Nipigon River Bridge West Abutment Technical · PDF filePublication Title Nipigon River Bridge West Abutment Bearing Technical Investigation Author(s) Kris Mermigas , Walter Kenedi,

Ministry of Transportation Highway Standards Branch

Bridge Office Report

Nipigon River Bridge West Abutment Bearing Technical Investigation

BRO-059

Publication Title


Author(s) Kris Mermigas, Walter Kenedi, David Lai, Ontario Ministry of Transportation

Originating Office Bridge Office, Highway Standards Branch, Ontario Ministry of Transportation

Report Number BRO-059; ISBN: 978-1-4606-8708-6

Publication Date September 2016

Ministry Contact Bridge Office, Highway Standards Branch, Ontario Ministry of Transportation

301 St. Paul Street, St. Catharines, Ontario, Canada L2R 7R3

Tel: (905) 704-2406; Fax: (905) 704-2060

Abstract On January 10, 2016 at 3:05 pm, the Nipigon River Bridge was closed to traffic. The bridge became impassable after the failure of 40 (7/8” ASTM A490) bolts at the northwest bearing caused the bridge to lift approximately 600 mm at the northwest corner. This report summarizes the Ministry of Transportation’s technical investigation into the cause of failure, including bolt testing. Factors dealing with management of the project are not the subject of this report.

The structural analysis of the bearing and its connections to the bridge revealed that the failure was caused by prying of the flexible shoe plate the bearing’s inability to accommodate rotation, combined with improper installation of the bolts on site (snug-fit tightening of nuts without bevelled washers). Additional factors which contributed to and accelerated the failure include local bending of the bolts and yielding of the shoe plate.

Key Words Cable-stayed bridge; uplift bearing failure; bolt failure

Distribution Unrestricted technical audience.

Technical Report Documentation Page

Ministry of Transportation Highway Standards Branch

Bridge Office Report

BRO-059

Nipigon River Bridge

West Abutment Bearing Technical Investigation

19 September 2016

Prepared by

Bridge Office Ontario Ministry of Transportation

301 St. Paul Street,

St. Catharines, Ontario, Canada L2R 7R3 Tel: (905) 704-2406; Fax (905) 704-2060

Published without prejudice as to the application of the findings. The Ministry of Transportation for Ontario (MTO) wishes to advise that this Report is being released to provide the factual background and potential causes for the January 10, 2016 occurrence involving the Nipigon River Bridge on Highway 11/17. These Reports are not intended to ascribe fault or liability to any particular party nor are the findings intended to be definitive. It will be necessary for the MTO to carry out further investigation in order to finally determine both causation for the occurrence and the party or parties responsible for the damages incurred as a result of the occurrence.

Crown copyright reserved.

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Table of Contents List of Figures .............................................................................................................................................. ii

List of Tables .............................................................................................................................................. iii

Appendices ................................................................................................................................................. iv

1. Executive Summary........................................................................................................................ 1

2. Introduction ..................................................................................................................................... 2 2.1. Purpose of the Investigation ............................................................................................................. 2 2.2. Methodology ..................................................................................................................................... 2

2.3. Qualifications of the Authors ............................................................................................................. 3

3. Bridge Behaviour ............................................................................................................................ 4 3.1. Global Behaviour of Nipigon River Bridge ........................................................................................ 5 3.2. Load Path from Back Stays to West Abutment ................................................................................. 7

4. Bearing Requirements ................................................................................................................... 9 4.1. Bearing Drawings ............................................................................................................................. 9 4.2. Bearing Forces ............................................................................................................................... 11

4.3. Shoe Plate ...................................................................................................................................... 12 4.4. Bearing Assembly ........................................................................................................................... 13

5. Observations and Testing ............................................................................................................ 16

5.1. Physical Observations .................................................................................................................... 16

5.2. Examination and Testing of Fractured Bolts ................................................................................... 20

5.2.1. Fractured Bolts from NW bearing ................................................................................................... 21 5.2.2. Intact Bolts from Centre-West Bearing ........................................................................................... 27 5.2.3. Conclusion Based on Test Results ................................................................................................. 30

5.3. Differences between Contract Drawings and Supplied Bearings .................................................... 31 5.3.1. Bolt Pre-Tension ............................................................................................................................. 31 5.3.2. Shoe Plate Material ........................................................................................................................ 31

5.3.3. Bolt Pattern between the Girder to the Shoe Plate ......................................................................... 32 5.3.4. Bolt Length and Washers ............................................................................................................... 33

6. Evaluation of Bearing and Attachments ..................................................................................... 35 6.1. Evaluation of the Bearing ................................................................................................................ 35 6.1.1. Design ............................................................................................................................................ 35

6.1.2. Rotation .......................................................................................................................................... 35 6.1.3. Uplift Guide Bar PTFE Evaluation .................................................................................................. 40 6.1.4. Uplift Guide Bar Capacity Evaluation .............................................................................................. 41 6.1.5. Shoe Plate Attachment to Top of Bearing ....................................................................................... 41 6.2. Evaluation of Shoe Plate ................................................................................................................ 42 6.3. Evaluation of Bolts .......................................................................................................................... 43 6.4. Evaluation of Prying in Shoe Plate ................................................................................................. 45

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6.4.1. Model of Shoe Plate Prying ............................................................................................................ 47 6.4.2. Analysis Results ............................................................................................................................. 51 6.4.3. Prying of Outer 1” Bolts Connecting the Shoe Plate to the Bearing ................................................ 51 6.4.4. Prying of the 7/8” Bolts in Phase 1 ................................................................................................. 53

6.4.5. Prying of the 7/8” Bolts At Centre-West Bearing Design Loads ...................................................... 55 6.5. Evaluation of Bolt Fatigue ............................................................................................................... 57

7. Discussion .................................................................................................................................... 61 7.1. Design and Construction Requirements ......................................................................................... 62

8. Conclusions .................................................................................................................................. 68

9. References .................................................................................................................................... 69

List of Figures Figure 1. Response of cable-stiffened, girder-stiffened, and tower-stiffened cable-stayed bridges,

shown with typical proportions of span and girder. ........................................................................... 4 Figure 2. Typical cable arrangements in cable-stayed bridges. ..................................................................... 5 Figure 3. Structural behaviour due to dead load: a) at end of balanced cantilevering, and b) at the end

of construction. ................................................................................................................................. 6 Figure 4. Structural behaviour due to the passage of a truck over a) east span, and b) west span. .............. 7

Figure 5. Load path from Stays to West Abutment in elevation and section. ................................................. 8 Figure 6. Rotations on bearing. ................................................................................................................... 10 Figure 7. Rotational Bearing Design Data Table from the Contract Drawings. ............................................ 11

Figure 8. Shoe plate bolts specified on A-10, revision C (sheet 218-1-R2). ................................................ 12

Figure 9. Shoe plate bolts revised within RFC-176, dated January 31, 2014. ............................................. 13 Figure 10. Components of the west abutment bearing assembly. ............................................................... 14 Figure 11. Northwest Bearing installation on October 5, 2015: a) sliding the sole plate of the bearing

into place after the shoe plate has been connected to the bearing flange and b) washers stacked 3 high over west half of shoe plate and 4 high over east half of the shoe plate. ................ 16

Figure 12. Northwest Shoe Plate with Failed Bolts of Shoe Plate Failed Bolts Looking East ...................... 17 Figure 13. Northwest Shoe Plate with Failed Bolts – note gap between shoe plate and bearing top

plate towards the middle ................................................................................................................. 17 Figure 14. Bolt numbering and premature bolt failure (north is up). Red indicates confirmed early

failure, blue indicates possible early failure. ................................................................................... 18 Figure 15. Top Surface of the Shoe Plate with Failed Bolts (looking down, north is up) .............................. 19 Figure 16. Northwest Bearing, Northwest corner of uplift restraint with PTFE crushed but still attached

to the bearing masonry plate (top steel surface of the photo) ......................................................... 19 Figure 17. Northwest Bearing, Southwest corner of uplift restraint with PTFE crushed and projecting

beyond the masonry plate (top steel surface of the photo) ............................................................. 19 Figure 18. Bolt sample from southwest bearing. .......................................................................................... 20

Figure 19. Fractured Surface of Failed Bolts with Visible Corrosion Product ............................................... 22 Figure 20. Northwest bearing looking East, after failure on January 10, 2016. ............................................ 23 Figure 21. Transverse crack on elongated side of bolt Z. ............................................................................ 23

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Figure 22. Bolt at the southeast corner of the shoe plate (location 40) which failed at the shank, with steel-to-steel rub markings on opposite sides of the shank. ........................................................... 24

Figure 23. Cylindrical specimens cut from a bolt shank by EDM at SSW. ................................................... 25 Figure 24. Images of intact bolts with distorted threads. .............................................................................. 28

Figure 25. Shoe plate bolts specified in RFC 176. ....................................................................................... 32 Figure 26. Bottom view of shoe plate with bolts as supplied per drawing Bearing Details-7, revision 8. ..... 32 Figure 27. Counterbore specified on A-10, revision C (sheet 218-1-R2). .................................................... 33 Figure 28. Counterbore detailed per drawing Bearing Details-7, revision 8. ................................................ 33 Figure 29. Separation of guide bars due to longitudinal rotation. ................................................................. 36 Figure 30. Vertical reaction and rotation at northwest bearing due to the CL-625-ONT Truck..................... 37 Figure 31. Amplification of bolt force due to uplift and longitudinal rotation. ................................................ 38

Figure 32. Separation of guide bars bearings due to a rotation of 0.8° in the axis of the girder. ................. 39

Figure 33. Shear Force and Moment Diagrams for Shoe Plate; a) assuming load evenly shared by all bolts, b) with prying force, and c) assuming all load through exterior line of bolts. ......................... 43

Figure 34. Local Prying of bolt due to lack of beveled washer. .................................................................... 44 Figure 35. Classical Prying of flexible plate in tension connection (Kulak, et al., 1987). .............................. 45 Figure 36. Nipigon River Bridge northwest bearing shoe plate prying under uplift. ...................................... 46 Figure 37. 3-dimensional finite element model of girder end and bearing in a) isometric view, and b)

sectional view. ................................................................................................................................ 47

Figure 38. Deformed bearing assembly due to longitudinal rotation combined with uplift. ........................... 48 Figure 39. Deformed bearing assembly due to transverse rotation combined with uplift. ............................ 49 Figure 40. Shoe plate strip model - properties. ............................................................................................ 49

Figure 41. Shoe plate strip frame model for pretensioned 7/8” bolt behaviour. ........................................... 50 Figure 42. Force in 1" bolts under phase 1 and design loads. ..................................................................... 52

Figure 43. Forces in 7/8" bolts for phase 1 (failure loads at northwest bearing). ......................................... 54 Figure 44. Forces on 7/8" bolts under design loads (centre-west bearing). ................................................. 56 Figure 45. Possible force distribution in bolts, after plasticity. ...................................................................... 59

List of Tables Table 1. Bearing Contract Drawing Revisions ............................................................................................... 9

Table 2. Bearing Reactions for Phase 1 (Failure Reactions) ....................................................................... 12 Table 3. Shoe Plate Deformation (Maximum Gap Measured between the Shoe and Sole Plates of

West Abutment Bearings and Maximum Deformation after Shoe Plates Removed) ...................... 17 Table 4. Tensile Test Requirements and Results of Specimens Machined from Fractured Bolts ................ 25 Table 5. Charpy Impact Test Results of Fractured Bolts ............................................................................. 26

Table 6. Summary of Tensile Test Results of Full Size Bolts ...................................................................... 29 Table 7. Tensile Test Results of Machined Specimens from Intact Bolts of Centre Bearing........................ 30 Table 8. Design Contact Pressure for PTFE Sliding Surfaces of Uplift Restraint Guide Bars ...................... 40 Table 9. Comparison between Shoe Plate Specified and Shoe Plate Provided .......................................... 42

Table 10. Bolt Forces and Capacities .......................................................................................................... 45 Table 11. Bolt Forces from Structural Model Based on Elastic Analysis (kN) .............................................. 51

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Table 12. Design and Construction Requirements for Each Component of Work Related to the West Abutment Bearings of the Nipigon River Bridge ............................................................................. 63

Appendices

Appendix A: Contract Drawings ..................................................................................................................... A Appendix B: Bearing Working Drawings ........................................................................................................ B

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1. Executive Summary On January 10, 2016 at 3:05 pm, the Nipigon River Bridge was closed to traffic. The bridge became impassable after the failure of 40 (7/8” ASTM A490) bolts at the northwest bearing caused the bridge to lift approximately 600 mm at the northwest corner. This report summarizes the Ministry of Transportation’s technical investigation into the cause of failure. Factors pertaining to the management of the project are not the subject of this report. The technical investigation into the failure involved;

1. testing of the bolts, 2. structural analysis of the northwest bearing and the associated connections to the bridge

girders and abutments, 3. evaluation of components in the load path from the girder to the west abutment

according to the Canadian Highway Bridge Design Code (CHBDC) requirements. The bolt testing was carried out at two independent laboratories, the National Research Council (NRC) in Ottawa Ontario and Surface Science Western (SSW) at Western University in London Ontario. Each laboratory issued comprehensive reports and their findings were reviewed as part of this report. The testing revealed that the bolts met the requirements of applicable standard ASTM A490 and the CHBDC requirements for use of steel in cold weather and were therefore not the reason for the failure at the northwest bearing. Detailed examinations of the bolt failure surfaces by the above laboratories, as well as visual inspection by the Ministry, found that the failure surfaces had striations consistent with low-cycle high-stress bolt failure. In addition, corrosion was observed on some of the failure surfaces, indicating that the failure was progressive and began prior to January 10, 2016. The structural analysis of the bearing and its connections to the adjacent components of the bridge revealed that the failure was caused by;

1. prying effects due to the flexible shoe plate leading to higher forces in the exterior line of bolts,

2. the bearing’s inability to accommodate rotation leading to higher forces in the end rows of bolts,

3. the lack of pretensioning of the bolts and lack of bevelled washers that lead to high fatigue stresses and a high-stress, low-cycle fatigue failure.

Each of these of factors on its own is significant and could have led to a failure, but combined they made failure inevitable. Other factors which also contributed to and accelerated the failure include local bending of the bolts and yielding of the shoe plate. The evaluation showed that the shoe plate, bolted connection between shoe plate and girder, bolted connection between shoe plate and bearing, and bearing design all failed to meet the requirements of the CHBDC.

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2. Introduction The existing two lane Nipigon River Bridge, located on highway 11/17 east of Thunder Bay, is being replaced with a new four lane cable-stayed bridge. In 2013, a $106 M dollar contract was awarded to Bot-Ferrovial Nipigon Joint Venture (BFNJV). On November 29, 2015, the newly built westbound lanes were opened to two way traffic and removal of the existing bridge began. This crossing is a strategic link in the Trans-Canada highway system. On January 10, 2016 at 3:05 pm, the Nipigon River Bridge was closed to traffic. The bridge became impassable after the failure of 40 bolts at the northwest bearing which caused the bridge to lift approximately 600 mm at the northwest corner. Ministry staff and the contractor worked overnight to level the bridge surface and the structure. Temporary concrete barriers (TCBs) were placed in the westbound lane, close to the expansion joint, as ballast to bring the superstructure back to its initial position. The bridge was opened to one lane of traffic the morning of January 11.

2.1. Purpose of the Investigation

The Assistant Deputy Ministers Office requested that the MTO’s Bridge Office review the design and construction to determine the cause of the failure, and determine factors that could have contributed to the failure of the northwest bearing. The objectives of the investigation are the following.

1. Establish the cause of the northwest bearing failure. 2. Evaluate the ability of all components, in the load path from the girder to the west

abutment, to meet the design requirements.

2.2. Methodology

The report summarizes the findings of the structural investigation into the cause of the failure, and assesses the ability of the remaining bearings to meet the design requirements. The design consultant has changed names due to corporate acquisitions during the project. This report will refer to the design consultant as Marshall Macklin Monaghan Limited (MMM), formerly McCormick Rankin Corporation. Similarly, the Contractor’s erection engineering firm has changed names and will be referred to as McElhanney Consulting Services Limited (McElhanney), formerly Infinity Engineering. As part of this investigation, failed bolts from the northwest bearing and intact bolts from the centre-west bearing were tested to determine compliance with the contract requirements. Northwest Region delivered specimens of the failed bolts to the Bridge Office on January 15, 2016. The Bridge Office examined the bolts the week of January 18. During the same week, the MTO retained Surface Science Western (SSW) at Western University and the National Research Council (NRC) Canada to conduct testing of the bolts. The testing covered physical tests including tensile yield and ultimate strength tests, impact notch-toughness testing,

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microscope examination of the failure surfaces, and chemical testing of the bolt material. The report is organized according to the Professional Engineers of Ontario Guideline on Forensic Engineering Investigations. The report is based on evidence found in drawings, addenda, specifications, bid enquiries, Working Drawings, correspondence and records, except where noted otherwise.

2.3. Qualifications of the Authors

Kris Mermigas, P.Eng. prepared this report. Kris is Head of Bridge Management with the MTO’s Bridge Office, is the Chair of the MTO’s Expansion Joint and Bearing Working Group, and is a technical subcommittee member of Section 11 (Joints and Bearings) of the Canadian Highway Bridge Design Code (CHBDC). Kris joined the MTO in 2012 as an Engineer in the Rehabilitation Section of the Bridge Office. Prior to joining the MTO, Kris worked for AECOM and LEA Consulting where he designed over 40 bridges, inspected over 400, and evaluated many structures. Kris completed his Master of Applied Science at University of Toronto in 2008. His thesis entitled Behaviour and Design of Extradosed Bridges, explores the influence of different geometric parameters such as tower height, girder depth, and pier dimensions on the structural behaviour, cost, and feasibility of extradosed bridges. Extradosed bridges can be classified as a subset of cable-stayed bridges. Walter Kenedi, P.Eng. completed a refined analysis of the girder end including the bearing. Walter completed his B.A.Sc. and M.A.Sc degrees at the University of Toronto. His undergraduate thesis studied shear lag in bolted connections, while the graduate thesis studied concrete filled HSS connections. Walter is Head of Design with the MTO’s Bridge Office, and is a technical subcommittee member of Section 5 (Methods of Analysis) and Section 14 (Evaluation) of the Canadian Highway Bridge Design Code. As former Head of Evaluations, Walter has evaluated over 200 bridges, and inspected over 500. He was the lead analyst in the structural evaluation of the causes and sequence of failure on the Sgt Aubrey Cosens Memorial (Latchford) Bridge. David Lai, P.Eng. lead the bolt testing investigation. David Lai completed his undergraduate and graduate study in civil engineering at McGill University. After graduation, he had worked on many well-known civil/structural engineering projects, including the design of the BCE Place in Downtown Toronto, the Scarborough Consilium, the reinforced concrete dome roof for the Metro Toronto Council Chamber, and the supervision of the subway construction in Singapore. He worked for T.Y.Lin Consulting Engineers in Singapore, and was an associate with Yolles Partnership in Toronto prior to joining the Ministry of Transportation of Ontario in 1990. Mr. Lai is currently the Head of the Bridge Rehabilitation Section of MTO responsible for the development of all provincial policies and standards for bridge rehabilitation and durability. He is the current chair of Section 15 (Rehabilitation) of the Canadian Highway Bridge Design Code, and a committee member of Section 16 (FRP). He is also a committee member of CSA S807 and S808.

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3. Bridge Behaviour There are three different approaches to providing stiffness in cable-stayed bridges, as shown in Figure 1, which determines how stability of the bridge is assured under live load. Each approach provides stiffness primarily in one of the three load-bearing elements of the cable-stayed bridge: the stays, the deck, or the towers.

Figure 1. Response of cable-stiffened, girder-stiffened, and tower-stiffened cable-stayed bridges,

shown with typical proportions of span and girder.

Since the 1980s, almost all cable-stayed bridges have been built as ‘multiple-stay’ bridges with cables spaced at the deck level less than 10 m apart. Combined with an increased understanding of aerodynamic stability and buckling safety of slender girders, this has led to slender girders. There are three main cable arrangements in cable stayed construction – fan, harped, and semi-fan, as shown in Figure 2. The fan cable configuration has cables anchored at a single point at the top of the tower, and loads the tower in axial compression only, with backstay cables to stabilise the tower and control girder deflections due to live load. With stiffer towers, the backstay cable forces become less significant. A harp cable configuration has roughly parallel cables with anchorages along the height of the tower and favours stiff towers, since live load at the quarter points of the main span will cause significant bending in towers, regardless of tower stiffness. The semi-fan cable arrangement captures the efficiency of a fan cable arrangement, with a large lever arm between the tension in the cable and the compression in the deck, while allowing cables to be anchored in a spaced out arrangement within a hollow tower. The tower is proportioned to provide stability during cantilever construction.

Stiffness from Backstay Cable Stiffness in Deck Stiffness in Tower

LL resisted by backstay LL resisted by bending in deck

LL offset by dead load in main span. Short back span prevents

backstay from going slack.

LL resisted by bending in deck

LL resisted by bending in the tower

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Figure 2. Typical cable arrangements in cable-stayed bridges.

