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Tampere University Dissertations 494
Derivation of CPTu Cone Factors for
Undrained Shear Strength and OCR in Finnish Clays
JUHA SELÄNPÄÄ
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ACKNOWLEDGEMENTS
I am beginning to feel relief now that the study has progressed this point where acknowledgements can be written. Digging for the answers to clay puzzles is difficult as clay does not give an unambiguous solution. Due to this many frustrating moments have been experienced in the along the years I have worked on this study. I am deeply grateful to the people around me for helping me through the difficulties. Sometimes just sharing the problem out loud with someone helps put things in perspective.
First, I would like to give immense thanks to the Finnish Transport Infrastructure Agency for funding this dissertation and especially the people there Erkki Mäkelä, Jaakko Heikkilä and Panu Tolla for giving their opinions in our meetings. Unfortunately, studies, including this one, do not offer overall answers and leave many question marks. Thus, finances will be needed in future too and hopefully, more sources for funding will be found.
Funding does not come without a visionary behind it. That visionary is my supervisor Prof. Tim Länsivaara who receives all my gratitude and thanks for all the assistance, guidance and most of all patience he has given me, during these years. Prof. Tim Länsivaara has done a great deal of significant work that has added to the understanding of soft soil behaviour.
Valuable comments have been received from my pre-examiners Prof. Steinar Nordal and Prof. Jelke Dijkstra, and my opponent Prof. Leena Korkiala-Tanttu for which I give my sincere gratitude. Their expertise has raised the dissertation to next level.
I was honoured to meet Prof. Paul Mayne and receive his advices during my visit to Georgia Institute of Technology. He has done amazing work in the field of understanding CPTu testing and other testing methods.
Thanks also go to Rolf Sandven for the help he gave us in taking the first steps with CPTu. Unfortunately, the steps ended too soon.
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During this study, I have been fortunate enough to get married to my beautiful wife Aino Lampela who is my greatest supporter. Thank you for tolerating me during the stressful times. Also, I would like to give my thanks to Aino’s parent for taking me as a family member. At the end of 2019, our beloved daughter Venla was born bringing a new perspective for our life.
For allowing me to become what I am now, the greatest thanks should be given to my own parents Aila and Timo Selänpää. They have let me make my own choices in life but were still there watching my back. I have always felt supported. Thanks also to my siblings Jutta and Janne, you both have influenced me throughout the years.
Thanks to my friends, often we have found relaxing moments drinking a few pints.
To all the people who have somehow been involved in this study I give my grateful thanks. Here I would like to mention a few: Bruno Di Buó for suffering with me, Joonas Mäenpää and Markus Haikola for helping with in-situ and laboratory testing, Niko Levo and Nuutti Vuorimies for guiding in laboratory, participants of our GEO-Group during years Juho Mansikkamäki, Ville Lehtonen, Marco D´Ignazio, Ali Vatanshenas, Mohammadsadegh Farhadi, and a special thanks to the landowners of the testing sites.
Tampere, August 2020
Juha Selänpää
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ABSTRACT
Information about soil strength is vital in many geotechnical problems, such as stability, bearing capacity and the design of deep excavations. Its false assessment could cause both overdesign and expensive solutions or an increased probability of failure. In Finland, undrained shear strength is commonly measured using the field vane test. Recent studies have shown that the undrained shear strength of clays can be significantly underestimated when old field vane devices with poor execution has been used.
The CPTu test is very widely used soil investigation method in many countries. In Finland, few years back it was very rarely used, especially for assessing soil properties, even though CPTu provides almost continuous measurement data with high accuracy. During CPTu test, cone resistance, sleeve friction and pore pressure are measured. The strength of the soil is not measured directly by CPTu, but the cone resistance is highly dependent on it. Thus, many correlations and theoretical approaches have been proposed to assess the soil strength from cone resistance. Empirical correlations work generally for the local conditions they have been derived to. Thus, proposed empirical correlation should be calibrated for Finnish soil conditions or new empirical correlation should be derived using own high-quality data. Due to the complexity of the cone penetration, theoretical approaches for assessing soil strength may for the moment not solely offer accurate solutions.
The aim of this study is to create practical and sufficiently accurate solutions for interpreting undrained shear strength from CPTu results. In addition, the study aims at evaluating, for the first time, the anisotropic nature of the undrained shear strength of Finnish clays.
