190.5 191.3

189.9 153.5

151.9 152.7

152.9 152.4

178.8 179.7

177.6

141.8 142.4

173.2 172.1 171.5 171

125.4 I26

165.7 165.1 125.2 124.3 123.5

101.9 102.4

101.6 147.3

146. I 146.4

102. I 101.9 101.6 133 X 133.4 1 0 1 .h 1 0 1 .h XX.Y 7h.2

- -

177.8

166.8

T - Th

We1 t

mn

21.1 19.l 19.1 17.:

15.; 15.l

14.; 16.1

14 15.t

12.5 14.:

13.; 14.f

12.1 11.; 18.t 14.1 11.5 13.1 11.5 11.; 1O.f 12.t 11.t 10.5 10.2 9.6

1 1 . 4 10.6 9.9 9.1 8.5

10.7 9.9 9.1 8.0 7.6 9.1 8.8 7.8 7.6 6.9 6.3 9.1 8 7.3 6.9 6.5 5.9 7.7 6.7 6 . I 9.9 8 7.2

6.1 6.6

5.8 7.3

h. 1 6.4

6. I 6.4

5.8 6.3 5.x 5.2 4.7 4.6 4.2

- -

-

ness %ng

T mm

-

-

T

36.6 32.0 32.0 27.9 23.9 20.2 26.8 21.7 18.8

21.6 25.4

17.5 23.7 21 .o 19 .o 16.2 31.4 23.6 19.7 22.1 19.6 17.3 14.8 21.3 18.8 17.4 15.6 13.2 19.6 17.7 16.0 14.5 12.7 18.9 17.0 15.0 13.3 10.9 16.0 14.3 12.8 10.9 11.2 8.6

15.7 13.0 11.5 9.7

10.7 8.5

13.7 11.8 10.2

12. I 14.0

10.7 1o.x 8.9 6.8

12.7 I0.Y X.6

1 0 . 0

0. X x.4

Y . 6 7.x 0.3 7.Y

7.6 7.7

Area of

section m 2

437.4 494.4

368.8 322.8 285.2 256.4

288.7 247.1 224.1 250.7 220.4 188.0 216.5 193.8 178.6 159.6 303.7 227.9 190.1 178.3 159.5 144.4 129. I 155.7 138.5 129.7 117.7 104.4

113.9 125.2

104.5 94.98 85.44

94.99 104.4

85.41

66.49 75.93

85.49 94.95

76.01 68.42

49.40 58.96

85.42 72.18 64.58 56.96 49.40 41.83 68.38 58.90 51 S O 60.83

47.47 41.77 36.30 31.39 55. I O 47.45 40.00 36.19 32.17 28.42 38.00 32.31 29 24.2 20.5 16.8

- -

53.18

Section factor H,/A (AN) I Box Profile

3 sides

m-’ 60 70 75 85 95

105 85

100 110

105 90

120 95

110 115 130

90 70

1 IO 105 115 I30 I45 110

130 120

I40 155

130 120

140 155 170 130

155 140

175 200

155 I40

I90 I75

205 240 140 165 185 210 215 250 1 6 0

210 185

1 6 0 180 200

245 215

285 I70

230 I95

220 245 275 210 240 23s 265 270 275

4 sides

70 80 80 95

105 I15

115 95

125 100

135 115

110 120 130 145

105 80

125 120 130

160 145

120 135

160 145

175 135 145 160 175 190 145 155

195 I75

220 160 175

215 195

230 270 160 I90 210 240 240 280

210 185

240 180 205 225 240 275 315

250 280 315 245 285 270 305 310 320

t 3 sides

m-l 45 50 60 65 75 80 70

90 50

80 70

95 75 85 90

100 50

80 70

90 80

100 1 IO 85 95 l00 110 120 90

100 105 1 l5 130 105 1 l5 125 140 1 6 0 105 115 130 145

190 160

105 125 135 I55 I70 195 I I5 130 150

140 I25

155

200 I75

225 120

1 6 0 140

170

215 1 9 0

I45 I 65 I75 1 9 0 1 9 0 195

4 sides

m-l

60 55

65 75

95 85

80 90

100 85 95

110 90 95

105 1 l 5 60 80 95 95

105 115 130 95

1 IO 1 l5 125 140 105 115

135 125

150 120 130 145

180 160

125 140 155 170 185 220 125 145 165 185

225 195

140 160 180 145 160 180 200 225 260

210 1 8 0

210 230 235 240


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Table B.2 Universal columns

mign serial Size

mm 356 x 406

356 x 368

305 x 305

254 x 254

203 x 203

152x 152

ion

pc metre - kg 634

467 55 l

340 393

287 235

177 153 129 283 240 I 98 158 137 118 97 I67 132 107 89 73 86 71 60 52 46

37 30 23

202

T Depth of

d o n D mm 414.7 455.7 436.6 419.1 406.4 393.7 381.0 374.7

362.0 368.3

355.6 365.3

339.9 352.6

327.2 320.5 314.5 307.8 289.1 276.4 266.7 260.4 254.0 222.3 215.9 209.6 206.2 203.2 161.8 157.5 152.4