3.1. Global Behaviour of Nipigon River Bridge

The Nipigon River Bridge is cable-stayed bridge of two spans, three planes of cables, a single pier, and a 37.21 m wide deck. The main span is 139 m, while the back-span is 112.8 m. The pier consists of three prestressed concrete towers rigidly connected to prestressed concrete pier legs and a concrete box girder diaphragm below the deck. The deck consists of three structural steel girders aligned in the plane of the cables, supporting structural steel floor beams at 3.6 m spacing. The structural steel components are composite with a precast concrete deck. The General Arrangement drawing of the bridge is included in Appendix A. The Nipigon River Bridge is half of a classical three-span cable-stayed bridge. It is proportioned with a length of backspan of 82% of the main span (41% when considered as a classical three span cable-stayed bridge), a girder span to depth ratio of 90 (180), and a semi-fan cable-arrangement. Two-span cable-stayed bridges commonly have a backspan which is shorter than the main span. The Nipigon River Bridge is stiffened by the backstay cables at the west abutment, and by the tower. The deck is integral with the tower which provides additional stiffness to control deflections during construction and in the permanent condition. The Nipigon River Bridge was designed to be constructed in balanced cantilever. In each step, a 10.8 m length of girder is installed onto the east span, a stay cable is installed, and the concrete deck is installed and made composite with the steel girders, followed by stressing of the stay. The same operation is repeated to add 10.8 m to the west span, before moving on to the next step of cantilevering. The Nipigon River Bridge has three planes of cables (denoted as north, centre, and south) each supporting a girder. The bridge is built in two halves. In the first phase, the north and centre towers, girders, and cables are constructed to support the north half of the deck system. In the second phase, the south tower, girder and cables are constructed to support the south half of the deck system. In the permanent condition, the north half of the bridge supports the westbound traffic and the south half of the bridge supports the eastbound traffic. After construction of phase 1, one lane of traffic in each direction is allowed on the bridge. It is in this configuration that failure of the northwest bearing occurred. Since the deck is constructed in balanced cantilever, there is practically no reaction at the bearing when the deck reaches the west abutment. At that point in time, the bridge superstructure is balanced about the central pier, cantilevering out 116 m in each direction from the tower. The loads, reactions, and internal forces are shown in Figure 3a. It is at this point that

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the west abutment bearing is installed. As cantilevering of the east span continues, the dead load from the additional length of the east span is balanced by uplift at the west abutment bearings. When the east span lands at the east abutment, there is a permanent uplift at the west abutment bearings due to dead load of the east span, as shown in Figure 3b. Permanent uplift exists at the west abutments bearings which stabilise the bridge.

Figure 3. Structural behaviour due to dead load: a) at end of balanced cantilevering, and b) at the

end of construction.

The reaction at the west abutment bearings, due to the passage of a truck over the bridge, is both positive (compression) and negative (uplift). For a truck travelling westbound over the bridge, the truck first loads the east span of the bridge. The downwards force of the truck is resisted primarily by an upward component of the force in the nearest stay cables. The tension in the cable is balanced by tension in the back stay cables, compression in the tower, compression in the deck, and uplift at the west bearing. The upper portion of the tower bends to balance tension in the stays across the east span by the back stays of the west span. The back stays are the stiffest load resisting system of the bridge, and therefore resist most of the truck load. Figure 4a shows the loads, reactions and forces within the bridge when a truck loads the east span. The downward force of the truck is resisted primarily by the upward component of the force in the nearest east span stay cables. This force is carried to the tower, where the load is carried by tension primarily by the stiffest cable – the back stay that is anchored at the abutment. As the truck travels across the west span, the bridge resists load differently. The downwards force of the truck is resisted primarily by the upwards component of the force in the nearest west span stay cables. Since the cables of the east span are not directly anchored at the abutment, they are more flexible and cannot counterbalance the truck load on west span. Instead, the tension in the stays loaded by the truck is resisted by compression in the back stay, as shown in Figure 4b. Since the back stays have a large tension due to dead load, the compression due to the truck load relieves that tension by only a small amount. This results in a compression on the west abutment bearing due to the live load, although the bearing has a net uplift force due to all loads at serviceability limit states (SLS).

a) b)

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Figure 4. Structural behaviour due to the passage of a truck over a) east span, and b) west span.

3.2. Load Path from Back Stays to West Abutment

The structural stability of the cable-stayed bridge relies on transfer of uplift force from the superstructure to the west abutment. The vertical (uplift) force exerted on the bearing from the back stay cables is resisted by the weight of concrete in the abutment through the following load path.

1. The tension in the back stays of the west span is transferred into the girder through a fin plate inserted through the top flange of the girder, and welded directly to the girder’s web.

2. The force is transferred through the girder’s web to the bottom flange through fillet welds between the web and flange. Vertical bearing stiffeners are welded on both sides of the web and welded to the bottom flange, and assist in transferring the force to the bottom flange.

3. The bottom flange is bolted to a shoe plate. 4. The shoe plate is in turn bolted to the bearing. 5. The bearing is anchored to the abutment by prestressing bars, which are anchored deep

in the abutment in order to engage the weight of the abutment concrete. Figure 5 illustrates the load path from stay cables to abutment in both longitudinal section (view cut through the abutment) and section cut through the bearing facing the west abutment.

a) b)

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Figure 5. Load path from Stays to West Abutment in elevation and section.

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4. Bearing Requirements

4.1. Bearing Drawings

The bearing requirements at the west abutment are described on Contract Drawings A-10 and A-11, included in Appendix A. The drawings were revised through tender and construction as described in Table 1. The bearing drawings were largely unchanged after Addendum 3 issued during tender. Table 1. Bearing Contract Drawing Revisions

Contract Drawing Revision Date Description of Changes

A-10 (Sheet 218) February, 2013

Drawings not sealed.

Initial tender drawing.

A-10 (Sheet 218-A), Addendum 2 April 5, 2013

Drawings not sealed.

Included requirements for post-tensioned anchor rods at the east abutment. Added shoe plate size and thickness, shown welded to bottom flange of girder.

A-10 (Sheet 218-B), Addendum 3

A-11 (Sheet 218-1), Addendum 3

April 30, 2013 Drawings not sealed.

Bearing information split into two drawings. Reactions at fatigue limit states added to the bearing design data table. Size and thickness of shoe plate increased at both abutments and bolted connection between west abutment shoe plate and girder added. East abutment bearing connection to abutment reverted back to anchor rod grouted in formed hole.

June 5, 2013 Tender Opening

A-10 (Sheet 218-B)

A-11 (Sheet 218-1-R1)

July 9, 2013

Seals applied.

Drawings reissued as part of full drawing set for construction, with seals applied. On drawing, A-11 west abutment minimum shoe plate thickness at one end changed from 54 to 52 mm.

A-10 (Sheet 218-B-R1),

A-11 (Sheet 218-1-R2), Instruction Notice #71

December 12, 2013 East abutment bearing anchor rod details changed and east abutment bearing grade of material for anchor rod changed (Note 11). The drawing contains two seals but the checker’s stamp is dated July 11, 2013, prior to the final revision – therefore invalidating it for this revision. Thus, this drawing is technically sealed by only the design engineer, whose seal is dated June 9, 2014.

CAN/CSA-S6-06: Canadian Highway Bridge Design Code (CHBDC) Section 11.6.1 General requires that specific design information be shown on the drawings. The bearing drawings of the Nipigon River Bridge convey the information required by the CHBDC. The drawings show the minimum and maximum loads corresponding to the critical combinations at serviceability limit

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states (SLS) and ultimate limit states (ULS), including dead load, total load, lateral loads, rotations, and translations. Drawing A-10 states rotation requirements of the bearings, in all three axes. Although the table does not state the direction in which the bearing shall rotate in the horizontal axis, it is understood that live load on the main span will cause an upwards rotation at the front of the bearing, and maximum uplift force. Live load on the back span will cause a downwards rotation at the front of the bearing, with a smaller uplift force, since the truck positioned in the back span will apply a downwards force at the west abutment. There is also out-of-plane rotation due to deflection of the end floor beam when the truck is near the end of the bridge. Note 14 on drawing A-10 states that the horizontal rotation shall be about the horizontal axis in all directions (i.e. longitudinal and transverse rotation), which encompasses the aforementioned rotations. Rotation about the vertical axis is the rotation of the bearing in plan. The bearing rotations are shown in Figure 6.

Figure 6. Rotations on bearing.

The CHBDC further states that for bearings other than elastomeric bearings, the bearings shall be designed to accommodate the rotations at ULS plus tolerances in fabrication and installation, plus an additional 1°. This additional tolerance is not referred to on the design drawings, but is covered by the specification OPSS 1203. The rotational bearing data tables for the northwest and centre west bearings are shown in Figure 7.

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Figure 7. Rotational Bearing Design Data Table from the Contract Drawings.

There is a large range of force specified for the reaction under dead load. At SLS, the uplift varies from a minimum of 620 kN to a maximum of 3530 kN for the centre-west bearing and 1320 kN to 1900 kN for the northwest bearing. This accounts for forces from construction of the south half of the bridge, changes in force from before and after the addition of the asphalt wearing surface, and changes due to long-term time-dependent material properties. The centre-west bearing supports the central plane of cables which are loaded with nearly double the weight of the exterior planes of cables. For these reasons, the force demands of the northwest and centre-west bearing are higher than those at the southwest bearing. Note 8 states that: “bearings shall be supplied by one of following manufacturers: Goodco Z-Tech, RJ Watson, Watson Bowman Acme, and Wercholoz Canada.”

4.2. Bearing Forces

The north half of the bridge opened to traffic in November, 2015. The reactions on the bearings for phase 1, when the north half was open to one lane of traffic in each direction, are provided in Table 2. MMM provided ULS total load reactions in an email dated January 25, 9:51 am, 2016. McElhanney provided as-built SLS dead load reactions in an email dated January 13, 9:20 pm, 2016. The remaining reactions were provided by MMM in an email dated February 7, 5:01 pm, 2016.

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Table 2. Bearing Reactions for Phase 1 (Failure Reactions)

Load Combination Centre-West Bearing Northwest Bearing

Serviceability Limit State Dead Load -1077 kN -1635 kN

Serviceability Limit State Total -1625 kN -2045 kN

Ultimate Limit States Dead Load -1185 kN -1800 kN

Ultimate Limit States Total Load -2200 kN -2650 kN

FLS Range -194 to 441 kN

(-636 to -1271 kN reaction)

-150 to 413 kN

(-1222 to -1785 kN reaction)

Note: a negative sign indicates uplift on the bearing.

Neither the design bearing data from the Contract Drawings, nor the forces provided for phase 1, were independently verified as part of this report. McElhanney and MMM compared force effects during construction and there was generally agreement in the values.

4.3. Shoe Plate

The shoe plate connects each bearing to the structural steel girder above it. The shoe plate and connection to the girder are designed by the design engineer (MMM) and detailed in the Contract on Drawing A-10 (Sheet 218-B-R1). A common shoe plate design and bolted connection to the girder is specified for all west abutment bearings. The shoe plate is the same size and has the same number of bolts for south, centre and north bearings despite the reaction at the centre girder being 73% larger than the north girder at ULS, and 40% larger at FLS. The bearing shoe plate is specified as 1000 mm long, 800 mm wide, with a thickness varying between 52 and 60 mm from end to end. Since the bearing sits level on the abutment, the shoe plate is bevelled to accommodate the 0.8% longitudinal slope of the roadway (and consequently, the girder bottom flange) to provide a level surface at the top of the bearing.

There are a total of 32, A325 (ASTM Standard A490, 2008), 22 mm diameter high strength bolts specified on the Contract Drawings which attach the bearing shoe plate to the bottom flange of the girder, as shown in Figure 8. The CHBDC Clause 10.18.4.2 allows for the use of 7/8” bolts interchangeably with 22 mm diameter (M22) bolts in 25 mm diameter holes. Drawing A-10 (Sheet 218-B-R1) note 1 specifies that all structural steel shall conform to CAN/CSA G40.21-M04 and shall be Grade 350W. Exposed metal surfaces are specified to be hot dipped galvanized.

Figure 8. Shoe plate bolts specified on A-10, revision C (sheet 218-1-R2).

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The number of bolts was increased through Request for Clarification (RFC) 176 from BFNJV, dated January 28, 2014. The RFC was submitted to address a note on drawing S-1 concerning the compensation of deflections in the longitudinal direction of the bridge. The submission notes an additional row of bolts at front and back of the bearing (10 rows total), at the same 100 mm bolt pitch as the original 8 rows of bolts. The additional bolts were added in plan and annotated with “additional set of bolts on each side allow field adjustment”, as shown in Figure 9. Based on the RFC, it appears that the 2 additional rows of bolts were for tolerance, not for additional structural capacity. As such, it may not have been the intent that all 40 bolts would be installed; however in final construction they were. The RFC was accepted by the MTO on January 31, 2014. This change was made before the initial bearing shop drawings were prepared. The first date on the shop drawings is August 4, 2014. It is unclear when the bolts were switched from grade A325 to A490.

Figure 9. Shoe plate bolts revised within RFC-176, dated January 31, 2014.

The bolts are shown to be installed head down in holes couterbored into the shoe plate. The counterbore is specified on the Contract Drawings as a 60 mm diameter circular recess, 18mm deep. It is not clear how the Contractor would have gripped the bolt head in the design on the Contract Drawing, except for installing the shoe plate onto the girder prior to installing the bearing. As described in Section 5.3.3, this counterbore was later changed.

4.4. Bearing Assembly

The bearing assembly, as supplied, consists of the rotational bearing, the shoe (or top) plate, a sole (or upper) plate, and the masonry (or bottom) plate sitting on the concrete abutment seat. The sole plate and masonry plates have interlocking guide bars with stainless steel or Polytetrafluoroethylene (PTFE) surfaces to resist uplift loads and allow translation. These components of the bearings are shown in Figure 10. OPSS 1203 clause 1203.04.01.06 states that “the top and bottom plates that are permanently attached to the structure shall be provided with the bearings.” For rotation bearings, the shoe plate, or top plate as referred to in OPSS 1203, is supplied with the bearing assembly, although it is designed by MMM as mentioned previously.

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Figure 10. Components of the west abutment bearing assembly.

The shoe plate acts as an intermediary component between the structural steel girder and the bearing and is supplied with the bearing assembly for rotational bearings. CAN/CSA-S6-06 clause 11.6.1.1 requires that bearings be “replaceable without damage to the structure or removal of anchorages permanently attached to the structure.” OPSS 1203, clause 12.04.01.09 Bearing Assembly Replacement states:

“The entire bearing assembly, except for the top plate used to attach it to the superstructure and the base plate used to anchor it to the substructure but including both contact surfaces of the sliding interface, shall be replaceable without damage to the structure and without removal of any concrete, welds, or anchorages permanently attached to the structure and without lifting the superstructure more than 5 mm. Bearings shall not be recessed into plates that are permanently attached to the structure.”

The shoe plate, or top plate as it is referred to in OPSS 1203, should be designed to stay in place while the bearing is replaced (although in most bearing replacements, the shoe plate is also removed and replaced with the bearing assembly owing to differences in dimensions and uncertainty about connection details between the bearing and top plate). In the design of typical highway bridge in Ontario, it is standard practice to provide a bolted connection between the bearing and the shoe plate, while the connection between the shoe plate and girder is typically welded. The practice in Ontario is that the connection between the girder and the shoe plate is specified by the designer, while the connection between the shoe plate and the bearing is designed by the bearing supplier. The detailing of the shoe plates at the Nipigon River Bridge follows the practice described above, although the less common bolted attachment of the bottom flange and shoe plate is used. The connection between the girder and shoe plate is clearly shown on the design drawings, while the connection between the shoe plate and the bearing is designed by the bearing supplier. The drawings do not provide any direction to the supplier on how to connect the bearing top plate to the shoe plate, nor do they require the bearing supplier to submit the connection details to the designer for approval. One feasible way to connect the bearing to the shoe plate is bolting at the edges of the shoe

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plate. Section 1 of drawing A-11, revision C (see Appendix A) depicts the west abutment bearing approximately 750 x 750 mm in plan which is smaller than the shoe plate. Working with those dimensions, it would be practically impossible to bolt the shoe plate to the bearing at the edges. A connection distributed across the shoe plate is the alternative to a connection at the edges. For the shoe plate dimensions and bolting pattern shown on the design drawings, it is difficult to envision connecting the shoe plate to the bearing within the central area of the shoe plate. Below the bearing, the connection between the bearing masonry plate and the abutment seat is shown schematically on the design drawings, but the final design is left to the Contractor. Post-tensioning anchorages are shown connecting the bearing to the abutment on drawing A-11 with a note stating the “bearing anchorage assembly to be designed by the Contractor.” Note 10 on drawing A-10 describes the materials to be used and level of prestressing to be adopted, but the geometry and details of the connection are left to the bearing supplier. Standard practice in Ontario is for the Contractor to design the anchorage of bearing masonry plates into the substructure. The detailing of the masonry plate connection on the Nipigon River Bridge follows the standard practice for rotational bearings. The connection is shown schematically on the design drawings, while the detailing is left to the Contractor.

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5. Observations and Testing The observations are based on information and correspondence provided by Northwestern Region. Evidence includes photos and videos taken at the site from January 10 to January 16, 2016. The report also documents observations made by engineers of the Bridge Office during their site visit of January 14, 2016. The Bridge Office requested all design calculations from MMM, and reviewed all the calculations provided. Calculations for the shoe plate design were dated after the failure. The review did not find calculations for the design of the shoe plate or for the bolted connection between the girder and shoe plate dated prior to failure.

5.1. Physical Observations

Verbal accounts from multiple sources affirmed that the bolts between the shoe plate and girder bottom flange were not pretensioned at the time of installation. This appears to be confirmed by the installation of nuts on stacked washers as shown in Figure 11b, apparently awaiting fabrication of washer plates. There is also e-mail correspondence between the Contractor and the bearing supplier on Tuesday, January 12, 2016 3:03 PM and Tuesday, January 12, 2016 4:58 PM, where there is discussion about the bolts not being pre-tensioned, but some disagreement on the reasons. The bolts likely remained in that condition leading up to their failure on January 10, 2016. Finally, on the day after the failure, MTO staff reported that the centre-west bearing bolts were loose and required tightening just to achieve a snug tight condition. The west abutment bearings were installed on October 5, 2015. Figure 11 shows the installation of the bearing sole plate after the shoe plate had already been attached.

Figure 11. Northwest Bearing installation on October 5, 2015: a) sliding the sole plate of the

bearing into place after the shoe plate has been connected to the bearing flange and b) washers stacked 3 high over west half of shoe plate and 4 high over east half of the shoe plate.

b) a)

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Photos of Figure 12 and Figure 13 were taken on January 10, 2016 after failure of the bridge. The bolts are fractured at the threads in all but a single bolt (the most southwesterly bolt). The shoe plate is noticeably bent upwards and separated from the bearing sole plate towards the interior lines of bolts. Northwest region staff measured the gaps between shoe plate and the sole plate for both the northwest and centre-west bearings. The maximum gaps are summarized in Table 3. Northwest region staff measured the deformations of the shoe plates after they were removed from the bridge. Based on the measurements, the northwest bearing shoe plate appears yielded along the interior bolt lines, whereas the centre-west bearing shoe plate is deformed primarily on the north side, with a maximum deformation along the exterior bolt line on the north side of the girder. Table 3. Shoe Plate Deformation (Maximum Gap Measured between the Shoe and Sole Plates of West Abutment Bearings and Maximum Deformation after Shoe Plates Removed)

Bearing Measurement West Side (back) East Side (front)

Northwest Bearing

Gap measured on site, mm 5 5

Deformation of shoe plate, mm 5.7 7.5

Centre-West Bearing Gap measured on site, mm 2 0

Deformation of shoe plate, mm 1.6 1.1

Figure 12. Northwest Shoe Plate with Failed

Bolts of Shoe Plate Failed Bolts Looking East

Figure 13. Northwest Shoe Plate with Failed Bolts – note gap between shoe plate and

bearing top plate towards the middle

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Figure 14 shows the numbering assigned to each bolt location and identifies bolt with confirmed or potential early failure, based on initial site observations. At least 2 bolts were immediately observed to have light corrosion of the failure surface, indicating they failed in advance of the girder lifting off the shoe plate on January 10. The bolts with indication of early failure were in the south exterior gage line of bolts. Figure 15 shows a plan view of the bolts through the shoe plate of the northwest bearing. Necking of the central fractured surface can be seen on many bolts, and beach marks, indicating fatigue type failure, are evident and oriented transversely to the bridge centerline.

Figure 14. Bolt numbering and premature bolt

failure (north is up). Red indicates confirmed early failure, blue indicates possible early failure.

The bolts were cut on January 10 in order to allow the girder to be lowered back down to bear on the shoe plate. The exact location of each bolt on the shoe plate was not recorded prior to cutting the bolts at their bases with a grinder. The bolts were assigned a letter code as an identifier. The letter codes were correlated to the above numbering where possible. Several bolts were noticeably bent along the shaft.

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Figure 15. Top Surface of the Shoe Plate with Failed Bolts (looking down, north is up)

Figure 16. Northwest Bearing, Northwest corner

of uplift restraint with PTFE crushed but still attached to the bearing masonry plate (top steel

surface of the photo)

Figure 17. Northwest Bearing, Southwest corner of uplift restraint with PTFE crushed and projecting beyond the masonry plate (top steel

surface of the photo)

Figure 16 and Figure 17 show damage to the PTFE surfaces at the back (west side) of the bearing, consistent with excessive contact pressures.

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The bearing Working Drawings specify 7/8” bolts 6” long for the connection of the shoe plate to the girder. As a point of reference, the Handbook of Steel Construction (Canadian Institute of Steel Construction, 2008) lists a 150 mm bolt as appropriate for a minimum grip length of 105 mm and the maximum grip length of 124 mm. The required grip length for this application varies from 98 mm to 106 mm along the length of the bevelled shoe plate. As measured from a spare bolt obtained from site, and shown in Figure 18, the bolt thread started approximately 113 mm from the bolt head – which exceeds the grip for all locations. Based on the actual range of grip lengths required for this connection, the length of bolts specified on the Working Drawings, which were the bolts supplied, were too long for the application. On October 8, 2015, BFNJV issued non-conformance report (NCR) 2013-6000-224 which identified that the RJ Watson shop drawing specified bolts that were too long.. As a corrective action, RJ Watson proposed a 16 mm thick structural steel washer plate to be installed under each quadrant of 10 bolts.