This study is focused on soft sensitive soils where disturbance of the soil could cause a great loss of initial strength in the soil. Hence, the sampling process should be of very high quality, in order to receive reliable results for comparison from undisturbed soil samples in the laboratory. A new tube sampler was developed and tested during this study. The developed tube sampler was capable of obtaining samples that could be classified to be of the highest class.
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Piezocone (CPTu) and field vane tests were conducted on five test sites. The laboratory tests contained CRS, DSS, triaxial extension and compression tests and determination of index properties. The five sites can be categorized as non-organic clay sites having a plasticity index between 15 and 60 and a sensitivity between 15 and 100.
The CPTu and field vane tests were performed with high accuracy equipment. The sensitive CPTU cone had a maximum capacity of 7.5 MPa, and the cone resistances varied typically between 150 – 300 kPa. The field vane tests were conducted with apparatus that rotated and measured the torque right above the vane. A large vane size (75mm x 150mm) was used to improve the accuracy by measuring the torque at a higher loading range.
In the laboratory, the in-situ undrained shear strength was determined by several means. To avoid loss of initial strength, samples were consolidated close to in-situ stress, securing that the initial yield stresses were not exceeded. From the results, it was possible to assess the anisotropy of Finnish clays by comparing the compression strengths to DSS strengths and extension strengths. SHANSEP (Stress History and Normalized Soil Engineering Properties) parameters were defined based on the results. The influence of the index properties to the anisotropy ratio was determined.
As CPTu does not directly measure the strength of the soil, empirical correlations based on high quality results were defined. Additionally, an empirical correlation for OCR (overconsolidation ratio) was also determined. Based on the results, the correlations seem to be accurate as the RSE (Relative Standard Error)values are mainly under 10 %.
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TIIVISTELMÄ
Tieto maan lujuudesta on oleellista monessa geoteknisessä ongelmassa kuten stabiliteettilaskennassa, kantokestävyydessä ja syvien kaivantojen mitoituksessa ja sen takia sen virheellinen arviointi voi johtaa ylimitoitukseen ja kalliisiin ratkaisuihin tai lisääntyneeseen murron todennäköisyyteen. Suomessa, suljettua leikkauslujuutta on yleensä mitattu siipikairauskokeella. Viimeaikaisissa tutkimuksissa on havaittu, että saven suljettua leikkauslujuutta on voitu selvästi aliarvioida, kun siipikairauskoe on suoritettu vanhalla laitteistolla huonoja toimintatapoja käyttäen.
CPTu kairauskoe on laajasti käytetty maaperän tutkimustapa. Suomessa sen käyttö muutama vuosi sitten oli harvinaista varsinkin maan ominaisuuksien arvioinnissa, vaikka CPTu kairauskoe tarjoaa lähes jatkuvia mittaustuloksia suurella tarkkuudella. CPTu kokeen aikana mitataan kärkivastusta, vaippakitkaa ja huokospainetta. Maan lujuutta ei mitata suoraan CPTu kokeella, mutta kärkivastus on hyvin riippuvainen siitä. Täten, monia korrelaatioita ja teoreettisia ratkaisuja on esitetty maan lujuuden arvioimiseen kärkivastuksesta. Empiiriset korrelaatiot toimivat paikallisissa olosuhteissa, mistä data korrelaatiolle on kerätty. Täten, muualle esitetty empiirinen korrelaatio pitäisi kalibroida suomalaisille maille tai empiirinen korrelaatio perustuen omiin korkealaatuisiin tuloksiin tulisi luoda. Johtuen kärjen tunkeutumisen kompleksisuudesta, teoreettiset ratkaisut maan lujuuden arvioimiseen eivät välttämättä vielä tarjoa tarkkaa ratkaisua.
Työn tarkoituksena on luoda käytännöllinen ja riittävän tarkka ratkaisu suljetun leikkauslujuuden tulkintaan CPTu tuloksista. Lisäksi tutkimus pyrkii arvioimaan ensimmäistä kertaa Suomessa suljetun leikkauslujuuden anisotropiaa.
Tämä tutkimus keskittyy pehmeisiin sensitiivisiin maihin, joissa maan häiriintyminen voi aiheuttaa suuren initiaalilujuuden menetyksen. Täten näytteenottimen pitää olla sovelias, jotta näytteenotin ei aiheuta maan häiriintymistä. Uusi putkinäytteenotin kehitettiin ja testattiin tämän tutkimuksen aikana. Tulosten perusteella näytteet voidaan laadun osalta luokitella useimmiten parhaimpaan luokkaan.