Width of

d o n B mm 424.1 418.5 412.4 407.0 403.0 399.0 395 .O 374.4

370.2 372.1

368.3 321.8

314.1 317.9

310.6 308.7 306.8 304.8 264.5 261 .O 258.3 255.9 254.0 208.8 206.2

203.9 203.2 154.4 152.9 152.4

m .2

-

- m Web

t

- mm 47.6 42.0 35.9

26.5 30.6

22.6 18.5 16.8 14.5 12.6 10.7 26.9 23.0 19.2 15.7 13.8 11.9 9.9 19.2 15.6 13.0 10.5 8.6 13.0 10.3 9.3 8.0 7.3

6.6 8.1

6.1 -

- ness npne

T

- mm 77.0 67.5 58.0

42.9 49.2

36.5 30.2

23.8 27.0

17.5 44.1 37.7 31.4 25.0 21.7 18.7 15.4 31.7

20.5 25.3

17.3 14.2 20.5 17.3 14.2 12.5 11.0 11.5 9.4 6.8

20.7

-

Area of

seaion m* 808.1 701.8 595.5

432.7 500.9

366.0 299.8 257.9 225.7 195.2 164.9 360.4

252.3 305.6

201.2 174.6 149.8 123.3 212.4 167.7 136.6 114.0 92.9 110.1 91.1 75.8 66.4 58.8 47.4 38.2 29.8

Section factor HJA (AIV) PI

3 sides

25 30

40 45 50 65 70 80

LO5 90

45 50 60 75 85

100 l20 60 75

I10 90

130 95

35

110 130 150 165 160

245 195

ile I 1 4 sides 3 sides

I

m-1 I m-'

40 E l 2 0 20 I5

:i I 30 25

65 30 75 I 40

.~

95 85

110 I 30 55 60 75 90

75 90

I 1 0 130 1 6 0 I 1 0 135 160

200 1 8 0

235 l90

300

45 50 55 65 30

40 35

50

60 55

75 40

60 50

70 80 60 70 80

105 95

100 120 155

l

K

4 sides

m -I

20 25 30

35 35

45 50 60 65 75 90 40 45 50 65 70

100 85

50 65 75

110 80

90

95 110 r25 140

~~

135 1 6 0 205


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Table B.3 I Section ~ ~-

Circular hollow 0 I E;; HpIA sections

Profile or Box

Desig

Outside liameter

D mm

21.3 26.9 33.7

42.4

48.3

60.3

76. I

88.9

114.3

139.7

168.3

193.7

219.1

3.2 3.2 2.6 3.2 4.0 2.6 3.2 4.0 3.2 4.0 5.0

4.0 3.2

5.0

4.0 3.2

5.0 3.2 4.0 5.0 3.6 5.0 6.3 5.0 6.3 8.0

10.0 5 .o 6.3 8.0

10.0 5 .o 6.3 8.0

10.0 12.5 16.0 5.0 6.3 8.0

10.0

16.0 12.5

20.0

370 355 415

285 345

410 340 275 335 270 22 5 330 270 220 325 265 215 325 260 210 285 210 I70 205 165 135 I I O 205 I65

105 I30

205 165

105 I30

85 70

205 165 130 105

65 85

55

Table B.3 continued

Desil

Outside diameter

D mm

244.5

273.0

323.9

355.6

406.4

457.0

508.0

6.3 8.0

10.0 12.5

20.0 16.0

25.0 6.3

10.0 8.0

12.5 16.0 20.0 25.0 6.3 8.0

10.0

16.0 12.5

20.0 25.0 8.0

10.0 12.5 16.0 20.0 25.0 10.0 12.5 16.0 20.0 25.0 32.0 10.0 12.5 16.0 20.0

32.0 25.0

40.0 10.0 12.5 16.0 20.0 25.0 32.0 40.0 50.0

<nation ~ Mass

metre Thickness per

-- --

41.4 52.3 64.9 80.3 101 125 I53

68.6 85.2

1 0 6 I34 166 204

97.8 I2 I

191 I54

235 295 I IO 137

216 174

266 335 41 I 123

1 9 4 I53

24 I 298 376 462 565

Area of

section

cm2

47. I 59.4 73.7 91.1 I15

172 141

52.8 66.6 82.6 102 129 I59 I95

62.9 79.4 98.6 122 I55 191 235

87.4 1 0 9 135 171 21 I 260 125 I55

243 196

300 376 140 I75 222 275

427 339

524 I56

247 195

3 79 307

479 588 719

Section factor HpIA W V )