Figure 18. Bolt sample from southwest bearing.

There were several shortcomings of RJ Watson’s proposed solution to the excess bolt length. The drawings are not sealed, they do not provide any indication of a bevel in the washer plates to account for the slope of the bevelled shoe plate, and they do not specify the grade of material for the washer plate. The drawings do not identify an installation method for these plates. Installation of such a plate would necessitate removal of the capacity of 10 bolts at a time, with a corresponding decrease in capacity of the connection of at least a quarter. Despite these uncertainties, the proposal of the NCR was accepted by the MTO on October 21, 2015. However, the washer plates were never installed.

5.2. Examination and Testing of Fractured Bolts

Except for one bolt that fractured immediately adjacent to the head, all bolts fractured immediately adjacent to the nut and therefore had to be cut at the top of the shoe plate in order to be removed. All the 40 bolts were first delivered to the Bridge Office of the MTO Highway Standards Branch in St. Catharines, for visual examination and for a photographic record of their conditions and marking. David Lai of the Bridge Office inspected the bolts on January 18 and January 19. The Ministry submitted bolts to the NRC and SSW for chemical composition analysis, fracture surface analysis, tensile testing of machined specimens, toughness testing (Charpy impact testing), and corrosion product analysis. NRC and SSW were first given the opportunity to examine the bolts separately and select their bolts for testing. The bolts were divided into three groups with more or less equal number. NRC and SSW were each provided 14 bolts while MTO

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kept 12 bolts for reference and possible future actions. A complete description of the testing and investigation by NRC is contained in Makar (2016). A complete description of the testing and investigation by SSW is contained in Ramamurthy et al. (2016). 5.2.1. Fractured Bolts from NW bearing

Visual Examination by the Bridge Office 5.2.1.1.

Based on visual examination with the aid of only a hand-held magnifying glass, the following observations could be made:

Nine bolts had a varying degree of brown coloured corrosion product at the fracture surface, indicating that these bolts might have failed earlier than the rest. It is difficult to determine the time difference between the first bolt fracture and the final total failure of the bearing, however, it is quite clear that not all bolts failed at the same time. Figure 19 shows the fracture surfaces of the bolts which had corrosion product.

Most of the bolts exhibited striations at the fracture surface that is typically associated with cyclic loading, and as can be seen in Figure 15, Figure 19, and Figure 20, the fracture surfaces of many bolts have a striated zone at opposite sides of the cross-section that is consistent with cyclic rotation in the east-west direction, that is, in the longitudinal direction of the bridge.

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Bolt No. C - Location 29

Bolt No. I - Location 4

Bolt No. AA - Location 34

Bolt No. G - Location 23

Bolt No. K - Location 6

Bolt No. CC - Location 33

Bolt No. H - Location 21

Bolt No. Z - Location 35

Bolt II - Location 31 (unverified)

Figure 19. Fractured Surface of Failed Bolts with Visible Corrosion Product

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Figure 20. Northwest bearing looking East, after

failure on January 10, 2016. Figure 21. Transverse crack on elongated

side of bolt Z.

Bending of the threaded portion appears to be an important contribution to the failure of many of the bolts. Some samples showed the threads are compressed on one side and elongated on the opposite side, while some also have a transverse crack on the elongated side, as observed in Figure 21.

Only one bolt (EE from location 40) failed immediately adjacent to the head and so it was possible to retrieve a much longer shank of the bolt after the failure. As can be seen in Figure 22, there are steel to steel rub markings on opposite sides of the shank, the upper marking is within the bottom flange of the girder while the lower marking is within the shoe plate.

Crack with rust stain

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Figure 22. Bolt at the southeast corner of the shoe plate (location 40) which failed at the shank,

with steel-to-steel rub markings on opposite sides of the shank.

Chemical Composition Analysis 5.2.1.2.

Both NRC and SSW conducted chemical composition analysis of the fractured bolts, the test results meet the requirement of ASTM A490 Type 1 bolts and compare well with the mill certificate submitted by the Contractor.

Fracture Surface Analysis 5.2.1.3.

The fracture surfaces were examined under scanning electronic microscope (SEM) at high magnification. The following conclusions are drawn from the combined observations of NRC and SSW.

Most of the fracture surfaces show fatigue striations.

The number of striations are relatively low ranging from 50 to 140, indicating that the bolts failed due to low cycle (ductile) fatigue; the load levels in service were high enough to cause plastic deformation on each cycle.

There are three types of failure indicated by the fracture surfaces: 1. One sided crack growth caused by either direct tensile loading or in combination

with bending, but the cyclic bending does not reverse much in magnitude beyond the neutral position

2. Two opposite-sided crack growth caused by unidirectional cyclic bending 3. Cyclic tensile loading with other factors

In all cases, cracks seem to have originated from the root of the threads and grew towards the bolt centre; the outer crack growth zones are usually much flatter with

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striations, and the microstructure exhibits limited ductility compared with the centre zone where final ductile failure occurred.

Tensile Test of Machined Specimens 5.2.1.4.

Since all the bolts were cut at the top of the shoe plate, the bolt heads were removed and so the remaining specimens were not suitable for full size bolt tensile testing. The bolt specimens were typically less than 70 mm long. ASTM A490 (ASTM Standard A490, 2008) states that when the length of bolts makes full size bolt testing impractical, then machined specimens shall be tested for tensile properties and shall meet the stipulated yield and tensile strength, elongation and reduction in area; the testing shall be in accordance with ASTM F606 and E8. Figure 23 shows four cylindrical specimens cut from one bolt shank by the Electric-Discharge Machining Technique (EDM) at SSW. This technique does not generate enough heat to alter the properties of the bolt and causes very little wastage of material.

Figure 23. Cylindrical specimens cut from a bolt shank by EDM at SSW.

Table 4. Tensile Test Requirements and Results of Specimens Machined from Fractured Bolts

0.2% Offset Yield Stress, MPa

Ultimate Tensile Stress, MPa

Elongation at Fracture, % (based on 50 mm gauge length)

Reduction of Area at necking, %

Remarks

A490 Specification Requirements

896 min. 1034 to 1192 14 min. 40 min.

SSW Test Results 1046 1120 15.85* 57.6 Mean value

NRC Test Results 1043 1089 18 58 Mean value

Note : * Strain at fracture based on 20 mm gauge length

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Table 4 summarises the tensile test results from both laboratories and in comparison with the ASTM A490 requirements. As can be seen in Table 4, the tensile test results from the two laboratories are close to each other and all the test specimens met the requirements of ASTM A490. The only uncertainty is that due to insufficient length of the specimens, a 50 mm gauge length cannot be achieved for the testing at SSW and the reported elongation is based on a 20 mm gauge length. The reduction of area at the necking fractured section is well exceeding the specification requirement indicating a very ductile behaviour. Prior to the testing, a concern was raised whether the tensile straining of the bolts to failure in the field has changed the mechanical properties of the bolts such that the test results do not truly represent the original properties. Generally, steel material that has been strained to beyond yield could be into strain-hardening and subsequently the yield point could be higher while losing some ductility. It should be noted that the tensile specimens were taken from the solid shank of the bolts and the failure is in the threaded area; the stress in the shank area is only around 75% of the stress in the threaded area and therefore likely would not have gone beyond yield. Therefore, the test results should be representative of the original properties of the bolts.

Cold Temperature Charpy Impact Tests 5.2.1.5.

There was speculation that the in-service cold temperature might have rendered the bolt material to be brittle, and therefore not absorb the impact energy due to uplift at the bearing. Although ASTM A490 does not specify any cold temperature impact test requirements, these bolts have been used in cold weather environments when additional testing is specified. It is prudent, as part of the forensic analysis, to conduct the cold temperature impact tests and see how the bolt material performs comparing with other high quality structural steel. CHBDC stipulates that for fracture critical components subject to minimum in-service temperature above -30º C, the test temperature shall be -20º C and the minimum impact energy required is 27 Joules. Table 5 shows the test results reported by the two laboratories. It can be seen from the table that the bolt material absorbed impact energy far greater than that specified for fracture critical bridge components according to CHBDC. Therefore, these A490 bolts are considered suitable for use in cold temperature. Table 5. Charpy Impact Test Results of Fractured Bolts

Laboratory Test Temperature Absorbed Energy

Cambridge Materials Testing Limited (CMTL)* -20 C 50 Joules

CMTL -30 C 45 Joules



NRC -28 C 50.6

NRC +21 C 77.9

* CMTL was retained by SSW for testing.

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Corrosion Product Analysis 5.2.1.6.

The fracture surface of bolts exhibiting corrosion products were analysed by SSW to determine the chemistry of the products; the following conclusions could be made:

There are three distinct regions on the fracture surface: a grey-coloured region, white deposit region, and a heavily corroded region in brown colour.

The grey-coloured regions are covered with a mixture of iron oxides and smaller amount of iron hydroxides.

Some locations within the grey-coloured regions were covered with only a thin layer of oxide/hydroxide since the metallic iron substrate was detected. This implies the areas had different time of exposure to the weathering.

The white deposits were enriched in carbonate, mostly present as sodium carbonate; the source is undetermined.

The heavily corroded regions exhibit greater accumulations of iron hydroxides compared with iron oxides. The hydroxide forms are consistent with those expected to form in a chloride environment.

5.2.2. Intact Bolts from Centre-West Bearing

Distribution of Samples 5.2.2.1.

A total of 40 intact bolts were first delivered to the Bridge Office for visual examination and for a photographic record of their conditions and marking. NRC and SSW were given the opportunity to examine the bolts separately and select their bolts for testing. NRC and SSW were each provided 10 bolts while MTO kept 20 bolts for reference and possible future actions. The bolt numbering does not follow the same pattern as the bolts from the northwest bearing.

Visual Examination 5.2.2.2.

Out of the 40 bolts, 11 of them were cut with the grinder in order to remove the nut and free the bolt from the bearing shoe plate, and 5 others have distorted threads such that the nut could not be put back on properly. Hence, there are only 24 bolts available for full size bolt tension test. The identification marking on the bolt head clearly shows A490 Type 1. Figure 24 shows some examples of bolts with distorted threads. Although none of these bolts fractured in service, the loading condition and the installed detail might have caused significant straining and distortion of the threaded portion, and bending of the bolts in the threaded portion was observed.

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Bolt 1 Bolt 2 Bolt 13

Figure 24. Images of intact bolts with distorted threads.

Tensile Test of Full Size Bolts 5.2.2.3.

Tensile testing of full size bolts was conducted by both laboratories according to ASTM F606/F606M-14a using a 10 degree wedge at the head. In addition, NRC tested one bolt with a 10 degree wedge at the nut which is a more severe condition not required by the ASTM F606. On the other hand, SSW tested some bolts with a 1 degree wedge at the nut to simulate the field condition at the west abutment of the Nipigon River Bridge. SSW also performed the proof load test on 4 bolts to 246 KN according to the requirements of ASTM A490 and they all passed. NRC did not perform the proof load test. Table 6 summarizes all the tensile test results of full size bolts.

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Table 6. Summary of Tensile Test Results of Full Size Bolts

Laboratory Bolt No. Max Tensile Load with 10 degree wedge at head (KN)

Max Tensile Load with 1 degree wedge at nut (KN)

Max Tensile Load with 10 degree wedge at nut (KN)

ASTM A490

Min. 308 KN

Max. 356 KN

SSW B3 346.9 Passed

SSW B10 346.0 Passed



SSW B6 338.5 Passed

SSW B8 343.8 Passed



NRC B7 346 Passed

NRC B11 343 Passed

NRC B17 342 Passed

NRC B22 340 Passed

As shown in Table 6, all the tensile test results met the requirements of ASTM A490. The results from the two laboratories are very consistent with a small coefficient of variation.

Tensile Test of Machined Specimens 5.2.2.4.

Although the intact bolts are long enough to be tested as full size bolts, it is advisable to test some machined specimens from them for comparison with those from the fractured bolts. The following Table 7 shows the tensile test results of the machined specimens from both testing laboratories. It can be seen that they all meet the yield stress, ultimate tensile strength, and reduction in area requirements of ASTM A490 and compares well with those from the fractured bolts. There is however an issue with the elongation value at fracture for the SSW specimens; the diameter of the specimens is 4 mm and the gauge length to diameter ratio is therefore much bigger than the specified ratio of 4 according to ASTM F606. NRC’s specimens have a diameter of 12.5 mm and therefore the aspect ratio is 4, resulting in an elongation well in excess of 14%. It is conceivable that a smaller diameter specimen does not have the same volume of material to undergo necking and therefore the corresponding longitudinal deformation is smaller. Table E-1 in the SSW report does show improving elongation with smaller gauge lengths and much closer to meeting the 14% requirement when the aspect ratio is close to 4.

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Table 7. Tensile Test Results of Machined Specimens from Intact Bolts of Centre Bearing

0.2% Offset Yield Stress, MPa

Ultimate Tensile Stress, MPa

Elongation at Fracture, % (based on 50 mm gauge length)

Reduction of Area at necking, %

Remarks

A490 Specification Requirements

896 min. 1034 to 1192 14 min. 40 min.

SSW Test 1059 1124 6.2 57.5 Mean value

NRC Test 1031 1103 17.67 58.0 Mean value

Fractured Surface Analysis 5.2.2.5.

The fracture surfaces after the full size tensile test were examined using the SEM and it is concluded that fracture usually initiates at the root of the first thread below the nut and progresses across the bolt either on a flat surface or at an angle to a thread root below the nut. All regions of the fracture surfaces display characteristics of ductile fracture. Both NRC and SSW sectioned some bolts longitudinally to examine them under microscope and cracks of length 100 to 400 microns have been observed throughout the threaded portion, including areas where the bolts have not experienced any load. These cracks are also on the top side of the thread crests and are heavily oxidized. Since the observed fracture planes always originate from the root of the thread, it is likely that these cracks are pre-existing due to the manufacturing process of the bolts and did not play a role in the failure of the bolts. 5.2.3. Conclusion Based on Test Results

The following conclusions are drawn from the test results from the two independent laboratories.

The test results from the two laboratories agree very well with each other with no outlying data, the differences were within the order of accuracy of the test methods and the normal variability of material.

The bolts met all the requirements of ASTM A490.

The bolts exhibit good ductility and impact energy absorption under cold temperature comparable to high quality structural steel normally used for bridge construction.

The bolts failed at different times and not simultaneously based on the appearance of the fracture surfaces and the development of corrosion products on some; it can therefore be construed that the bolts were subjected to uneven loading and those with higher loading would fail first, then the remaining ones had to carry a higher share of load than before leading to progressive failure.

The majority of bolts failed by low cycle high strain fatigue, the load levels in service were high enough to cause plastic deformation on each cycle.

The bolts were likely not tightened according to normal requirements for bridge construction since most of the fracture surfaces exhibit fatigue striations on opposite ends of the fracture surface; indicating the cycle rotation is in the longitudinal direction of the bridge.

Bending is a contributing factor for the failure of many bolts based on the deformation of

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the threaded portion. However, the laboratory static tests for the full size bolts incorporated inclined surfaces (a 10 degree wedge under the bolt head or a 1 degree tapered washer under the nut) to create a bending effect in combination with tensile load, and the ultimate tensile capacity for either case still met the specification requirement. Hence, the failure of the bolts is due to overloading in service beyond what each bolt is meant to carry.

5.3. Differences between Contract Drawings and Supplied Bearings

There are a number of notable differences between the bearing requirements of the Contract Drawings and the bearings supplied according to the Working Drawings (included in Appendix B), as listed below and described in the following sections.

1) Based on the submitted Working Drawings, the bearing could not accommodate rotation (see Section 6.1.2 for further discussion).

2) The bolts connecting the shoe plate to the girder bottom flange were specified on the Contract Drawings as A325, but were supplied as A490 (see Section 4.3).

3) The bolts were not adequately pretensioned. 4) The shoe plate material specified in the Working Drawings and supplied was a different

material and grade than specified on the Contract Drawings. 5) The bolt configuration and counterbore details supplied were different than on the

Contract Drawings and somewhat different from the submitted change proposal. 6) The bolts specified in the Working Drawings, and those supplied, did not have bevelled

washers and were too long for the actual grip length. 5.3.1. Bolt Pre-Tension

The connection of the shoe plate to the girder bottom flange was specified on the Contract Drawings and is covered by the CHBDC as referenced by OPSS 922 and OPSS 1203. CHBDC clause 10.18.2.1 states that all high-strength bolts shall be pretensioned in accordance with Clause 10.24.6.3. Clause 10.24.6.3 further refers to Clause 10.24.6.6 which mandates the use of turn-of-nut pretensioning for bolts. Based on the evidence identified in Section 5.1 of this report, the bolts were not properly pre-tensioned in accordance with the CHBDC. The CHBDC Clause 10.17.2.6, in the fatigue section, also states that high strength bolt shall be pretensioned in accordance with clause 10.24.6.3. Makar (2016) describes and illustrates the effect of pretension on the cyclic stress range in the bolts. CHBDC Clause 10.24.2.2 states that the erection procedures shall include the bolt installation requirements. Since the erection procedures for the installation of the bearings did not identify the tightening requirements, the submission of the erection procedures was not in conformance with the requirements of the CHBDC. 5.3.2. Shoe Plate Material

The size and thickness of the bearing shoe plate supplied matches the shoe plate specified on the design drawings, but the grade of material is different. Drawings A-10 note 1 specifies that all structural steel shall conform to CAN/CSA G40.21-M04 and shall be Grade 350W. The

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bearing assemblies as supplied for the northwest and centre-west bearings respectively are detailed on drawings titled DB440MR Type I Assembly, revision 8, and DB440MR Type II Assembly, revision 8. The material for the shoe plates, sole plates, and masonry plates is specified as ASTM A36 which has a yield strength of 248 MPa as compared with the 350 MPa yield strength specified on the design drawings. 5.3.3. Bolt Pattern between the Girder to the Shoe Plate

As stated in Section 4.3, there were a total of 32, A325 22 mm diameter (M22) high strength bolts specified on the Contract Drawings which attach the bearing shoe plate to the bottom flange of the girder, as shown in Figure 8. Request for Clarification (RFC) 176 increased the number of bolt holes to 40 and is shown in Figure 25. The bearing shoe plate is specified as 1000 mm long, 800 mm wide, with a thickness varying between 52 and 60 mm from end to end in order to accommodate the longitudinal slope of the roadway. The supplied shoe plate is the same overall dimensions as specified, but is connected to the bottom flange with 40, A490 7/8” bolts as explained in Section 4.3 and as shown in Figure 26. There were differences in the spacing of the bolts, perhaps because stiffeners make the exact location of the additional bolts shown in the RFC infeasible.

Figure 25. Shoe plate bolts specified in RFC 176.

Figure 26. Bottom view of shoe plate with bolts as supplied per drawing Bearing Details-7,

revision 8.

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Figure 27. Counterbore specified on A-10, revision C (sheet 218-1-R2).

Figure 28. Counterbore detailed per drawing Bearing Details-7, revision 8.

The bolts are shown on the Contract Drawings, head side down, in holes couterbored into the shoe plate. The counterbore is specified on the Contract Drawings as a 60 mm diameter circular recess, as shown in Figure 27. As supplied, the counterbore is machined into the shoe plate as a rectangular recess with rounded corners, as shown in Figure 28. The recess is longer in one direction than the other with the intent to accommodate the hexagonal bolt head and prevent it from rotating. However, the bolt specifications allows the corner to corner dimension of the bolt head to be from 39.5 mm to 41.5 mm, meaning that a bolt on the low end of tolerance would still be able to freely turn in the counterbore. The effective width of the plate as supplied, on a line through the centres of the 10 bolts, is 400 mm, as compared with 480 mm in the Contract Drawings. The width of shoe plate material removed by the counterbore at the bolt centrelines, on a line through 10 bolts, is 176 mm compared with 240 mm specified in the Contract Drawings. It appears that the change in the counterbore did not affect the shoe plate capacity, and may have actually assisted with it when considered separately from the material strength of the shoe plate. The evaluation of this effect is covered in Section 6.2. The size and thickness of the bearing shoe plate supplied per the Working Drawings matches the shoe plate specified on the design drawings (and RFC 176). There were minor changes to the bolt spacings and counterbores which did not appear to negatively affect the strength. 5.3.4. Bolt Length and Washers

Detail A of design drawing A-11 specifies beveled washers to be installed below the nut of the A490 bolts between the shoe plate and bottom flange, as shown in Figure 27. This is a requirement of CHBDC clause 10.24.6.5 to compensate for any lack of parallelism between the outer clamped surfaces. In the Working Drawings, a hardened washer was specified, but it is not specified to be beveled. Therefore, the bolt assembly shown on the Working Drawings does not comply with the requirement of the design drawings for a beveled washer, and does not meet the requirements of CHBDC clause 10.24.6.5.

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CHBDC Clause 10.24.6.4 (b)(ii) specifies that hardened washers be provided under the head and nut when steel with a minimum yield strength of less than 280 MPa is specified. This is a requirement to prevent galling and indentation of the steel (Kulak & Grondin, 2006). The Working Drawings do not specify washers between the bolt head and the A36 shoe plate, and therefore do not meet this requirement of the CHBDC for the grade of material. From the time of installation of the bearings until the failure of the northwest bearing on January 10, 2016, the bolts were installed with 3 or 4 stacked washers under the nut in order to compensate for a bolt with grip length in excess of the required grip length. The bolts detailed on the Working Drawings were too long for the grip length of the steel plates being connected. Revision 8 (as-built) Working Drawings are sealed November 24, 2015 and show the bolts with single washers. Therefore, the as-built Working Drawings for the bearings do not reflect the actual installation with respect to the bolts connecting the girder to the shoe plate.