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Kenttäkokeina käytettiin puristinkairausta huokospainemittauksella (CPTu) sekä siipikairausta. Näytteille tehtiin laboratoriossa CRS-, DSS-, kolmiaksiaalisia veto- ja puristuskokeita sekä lukuisia indeksikokeita. Tutkimuskohteiden määrä on viisi. Nämä viisi kohdetta ovat epäorgaanisia savimaita, joissa plastisuus vaihtelee mittausten perusteella 15 ja 60 välillä sekä sensitiivisyyden osalta välillä 15–100.
CPTu- ja siipikairauskokeet suoritettiin laitteistoilla, joiden mittaustarkkuus on hyvin suuri. CPTu-laitteiston osalta suuri mittaustarkkuus on saavutettu rajoittamalla kärjen kuormituskapasiteetti 7,5 MPa. Tyypillisesti kärkivastukset vaihtelivat mittauksissa välillä 150–300 kPa. Siipikairaukset suoritettiin laitteistolla, jossa siiven pyöritys ja momentin mittaus tapahtuu läheltä siipeä. Isoa siipikokoa (75 x 150 mm) käytettiin, jotta momentin mittaus tapahtuu tarkalta mittausalueelta.
Laboratoriossa näytteistä mitattiin initiaalilujuus suljetussa tilassa. Initiaalilujuuden menetyksen välttämiseksi näytteet konsolidoitiin myötöjännitystä pienempään jännitystilaan. Tuloksista voitiin arvioida lujuuden anisotropiaa suomalaisissa savissa vertailemalla puristuslujuutta DSS:llä määritettyyn lujuuteen sekä vetokoelujuuteen. Tuloksista määritettiin SHANSEP (Stress History And Normalized Soil Engineering Properties) -parameterit. Indeksikokeiden tulosten vaikutusta arvioitiin normalisoituihin tuloksiin.
Koska CPTu kokeella ei mitata suoraan maan lujuutta, empiiriset korrelaatiot määritettiin korkealaatuisten tulosten välille. Lisäksi esikonsolidaatiolle määritettiin myös korrelaatio. Tulosten perusteella korrelaatiolla määritetyt tulokset ovat hyvin tarkat sillä RSE (Relative Standard Error) -arvot voivat olla alle 10 %.
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CONTENTS
acknowledgements ....................................................................................................................... iii
abstract ...........................................................................................................................................v
Tiivistelmä ...................................................................................................................................vii
contents ........................................................................................................................................ ix
notation ....................................................................................................................................... xiii
1 Introduction.....................................................................................................................23
1.1 Background on the research issue .....................................................................23
1.2 Aim and objectives of the study ........................................................................24
1.3 Defining the problem framework ......................................................................25
2 Undrained shear strength ...............................................................................................27
2.1 Definition of shear strength ...............................................................................27
2.2 Factors influencing the undrained shear strength ............................................28
2.2.1 Stress history ......................................................................................28
2.2.2 Anisotropy..........................................................................................29
2.2.3 Viscous effects ...................................................................................31
2.2.4 Temperature effect ............................................................................32
2.2.5 Structure effects .................................................................................33
2.2.6 Sample disturbance ............................................................................34
2.2.7 Softening ............................................................................................35
2.3 Estimation of undrained shear strength by SHANSEP method ....................37
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2.4 Estimation of undrained shear strength using the concept of Critical State Mechanics ............................................................................................................39
3 Principles of used laboratory and field tests to assess the behavior of soil ...............44
3.1 Triaxial compression test ....................................................................................44
3.2 Triaxial extension test .........................................................................................46
3.3 Direct simple shear .............................................................................................46
3.4 Oedometer test ....................................................................................................48
3.5 Field vane test ......................................................................................................49
3.6 Piezocone test ......................................................................................................51
3.6.1 Tip resistance .....................................................................................52
3.6.2 Pore pressure .....................................................................................53
3.6.3 Sleeve friction ....................................................................................54