Profile or Box

I65

105 I30

85

160 I30 105 85

1 0 0 80 65 55 45 35

105 80 65

40 50

35 25

1 0 0 80 65 50


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Table B.4 IC '! *I

hollow sections ~1 -m I

Rectangular

(square) __ l-

40x40

50x50

60x60

70x70

80x80

90x90

l00X l o o

l2OXl20

140x140

2.5 3.0 3.2 4.0 5.0

3.0 2.5

4.0 3.2

6.3 5.0

3.0 3.2 4.0 5.0 6.3

3.0 3.6

6.3 5.0

3.0 3.6 5.0 6.3 8.0 3.6 5.0 6.3 8.0 4.0 5.0 6.3

10.0 8.0

5.0 6.3

10.0 8.0

12.5

6.3 5.0

x.0

8.0

Designation

Size

mm mm

t DXD Thickness

,

10.0 8.0

12.5

Mass

metre W

kg 2.92 3.45

4.46 3.66

5.40 3.7 1 4.39 4.66

6.97 5.72

8.49 5.34 5.67 6.97 8.54

10.5 12.8 6.28

10.1 7.46

12.5 15.3 7.22 8.59

11.7 14.4 17.8 9.72

13.3 16.4 20.4 12.0 14.8 18.4 22.9 27.9

22.3 18.0

27.9 34.2 41.6 21.1 26.3 32.9 40.4 49.5

Area of

section

cmz

3.72 4.40 4.66 5.M 6.88 4.72 5.W 5.94 7.28

10.8 8.88

6.80 7.22 8.88

10.9 13.3 16.3 8.00 9.50

12.9

19.5 15.9

9.20 10.9 14.9

22.7 I 8.4

12.4 16.9 20.9 25.9 15.3

23.4 18.9

29. I 35.5 22.9 28.5 35.5 43.5 53.0 26.9

41.9 33.5

63.0 51.5

1 Section factor Hp/A

m-'

275 325

260 210 175 320 270 255 205 I70 140 265 250 205 165 135 I IO 265 220 I65 I30 I I O 260 220 l 6 0 I30 I 05 220 160 I30 I 05 I95 l 6 0 I30 105 85

I55 I25 1 0 0 85 70

I 55 125 1 0 0 80 65

430 365 345 280 235 425 355 335 275 225 185 355 330 270 220

I45 I80

3 50 29.5 215 175 I45 350

215 295

175 1 4 0 290 21s 170 1 4 0 260 210

135 I70

I 15 210 170 135 I10 90

210 l65

1 IO I35

90

t

Table B.4 continued

Designation

ISOX150 6.3

10.0 8.0

12.5 16.0

160x160 5.0 6.3

12.5 180X 180 6.3 I 1:::: 1 16.0

123

200x200 5.0 6.3 l 1:::: I 16.0

12.5

250x250

10.0 8.0 6.3

16.0 12.5

300 x 300 6. 3 x.0

10.0 I 12.5 16.0

350x350 10.0 8.0

12.5 16.0

4oox400 I 1::: 1 20.0 16.0

- Mass

metre Per

kg 22.7 28.3

43.6 3 5.4

66.4 53.4

24.2 30.3

46.7 37.9

57.3 34.2 43.0 53.0 65.2 81.4 30.5 38.2 48.0

73.0 59.3

91.5 48. I 60.5 75.0 92.6 117

57.9 73.1

I I2 I42

85.7 1 0 6 l32 I67

I 52 I22

I92 237

90.7

Section factor HpIA ( A m

28.9

45. 1 36.0

68.0 55.5

84.5

38.5 30,9

48.3

73.0 59.5

43.6 54.7 67.5 83.0 1 0 4

38.9 48.6 61.1

93.0 75.5

l l7 61.2

95.5 77. I

I18 I49

93.1 73.8

116 143 181 1 0 9 136

213 168

I56

245 193

302

I 55 125 100 80 65 5 5

I 5s 125 1 0 0 x0 65

125 1 0 0

65 50

I 55 125 100 80 65 50

125 95 80 65 50

I20 95 80 65 5 0 9s 75 65 50 75 60

40 50

8n

4 sides r - - - - - l $01 L- - - ---J

m-'