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6. Evaluation of Bearing and Attachments The evaluation follows all components in the load path from the girder to the abutment. Simple calculations and models were used to evaluate the forces in each of the components under two load cases: the loads at failure of the northwest girder (phase 1 when only the north half of the bridge was opened), and the loads specified on the design drawings. In some cases the components are inadequate to accommodate the loads at failure, and by extrapolation cannot accommodate full design loads specified on the drawings. Refined analysis validated the findings of the simple models. This section discusses the capacity of individual components followed by the behaviour considering the interaction between components.

6.1. Evaluation of the Bearing

6.1.1. Design

The bearing design consists of a disc bearing to transfer compression (positive reaction), surrounded by a large masonry plate and sole plate each with guide bars that interlock to resist uplift, as shown in Figure 5 and Figure 10. When subjected to uplift, the bearing slides along the plane of interlocking guide bars. The lower guide bar is laminated with a stainless steel surface, while the upper guide is laminated with a PTFE sheet filled with glass fibre. The guide bars accommodate transverse displacements and longitudinal displacements by sliding between the stainless steel and PTFE surfaces. When in uplift, the disc bearing sees no load and does not participate in the function of the bearing. The bearings are not designed for rotation in uplift, neither longitudinal nor transverse rotation. Rotation is usually accommodated with convex and concave surfaces that slide against each other – something that did not exist on this bearing. The reactions on the design drawings (Figure 7) indicate the west abutment bearings are to be in uplift at all times at SLS with corresponding rotations clearly specified. Note 14 on drawing A-10 requires that the bearing accommodate the horizontal rotations in all directions (longitudinal and transverse). These same requirements exist in phase 1, although as shown in Table 2, the loading is lower, but still uplift at all times. The only way the bearings allow rotation in uplift is for the guides to separate, as explained in Section 6.1.2. 6.1.2. Rotation

The rotation specified on the design drawings is a maximum of 0.8° (0.014 radians) at SLS, and as stated in Section 5, the bearing must be designed for an additional 1°. The only way the bearing can accommodate an imposed rotation is for the guide bars to separate. There is uplift across the guide bars due to permanent loads. The uplift creates a restraining moment across the bearing which has to be overcome in order for the plates to separate. The bearing is fixed against rotation until the restraining moment is overcome by an applied moment of larger magnitude. As the applied moment increases, the stresses increase at one end of the bearing and decrease at the other end until the one end separates and the bearing rotates about the far end.

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Any separation of the uplift restraint guide bars at SLS occurs simultaneously with high contact pressures at the other end of the bearing. This is unacceptable and constitutes failure because the pressure on the PTFE surface exceeds the limits allowed by the CHBDC. This is explained further in Section 6.1.2.1 and Section 6.1.2.2.

Longitudinal Rotation 6.1.2.1.

Longitudinal rotation occurs from deflection of the main girder. The imposed rotation of 0.8° (0.014 radians), as specified in the design, causes both guide bars to rock up onto the ends of the guide, as shown in Figure 29. The uplift force is transferred across a small contact area between the interlocking guide bars, at the front or back of the bearing. The overlap of interlocking guide bars is at least 785 mm. A rotation of 0.8° corresponds to a separation of 11 mm at one end. The PTFE has some ability to distribute the force over a longer length, but over time the PTFE will crush and deteriorate since the pressure at each end of the guide bar will exceed the allowable pressure. Figure 16 and Figure 17 show photographic evidence of crushing of the PTFE.

Figure 29. Separation of guide bars due to longitudinal rotation.

Live load on the bridge imposes a rotation at the west abutment bearings. On February 26, 2016, McElhanney Consulting Services Inc. provided information about the forces and rotations at the west abutment bearings from the model of the north half of the bridge, representative of the bridge’s state of completion at failure. Figure 30 describes the rotation of the northwest bearing (if it were free to rotate) as a CL-625-ONT design truck travels across the bridge. The model does not include dynamic load allowance. The northwest bearing is most heavily influenced by a truck in the westbound lane, and therefore the truck travels in the westerly direction from right to left. The maximum counterclockwise rotation of 0.06° (0.001 radians) occurs when a truck is in the east span 80 m from the tower, while the maximum clockwise rotation of 0.16° (0.003 radians) occurs when the truck is 20 m from the west abutment. While these rotations are much smaller than the SLS requirement of 0.8° (0.014 radians), the maximum clockwise rotation of 0.16° corresponds to a separation of 2 mm at the guide bars. This rotation occurs when the truck causes a positive (compressive) reaction at the bearing, although due to the pre-existing dead loads, the bearing remains in net uplift. Even before separation occurs, the rotation causes the PTFE and bolts at the front or back of the bearing

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(depending on direction of rotation) to take most of the load and the opposite end with reduced contact pressure to take virtually none.

Figure 30. Vertical reaction and rotation at northwest bearing due to the CL-625-ONT Truck.

Temperature effects also impose rotations on the bearing in all three axes. Rotations occur due to temperature range effects (overall temperature drop or rise applied to the entire bridge), temperature gradients in the deck (deck slab surface at a higher temperature than steel), and differential temperature between the deck and cables. Differential temperature accounts for the cables warming up at a faster rate than the deck and girders, due to their smaller relative thermal mass, and their exposure above the deck. As the cables elongate, the deck deflects downwards under permanent loads, imposing a clockwise rotation at the northwest bearing. The magnitude of this rotation is enough to cause the bearing to engage more at the back side of the guide bars during a sunny winter day. Similarly, when the temperature rapidly cools, the cables shorten and engage more on the front side of the guide bars. MMM provided a breakdown of rotations about the three axes of the bearing in an email of March 2, 2016. At ULS, the magnitude of rotation due to combined temperature effects is smaller than the rotations due live load, but the live load rotations represent a rare occurrence where one span of the bridge is fully loaded with trucks while the other is completely unloaded. In service, the rotations due to live load are lower than those due to temperature. Figure 30 ignores any temperature effects (and dynamic load allowance), thus underestimating the force effects caused by separation by a fair amount. With rotation of the girder, separation occurs at one end of the bearing guide bars as previously shown in Figure 29. This means that the highest force is transferred between the bearing and the girder at the end where the contact pressure is highest, varying linearly to almost no force transfer at the end where the separation occurs. For example, for a clockwise rotation, the bolt forces are illustrated in Figure 31.

-0.1

0.0

0.1

0.2

-150

-50

50

150

250

350

-140 -120 -100 -80 -60 -40 -20 0 20 40 60 80 100 120 140

Rota

tion a

t N

ort

hw

est

Bearing,

degre

es

Reaction a

t N

ort

hw

est

Bearing,

kN

Position of Lead Axle Relative to the Tower, m

East Abutment

West Abutment

Tower

Lead axle position when last truck axle leaves bridge

Rotation atNW Bearing

Reaction atNW Bearing (uplift is positive)

Direction of Travel

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Figure 31. Amplification of bolt force due to uplift and longitudinal rotation.

Bolt forces were calculated based on a linear distribution across the bolted connection, similar to design of a bolted web field splice. All the bolts across the connection see tension, the highest of which is at the end where the guide bars are in contact. At the end where the guide bars are separated, there is compression at the contact surface between the shoe plate and the girder. In this simplified elastic analysis in isolation of transverse effects, the front row of bolts sees a force which is 2.8 times larger than the average force in a bolt row. With deflection of the shoe plate and sole plate, the front row of bolts could be subject to an even higher force. The effects are further described in Section 6.4. Similarly, the guide bars, sole plate, and shoe plate also see similar increases in load effects.

Transverse Rotation 6.1.2.2.

Transverse rotation occurs from deflection of the transverse floor beam. About the horizontal axis aligned with the girder, an imposed rotation of 0.8° (0.014 radians), as given in the Contract Drawings (see Section 4.1) would cause one guide bar to disengage and one guide bar to resist the entire uplift force. The guide bars are spaced 580 mm centre-to-centre. A transverse rotation of 0.8° corresponds to a separation of 8 mm at one of the guide bars, as shown in Figure 32. In reality, that scenario is implausible in service since the girder would have to rotate 17 mm out-of-plumb in order to impose such a rotation. However, the girder does rotate some amount along its axis due to deflection of the floor beams under dead and live load, as well as torsion imposed by the cantilevered sidewalk.

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Figure 32. Separation of guide bars bearings due to a rotation of 0.8° in the axis of the girder.

A built-in rotation may also exist due to the construction sequence. As each section of cantilevered construction progresses, the 10.8 m length of girder is installed followed by the floor beams. The deck panels load the floor beams causing them to deflect downwards. In turn, the floor beams impose a rotation at their connection with the girders, as they are connected with a moment connection. The construction sequence leads to the potential for unequal distribution of tensile force between the south and north interlocking guide bars due to the transverse rotation. Even though the shoe plate was attached to the girder bottom flange, the west abutment bearings were only fully installed after placing the structural steel girders and floor beams (see Figure 11), but before placing the deck panels. It is therefore possible that the level plane of the bearing does not match the plane of the bottom flange which may not be level. The distribution of the force to both guide bars may not be perfectly uniform. Similar to longitudinal rotation, when separation of the guide bars occurs, all the force is transferred through the remaining guide bar that is in contact. Accordingly, the bolts on the side in contact are required to carry the full uplift force. This simplified analogy indicates that one line of the bearing sole plate to shoe plate 1” bolts (plus the 7/8” shoe plate to flange bolts on this side) sees a force which is two times larger than the average force. Similarly, the guide bars, sole plate, and shoe plate also see similar increases in load effects. On January 10, the failure of the northwest bearing imposed a large transverse rotation on the centre-west bearing. Additionally, the uplift reaction at the centre-west bearing when the northwest corner of the bridge was elevated would have been higher than in service. The centre-west shoe plate is yielded along the line of exterior bolts closest to the north guide bar which is at the side of the bearing where the guide would be engaged to transmit the full uplift reaction. The deformed shape of the shoe plate at the centre-west bearing provides evidence that the south guide bar disengaged and the reaction on the north guide bar increased substantially. Although not indicative of the performance of the bearing in service, this illustrates that the uplift reaction is not shared equally between guide bars when subjected to an imposed transverse rotation.

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Plan Rotation 6.1.2.3.

The bearing demand requirements for plan rotation were +/-1.0°. Bridge Office was not provided with any plan rotation values under the phase 1 loading. Plan rotation was not considered to have contributed to the failure and no analysis was done for it. It is noted that geometrically, there would be contact between the metal surfaces of guidebars at this magnitude of plan rotation. 6.1.3. Uplift Guide Bar PTFE Evaluation

The uplift restraint guide bars are laminated to allow longitudinal translation along the bridge through sliding of the guide bars. The upper guide bars are laminated with PTFE, while the lower guide bars are laminated with stainless steel sheet. The PTFE specified on the bearing Working Drawings is 52 mm wide x 785 mm long. The dimensions of PTFE are the same for both centre-west and northwest bearings. The PTFE is 15% glass filled sheet, mechanically anchored, attached to the top guide bar and is unconfined. CHBDC Clause 11.6.3.6 establishes clear limits on the maximum pressure which can be applied on PTFE. OPSS 1203 requires that the PTFE be unfilled sheet, recessed in the backing material. The unconfined, filled sheet PTFE supplied for the uplift bearings at the Nipigon River Bridge does not meet this requirement. Table 8 summarizes the average contact pressure across the PTFE, assuming the reactions across the bearings are distributed uniformly to both guide bars, and uniformly across the area of surface of the PTFE, and compares the design to the CHBDC limits. Table 8. Design Contact Pressure for PTFE Sliding Surfaces of Uplift Restraint Guide Bars

Criteria Centre-West Bearing Northwest Bearing

SLS Dead Load Uplift 3530 kN 1900

Average contact pressure on PTFE 43.2 MPa 23.3 MPa

Maximum average contact pressure (CHBDC Table 11.3)

30 MPa 30 MPa

Assessment The average pressure exceeds the CHBDC limit by 44%.

Meets the CHBDC requirements.

SLS Total Uplift 4410 kN 2540 kN

Average contact pressure on PTFE 54.0 MPa 31.1 MPa

Maximum average contact pressure (CHBDC Table 11.3)

45 MPa 45 MPa

Assessment The average pressure exceeds the CHBDC limit by 20%.

Meets the CHBDC requirements.

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The average contact pressures for the northwest bearing would be within the CHDBC limits if the contact pressure was uniform along the length of both guide bars. The PTFE contact pressures of the centre-west bearing exceed the CHBDC limits even when a uniform pressure was assumed. The CHBDC allows the maximum contact pressure at the edge of the PTFE to exceed the limits in the table by 20%. It is clear that any separation of a guide bar to allow rotation only increases the contact pressure beyond the limits of the CHBDC. Any separation of the guide bar would result in double or triple the average pressure values at the heavily loaded guide bar end, which would mean that neither bearing would meet the requirements of CHBDC Clause 11.6.3.6. Any separation of guide bars due to an imposed rotation represents failure to meet the SLS requirements of the CHBDC. Any larger separation would cause the pressures to be exceeded by a greater extent, or at lower loads, including the loads for phase 1. Although the failure of the PTFE violates the CHBDC, it is expected to lead to local damage to the PTFE, steel on steel contact, and increased friction, but likely not a significant contribution to the January 10 failure that occurred. 6.1.4. Uplift Guide Bar Capacity Evaluation

Uplift is provided by interlocking guide bars between the sole plate and the masonry plate of the bearing. Assuming that uplift force across the bearing is shared equally between the two guide bars, and no rotation on the bearing, the uplift at each guide bar imposes a moment on that guide bar. The factored moment at ULS on each guide bar is 123 kNm (corresponding to the maximum uplift force of 3370 kN shared equally between both guide bars and accounting for the transverse displacement across the bearing), whereas the factored resistance of the guide bar is 251 kNm. The two top and bottom guide bars are all of the same dimensions and therefore have equal capacities. The guide bars of the northwest bearing are adequately detailed to resist uplift without rotation. With the same assumptions but higher uplift reaction of 5300 kN at ULS, the guide bars of the centre-west bearing are adequately detailed to resist uplift without rotation albeit with less reserve capacity. Upon longitudinal rotation, the guide bar plates separate (or have reduced pressure at one end) and the uplift force is transferred at the front or back of the guide bars. The full length of the guide bar is not effective in resisting this force, and therefore the capacity of the guide bars is less than the aforementioned value. Similarly, with transverse rotation, the right and left guide bars (depending on the direction of rotation) will not be evenly loaded and one guide bar will have an applied factored moment greater than the aforementioned value. With these higher factored moments, local bending and yielding of the guide bars could have occurred. The evaluation of guide bars shows that they were designed assuming uniform loading from a bearing that can properly rotate. The guide bars do not have the capacity to accommodate any separation of the guide bars, and therefore are not designed to accommodate any rotation in uplift as it actually occurred. 6.1.5. Shoe Plate Attachment to Top of Bearing

The Contract Drawings do not provide any direction to the Contractor on how to connect the bearing upper plate to the shoe plate; such as schematic diagrams, notes, or requirements to

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obtain approval for the connection from the designer. The Working Drawings adopt a bolted connection along the outer edges of the shoe plate, utilizing one row of 12, 1” diameter A490 bolts along each edge of the shoe plate. The bearing connection with the shoe plate interacts with the shoe plate connection to the girder. The shoe plate pries against the outer edge of the bearing sole place, and pries against the central contact surface between the shoe plate and the girder bottom flange right below the web. An analysis of prying between the flange and shoe plate, and shoe plate and sole plate, are provided in Section 6.4.

6.2. Evaluation of Shoe Plate

As explained in Section 5.3, the bearing supplier supplied; 1) a different material grade of the shoe plate, 2) a different bolt arrangement, and 3) different counterbore details, than shown on the design drawings, and in RFC 176. The tapered shoe plate was intended to be connected to the bottom flange with pretensioned bolts. However, even if they were pretensioned, it was found that the applied uplift force reduces the contact pressure between the shoe plate and flange to such a degree that two plates could not transfer adequate shear forces between them to act compositely for the working, serviceability or ultimate loads. Table 9 summarizes the section properties and capacity of the shoe plate through a line of 8 or 10 bolts, on the Contract Drawings and on the Working Drawings. Table 9. Comparison between Shoe Plate Specified and Shoe Plate Provided

Property 8 bolts, per design drawing A-11, revision B

10 bolts, per as-built bearing Working Drawings

Area, mm² 40,060 39,150

Moment of Inertia, mm4 9,868,650 9,950,440

Material G40.21-M04, Grade 350W ASTM A36

Yield Strength, MPa 350 248

Applied ULS Moment at Outer Line of Bolts (assuming case a) or c) in Figure 33), kN

158 (phase 1)*

317 (design load, centre-west)*

Nominal Moment Resistance at Yield, kNm 123.7 87.0

Factored Moment Resistance (Mr), kNm 175.5 125.7

* After failure of the outer line of bolts, moments would be much larger.

The net area and the moment of inertia of shoe plate at the bored sections is approximately the same in the design and Working Drawings. However, the bending resistance of the shoe plate supplied per the Working Drawings is considerably lower owing to the lower strength of material. Adding bolts to the connection did not weaken the capacity of the shoe plate, but changing the material in the shoe plate reduced its capacity by 29% due to the lower yield strength. Determining the bending moment in the shoe plate is not simple and depends on the flow of forces and any prying action in the bolted connection. The values are presented in Section 6.4

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and the values can only be determined with a detailed analysis. For reference, schematic shear force and moment diagrams of the shoe plate based on all bolts remaining elastic, for three scenarios, are shown in Figure 33. The first case has the uplift force shared evenly by all 7/8” bolts with no prying. The second has a prying force applied at the edge of the shoe plate and at the centre of the web. Details of the exact distribution of this force are provided in Section 6.4. The third case has the uplift force resisted entirely by the outer line of 7/8” bolts. These diagrams show the same value for the moment at the outside line of bolts for the first and third cases, and a somewhat reduced moment for the realistic second case where prying is considered. Figure 33b is the realistic representation of the forces on the shoe plate. These scenarios illustrate that prying does not increase the forces in the shoe plate compared with simplified assumptions that may have been used to design the shoe plate.

Figure 33. Shear Force and Moment Diagrams for Shoe Plate; a) assuming load evenly shared by

all bolts, b) with prying force, and c) assuming all load through exterior line of bolts.

6.3. Evaluation of Bolts

As explained in the Section 5.1, the Contractor installed the bolts without bevelled washers and did not pretension the bolts. The bolts are loaded eccentrically as one side of the nut contacts the steel surface which is sloped at 0.8% in the longitudinal direction, resulting in local bending of the fastener, referred to in literature as local prying. This effect is well-documented and explains the scatter in tests of bolt prying (Kulak, et al., 1987). The local prying of the bolt head imposes a bending stress in the bolt which adds to the axial tensile stress in the bolt (see Figure 34). Premature failure of the bolt, at lower than ultimate tensile strength, is then possible. The local prying effect is pronounced when the distance between the prying surface and the bolt is much larger than the distance between the bolt and the applied load, as is the case for the shoe plate to girder connection at the west bearings.

b) a) c)

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Figure 34. Local Prying of bolt due to lack of beveled washer.

The stress on a uniformly loaded bolt nut is determined from Force / Area (P/A). For an eccentrically loaded bolt nut, the extreme stress also has a bending component ( P/A + Pe / S). Assuming the resultant force of a 7/8” A490 bolt loads the plate at half of the bolt diameter (i.e. the bolt is loaded with an eccentricity of d/2), an axial force of 75 kN results in a stress of about 195 MPa when evenly loaded, and 5 times that value (985 MPa) with the eccentric loading. This magnitude of stress is likely to cause yielding of bolt (which has specified yield strength of about 940 MPa). In the extreme, the eccentricity could be as large as ¾ of the diameter (at the edge of the nut). This would result in the stress with eccentric loading being 7 times the uniformly loaded value. Although the 7/8” A490 bolts can be expected to yield at less than 100 kN due to this local prying, they do not necessarily fail under static load. Bolts with sufficient ductility would be expected to deform to the extent that the contact pressure between the bolt-head (or nut) and the prying plate becomes more uniform. Therefore, the local effect of bolt prying can relieve itself through deformation of the bolt shank. ASTM A490M – 09, Standard Specification for High-Strength Steel Bolts, clauses 10.9 and 10.9.3, for Structural Steel Joints (Metric), requires tensile, proof load, and hardness tests be conducted with Test Method F606M. A490M bolts require a wedge test (ASTM F606/F606M) of full scale specimens for 7/8” bolts. The wedge test requires the bolt head be tested in contact with a surface bevelled at 10° to the bolt head. For 7/8” bolts, a minimum tensile load of 308 kN is specified. The requirements of this test impose considerable local bending and yielding of the bolt prior to achieving the full tensile load. The test requires that the fracture occur through the threads and not at the junction of the bolt head to shank. Bolt testing conducted at both NRC and SSW, as discussed in Section 5.2.2, confirmed that under static load the ultimate strength is not affected by eccentricity. The increased stress due to eccentric loading, and the ASTM test described above are true for bolts with large differences in slopes between the steel and bolt nut surfaces, and consequently a large gap between the nut and flange at one end. With the 0.8% grade, the gap between the nut and the flange is only about 0.3 mm. This means that the first amount of force goes towards elongating and bending the bolt to match the slope of the bottom flange. The result is that the

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first amount of force on the bolts goes towards a locked-in bending stress of about 160 MPa (with compression on one side, tension on the other) to get the surfaces to mate. This effect is discussed farther in Section 6.5 on fatigue in bolts. With longitudinal girder rotation alternating depending on which span the load is in, this creates alternating compression and tension on opposite sides of the bolts – which is similar to the failure surface described in Section 5.2 The factored bolt resistance for 1” bolts in tension, with their 1040 MPa ultimate strength, determined from CHBDC clause 10.18.2.2.1, is 316 kN. The yield stress (using the 2% offset method as per ASTM A490) of 940 MPa results in a yield force (and therefore the force due to pre-tensioning), of 286 kN (ASTM A490 gives a value of 351 kN for the yield strength, although it does not include a material reduction factor, ϕb). The failure load of the bolt is 404 kN (ASTM). For 7/8” bolts, the factored ultimate strength is 242 kN, the yield strength is 219 kN and the failure load is 308 kN (from ASTM). The bolt forces for 1” and 7/8” bolts are summarized in Table 10. Table 10. Bolt Forces and Capacities

1” Bolt 7/8” Bolt

Pre-Tension Force 286 kN 219 kN

Factored Resistance 316 kN 242 kN

Ultimate Strength (Minimum Tensile Load specified in ASTM A490)

404 kN 308 kN

6.4. Evaluation of Prying in Shoe Plate

Prying is the phenomenon in structural connections where the tensile bolt forces are magnified due to bending of the plates and the resulting concentrated contact pressure forces between the plates (Kulak, et al., 1987). A classical prying situation is shown in Figure 35. The magnitude of the prying is related to the distances between bolt lines and the edges of plates, as well as the thickness of the plates.