3.6.4 The corrections for the measurements ............................................54
4 Analytical theories behind the interpretation of undrained shear strength from cone penetration..............................................................................................................56
4.1 Bearing capacity theory .......................................................................................56
4.2 NTH method .......................................................................................................58
4.3 Cavity expansion theory .....................................................................................60
4.4 Strain path method ..............................................................................................63
4.5 Estimation of undrained shear strength from excess pore pressure...............65
4.6 Spherical cavity expansion theory combined with critical state soil mechanics (SCE-CSSM) to assess OCR ............................................................66
4.7 Considerations related to evaluations ................................................................68
5 Test procedures used in the study and quality classifications .....................................71
5.1 Undrained triaxial tests .......................................................................................71
5.2 Direct simple shear test ......................................................................................73
5.3 CRS oedometer ...................................................................................................74
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5.4 Field vane .............................................................................................................75
5.5 Piezocone test ......................................................................................................77
5.6 Sample and test quality classification .................................................................82
6 Sampling and the test results ..........................................................................................84
6.1 Improvement of sample quality by large tube sampler ....................................84
6.2 The cutting of the tube sample ..........................................................................90
6.3 Influences on relative void ratio changes at reconsolidation...........................91
7 Testing sites .....................................................................................................................96
7.1 Perniö, Salo ....................................................................................................... 100
7.2 Masku ................................................................................................................ 104
7.3 Sipoo.................................................................................................................. 107
7.4 Paimio................................................................................................................ 110
7.5 Lempäälä ........................................................................................................... 114
8 Measured results ........................................................................................................... 117
8.1 Some correlations between the index test results for Finnish clay .............. 117
8.2 Field vane comparison between uphole and downhole devices .................. 124
8.3 Comparison between anisotropic and isotropic consolidation in triaxial tests .................................................................................................................... 127
8.4 Normalized undrained shear strength and its relationships to the index properties .......................................................................................................... 130
8.5 Anisotropy on Finnish soft soils and relationship to index properties ....... 138
8.6 Effective strength failure criteria..................................................................... 142
8.7 Stress-strain behaviour ..................................................................................... 149
8.7.1 Field vane stress-rotation curves ................................................... 150
8.7.2 DSS Stress-strain curves ................................................................ 151
8.7.3 Triaxial compression stress-strain curves ..................................... 153
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8.7.4 Triaxial extension stress-strain curves .......................................... 155
8.8 Stress paths of triaxial tests.............................................................................. 157
8.9 Oedometer results ............................................................................................ 159
9 Interpretation of the CPTu measurements ................................................................ 161
9.1 Estimation of the over consolidation ratio by means of the index properties .......................................................................................................... 163
9.1.1 Improvement of correlation using other variables ...................... 168
9.1.2 Consideration of the influence of the rigidity index on the cone factor of the net cone resistance .......................................... 168
9.1.3 Consideration of influence of stress exponent ............................ 174
9.2 Estimation of undrained shear strength ......................................................... 176