210 165 135 I10 90 70

205 I65 135 I 10 90

I65 I30 105

70 205 165

10s I30

x5 70

165 I30 I 05 85 65

I65 I30 105 85 65

I30 105 85 65

l 05 85 65 55

x5

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T Table B.5 Rectangular hollow sections

Designation

DXB Size

mm 50X 30

60x40

80x40

90x50

l00X 50

100x60

120x60

120x80

l50X 1 0 0

rhickness t

mm

3.0 2.5

4.0 3.2

5 .O 2.5 3 .O

4.0 3 2

.50 6.3 3.0

4.0 3.2

5.0 6.3 8.0 3.0 3.6 5 .O 6.3 8.0 3.0

4.0 3.2

5.0 6.3 8.0 3.0 3.6

6.3 5.0

8.0 3.6 5.0 6.3 8.0

6.3 5.0

10.0 8.0

5.0 6.3

10.0 8.0

12.5

Mass per

metre

kg

3.45 2.92

4.46 3.66

5.40 3.71 4.39 4.66

6.97 5.72

8.49 5.34

6.97 5.67

8.54 10.5 12.8 6.28 7.46

10.1 12.5 15.3 6.75 7.18 8.86

10.9 13.4 16.6

8.59 7.22

11.7 14.4 17.8

13.3 9.72

20.4 16.4

14.8 18.4 22.9 27.9 18.7 23.3 29.1 35.7 43.6

Area of

section

cmz 3.72 4.40 4.66

6.88 5.68

4.72 5.60

7.28 5.94

8.88 10.8 6.80 7.22 8.88

10.9 13.3 16.3 8.00 9.50

12.9 15.9 19.5 8.60 9.14

11.3 13.9

21.1 17.1

10.9 9.20

18.4 14.9

22.7 12.4

20.9 16.9

25.9 18.9 23.4 29.1 35.5

29.7 23.9

37. I 45.5 55.5

Section factor HpIA (AIV) l 4 sides

I

295 350

280 230 1 9 0 340

270 285

220 180 I50 295 275 225 185 150 125 290 240 I80 145 120 290 275 220 180 145 I20 285 240 175 140 115 240 I80 145 115 1 70 135 110 90

I65 135 I10 90 70

'S

m-' 295 250 235 195 1 6 0 295 250 235 1 9 0 1 6 0 130 235 220 I80

I20 I45

1 0 0 240 200 145 120 95

235 220

145 I75

115

240 95

200

I20 I50

95 195 140 115 95

I50 I20 95 80

145 l 20 95 75 65

m-#

430 365 345 280 235 425

335 355

225 275

185 355 330 270 220 180 145 350 295 215 I75 I45 350 330 265 215 175 140 350 295 215 I75 140 290 215 170 1 4 0 210 170 135 115 210 170 I35 I10 90

Table B.5 continued overleaf


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Section factor HpIA (A/W Table B.5 3 sides 4 sides

continued

Ih0X80 5.0 6.3 8.0

10.0 12.5

200x 1 0 0 6.3 5.0

8.0 10.0 12.5 16.0

200x I20 5.0 6.3

10.0 8.0

12.5 250X 150

6.3 5.0

10.0 8.0

12.5 16.0

300x200 6. 3

10.0 8.0

12.5 16.0

10.0 12.5 16.0

500x200 10.0 8.0

12.5 16.0

500x300 10.0 12.5 16.0 20.0

175 1 4 0 115 90 75

175 1 4 0 I 10 90 75 h0

I70 135 I IO 85 70

I 65 135 105 X5 70 55

I30 105 X5 70 5 5

105 X 5 70 5 5

105 X5 70 55

I I O 90 70 55 X5 65 55 45

t 1 4 0 I10 90 75 m

1 4 0 I10 90

h0 70

45 1 4 0 I15 90 75 h0

1 4 0 I15 90 75 m 45

115 90 75 60 45 X5 70 55 45 8s 70 55 45

65 55 40 70

45 55

35

xs

210 I70 135 1 IO 90

210 I65 I35 I IO 90 70

205 165 I35 I10 90

205

I30 165

105

70 85

165 I30 105 R5 65

I20 105

65 85

I20 I os 85 65

I30 105 85 65

105 x5 65 55


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