Figure 35. Classical Prying of flexible plate in tension connection (Kulak, et al., 1987).

For the Nipigon Bridge, the shoe plate connection to the bottom flange consists of two gage lines of 7/8” bolts, arranged symmetrically about the girder web. The shoe plate connection to

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the bearing sole plate is by means of one line of 1” bolts at each side of the shoe plate. The girder exerts an upward force on the shoe plate, which in turn is “pulled down” at the edges by the 1” bolts to the bearing, resulting in the deformed shape shown in Figure 36. There is minimal upwards bending of the bearing sole plate due to the distance between the guide bars and the 1” bolt line. Two gage lines of bolts are generally not advisable for T-stub connections because the majority of the load is resisted by one line of bolts, unless the plates are stiffened (Kulak, et al., 1987). The bottom flange was stiffened with the three vertical bearing stiffeners, however, the shoe plate cannot be stiffened and relies on its thickness alone to provide stiffness. The deformed shape of the shoe plate at the Nipigon River Bridge, collaborates the theory and analysis that shows the interior gage lines of bolts connecting the girder to the shoe plate are less effective and the exterior gage lines are subject to considerably higher than average forces.

Figure 36. Nipigon River Bridge northwest bearing shoe plate prying under uplift.

There are two potential locations where prying can occur. The first is at the tip of the shoe plate where it deflects and causes a high concentrated contact pressure against the bearing sole plate. This is shown as force Q25 in Figure 36. The magnitude of this prying was determined using the structural model described in Section 6.4.1. Looking at a free body diagram, the two Q25 forces are added to the uplift reaction as the force that must be resisted by the 1” bolts. The second location where prying could occur is at the centreline of the girder web (and some distance on either side), where the shoe plate may deflects upwards and react against the flange. This is shown as force Q22 in Figure 36. Looking at a free body diagram, the force Q22 is added to the uplift reaction as the force that must be resisted by the 7/8” bolts. The magnitude of this prying was determined using the structural model described in Section 6.4.1. The magnitudes of the forces in the bolts are discussed in the subsequent sections. The CHBDC Clause 10.17.2.6 states that connected parts shall be arranged so that prying forces

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are minimized and the prying force shall not exceed 30% of the external load. The CHBDC’s wording has some uncertainty as to whether this prying force is only the additional force due to contact pressure, or whether it includes the uneven force distribution between two lines of bolts. For a single bolt line, as likely envisaged by the Code writers, the latter component would not exist. Referring to the reference in the CHBDC (Kulak, 2005), the intent of the prying limit is to ensure evenness in the bolt loads and the prying factor can be taken as the ratio of the maximum bolt force to the nominal bolt force (total force divided by 40 bolts). The subsequent sections will show that this amount had been exceeded for both the 1” and 7/8” bolts. 6.4.1. Model of Shoe Plate Prying

The Bridge Office created an elastic finite element model of the beam end and the components in the bearing. It consisted of shell elements to represent the various steel plates in the girder and bearing, along with tension only member to represent the bolts and compression only members to represent the contact pressure between connected plates. A constant thickness of 56mm was used for the shoe plate, which is the average thickness, and the bolt holes and counterbores were also modelled as thinner plates. Shell elements beyond the shoe plate connection have a coarser grid. The model is shown in Figure 37.

Figure 37. 3-dimensional finite element model of girder end and bearing in a) isometric view, and

b) sectional view.

Four types of loading were considered as independent and were never combined;

1. concentric uplift (representing a bearing that is free to rotate without creating uneven loading of bolts) with snug tight bolts,

2. longitudinal rotation of 0.8 degrees (0.014 radians) with uplift, with snug tight bolts, 3. transverse rotation with uplift of 0.8 degrees, with snug tight bolts,

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4. concentric uplift with bolts that were pre-tensioned to represent turn-of-nut tightening specified in the Contract.

The result of the analysis are described and shown graphically in the subsequent sections. The forces in the bolts where determined at the loads that occurred under phase 1, as well as the final design loads. The deformed shape of the bearing assembly and girder end is shown in Figure 38 and Figure 39 for an imposed longitudinal rotation of 0.8 degrees and an imposed transverse rotation of 0.8 degrees, respectively.

Figure 38. Deformed bearing assembly due to longitudinal rotation combined with uplift.

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Figure 39. Deformed bearing assembly due to transverse rotation combined with uplift.

Calibration Model 6.4.1.1.

As a calibration to the 3 dimensional model, prying action was also investigated with simple frame models of a strip of the shoe plate (800 mm length with an 83 mm wide section). The section properties are calculated for the minimum shoe plate thickness of 52 mm, and reflect the reduction in shoe plate material at bolt holes and counter bores. Figure 40 shows the representative strip of shoe plate.

Figure 40. Shoe plate strip model - properties.

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The simple model makes several assumptions. The girder bottom flange, which is stiffened with transverse stiffeners at the centre of the shoe plate and at either end of the shoe plate, can be modelled as a rigid plane. The 1” and 7/8” bolts are modelled as members with bolt area and length of bolt shank from bolt head to nut, assuming a single washer at each bolt according to the design drawings. Two models were used to represent snug-fit and pretensioned bolt installation. In an initial model for snug-fit bolts, an imposed displacement is applied equally to all 7/8” bolts, which represent uplift from the bottom flange. The same displacement is applied to the centreline of the shoe plate since its upwards deflection is restricted by the bottom flange. Reactions were placed at the tips of the shoe plate to represent the restraint offered by the sole plate from below. This initial model quantified prying at both 7/8” and 1” bolts for snug fit installation (initial assumption of no tension in bolt). The upper reaction at the centre of the plate is the prying force Q22 while the lower reactions at the tips of the shoe plate are the prying forces Q25 from Figure 36. In a subsequent model load was applied in a downwards direction at the 1” bolts, neglecting prying of the 1” bolts, and an incremental analysis was performed to accumulate the forces in the elements representing the bolts as they change from a prestressed compression element to a bolt in tension. These are the bolt forces B22 from Figure 36. The frame model, with pretensioned bolt behaviour, is shown in Figure 41.

Figure 41. Shoe plate strip frame model for pretensioned 7/8” bolt behaviour.

The results of the refined and simple models were in agreement. The simple model slightly underestimates the magnitude of prying and the corresponding bolt forces since it assumes the girder flange and bearing sole plate are infinitely stiff. The refined model accounts for the actual stiffness of the bottom flange and the sole plate of the bearing.

CONTACT BETWEEN SHOE PLATE AND FLANGE PLATES

(COMPRESSION ONLY)

7/8” BOLTS (AREA OF CONTACT SURFACE BETWEEN SHOE PLATE AND

FLANGE PLATE IN COMPRESSION, BOLT AREA IN TENSION)

1” BOLTS

LOADING APPLIED AT 1” BOLTS

(PRETENSIONED BOLT BEHAVIOUR)

VERTICAL

SUPPORT

SHOE PLATE HORIZONTAL SUPPORT

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6.4.2. Analysis Results

The analysis is summarized in several charts showing the uplift load and the maximum force in the bolt. Figure 42 shows the chart for the force in the 1” bolt under design loads. Figure 43 shows the chart for the force in the 7/8” bolt under loading in phase 1. Figure 44 shows the force in the 7/8” bolts under design loads. The values from these figures are summarized in Table 11 below. Table 11. Bolt Forces from Structural Model Based on Elastic Analysis (kN)

Phase 1 (Failure) Loading Design Loading

FLS SLS ULS FLS SLS ULS

1” B

olt

s

Force with even load distribution on bolts

286 - 286 286 286 286 - 286 286 286

Bolt Force with no rotation 286 - 286 286 286 286 - 286 302 *351

Bolt Force with Longitudinal Rotation 286 - *330 #385 #445 315 - #410 #700 #823

Bolt Force with Transverse Rotation 286 - 286 286 286 286 - 286 *395 #465

Bolt Force with Pre-Tensioned 1” Bolts and no rotational effect

286 286 286 286 302 *351†

7/8”

Bo

lts

Force with even load distribution on bolts (uplift divided by 40 bolts, snug-fit installation, neglects prying)

31 - 45 51 66 44-61 110 133

Bolt Force with no rotation 110-160 184 238 157-220 #396 #476

Bolt Force with Longitudinal Rotation #320 - #440 #495 #630 #410 - #495 #1156 #1156

Bolt Force with Transverse Rotation *265 - #365 #405 #515 #310 - #480 #702 #790

Bolt Force with Pre-Tensioned 7/8” Bolts and no rotational effect

219 - 219 220 *249 219 -*239 #396 #476†

* #

†

Bolts with Force Exceeding Factored Resistance.

Bolts with Force Exceeding Ultimate Strength – likely some plastic deformation would occur prior to failure.

Due to limitations in the modelling, the forces in the pretensioned bolts could be higher than stated.

6.4.3. Prying of Outer 1” Bolts Connecting the Shoe Plate to the Bearing

Figure 42 shows that the maximum force in the 1” bolts. As described below, the 1” bolts were adequate under the loading for phase 1, and would have been somewhat deficient under the full design loads.

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Figure 42. Force in 1" bolts under phase 1 and design loads.

1) Pure Uplift (no rotation)

The detailed analysis considers scenarios for 1” bolts both snug-tight and pre-tensioned. The detailed analysis found that there was a prying force on the 1” bolts for the uplift without rotation. The bolt force ranged from 154% to 140% of the average bolt force (i.e. total uplift divided by 24 bolts), being slightly larger near the locations of the vertical stiffeners. Using the simple model, the value was 128%. The difference can be explained by the bending of the bearing sole plate and of the bottom flange – both which were assumed to be infinitely stiff in the simple model.

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For calibration, when the flange and sole plate were made infinitely rigid in the 3-D model, the value of bolt force amplification was 126%. From Figure 42, it can be seen that the 1” bolts would have been adequate for the full loading in phase 1, and for SLS design loads. The 1” bolts would have been overloaded for the ULS design loads.

2) Uplift With Rotation With the full design rotation, there is separation of the guide bars and the uplift force is concentrated on only part of the bearing. This leads to only some of the 1” being partly or even minimally being involved in resisting the uplift force. Under the normal fatigue range (1750 kN to 2450 kN) of design loads, the force in the 1” bolt would be at or exceeding the bolt resistance for longitudinal rotation. For the SLS and ULS design loads, the 1” bolts would likely exceed their ultimate strength for both longitudinal and transverse rotation. For the fatigue range and SLS loading in phase 1 (not shown in Figure 42), the 1” bolt forces would have also been close to the ultimate capacity of the bolt under longitudinal rotation. Under transverse rotation, the 1” bolts would have adequate capacity in phase 1. However, the 7/8” bolts would have been even more overloaded and their yielding, plastic deformation and load sharing would have reduced the forces on the 1” bolts as well compared to the above linear analysis. It is also possible that the actual rotations were not quite this high as there was no physical indication of damage to the 1” bolts.

3) Uplift with Pre-tensioned Bolts (no rotation)

At low uplift force values, the bolt force remains constant at the pre-tension value of about 286 kN. At high uplift forces, the contact pressure induced by the pre-tensioning is fully relieved and the behaviour of the connection is similar to the case where there was no pretension. From Figure 42, it can be seen that the bolts would have remained fully pre-tensioned for the fatigue cyclic loads and for the ULS loads in phase 1. At about 3500 kN, separation of the plates would begin to occur. At the design SLS load of 4410 kN, the bolt force would be just over 300 kN. At the ULS design load of 5300 kN, the force would be similar to the case without pre-tension, and would be in excess of the bolt factored resistance, although still below a value that would cause likely failure. The 1” bolts were under-designed for the design loads, but this deficiency likely played no part in the failure. 6.4.4. Prying of the 7/8” Bolts in Phase 1

Figure 43 shows that the maximum force in the 7/8” bolts under the loading in phase 1. As described below, the 7/8” bolts and shoe plate would have been adequate or marginally adequate for the loading in phase 1 if the bearing could accommodate rotation. The bolts would, however, be subject to large cyclic loading if they were not pre-tensioned, and small cyclic loads (which is still undesirable) if they were pre-tensioned. Without the ability of the bearing to properly rotate, the bolt forces and cyclic forces would be excessive, the exact magnitude depending on the actual values of rotation that was imposed.

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Figure 43. Forces in 7/8" bolts for phase 1 (failure loads at northwest bearing).


The maximum force amplification for the exterior lines of bolts was found to be 350% of the average bolt force (i.e. total uplift divided by 40 bolts). For the simple calibration model, the magnitude of the force amplification was found to be 286%. To isolate this difference, the flange stiffness was increased in the 3D model and the bolt amplification was found to be 290%, which was very close to the simple model. Thus the differences can be explained by the longitudinal variation of the flange stiffness due to the vertical bearing stiffeners. There were two other factors that contributed to the bolt force amplification. The first is the true prying force, caused by contact pressure near the girder centreline from the upward deflection of the shoe plate. The

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second is because the stiffer load path is through the outer line of bolts since it involves less bending of the shoe plate in the load path. The contribution of this second component could be isolated by removing the contact pressure elements from the 3D model. It was found that the stiff load path was the largest factor and accounted for about 50% of the total bolt force amplification, while true prying accounted for about 30% and varying stiffness of the bottom flange accounted for about 20%. The inner line of bolts carried virtually no tension. This varied from zero tension away from the vertical stiffener plates, to a force of about 10% of the average bolt force near the stiffener plates. Under ULS loading in phase 1, the maximum bolt force would be roughly equal to the factored capacity. The weaker A36 shoe plate would have exceeded its yield capacity at a load between the SLS and ULS load.

2) Uplift With Rotation

With the full design rotation, the force in the bolt would have exceeded the capacity of the 7/8” bolt at loads below the normal fatigue range. At the SLS load, the bolt force would have exceeded its ultimate strength. The shoe plate would have also yielded at loads within the normal fatigue range. Failure of the bolts and shoe plate did not happen immediately after opening since the actual rotations were likely less than the design rotation and since plastic deformation of the bolts allowed some load sharing and load redistribution.


At low uplift force values, the bolt force remains constant at the pre-tension value of about 219 kN. At high uplift forces, the contact pressure induced by the pre-tensioning is fully relieved and the behaviour of the connection is similar to the case where there was no pretension. At the normal fatigue load range in phase 1, the bolt force would remain unchanged at the pre-tension value. At roughly the SLS load, the shoe plate would begin to separate at the most heavily loaded bolt. Somewhat prior to the ultimate load, the A36 shoe plate would yield. At the ULS load, the bolt force would roughly equal the factored bolt capacity. 6.4.5. Prying of the 7/8” Bolts At Centre-West Bearing Design Loads

Figure 44 shows that the maximum force in the 7/8” bolts under the full design loading at the centre-west bearing. As described below, the 7/8” bolts and shoe plate would not have been adequate for the SLS or ULS design loading. The bolts would have also been subject to cyclic loading since the flange and shoe plate would separate under fatigue loading.

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Figure 44. Forces on 7/8" bolts under design loads (centre-west bearing).


Under ULS design loading, and even SLS loading, the maximum bolt force would greatly exceed the factored capacity and even the ultimate capacity. The weaker A36 shoe plate would have exceeded its yield capacity at a load at the top of the fatigue range, while even a shoe plate with 350A steel would have yielded before the SLS load.

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2) Uplift With Rotation

This condition was not explored for the design loads since it became obvious that even under the reduced loading in phase 1, the bolts and shoe plate would have had their capacities greatly exceeded.


The bolt pretension would have been overcome near the top of the normal fatigue range. The bolts would have not failed, but been subjected to fatigue loading, under which they perform very poorly. The A36 shoe would have also yielded with a load near the top of the fatigue range. Before the SLS load, the 7/8” bolts would have exceeded their ultimate capacities and even the 350A shoe plate would have yielded.

6.5. Evaluation of Bolt Fatigue

The following factors contribute to fatigue of the 7/8” A490 bolts connecting the bottom flange with the shoe plate.

1. Axial tension of the bolts due to the force range on the bearings in uplift, which are amplified by prying of the shoe plate since the 7/8” bolts are not tightened.

2. Local prying of the bolts due to eccentricity of bearing on the nut, due to lack of bearing in the longitudinal direction of the bridge (no bevelled washer and lack of pretension of bolts). As described in Section 6.3, this amounts to about 160 MPa.

3. Local prying of the bolts due to eccentricity of bearing on the nut, in the transverse direction, due to prying and bending of the shoe plate with lack of pretension of bolts.

4. Amplification of applied tension of the bolts due to inability to accommodate rotation. This line in shown in the preceding charts and results in greatly increased bolt forces.

The first item, prying of the shoe plate, is discussed in Section 6.4. At failure, the force range at the northwest bearing due to FLS was 563 kN (see Table 2) which corresponds to an average bolt fatigue force range with no rotation of about 50 kN (see Figure 43) and a stress range fsr of 142 MPa. Without prying, and having the FLS force divided evenly on the 40 bolts, the stress range would be 40 MPa. At FLS, the CHBDC requires that average fatigue stress range, 0.52 fsr, (74 MPa, or 20 MPa for evenly loaded bolts) is less than the fatigue stress range resistance Fsr. The fatigues stress range resistance is calculated based on the number of cycles of live load or the constant amplitude threshold stress range. A490 bolts have a constant amplitude threshold stress range Fsrt of 262 MPa according to CHBDC, however implicit in this is the assumption that the bolts would be fully pre-tensioned. For simple threaded parts without pretension, the fatigue category is “E”, with a constant amplitude threshold stress range, Fsrt of 31 MPa. Therefore, based on infinite cycles, Fsr is 16 MPa (Fsrt/2). At failure, the number of trucks which crossed the bridge is estimated at 1300 trucks per day for 50 days. For the number of fatigue cycles at 50 days in service, the fatigue stress range resistance (Fsr= 177 MPa), exceeds the applied 0.52 fsr stress value of 74 MPa, therefore, fatigue of the bolts in axial tension alone from prying does not explain the failure at this early time. The resistance at an infinite fatigue life (16

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MPa), is less than the applied 0.52 fsr stress value (of 20 MPa) with full even load distribution, meaning that the lack of tightness could have eventually lead to failure even if prying did not exist or at least an inadequate amount of safety against fatigue failure. Since the bolts were not pretensioned, they would loosen after several load cycles, and there could be an additional impact added to the bolt loading not accounted for above. The second item, the lack of bevelled washer, is expected to add a cyclic stress of about 160 MPa to the extreme fibre of the bolt, as described in Section 6.3. This high value would be on the west face of the bolt, although it could exist to a lesser degree on the east face depending on which span is loaded and the resulting direction of girder rotation. With this added to the 74 MPa, the actual stress range is larger than the fatigue stress range resistance (160 + 74 = 234 > 177 MPa), and failure could be expected, or at least an inadequate amount of safety against fatigue failure. The third item relates to local prying of the bolts due to lack of pretension. As discussed previously, local prying causes local yielding of the bolts which has the potential to alleviate the local prying effects. Nevertheless, local prying is unpredictable and often leads to premature failure of bolts (Kulak, et al., 1987). At service loads, this is expected to be a smaller effect. The fourth item, amplification of applied tension due to inability of the bearing to accommodate rotation, increases the applied loading on the bolts at the front and back of the bearing considerably. Assuming an amplification factor of 3 (as is visible from Figure 43), the bolt fatigue stress range due the passage of a fatigue truck (3 x 74 = 222 MPa). This in itself exceeds the stress range resistance for the actual number of cycles at failure (222 > 177 MPa). In combination with the lack bolt tightness and lack of bevelled washer, the fatigue stress (222 + 160 = 382 MPa) would exceed the allowable at the time of failure (177 MPa) by about 115%. Finally, the bolts are beyond yield for the above noted amplification and the CHBDC equations are strictly for loads within the elastic range. These load levels would mean that the bolts would be subject to an alternating plasticity mechanism and not true fatigue. This is also referred to as or low cycle fatigue , since the steel can tolerate significantly less stress cycles than when within the elastic range. Figure 19 shows beach marks on the fracture surface which are consistent with high stress, lower cycle loading. At the northwest bearing which failed, the capacity of the bolted connection between the girder bottom flange and shoe plate is weaker than the applied forces required to separate the guide bars of the bearing. Simply put, the bolts could have exhibited plastic behaviour before separation occurred.