9.2.1 Undrained compression strength .................................................. 177
9.2.2 Undrained extension strength ....................................................... 181
9.2.3 DSS strength ................................................................................... 185
9.2.4 Vane strength .................................................................................. 188
9.3 Selections of the interpretation equation for the OCR and undrained shear strength.................................................................................................... 193
10 Discussion..................................................................................................................... 204
11 Conclusion .................................................................................................................... 210
12 Further research ........................................................................................................... 214
References ................................................................................................................................ 215
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NOTATION
Latin letter
a Henkel’s parameter
a Correction factor to correct cone resistance for pore pressure effects
a´ Effective attraction [kPa]
an Correction factor to correct cone resistance for pore pressure effects
As Area of friction sleeve
Asb Cross sectional area of the bottom of the friction sleeve
Ast Cross sectional area of the top of the friction sleeve and
b Correction factor to sleeve friction for pore pressure effects
bn Correction factor to sleeve friction for pore pressure effects
B Foundation width [m]
Bq Pore pressure ratio (Δu2 / qnet) [kPa]
c´ Effective cohesion [kPa]
Cc Compression index
d Diameter
D Diameter
e Void ratio
ef Void ratio in critical state
E Elastic modulus [kPa]
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fs Sleeve friction [kPa]
fT Corrected sleeve friction [kPa]
F Function of yield surface
Fr Normalized friction ratio (100 % · (fT / qnet)
gmax Normalized shear modulus
G Shear modulus [kPa]
Gmax Maximum shear modulus [kPa]
Gs Secant shear modulus [kPa]
H Height
Ic Material index
Ip Plasticity index [%]
Ir Rigidity index (G / su)
k Multiplier for Qt to estimate OCR
k Multiplier according to cylindrical or spherical expansions
K0 Lateral stress coefficient
l Length
m SHANSEP exponent
m´ Stress exponent
M Critical stress ratio of q and p´
Mc1 Critical stress ratio defined from maximum deviatoric stress
Mc2 Critical stress ratio defined from maximum ratio of q and p´
n Stress exponent
N Normalized cone resistance (qnet / σ´v0)
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Nc Bearing capacity factor for cohesion
Nkt Cone factor for net cone resistance-based interpretation
Nke Cone factor for effective cone resistance-based interpretation
NΔu Cone factor for excess pore pressure-based interpretation
Nu Bearing capacity factor in NTH method
Nq Bearing capacity factor for overburden
Nγ Bearing capacity factor for self-weight of soil
Qt Normalized net cone resistance (qnet / σ´v0)
p Mean total stress [kPa]
p0 (Mean) initial total stress [kPa]
pc (Mean) preconsolidation pressure [kPa]
Pw Shear induced pore pressure
p´ Mean effective stress
p´0 Initial mean effective stress [kPa]
p´c Mean effective stress at preconsolidation pressure [kPa]
p´f Critical state of mean effective stress [kPa]
Pu Ultimate pressure [kPa]
q Deviatoric stress [kPa]
qc Cone resistance [kPa]
qeff Effective cone resistance (qT – u2) [kPa]
qf Critical state of deviatoric stress [kPa]
qnet Net cone resistance (qT – σv0) [kPa]
qT Corrected cone resistance [kPa]
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qult Ultimate bearing capacity [kPa]
r Radius
R Radius
R2 R-squared value
Rf Friction ratio (fs / qnet · 100 %) [%]
Ri Initial radius
Rp Radius of the elastic-plastic boundary
Ru Ultimate radius of cavity
S SHANSEP strength ratio
St Sensitivity
su Undrained shear strength [kPa]
su,Comp Undrained compression triaxial strength [kPa]
suC Undrained compression triaxial strength [kPa]
su,DSS Direct simple shear strength [kPa]
suD Direct simple shear strength [kPa]
su,Ext Undrained extension triaxial strength [kPa]
suE Undrained extension triaxial strength [kPa]
su,FVcorr Corrected vane strength [kPa]
su,FVmeas Measured vane strength based on conversion of maximum torque to shear stress [kPa]
t Thickness
u Pore pressure [kPa]
u0 Initial pore water pressure [kPa]
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u1 Pore water pressure measured on the cone [kPa]
u2 Pore water pressure measured behind cone base [kPa]
u3 Pore water pressure measured behind friction sleeve [kPa]
Δu Excess pore pressure [kPa]
Δu2 Excess pore pressure behind cone base (u2 – u0) [kPa]
Δuf Excess pore pressure at failure [kPa]
Δumean Pore water pressure caused by change of mean effective stress [kPa]
up The smallest displacement where yielding occurs during cavity expan-sion
ur Radial displacement
Δushear Pore water pressure changed by shearing [kPa]
U* Normalized excess pore water pressure (Δu2 / σ´v0)
w Natural water content [%]
wL Liquid limit [%]
wP Plastic limit [%]
Greek symbols
α Inclination
αf Cone roughness
αs Shaft roughness
β Angle of plastification
Δ Change
Δ Initial stress condition
γ Unit weight [kN/m3]
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γ´ Effective unit weight [kN/m3]
γ Shear strain
γref Reference shear strain
ε Strain
έ Strain rate
κ Slope of the unloading-reloading line in (ln p´, e) space
λ Slope of the normal consolidation line in (ln p´, e) space
μ Correction factor for su,FVmeas