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Figure 45. Possible force distribution in bolts, after plasticity.

a)

b)

c)

d)

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A plastic analysis was not performed on the shoe plate connection since the elastic analysis showed significant enough overload to conclude the deficient elements within the connection. Conceptually, it would be expected that the bolts at one end or the other would elongate when the girder is subject to longitudinal rotation since they are subject to the highest loading (see Figure 45a). This elongation would open a small gap between the shoe plate and flange. When the rotation switches directions, the bolts at the other end would elongate and a gap would open at that end (see Figure 45b). When the connection is then subject to pure uplift load (or minimal rotation), much of the load would be carried by bolts closer to the bearing centerline since they did not previously elongate during the girder rotations. These bolts could be carrying more load than calculated using the elastic analysis. This is shown in Figure 45c). When next subject to rotation and uplift, the gap would open some more, leading to plastic elongation of the extreme bolts. This plastic elongation would progress to not just the extreme bolt (west most bolt in Figure 45d), but also to the neighbouring bolts. Depending on the number of pure uplift load cycles and uplift with rotation, it could be either the bolts at the centerline of bearing, or the edge bolts, that fail in low cycle fatigue first. Once the first bolt fails, the load on other bolts increases, leading to a progressive failure of the connection.

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7. Discussion To summarize the pertinent results of the evaluation, there are 6 mechanisms that contributed to failure of the bolts at the northwest bearing, with the first 3 being the most significant.

1. Amplification of applied tension of the bolts due to prying of the flexible shoe plate. The prying of the shoe plate, the relative stiffness of the load path through the outside line of bolts, and the varying stiffness of the bottom flange, amplifies the bolt force in the exterior 7/8” bolt lines and renders the bolts of the interior 7/8” bolts almost completely ineffective. Prying action of the shoe plate is influenced by the detailing of the 1” bolts between the shoe plate the bearing, by the bearing supplier. However, it is not clear that another connection would have been feasible, and the designer specified the 7/8” bolt configuration between girder bottom flange and shoe plate without conditions. In simple terms, based on elastic analysis, the total effect of prying of the shoe plate leads to bolt forces which could be up to 3.5 times greater than the average bolt force (assuming uplift reaction divided by 40 bolts connecting the girder bottom flange to the shoe plate).

2. Amplification of applied tension of the bolts due to the bearing’s inability to

accommodate rotation. The bearing’s inability to accommodate rotation causes an amplification of the applied tension in the 7/8” bolts. In isolation of other mechanisms, the bolt force at front or back of the shoe plate could be amplified 2.8 times the average force at a 7/8” bolt row.

3. Local prying of the snug tight bolts due to eccentricity of bearing on the nut (loose bolt

and no beveled washer). Local prying increases the bolt tension on one side caused by bending of the bolt from the uneven contact. It was found that this would add a cyclic stress of about 160 MPa to the extreme fibre of the bolt. This effect would have been eliminated with properly pre-tensioned bolts. When combined with the effects described 1 and 2, local prying makes it more likely that the extreme fibre of the bolt would exceed yield stress at working and fatigue loads. With fatigue stress due to bending in addition to the direct cyclic axial tension on the bolt, the fatigue resistance was exceeded at a small number of cycles and lead to high fatigue stresses and a high-stress, low-cycle fatigue failure.

4. Local prying of the bolts due to eccentricity of bearing on the nut, in the transverse

direction due to bending and prying of the shoe plate. Similar to the local prying due to the lack of bevelled washer, this local prying is caused by uneven bearing of the nut against a shoe plate that bends. Since the shoe plate was expected to yield at some load levels, this local prying is expected to occur, although its magnitude is difficult to quantify. It is a factor that leads to unpredictable behaviour and

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potentially premature failure of bolts (Kulak, et al., 1987).

5. Premature yielding and inadequate bending capacity of the shoe plate as specified in the design, separate from its flexibility.

6. Premature yielding and inadequate capacity of the shoe plate due to a lower supplied material grade than specified. Premature yielding of the shoe plate before yielding of bolts can increase local prying, and lead to greater elongation and non-concentric loading of the bolts. The inadequate shoe plate could have been a catalyst of failure but is not likely a large cause of the failure. The capacity of the shoe plate would have been a larger factor in the final configuration if the bolts were properly pre-tensioned and the bearing could properly accommodate rotation. Even with a Grade 350A shoe plate, the capacity would have been exceeded under the design loads.

The llikely load on the northwest bearing was 2045 kN (based on SLS phase 1 load). The supplier used a common design for all bearings at the west abutment. The highest load specified at ULS at a west abutment bearing was 5300 kN, for the centre bearing. This means that the failure occurred at a load of only 40% of the load the bearing was designed to take. Another purpose of the report was to evaluate the ability of the components, in the load path from the girder to the west abutment, to meet the requirements of the CHBDC. The evaluation showed that the shoe plate, bolted connection between shoe plate and girder, bolted connection between shoe plate and bearing, and bearing design all failed to meet the requirements of the CHBDC. These are summarized in Table 12.

7.1. Design and Construction Requirements

Table 12 summarizes the design and construction requirements of each component of work related to the west abutment bearings, and compares the installed conditions to the contractual requirements. The comments explain the inter-relationships between components.

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Table 12. Design and Construction Requirements for Each Component of Work Related to the West Abutment Bearings of the Nipigon River Bridge

Component What are the requirements? Was the provided information complete? Were the requirements met?

Comments

Overall bridge design, calculation of reactions, movements, and rotations at the bearings

Bridge to be designed according to the CHBDC to satisfy SLS, ULS and FLS.

Yes – the provided design information was complete. The correctness of the design data was not verified by the MTO.

Conformance with the CHBDC was not independently verified by the MTO.

The Specialist Construction Engineer was able to design the erection respecting the limit states of the CHBDC. The as-built reactions provided by the Specialist Engineer are within the limits of the Contract Drawings.

Contract Documents (for bearings)

Contract Documents are complete and convey to the Contractor all the necessary requirements to the design.

Yes – the structural requirements for the bearings are completely defined in the Contract Drawings (loads, movements and rotations at all limit states). The shoe plate and bolted connection to the girder are detailed on the Contract Drawings.

Although the requirements are complete, the designer may not have included the appropriate requirements. The designer used OPSS 1203 in the Contract although the specification is written for supply of bearings already prequalified by the MTO and on the DSM, not for the supply bearings subject to permanent uplift. OPSS 1203 places the responsibility for the design of the shoe plate connection to the bearings with the Contractor. Meanwhile, the same connection between the bearing and shoe plate affects the forces on the shoe plate and the capacity of the bolted connection to the girder. The designer did not require the Contractor to verify the impact of their bearing design on the structure, and did not require the Contractor to submit the connection for approval.

Shoe plate Shoe plate to be designed to the CHBDC for the loads transferred from the girder to the bearings.

Yes – the shoe plate and the bolted connection to the girder are detailed on the Contract Drawings, indicating they were designed by the bridge designer.

No – as designed, the shoe plates do not meet the requirements of the CHBDC for the bearing reactions specified on the Contract Drawings.

The force effects on the shoe plate are affected by the way the shoe plate is connected to the bearing (to be designed by the Contractor). The responsibility for verifying the shoe plate was not transferred to the Contractor. No requirements were included in the Contract to require the Contractor submit the final connection to the Design Engineer for review or approval.

MMM did not provide calculations for the shoe plate design, or for the bolted connection between the girder and shoe plate that were dated prior to failure.

OPSS 1203 requires the shoe plate be supplied with the bearing. Contractor to supply the shoe plate as specified on the Contract Drawings.

No – the shoe plate supplied was a different grade of material.

The material specified on the Contract Drawings was CAN/CSA-G40.21 350W with a yield strength of 350 MPa. The shoe plate supplied was ASTM A36 steel with a yield strength of 248 kPa. As supplied, the shoe plate’s capacity was only 71% of the design on the Contract Drawings.

Bolted connection between the girder and shoe plate

Bolted connection to be designed to the CHBDC for the loads transferred from the girder to the bearings.

Yes – the shoe plate and the bolted connection to the girder are detailed on the Contract Drawings, indicating they were designed by the bridge designer.

No – the bolted connection does not have adequate strength to meet the requirements. Prying of the shoe plate amplifies the force on the bolts.

The capacity of the bolted connection is affected by the attachment of the bearing to the shoe plate. However, no requirements were included in the Contract to require the Contractor to evaluate the connection or submit the final connection or bearing design to the Design Engineer for review or approval.

Contract Drawings specified 32 - M22 A325 high strength bolts complete with bevelled washers.

Yes – the information provided by the Contractor was complete.

No – 32 - 7/8” A490 bolts detailed on Working Drawings but were too long and not detailed with bevelled washers.

Imperial 7/8” bolts supplied are an acceptable equivalent of M22 bolts. Bolts detailed and supplied were too long for the application. The Contractor issued a non-conformance report (NCR 213-6000-224) identifying the long bolts and proposing 16 mm structural steel washer plates. The proposal was accepted but never implemented. It is unclear why the supplied bolts were A490 instead of A325 as specified in the contract documents and it is unclear when the decision for the substitution was made.

Supply A490 high strength bolts compliant with the requirements of the ASTM A490 specification

Yes The Contractor submitted Bolt Certifications as required in the specifications. NRC and Surface Science Western conducted strength and notch toughness testing of failed bolt specimens. Tests on failed bolt specimens confirmed that the material requirements for the A490 bolts met the requirements.

Bolts to be pretensioned by turn-of-nut method per the CHBDC.

No – Bolts were not pretensioned Without properly pretensioned bolts, the moment in the shoe plate is 50% higher than with pretensioned bolts. The shoe plate yields in service.

Non-pretensioned bolts in this connection are subjected to local prying which causes uneven loading of the bolt head and nut, bending of the bolts along their shank, and local yielding of the bolts.

Non-pretensioned bolts are subject to cyclic loading, and subject to a large fatigue stress range in service.

The CHBDC requires that all high strength bolts in a bridge be pretensioned.

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Component What are the requirements? Was the provided information complete? Were the requirements met?

Comments

Bolted connection between the shoe plate and the bearing

OPSS 1203 requires the bearing be mechanically fastened to the shoe plate and the connection be designed for the loads on the Contract Drawings.

Yes/No – the bolted connection is adequate, but only if the bearing were able to rotate.

The bolted connection is adequate only if the bearing permits rotation in uplift. Since the bearing does not permit rotation in uplift, the bolted connection is inadequate considering the amplification of the bolt force due to inability to accommodate rotation in uplift.

Bearing design Bearing shall meet the design requirements of the Contract Documents.

No

The bearings do not allow rotation in uplift.

The bearings do not use the material and grade of steel specified in the Contract.

The bearing sliding surfaces of the uplift restraint guide bars at the centre-west bearing do not meet the requirements of CHBDC and OPSS 1203.

Uplift is restrained by interlocking guide bars between the sole plate and the masonry plate of the bearing which slide along a horizontal plane. In uplift, the bearing does not allow the rotations specified in the Contract Drawings. When forced to rotate, the guide bars separate and the uplift force is transferred at the front or back of the guide bars. The bearings were not designed to accommodate any separation of the guide bars. Therefore, the bearings are not designed to accommodate any rotation in uplift.

The material specified on the Contract Drawings was CAN/CSA-G40.21 350W with a yield strength of 350 MPa. The bearings were manufactured of ASTM A36 steel with a yield strength of 248 kPa.

The average contact pressures for the northwest bearing are within the CHDBC limits, but the contact pressures of the centre-west bearing exceed the CHBDC limits. The centre-west bearing PTFE sliding surface is inadequate, even for uniformly applied pressure.

Connection of the bearing to the abutment

The bearing to be prestressed down to the abutments with 120% of the ultimate uplift load indicated on the Contract Drawings. Addendum 2 requires the prestressing bars be detailed by the Contractor.

Yes Prestressing bars were detailed on the bearing drawings and meet the requirements stated on the Contract Drawings and the Bearings NSSP in Addendum 2.

Bars installed and prestressed to the level indicated on the bearing Working Drawings.

Yes Bar stressing report confirms stressing of 16 bars on October 6, 2015.The bars were stressed to 120% of the ultimate uplift load as required by the design.

Working Drawings OPSS 1203 requires the Contractor submit Working Drawings to the CA at least 1 week prior to commencement of fabrication. OPSS 1203 requires the Contractor submit a Certificate of Compliance to the CA upon completion of the fabrication and prior to installation of the bearings.

Yes – the Working Drawings and Certificate of Compliance were submitted.

Working Drawings revision No. 7 were submitted to the CA with two seals dated March 11, 2015. A Certificate of Compliance was written by Remisz Consulting Engineers Ltd. stating that the Disktron Bearing shop drawings were in general compliance with the Contract Drawings and Specifications, the applicable sections of the CAN/CSA-S6-06, CAN/CSA-G40.21-M04, OPSS 1203, and AASHTO LRFD Bridges Design Specifications for Highway Bridges, 6th Edition. Working Drawings and Certificate of Compliance were received by the MTO on March 30, 2015.

Per OPSS 1203, verify that the drawings are consistent with the Contract Documents and sound engineering practices.

No – the Engineer (working for the bearing supplier) sealed the Working Drawings even though the requirements were not met.

The bearings fail to meet the requirements of the Contract Documents. The incorrect grade of steel is used, the bearings do not accommodate rotation in uplift, the PTFE is overstressed, and the bolts are not equipped with bevelled washers. The Engineer sealed the drawings without properly verifying the Working Drawings were consistent with the Contract Documents.

Per OPSS 922, submit a Certificate of Conformance stating that the fabrication, installation and adjustments are in general conformance with the requirements of the layout and installation drawings, Working Drawings, and Contract Documents for the bearings.

No – the Quality Verification Engineer (QVE) submitted a Certificate of Conformance even though the requirements of the Contract Documents were not met.

The QVE issued a Certificate of Conformance, sealed November 24, 2015, stating the bearings were installed in general conformance with the stamped Working Drawings and the requirements of the Contract Documents. The Certificate of Conformance was received by the MTO on November 29, 2015. Despite this, the bearings fail to meet the requirements of the Contract Documents. The incorrect grade of steel is used, the bearings do not accommodate rotation in uplift, the bolts were not equipped with bevelled washers, and the bolts were not pretensioned upon installation. The QVE sealed as-built drawings.

Verify the submission contains all the submission requirements outlined in the Contract Documents, specifically OPSS 1203.

Yes – the information listed in OPSS 1203 appears on the Working Drawings. The Working Drawings were sealed

Although OPSS 1203 only requires that one Engineer affix his or her seal and signature on the Working Drawings, verifying that the drawings are consistent with the Contract Documents, the Working Drawings submitted contained 2 stamps.

- 65 -


BRO-059

8. Conclusions The structural analysis of the bearing and its connections to the adjacent components of the bridge revealed that the failure at the northwest bearing of the Nipigon River Bridge was caused by:

1. prying effects due to the flexible shoe plate leading to higher forces in the exterior line of bolts,

2. the bearing’s inability to accommodate rotation leading to higher forces in the end rows of bolts,

3. the lack of pretensioning of the bolts and lack of bevelled washers that lead to high fatigue stresses and a high-stress, low-cycle fatigue failure.

Each of these of factors on its own is significant and could have led to a failure, but combined they made failure inevitable. Other factors which also contributed to and accelerated the failure include local bending of the bolts, yielding of the shoe plate. Another purpose of the report was to evaluate the ability of the components, in the load path from the girder to the west abutment, to meet the requirements of the CHBDC. The evaluation showed that the shoe plate, bolted connection between shoe plate and girder, bolted connection between shoe plate and bearing, and bearing design all failed to meet the requirements of the CHBDC.

- 66 -


BRO-059

9. References ASTM Standard A490, 2008. Standard Specification for High-Strength Bolts, Classes 10.9 and 10.9.3, for Structural Steel Joints. West Conshohocken, PA: ASTM International.

Canadian Institute of Steel Construction, 2008. Handbook of Steel Construction. Tenth Edition ed. Toronto: Canadian Institute of Steel Construction.

Canadian Standards Association, 2006. CAN/CSA-S6-06: Canadian Highway Bridge Design Code. Mississauga, Ontario: CSA.

Kulak, G. L., 2005. High Strength Bolting for Canadian Engineers. 1st ed. Toronto: Candadian Institute of Steel Construction.

Kulak, G. L., Fisher, J. W. & Struik, J. H. A., 1987. Guide to Design Criteria for Bolted and Riveted Joints. New York: John Wiley and Sons.

Kulak, G. L. & Grondin, G. Y., 2006. Limit States Design in Structural Steel. Eighth Edition ed. Toronto: Canadian Institute of Steel Construction.

Makar, J., 2016. Evaluation of Failed Nipigon River Bridge West Abutment Bolts, May 16 2016, Ottawa: National Research Council Canada.

Ramamurthy, S. et al., 2016. Report on the Evaluation of the Bolts Provided by MTO, 18 April 2016, London: Surface Science Western.


BRO-059

Appendix A: Contract Drawings Drawings G-1, sheet 190-A-R2, General Arrangement Drawing A-1, sheet 208-C-R2, West Abutment Details Drawing A-10, sheet 218-B-R1, Bearing Plan and Design Data Drawing A-11, sheet 218-1-R2, Bearing Details Drawing S-3, sheet 292-B-R1, Structural Steel West Span North & South Girder Drawing S-7, sheet 296-B-R1, Structural Steel Girder Stiffener Details Drawing S-13, sheet 302-C-R1, Structural Steel Typical Floor Beam

PCRN

Stamp

mal53583

Typewritten Text

mal53583

Typewritten Text

mal53583

Typewritten Text

RECEIVED OCTOBER 10, 2014 - RFC #310

PCRN

Stamp

mal53583

Typewritten Text

INSTRUCTION NOTICE #71 - JUNE 11, 2014

PCRN

Stamp

mal53583

Typewritten Text

INSTRUCTION NOTICE #71 - JUNE 11, 2014

PCRN

Stamp

mal53583

Typewritten Text

RECEIVED OCTOBER 9, 2013 - RFC#012

PCRN

Stamp

mal53583

Typewritten Text

RECEIVED OCTOBER 9, 2013 - RFC#012

mal53583

Typewritten Text

RECEIVED NOVEMBER 13, 2013


BRO-059

Appendix B: Bearing Working Drawings Bearing Working Drawings, RJ Watson Inc., Revision No. 8, As-built, October 21, 2015 Sheet 1 – Notes and Revisions Log Sheet 2 – Bearing Installation Sheet 3 – DB440MR Type I Assembly Sheet 4 – DB2860MR Assembly Sheet 5 – DB1860MR Assembly Sheet 6 – DB440MR Type II Assembly Sheet 7 – Bearing Details-1 Sheet 8 – Bearing Details-2 Sheet 9 – Bearing Details-3 Sheet 10 – Bearing Details-4 Sheet 11 – Bearing Details-5 Sheet 12 – Bearing Details-6 Sheet 13 – Bearing Details-7 Sheet 14 – Bearing Details-8 Sheet 15 – As Built Measurements

SCALE:

PLAN

NONE

BEARING LOCATIONS

UPSTATION

DENOTES THICK END

TYP=TYPICALTPI= THREADS PER INCH

DENOTES QTY.##

CLOUD NOTES ARE TO BEPERMANENTLY MARKED IN

POSITIONS SHOWN ON DETAILS

BEARING #1

BEARING #2

BEARING #3

BEARING #4

BEARING #5

BEARING #6

CL WEST ABUTMENT CL EAST ABUTMENT

CL NORTH GIRDER

CL CENTER GIRDER

CL SOUTH GIRDER

SERIAL NO: 13003-*; (*)=BEARING NUMBER

: 10/21/15

6

DATE: 1/28/15

RELEASED FOR FABRICATIONR.J. WATSON, INC

1. ALL OTHER WELDING SHALL BE PERFORMED IN ACCORDANCE WITH CAN/CSA-W59-M. ALL WELDING SHALL BE DONE WITH ELECTRODES CERTIFIED BY THE CANADIAN WELDING BUREAU TO THE REQUIREMENTS OF CAN/CSA W48.

2. ALL STAINLESS STEEL WELDING SHALL BE IN ACCORDANCE WITH AWS D1.6 WELDING

CODE-STAINLESS STEEL.

3. WELDING TO BEARING PLATES AFTER ASSEMBLY IS PERMITTED PROVIDED WELDING PROCEDURES RESTRICT THE MAXIMUM TEMPERATURE IN THE AREAS OF THE POLYTRON DISC AND PTFE TO NO MORE THAN 107 C AS DETERMINED BY USE OF TEMPERATURE INDICATING WAX PENCILS. PROTECTIVE MATERIAL SHALL BE PLACED OVER THE BEARING ASSEMBLY TO PROTECT AGAINST SPARKS AND FLASH. AFTER WELDING IS COMPLETE, EXPOSED STEEL SHALL BE TOUCHED UP IN ACCORDANCE WITH THE SPECIFIED COATING SYSTEM.

NOTES:GENERAL1. MARK CENTERLINES ON BEARING MASONRY PLATE, BEVELED SHOE PLATE, AND

SOLE PLATE EDGES. THESE IDENTIFICATION MARKS WILL BE USED TO MEASURE OFFSETS IN THE FIELD. USE INDELIBLE INK TO PLACE THESE MARKS.

2. MARK EACH BEARING WITH THE NAME OF THE MANUFACTURER, MODEL NUMBER, INSTALLATION LOCATION, AND DATE OF MANUFACTURE.

3. ALL WORK SHALL BE IN ACCORDANCE WITH CSA STANDARD CAN/CSA G40.21-M04, OPSS 1203, AASHTO LRFD BRIDGE DESIGN SPECIFICATIONS FOR HIGHWAY BRIDGES, 6TH EDITION, AND THE SPECIAL PROVISIONS.

MATERIALS1. STEEL PLATES SHALL BE AS NOTED ON BILL OF MATERIALS. ASTM A36 SHALL

BE SUPPLIED AS ASTM A36 OR BETTER BASED ON MATERIAL AVAILABILITY. GRADE50 (345) STEEL SHALL BE SUPPLIED AS EITHER ASTM A572 GR 50 (345) AND/ORASTM A588 BASED ON MATERIAL AVAILABILITY.

2. ALL STAINLESS STEEL SHALL BE 12 GAUGE ASTM A240, TYPE 304 AND HAVE A NO. 8 BRIGHT MIRROR FINISH UNLESS OTHERWISE NOTED.

3. PTFE SHEET SHALL CONFORM TO THE CANADIAN HIGHWAY DESIGN CODE, SECTION 11.6.3, WITH SPHERICAL RESERVOIRS FOR LUBRICATION WHICH ARE 8MM DIAMETER AND 2.5MM DEEP. THE RESERVOIRS SHALL BE EVENLY DISTRIBUTED ACROSS THE SURFACE AND SHALL OCCUPY 20 TO 30% OFF THE SURFACE. ALL SHEET PTFE SHALL BE ETCHED ON ONE SIDE PRIOR TO BONDING. A UNIFORM LAYER OF CHEMGRIP OR RJ WATSON APPROVED EQUAL EPOXY ADHESIVE SHALL BE APPLIED FOR EFFECTIVE PTFE BONDING.

4. THE POLYTRON DISC SHALL BE IN ACCORDANCE WITH CAN/CSA-56-00 SECTION 11.6.7.2. THE TOP AND BOTTOM SURFACES OF THE POLYTRON DISC SHALL BE ROUGHENED. THE DUROMETER SHALL BE 62D 3 AND THE COLOR SHALL BE RED.

COATINGS1. REMOVE MILL SCALE PRIOR TO HOT-DIP GALVANIZING USING CHEMICAL CLEANING

METHODS.

2. HOT-DIP GALVANIZE IN ACCORDANCE WITH ASTM A123/A123-M STANDARD SPECIFICATION FOR ZINC (HOT-DIP GALVANIZED) COATINGS ON IRON AND STEEL PRODUCTS.

GALVANIZING SHALL NOT BE APPLIED TO THE FOLLOWING SURFACES:-AREAS DIRECTLY ABOVE, BENEATH OR WITHIN 1/4" OF THE POLYTRON DISC-AREAS OF PTFE BONDING-SRM AND SRM CLEARANCE HOLE-STAINLESS STEEL SHEET.

REPAIR GALVANIZED COATINGS USING GALVAGUARD BY TECK COMINCO LTD.

3. CONNECTION BOLTS, ANCHOR RODS, HEAVY HEX NUTS AND HARDENED WASHERS SHALL BE HOT-DIP GALVANIZED BY A METHOD CONFORMING TO ASTM A153/A153M.

4. ASTM A490 BOLTS SHALL RECIEVE A DACROMET COATING IN ACCORDANCE WITHASTM F1136.

WELDING

SHIPPING AND HANDLING1. PTFE AND STAINLESS STEEL SHALL BE PROTECTED FROM DAMAGE, AIRBORNE DEBRIS

AND CONTAMINANTS AT ALL TIMES. THESE SURFACES SHALL BE INSPECTED FOR SUCH DAMAGE AND DEBRIS PRIOR TO FINAL ASSEMBLY AND INSTALLATION.

2. COMPLETED BEARINGS SHALL BE INDIVIDUALLY BANDED IN THE UPRIGHT POSITION. BANDS SHALL PREVENT SEPARATION OF UPPER AND LOWER BEARING COMPONENTS AND SHALL NOT BE REMOVED PRIOR TO INSTALLATION.

3. BEARING ASSEMBLIES SHALL BE HANDLED BY THEIR BOTTOM SURFACES ONLY, AND SHALL NOT BE LIFTED BY THEIR TOPS, SIDES AND/OR SHIPPING BANDS.

4. BEARINGS SHALL BE STORED IN A CLEAN, DRY, AND UPRIGHT POSITION.

5. AT NO TIME PRIOR TO THE COMPLETION OF THE PROJECT MAY ANY BEARING BE DISASSEMBLED WITHOUT AUTHORIZATION FROM RJ WATSON, INC.

INSTALLATION1. CONTRACTOR SHALL TAKE SPECIAL CARE TO PROTECT STAINLESS STEEL AND PTFE

SURFACES FROM DAMAGE AND/OR DEBRIS INTRUSION DURING THE INSTALLATION OF BEARINGS AND ANCHORAGE.

2. BEARINGS SHALL BE INSTALLED IN ACCORDANCE WITH THE TOLERANCES PROVIDED IN AASHTO LRFD BRIDGE CONSTRUCTION SPECIFICATIONS, SECTION 18.3.5.

5. CONTRACTOR SHALL ADJUST PEDESTAL ELEVATIONS TO ACCOMMODATE FINAL BEARING HEIGHT.

6. BEARINGS ARE TO BE SET LEVEL.

REVISIONS

REV. DESCRIPTION SHEET NO. DATE

1 POST-TENSIONING SYSTEM ADJUSTMENTS 2-3, 6 9/3/2014

2 ADDED GROUT TUBE 2-3, 6 9/25/2014

3 FIXED PARTMARK LABEL 8-9 9/30/2014

4 CONNECTION BOLT QTY 3 10/8/2014

5 POST-TENSIONING SYSTEM 2-3, 6 10/21/2014

6 A490 DACROMET BOLTS, TOP ANCHOR ROD 1-3, 6 11/3/2014

7 CORRECTION 10 1/27/2015

8 CONTRACTOR COMMENTS 2-3, 6-8, 10-13 10/21/2015

DRAWING INDEX

SHEET NO. DESCRIPTION

1 NOTES AND REVISIONS LOG

2 BEARING INSTALLATION

3 DB440MR TYPE I ASSEMBLY

4 DB2860MR ASSEMBLY

5 DB1830MR ASSEMBLY

6 DB440M TYPE II ASSEMBLY

7 BEARING DETAILS-1

8 BEARING DETAILS-2

9 BEARING DETAILS-3

10 BEARING DETAILS-4





15 AS BUILT MEASURMENTS

BEARING NUMBER MODEL NUMBER BEARING LOCATION GIRDER

1 DB440MR TYPE I WEST ABUTMENT NORTH2 DB440MR TYPE II WEST ABUTMENT CENTER3 DB440MR TYPE I WEST ABUTMENT SOUTH4 DB1830MR EAST ABUTMENT NORTH5 DB2860MR EAST ABUTMENT CENTER6 DB1830MR EAST ABUTMENT SOUTH

8 DATEREV.NO.

AS PER AASHTO LRFD BRIDGE CONSTRUCTIONSPECIFICATIONS TABLE 18.1.4.2-1.

UNLESS NOTED OTHERWISE

PROJECT INFORMATION:

DO NOT SCALE DRAWING

TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:

OWNER:ALDEN, NEW YORK 14004

11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 1 TO 15

ORDER NO:

TITLE:

DATE:

DATE:ONTARIO MINISTRY OF TRANSPORTATION

MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

NOTES AND REVISIONS LOGTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01

NIPIGON RIVER BRIDGE REPLACEMENT

ALL DIMENSIONS IN MILLIMETERSALL WEIGHTS IN KG.

LONGITUDINAL

UPSTATION

DENOTES THICK END


DENOTES QTY.##



10.000 THICK, NON-SHRINKBEDDING GROUT

600EMBEDMENT

ELEVDB1830MR SHOWNDB2860MR INSTALLED SIMILAR

GIRDER

HEAVY HEX NUT (HC3)6 PLACES

HARDENED WASHER (HB3)12 PLACES

BEVELED SHOE PLATE (PA3)

SOLE PLATE (PE3)

CONNECTION BOLT (HA3)6 PLACES

HARDENED WASHER (HH3)8 PLACES

HEAVY HEX NUT (HJ3)8 PLACES

ANCHOR ROD (HG3)4 PLACES

MASONRY PLATE (PC3)

CONCRETE PEDESTAL

CONCRETE PIER

CL BEARING & GIRDER

10

10

LONGITUDINALONLY.SEE WELDINGNOTE #1

PLANDB1830MR SHOWNDB2860MR INSTALLED SIMILAR

UPSTATION

LONGITUDINAL

CL BEARING

CL GIRDER &BEARING

10.000 THICK, NON-SHRINKBEDDING GROUT

3300

EMBEDMENT

ELEVDB440MR TYPE I SHOWNDB440MR TYPE II INSTALLED SIMILAR

GIRDER

CONNECTION BOLT (HA1)32 PLACES

HEAVY HEX NUT (HC1)32 PLACES

HARDENED WASHER (HB1)32 PLACES

CONNECTION BOLT (HD1)24 PLACES

HEAVY HEX NUT (HF1)24 PLACES

HARDENED WASHER (HE1)48 PLACES

BEVELED SHOE PLATE (PA1)

UPPER BEARING BLOCK (PE1)

TOP ANCHOR ROD (HG1A)8 PLACES

HEAVY HEX NUT (HJ1)16 PLACES

HARDENED WASHER (HH1)8 PLACES

LOWER BEARING BLOCK (PC1)

CONCRETE PEDESTAL

CONCRETE PIER

CL BEARING & GIRDER

WASHER PLATE (HL1)8 PLACES

STOP-TYPE COUPLER (HM1)8 PLACES

1

INTERNAL GROUT TUBE (HN1)8 PLACES

BOTTOM ANCHOR ROD (HG1B)8 PLACES

5

TOP OF PILE CAP

6

PLANDB440MR TYPE I SHOWNDB440MR TYPE II INSTALLED SIMILAR

CL BEARING &GIRDER

CL BEARING

1

1

1

2

: 10/21/15

5

DATE: 1/28/15


8

8

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 2 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

INSTALLATION VIEWSTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



421+ 6.00.0

UNLOADED

477+ 6.00.0 AT CL

UNLOADED

WEIGHT:

ELEV

2967.81 KG.2

DB440MR TYPE I ASSEMBLY

SEE DETAIL A

PA1

PE1

PC1BA1MA1PD1

LONGITUDINAL

UPSTATION

PLAN

WEIGHT:

DB440MR TYPE I ASSEMBLY

2 2967.81 KG.

25 GAP2 PLACES

5 GAP 2 PLACES

20 GAP2 PLACES

DET GAP DETAIL

A

SEE MATERIALSNOTE #3

TT1

ST1

TL1

SL1

SS1


PB1

NOTE: DIMENSIONS IN BILL OF MATERIALS REFLECT THE DIMENSIONS OF RAWMATERIALS TO BE ORDERED. SEE DETAIL SHEETS FOR FINAL DIMENSIONSOF FABRICATED INDIVIDUAL PARTS.

NOTE: (+) LONGITUDINAL DISPLACEMENT REPRESENTSSUPERSTRUCTURE EXPANSION & (-) LONGITUDINAL DISPLACEMENT REPRESENTS SUPERSTRUCTURE CONTRACTION.

111

12

2

: 10/21/15

4

5

5

6

6

66

66

6

DATE: 1/28/15


8

8

ITEM NO.

Job Number Part Mark QTY. TotalQty Length Width Thickness Ø ID DESCRIPTION Material REMARKS

1 13003-01 PA1 1 2 1000 800 63.500 - - BEVELED SHOE PLATE ASTM A36 2 13003-01 PB1 1 2 820 232 44.450 - - UPPER BEARING PLATE ASTM A36 3 13003-01 PC1 1 2 1000 1232 304.8 - - LOWER BEARING BLOCK ASTM A364 13003-01 PD1 1 2 - - 114.300 232 - LOWER BEARING PLATE ASTM A36 5 13003-01 PE1 1 2 1000 800 304.8 - - UPPER BEARING BLOCK ASTM A366 13003-01 ST1 1 2 990 228 2.667 - - TOP STAINLESS STEEL ASTM A240 TYPE 3047 13003-01 SS1 2 4 990 25 2.667 - - SIDE STAINLESS STEEL ASTM A240 TYPE 3048 13003-01 SL1 2 4 990 91 2.667 - - LOWER STAINLESS STEEL ASTM A240 TYPE 3049 13003-01 BA1 1 2 75.620 - - 127 - SRM AISI 4000 SERIES Fy min = 105 ksi

10 13003-01 MA1 1 2 - - 16 182 130 POLYTRON DISC POLYETHER URETHANE 62D11 13003-01 TT1 1 2 785 182 4.760 - - TOP PTFE PTFE UNFILLED DIMPLED LUBRICATED12 13003-01 TS1 2 4 785 19 4.760 - - SIDE PTFE 15% GLASS FILLED SHEET13 13003-01 TL1 2 4 785 52 4.760 - - LOWER PTFE 15% GLASS FILLED SHEET14 13003-01 HA1 32 64 6.000 - - 0.875 - CONNECTION BOLT ASTM A490 DACROMET, MM EQUIV. 22 x 152.415 13003-01 HB1 32 64 - - - 0.875 - HARDENED WASHER ASTM F436 HDG DACROMET, MM EQUIV. 2216 13003-01 HC1 32 64 - - - 0.875 - HEAVY HEX NUT ASTM A563 HDG DACROMET, MM EQUIV. 2217 13003-01 HD1 24 48 8.000 - - 1.000 - CONNECTION BOLT ASTM A490 DACROMET, MM EQUIV. 25.4 x 203.218 13003-01 HE1 48 96 - - - 1.000 - HARDENED WASHER ASTM F436 HDG DACROMET, MM EQUIV. 25.419 13003-01 HF1 24 48 - - - 1.000 - HEAVY HEX NUT ASTM F436 HDG DACROMET, MM EQUIV. 25.420 13003-01 HK1 8 16 131.650 - - 4.000 - CORRUGATED SHEATH AASHTO M270 GR 36 MM EQUIV. 102 x 334421 13003-01 HG1A 8 16 24.500 - - 1.750 - TOP ANCHOR ROD ASTM A722 HDG MM EQUIV. 44.5 x 62022 13003-01 HG1B 8 16 127.800 - - 1.750 - BOTTOM ANCHOR ROD ASTM A722 HDG MM EQUIV. 44.5 x 324323 13003-01 HH1 8 16 - - - 1.750 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 44.5 R9F-1624 13003-01 HJ1 16 32 - - - 1.750 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 44.5 R73-1425 13003-01 HL1 8 16 200 200 22 44.500 - WASHER PLATE ASTM A36 HDG26 13003-01 HM1 8 16 8.500 - - 1.750 - STOP-TYPE COUPLER ASTM A29 GR C1045 MM EQUIV. 44.5 x 215.90 R72-1427 13003-01 HN1 8 16 - - - 0.500 0.375 INTERNAL GROUT TUBE - MM EQUIV. 12.7 O.D. X 9.525 I.D.

DB440MR TYPE I BEARING CAPACITY TABLE

VERTICAL SERVICE LOAD (kN) 440VERTICAL ULTIMATE LOAD (kN) 1160HORIZONTAL SERVICE LOAD (kN) 441

HORIZONTAL ULTIMATE UPLIFT LOAD (kN) 530ULTIMATE UPLIFT LOAD (kN) 5300FATIGUE UPLIFT LOAD (kN) 2450

TOTAL LONGITUDINAL DISPLACEMENT (mm) +75/-130TOTAL TRANSVERSE DISPLACEMENT (mm) 40.00

SERVICE ROTATION (rad) ±.0175STRENGTH ROTATION (rad) ±.0244

WEIGHT (KG) 2967.81APPROX. DEAD LOAD DEFLECTION (mm) 1.60

LOCATION WEST ABUTMENT, NORTH & SOUTH GIRDERSQUANTITY 2

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 3 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

DB440MR TYPE I ASSEMBLYTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01


ALL DIMENSIONS IN MILLIMETERSALL WEIGHTS IN KGS.

DENOTES THICK END


DENOTES QTY.##



162+ 6.00.0

UNLOADED

219+ 6.00.0 AT CL

UNLOADED

WEIGHT:

ELEV

774.19 KG.1

DB2860MR ASSEMBLY

SEE DETAIL B

PA2

PC2BA2 MA2PD2

PE2

LONGITUDINAL

UPSTATION

WEIGHT:

PLAN

774.19 KG.1

DB2860MR ASSEMBLY

5 GAP2 PLACES

5 GAP2 PLACES

11 GAP2 PLACES

DET GAP DETAIL

B

ST2

TT2

SS2

SL2

GB2B

GB2A

TS2

TL2

PB2





: 10/21/15

DATE: 1/28/15


ITEM NO. Job Number Part Mark QTY. TotalQty Length Width Thickness Ø ID DESCRIPTION Material REMARKS

1 13003-02 PA2 1 1 800 800 63.500 - - BEVELED SHOE PLATE ASTM 572 GR 502 13003-02 PB2 1 1 390 390 31.750 - - UPPER BEARING PLATE ASTM 572 GR 503 13003-02 PC2 1 1 800 1000 31.750 - - MASONRY PLATE ASTM 572 GR 504 13003-02 PD2 1 1 - - 31.750 390 - LOWER BEARING PLATE ASTM 572 GR 505 13003-02 PE2 1 1 800 800 31.750 - - SOLE PLATE ASTM 572 GR 506 13003-02 ST2 1 1 616 352 2.667 - - TOP STAINLESS STEEL ASTM A240 TYPE 3047 13003-02 SS2 2 2 616 25 2.667 - - SIDE STAINLESS STEEL ASTM A240 TYPE 3048 13003-02 SL2 2 2 616 29 2.667 - - LOWER STAINLESS STEEL ASTM A240 TYPE 3049 13003-02 BA2 1 1 78.620 - - 83 - SRM AISI 4000 SERIES Fy min = 105 ksi

10 13003-02 GB2A 2 2 642 76.200 76.200 - - LOWER GUIDE BAR ASTM 572 GR 5011 13003-02 GB2B 2 2 642 76.200 76.200 - - UPPER GUIDE BAR ASTM 572 GR 5012 13003-02 MA2 1 1 30 336 86 POLYTRON DISC POLYETHER URETHANE 62D13 13003-02 TT2 1 1 - - 4.760 336 - TOP PTFE PTFE UNFILLED DIMPLED LUBRICATED14 13003-02 TS2 2 2 336 21 4.760 - - SIDE PTFE 15% GLASS FILLED SHEET15 13003-02 TL2 2 2 336 19 4.760 - - LOWER PTFE 15% GLASS FILLED SHEET16 13003-02 HA2 6 6 5.500 - - 0.875 - CONNECTION BOLT ASTM A325 HDG MM EQUIV. 22 x 13917 13003-02 HB2 12 12 - - - 0.875 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 2218 13003-02 HC2 6 6 - - - 0.875 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 2219 13003-02 HG2 4 4 33.250 - - 1.375 - ANCHOR ROD ASTM F1554 GR 55 HDG MM EQUIV. 35 x 844.520 13003-02 HH2 8 8 - - - 1.375 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 3521 13003-02 HJ2 8 8 - - - 1.375 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 35

DB2860MR BEARING CAPACITY TABLE

VERTICAL SERVICE LOAD (kN) 2860VERTICAL ULTIMATE LOAD (kN) 4520HORIZONTAL SERVICE LOAD (kN) 286HORIZONTAL ULTIMATE LOAD (kN) 452

ULTIMATE UPLIFT LOAD (kN) 530TOTAL LONGITUDINAL DISPLACEMENT (mm) +115/-165TOTAL TRANSVERSE DISPLACEMENT (mm) 10.00



LOCATION EAST ABUTMENT, CENTER GIRDERQUANTITY 1

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 4 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

DB2860MR ASSEMBLYTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



157+ 6.00.0

UNLOADED

214+ 6.00.0 AT CL

UNLOADED

WEIGHT:

ELEV DB1830MR ASSEMBLY

750.52 KG.2

SEE DETAIL C

PA3

PE3

PC3MA3BA3PD3

LONGITUDINAL

UPSTATION

WEIGHT:

PLAN

750.52 KG.2

DB1830MR ASSEMBLY

20 GAP2 PLACES

25 GAP2 PLACES

5 GAP2 PLACES

DET GAP DETAIL

C

ST3

TT3


PB3

TS3SS3 GB3B

GB3A

SL3

TL3




: 10/21/15

DATE: 1/28/15


ITEM NO.

Job Number Part Mark QTY. TotalQty Length Width Thickness Ø ID DESCRIPTION Material REMARKS

1 13003-03 PA3 1 2 800 800 63.500 - - BEVELED SHOE PLATE ASTM 572 GR 502 13003-03 PB3 1 2 312 312 31.750 - - UPPER BEARING PLATE ASTM 572 GR 503 13003-03 PC3 1 2 800 1000 31.750 - - MASONRY PLATE ASTM 572 GR 504 13003-03 PD3 1 2 - - 31.750 312 - LOWER BEARING PLATE ASTM 572 GR 505 13003-03 PE3 1 2 800 800 31.750 - - SOLE PLATE ASTM A572 GR 506 13003-03 ST3 1 2 550 316 2.667 - - TOP STAINLESS STEEL ASTM A240 TYPE 3047 13003-03 SS3 2 4 550 25 2.667 - - SIDE STAINLESS STEEL ASTM A240 TYPE 3048 13003-03 SL3 2 4 550 59 2.667 - - LOWER STAINLESS STEEL ASTM A240 TYPE 3049 13003-03 BA3 1 2 70.620 - - 64 - SRM AISI 4000 SERIES Fy min = 105 ksi

10 13003-03 GB3A 2 4 600 76.200 95.250 - - LOWER GUIDE BAR ASTM 572 GR 5011 13003-03 GB3B 2 4 600 76.200 95.250 - - UPPER GUIDE BAR ASTM 572 GR 5012 13003-03 MA3 1 2 25.400 270 67 POLYTRON DISC POLYETHER URETHANE 62D13 13003-03 TT3 1 2 - - 4.760 270 - TOP PTFE PTFE UNFILLED DIMPLED LUBRICATED14 13003-03 TS3 2 4 270 19 4.760 - - SIDE PTFE 15% GLASS FILLED SHEET15 13003-03 TL3 2 4 270 19 4.760 - - LOWER PTFE 15% GLASS FILLED SHEET16 13003-03 HA3 6 12 5.500 - - 0.875 - CONNECTION BOLT ASTM A325 HDG MM EQUIV. 22 x 13917 13003-03 HB3 12 24 - - - 0.875 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 2218 13003-03 HC3 6 12 - - - 0.875 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 2219 13003-03 HG3 4 8 33.250 - - 1.375 - ANCHOR ROD ASTM F1554 GR 55 HDG MM EQUIV. 35 x 844.520 13003-03 HH3 8 16 - - - 1.375 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 3521 13003-03 HJ3 8 16 - - - 1.375 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 35

DB1830MR BEARING CAPACITY TABLE


ULTIMATE UPLIFT LOAD (kN) 480

TOTAL LONGITUDINAL DISPLACEMENT (MM) NORTH GIRDER: +115/-165, SOUTH GIRDER: +95/-165

TOTAL TRANSVERSE DISPLACEMENT (MM) 40.00SERVICE ROTATION (rad) ±.0175STRENGTH ROTATION (rad) ±.0244

WEIGHT (KG) 750.52APPROX. DEAD LOAD DEFLECTION (MM) 2.54

LOCATION EAST ABUTMENT, NORTH & SOUTH GIRDERSQUANTITY 2

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 5 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

DB1830MR ASSEMBLYTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



421+ 6.00.0

UNLOADED

477+ 6.00.0 AT CL

UNLOADED

WEIGHT:

ELEV

2967.81 KG.1

DB440MR TYPE II ASSEMBLY

SEE DETAIL D

PA4

BA4MA4PD4

LONGITUDINAL

UPSTATION

WEIGHT:

PLAN

2967.81 KG.1


7.5 GAP2 PLACES

5 GAP2 PLACES

5 GAP2 PLACES

DET

D

GAP DETAIL

ST4

TT4

PB4

SS4SL4

TL4



1



11

1

2

2

: 10/21/15

5

5

6

6

66

66

6

DATE: 1/28/15


8

8

ITEM NO.

Job Number

Part Mark QTY. TotalQty Length Width Thickness Ø ID DESCRIPTION Material Remarks

1 13003-04 PA4 1 1 1000 800 63.500 - - BEVELED SHOE PLATE ASTM A36 2 13003-04 PB4 1 1 820 232 44.450 - - UPPER BEARING PLATE ASTM A36 3 13003-04 PC4 1 1 1000 1232 304.8 - - LOWER BEARING BLOCK ASTM A36 4 13003-04 PD4 1 1 - - 114.300 232 - LOWER BEARING PLATE ASTM A36 5 13003-04 PE4 1 1 1000 800 304.8 - - UPPER BEARING BLOCK ASTM A36 6 13003-04 ST4 1 1 990 228 2.667 - - TOP STAINLESS STEEL ASTM A240 TYPE 3047 13003-04 SS4 2 2 990 25 2.667 - - SIDE STAINLESS STEEL ASTM A240 TYPE 3048 13003-04 SL4 2 2 990 91 2.667 - - LOWER STAINLESS STEEL ASTM A240 TYPE 3049 13003-04 BA4 1 1 75.620 - - 127 - SRM AISI 4000 SERIES Fy min = 105 ksi

10 13003-04 MA4 1 1 - - 16 182 130 POLYTRON DISC POLYETHER URETHANE 62D11 13003-04 TT4 1 1 785 182 4.760 - - TOP PTFE PTFE UNFILLED DIMPLED LUBRICATED12 13003-04 TS4 2 2 785 19 4.760 - - SIDE PTFE 15% GLASS FILLED SHEET13 13003-04 TL4 2 2 785 52 4.760 - - LOWER PTFE 15% GLASS FILLED SHEET14 13003-04 HA4 32 32 6.000 - - 0.875 - CONNECTION BOLT ASTM A490 DACROMET, MM EQUIV. 22 x 152.415 13003-04 HB4 32 32 - - - 0.875 - HARDENED WASHER ASTM F436 HDG DACROMET, MM EQUIV. 2216 13003-04 HC4 32 32 - - - 0.875 - HEAVY HEX NUT ASTM A563 HDG DACROMET, MM EQUIV. 2217 13003-04 HD4 24 24 8.000 - - 1.000 - CONNECTION BOLT ASTM A490 DACROMET, MM EQUIV. 25.4 x 203.218 13003-04 HE4 48 48 - - - 1.000 - HARDENED WASHER ASTM F436 HDG DACROMET, MM EQUIV. 25.419 13003-04 HF4 24 24 - - - 1.000 - HEAVY HEX NUT ASTM F436 HDG DACROMET, MM EQUIV. 25.420 13003-04 HK4 8 8 131.650 - - 4.000 - CORRUGATED SHEATH AASHTO M270 GR 36 MM EQUIV. 102 x 334421 13003-04 HG4A 8 8 24.500 - - 1.750 - TOP ANCHOR ROD ASTM A722 HDG MM EQUIV. 44.5 x 62022 13003-04 HG4B 8 8 127.800 - - 1.750 - BOTTOM ANCHOR ROD ASTM A722 HDG MM EQUIV. 44.5 x 324323 13003-04 HH4 8 8 - - - 1.750 - HARDENED WASHER ASTM F436 HDG MM EQUIV. 44.5 R9F-1624 13003-04 HJ4 16 16 - - - 1.750 - HEAVY HEX NUT ASTM A563 HDG MM EQUIV. 44.5 R73-1425 13003-04 HL4 8 8 200 200 22 44.500 - WASHER PLATE ASTM A36 HDG26 13003-04 HM4 8 8 8.500 - - 1.750 - STOP-TYPE COUPLER ASTM A29 GR C1045 MM EQUIV. 44.5 x 215.90 R72-1427 13003-04 HN4 8 8 - - - 0.500 0.375 INTERNAL GROUT TUBE - MM EQUIV. 12.7 O.D. X 9.525 I.D.

DB440MR TYPE II BEARING CAPACITY TABLE


ULTIMATE UPLIFT LOAD (kN) 5300FATIGUE UPLIFT LOAD (kN) 2450

TOTAL LONGITUDINAL DISPLACEMENT (mm) +75/-130TOTAL TRANSVERSE DISPLACEMENT (mm) 10.00



LOCATION WEST ABUTMENT, CENTER GIRDERQUANTITY 1

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 6 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

DB440MR TYPE II ASSEMBLYTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01


ALL DIMENSIONS IN MILLIMETERSALL WEIGHTS IN KGS.

LONGITUDINAL

DENOTES THICK END


DENOTES QTY.##



114.3

114.3

530.854+-0.80.0

INSIDE GUIDE BARS

265.427+-0.40.0

INSIDE GUIDE BARS

SEE DETAIL E

PC1

PD1

R6R19

19 X 45CHAMFER

12

1230

1074

537

1000

844

728.854+-0.80.0

INSIDE GUIDE BARS

364.427+-0.40.0

INSIDE GUIDE BARS

232

130 THRU PD1 ONLY W/ 3 x 45 CHAMFER

57.2 THRU8 PLACES

422

141 282

UPSTATION

1000

167.2

734.2

429.6

124.8

62.5

214.9

367.1

DET GUIDE BAR DETAIL

E

#10-32 TPIX 15.875 DEEP

6 PLACES97

65

196

73.5

19 X 45CHAMFER

R6

R19

31.75

31.75

486.854+-0.80.0

INSIDE GUIDE BARS

243.427+-0.40.0

INSIDE GUIDE BARS

GB2ASEE GUIDE BAR

DETAIL F

PC2PD2 10

8(8)

8(8)SEE WELDINGNOTE #1

576.854+-0.80.0

INSIDE GUIDE BARS

288.427+-0.40.0

INSIDE GUIDE BARS

800

1000

390


41.275 x 150SLOT THRU4 PLACES

438.5

877

41.275

284.14

568.28

150

LONGITUDINAL

UPSTATION

8

8

SEAL ENDS4 PLACES

73.97

642

286

143


F

#10-32 TPIX 15.875 DEEP

3 PLACES

70

25

19

35.5

R6

2

TOTAL LOWER GUIDE BARGB2A

WEIGHT:2 13.68 KG.

DB2860MR MASONRYPLATE ASSEMBLYASSY

WEIGHT:1 254.05 KG.

1

TOTAL MASONRY PLATEPC2

WEIGHT:1 198.34 KG.

1

TOTAL LOWER BEARING PLATEPD2

WEIGHT:1 28.36 KG.

DB440MR TYPE I LOWER BEARING BLOCK ASSEMBLYASSY

WEIGHT:2 1471.57 KG.

2

TOTAL

2

TOTAL LOWER BEARING PLATEPD1

WEIGHT:1 26.00 KG.

UPPER BEARING BLOCKPC1

WEIGHT:1 1445.57 KG.

: 10/21/15

DATE: 1/28/15


8

88

8

8

8

8

8

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 7 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

BEARING DETAILS-1THUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



31.75

31.75

466.854+-0.80.0

INSIDE GUIDE BARS

233.427+-0.40.0

INSIDE GUIDE BARS

GB3ASEE GUIDE BAR

DETAIL G

PC3 PD310

8(8)

8(8)SEE WELDINGNOTE #1

1000

800

596.854+-0.80.0

INSIDE GUIDE BARS

298.427+-0.40.0

INSIDE GUIDE BARS

312


41.275 x 150SLOT THRU4 PLACES

877

438.5

41.275

284.14

568.28

150

LONGITUDINAL

LONGITUDINAL

UPSTATION

UPSTATION

8

8

SEAL ENDS4 PLACES

600

220

11069.37


G

#10-32 TPIX 15.875 DEEP

3 PLACES

19 90

2555.5

R6

LENGTH

THREAD

LENGTH

1.5875 X 45CHAMFER2 PLACES

12 TPI

465.854+ 0.80.0

INSIDE GUIDE BARS

232.927+ 0.40.0

INSIDE GUIDE BARS

114.3

114.3

SEE GUIDE BARDETAIL ESHEET 7

PC4 PD4

R6R19

19 X 45CHAMFER

12

331.927+ 0.40.0

INSIDE GUIDE BARS

663.854+ 0.80.0

INSIDE GUIDE BARS

1000

844

282

422

141

1230

1074

537

232

130 THRUPD4 ONLYW/ 3 x 45CHAMFER

57.2 THRU8 PLACES

4

TOTAL LOWER GUIDE BARGB3A

WEIGHT:2 14.03 KG.

DB1830MR MASONRYPLATE ASSEMBLYASSY

WEIGHT:2 244.59 KG.

2

TOTAL

2

TOTAL

LOWER BEARING PLATEPD3

WEIGHT:1 18.19 KG.

MASONRY PLATEPC3

WEIGHT:1 198.34 KG.

LOWER BEARING PLATE

WEIGHT:

TOTAL

WEIGHT:

LOWER BEARING BLOCK

1

1471.57 KG.

1

DB440MR TYPE II LOWER BEARING BLOCK ASSEMBLY

TOTAL PD4

1 26.00 KG.1445.57 KG.

PC4

1

1

ASSY

WEIGHT:

: 10/21/15

3

DATE: 1/28/15


8

8

8

Part Mark Quantity Ø Length Thread

Length Weight (KG)

BA1 2 127 75.620 41.280 7.51

BA2 1 83 78.620 29.620 3.33

BA3 2 64 70.620 29.620 1.78

BA4 1 127 75.620 41.280 7.51

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 8 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



44.45

22.225 2.38 RECESS

59.60 KG.WEIGHT:1

PB1 UPPER BEARING PLATEPB4

2

TOTAL

33

820

735

430.2

125.4

367.5

410

215.1

62.7

232.000+-0.00.8

185.175 RECESS

788.175

RECESS

R12.7

127 x 12 TPI THRU

LONGITUDINAL

#10-32 TPIX 15.875 DEEP12 PLACES

31.75

15.875 2.38 RECESS

WEIGHT: 35.09 KG.1

PB2 UPPER BEARING PLATE

390.000+-0.00.8

390

286

143

339 RECESS

83 x12 TPI THRU

LONGITUDINAL

#10-32 TPIX 15.875 DEEP

6 PLACES

31.75

15.875 2.38 RECESS

WEIGHT: 22.51 KG.2

PB3 UPPER BEARING PLATE

312.000+-0.00.8

312

220

110

273 RECESS

64 x12 TPI THRU

LONGITUDINAL

#10-32 TPIX 15.875 DEEP

6 PLACES

: 10/21/15

DATE: 1/28/15


8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 9 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



WEIGHT: 747.80 KG.3

PE1 SOLE PLATE

DENOTES THICK END


DENOTES QTY.##



800

719

359.5

1000

911.1

745.4

579.8

414.14

248.48

82.8

455.55

500

372.7

289.9

207.07

124.24

41.4

28.6 THRU24 PLACES

LONGITUDINAL

UPSTATION

SEE GENERAL NOTES

#1 & 2 FOR MARKINGS

286.85+-0.80.0

INSIDE GUIDE BARS(TYPE I)

143.42+-0.40.0

INSIDE GUIDE BARS(TYPE I)

990

STAINLESS STEEL

128.43+ 0.40.0

INSIDE GUIDE BARS(TYPE II)

256.85+ 0.80.0

INSIDE GUIDE BARS(TYPE II)

STX

WEIGHT:

SL1 SL4

2

LOWER STAINLESS STEEL

4 1.92 KG.1085 KG.1 WEIGHT:

PE4 SOLE PLATE

4.82 KG.1 WEIGHT:

ST4 TOP STAINLESS STEEL

PE1

2

ST1

2

1094.72 KG.3 WEIGHT:

ASSYDB440 TYPE I & IISOLE PLATE ASSEMBLY

0.53 KG.2 WEIGHT:

SS4SS1

4

SIDE STAINLESS STEEL

3

TOTAL

TOTAL

3

6

TOTAL

6

TOTAL

7

SEE WELDINGNOTE #2

155

31

81

127.13BOTTOM OF GUIDE BAR TO

CL STAINLESS SEEL

SEE GUIDE BARDETAIL H

SLX

SSX

PEX

R19

R6

3 SIDES2 PLACESTACK 4TH SIDESEE WELDINGNOTE #2

1000

155


H

196

97

65 R6

R19

A

B C

D

E E

D

E

C

A

B

E

: 10/21/15

DATE: 1/28/15


8

8

8

8

8

8

8

TOP VIEWBOTTOM VIEW

SOLE PLATE THICKNESS TABLE

WEST ABUTMENT,

NORTH, CENTER &

SOUTH GIRDERS

POINT THICKNESSA 127B 127C 127D 127E 127

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 10 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



414.85+-0.80.0

INSIDE GUIDE BARS

207.427+-0.40.0

INSIDE GUIDE BARS

616

STAINLESS STEEL

ST2

SEE WELDINGNOTE #2

SEAL ENDS4 PLACES

800

722

361

800

722

361

25.4 THRU6 PLACESLO

NGITUDINAL

UPSTATION

SEE GENERAL NOTES

#1 & 2 FOR MARKINGS

75

54.075BOTTOM OF

GUIDE BAR TOCL STAINLESS STEEL

PE2

GB2BSEE GUIDE BARDETAIL J

SS2

SL2


8G

8SEE WELDINGNOTE #1

642

75


J

19

70

25

R6

A

E

B C

E

D

E

C

D

E

A

B


2

TOTAL

2

TOTAL SS2

WEIGHT:2 0.33 KG.

UPPER GUIDE BARGB2B

WEIGHT:2 13.82 KG.

SOLE PLATE ASSEMBLYASSY

WEIGHT:1 192.67 KG.

1

TOTAL

1

TOTAL

TOP STAINLESS STEELST2

WEIGHT:1 4.63 KG.

SOLE PLATEPE2

WEIGHT:1 158.98 KG.

0.38 KG.

TOTAL

WEIGHT:

SL2

2

LOWER STAINLESS STEEL

2

SOLE PLATEPE2

1 WEIGHT: 158.98 KG.

: 10/21/15

DATE: 1/28/15


8

TOP VIEWBOTTOM VIEW


EAST ABUTMENT, CENTER

GIRDER

POINT THICKNESSA 31.750B 31.750C 31.750D 31.750E 31.750

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 11 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



WEIGHT: 158.98 KG.2

PE3 SOLE PLATE


K

DENOTES THICK END


DENOTES QTY.##



800

722

361

800

722

361

25.400 THRU6 PLACES

LONGITUDINAL

UPSTATION

366.854+-0.80.0

INSIDE GUIDE BARS

183.427+-0.40.0

INSIDE GUIDE BARS

550 STAINLESS STEEL

ST3


4

TOTAL

4

TOTAL SS3

WEIGHT:2 0.29 KG.

UPPER GUIDE BARGB3B

WEIGHT:2 14.71 KG.

SOLE PLATE ASSEMBLYASSY

WEIGHT:2

2

TOTAL

2

TOTAL

TOP STAINLESS STEELST3

WEIGHT:1 3.71 KG.

SOLE PLATEPE3

WEIGHT:1 158.98 KG.

4

LOWER STAINLESS STEELTOTAL

2

SL3

WEIGHT: 0.69 KG.

SEE WELDINGNOTE #2

SEAL ENDS4 PLACES

75

54.075BOTTOM OF

GUIDE BAR TOCL STAINLESS STEEL

PE3

GB3BSEE GUIDE BARDETAIL K

SL3

SS3


8G

8SEE WELDINGNOTE #1

75

600

90

25

19

R6

A

E

B C

E

D D

C

E

B

A

E

SEE GENERAL NOTES

#1 & 2 FOR MARKINGS

: 10/21/15

DATE: 1/28/15


8

BOTTOM VIEWTOP VIEW


EAST ABUTMENT,

NORTH & SOUTH

GIRDERS

POINT THICKNESSA 31.750B 31.750C 31.750D 31.750E 31.750

8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 12 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



THICK EDGE: B-C

DENOTES THICK END


DENOTES QTY.##



800

719

359.5

1000

911.1

745.4

579.8

414.14

248.48

82.8

455.55

500

372.7

289.9

207.07

124.24

41.4

28.6 THRU24 PLACES

LONGITUDINAL

UPSTATION

333.48 KG.WEIGHT:1

PA4 BEVELED SHOE PLATEPA1

23

TOTAL

18 RECESS(TYP.)

480

160

240

80

840

680

520

360

420

340

260

180

44(TYP.)

40(TYP.)

R13(TYP.)

25.4 THRU40 PLACES

100

200

A

E

B C

E

D

E

D

C B

A

E

SEE GENERAL NOTES

#1 & 2 FOR MARKINGS

: 10/21/15

DATE: 1/28/15


8

8

8

8

8

8

BOTTOM VIEWTOP VIEW

BEVELED SHOE PLATE THICKNESS TABLE

WEST ABUTMENT, NORTH,

CENTER & SOUTH GIRDERS


8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 13 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



THICK EDGE: B-C

DENOTES THICK END


DENOTES QTY.##



800

722

361

800

722

361

25.400 THRU6 PLACES

LONGITUDINAL

UPSTATION

3

TOTAL

1

PA2

285.41 KG.WEIGHT:2

PA3 BEVELED SHOE PLATE

SEE GENERAL NOTES

#1 & 2 FOR MARKINGS

A

B C

D

E E

: 10/21/15

DATE: 1/28/15


BEVELED SHOE PLATE THICKNESS TABLE

EAST ABUTMENT,

NORTH, CENTER &

SOUTH GIRDERS


8 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 14 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON


WP NO. 124-90-01



DENOTES THICK END


DENOTES QTY.##



DENOTES THICK END


DENOTES QTY.##



465.854+ 0.80.0

INSIDE GUIDE BARS

232.927+ 0.40.0

INSIDE GUIDEBARS

116

114.3

187

197.8

99 663 98

1232

66 66

185

196.6

167167

1089.79 KG.WEIGHT:

PC4 MASONRY PLATE

26.00 KG.WEIGHT:

PD4 LOWER BEARING PLATE

1471.57 KG.WEIGHT:

ASSYDB440MR TYPE IIMASONRY PLATE ASSEMBLY

177.89 KG.WEIGHT:

GB4A LOWER GUIDE BAR

155

31

80

65

100 97

196 195

94 94

80643.5

65

802

255.5

154

747.80 KG.WEIGHT:

PE4 SOLE PLATE

4.82 KG.WEIGHT:

ST4 TOP STAINLESS STEEL

1094.72 KG.WEIGHT:

ASSYDB440 TYPE IISOLE PLATE ASSEMBLY

168.60 KG.WEIGHT:

GB4B UPPER GUIDE BAR

0.53 KG.WEIGHT:

SS4 SIDE STAINLESS STEEL

WEIGHT: 1.92 KG.

LOWER STAINLESS STEELSL4

4

5

9.5

10.5

8.612.1

NESE

E 2967.81 KG.


WEIGHT:

ELEV

4

4

8.5

9

3.16.9

SWNW

W 2967.81 KG.


WEIGHT:

ELEV

: 10/21/158 DATEREV.NO.





TOLERANCING:

LEGEND

CONTRACTOR:

CONSULTANT:


11035 WALDEN AVE,

BEARING TYPE:

CHK'D BY:

DTL'D BY:

SHEET: 15 TO 15

ORDER NO:

TITLE:

DATE:


MMM

BOT CONSTRUCTION

MR 8/4/2014

TWB 8/18/201413003

DISKTRON

AS BUILT MEASURMENTSTHUNDER BAY, CANADACONTRACT NO. 2013-6000

WP NO. 124-90-01