σ Total stress [kPa]
σ0 Initial total stress [kPa]
σ1 Total major principal stress [kPa]
σ1 Total intermediate principal stress [kPa]
σ3 Total minor principal stress [kPa]
σcell Cell pressure [kPa]
σh Total horizontal stress [kPa]
σh0 Initial total horizontal stress [kPa]
σv Total vertical stress [kPa]
σv0 Initial total vertical stress [kPa]
σr Radial stress [kPa]
σz Vertical stress [kPa]
σθ Hoop stress [kPa]
σ´ Effective stress [kPa]
σ´1 Effective major principal stress [kPa]
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σ´2 Effective intermediate principal stress [kPa]
σ´3 Effective minor principal stress [kPa]
σ´c Preconsolidation stress [kPa]
σ´n Effective normal stress [kPa]
σ´v Effective vertical stress [kPa]
σ´v0 Initial effective vertical stress [kPa]
σ´z Effective vertical stress [kPa]
σ´y Effective yield stress [kPa]
τ Shear stress [kPa]
τf Shear stress at failure [kPa]
μ1 Correction factor for measured field vane strength defined by the function of the liquid limit
μ2 Correction factor for measured field vane strength defined by the function of the ratio of measured field vane strength and initial effec-tive stress
φ´ Friction angle [°]
φ'qmax Friction angle defined from peak value(s) of deviatoric stress [°]
φ'MO Friction angle defined from maximum ratio of σ´1 / σ´3 (maximum obliquity) [°]
� void ratio under mean effective stress of 1 kPa
v Poisson’s ratio
Λ plastic volumetric strain ratio
Common subscripts
0 Initial state
xx
1 Major principal stress or strain
2 Intermediate principal stress or strain
3 Minor principal stress or strain
c Preconsolidation state
a Axial
Ave Average
Comp Compression
Ext Extension
h Horizontal direction
f Failure state
max Maximum
u Ultimate
ref Reference
peak The highest value
v Vertical direction
z Depth
Acronyms
CAUC Anisotropically consolidated, undrained compression triaxial test
CAUE Anisotropically consolidated, undrained extension triaxial test
CICU Isotropically consolidated undrained compression triaxial test
CIUC Consolidated isotropically undrained compression triaxial test
CRS Constant rate of strain
xxi
CSL Critical state line
CSSM Critical state soil mechanics
CPT Cone penetration test
CPTu Cone penetration test with pore water pressure measurement
DSS Direct simple shear test
ESP Effective stress path
FEM Finite Element Method
FC Fall cone
FV Field vane
IL Incremental loading
ISO the International Organization for Standardization
LI Liquidity index ((w – wP) / (wL – wP))
MCC Modified Cam-Clay
NGI Norwegian geotechnical institute
NTH Norwegian Institute of Technology (now NTNU)
NTNU Norwegian University of Science and Technology.
OCR Overconsolidation ratio
RSE Relative standard error
SBT Soil behaviour type
SCE Spherical cavity expansion
SD Standard deviation
SHANSEP Stress History And Normalized Soil Engineering Properties
SGI Swedish Geotechnical Institute
xxii
TSP Total stress path
TAU Tampere University
TUT Tampere University of Technology (now TAU)
TX Triaxial test
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1 INTRODUCTION
1.1 Background on the research issue
The cone penetration test is one of the most widely utilized methods in soil investigations in land and marine environments. The cone penetration test with a pore pressure measurement called the piezocone test was introduced in the 70s and it is a crucial part of cone penetration testing, especially in fine grained soils. Thus far, the utilization of the cone penetration test in Finland has been only marginal.
The piezocone test allows soil layers as well many geotechnical parameters to be assessed. One of the assessed parameters is soil strength. The soil strength is related to many geotechnical problems like stability and bearing capacity and due to that, its false assessment could cause both overdesign and expensive solutions or an increased probability of failure. The project aiming to classify Finnish railroads by introducing the European classification of railroad networks, has a huge potential for saving costs; the classification changes the dimensioning loads (Andersson-Berlin 2012). One of the classifications is related to the load capacity that railway lines are allowed to carry. This means that a low safety factor in the stability calculation may result in the decreasing of train loads on the railways, or the stability needs to be improved by soil improvement or structural improvements both of which will incur increased costs.
Previously the short-term stability calculations have been based on the corrected field vane results. Some doubts were expressed about the results when the vane test was equipped with a slip-coupling without casings (Mansikkamäki 2015). Based on the finding, it was felt that there was a need for another in-situ tool for estimating the soil strength. The CPTu was then seen as one promising alternative.
In the 90s, an effort was made to harness CPTu into daily use to assess undrained shear strength. However, the CPTu cones used then had more problems with accuracy especially for low load values. Nowadays, more accurate sensitive cones have been introduced for soft soils. Another problem in the 90s was the attempt to
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compare the CPTu results to vane measurements while the vane test itself was having several problems related to equipment and execution.
The attempts in the 90s were very useful to understand the demands required of the equipment, the executions and the calibration tests while the theoretical considerations and comparisons were not sufficient to convince the geotechnical community in Finland. To verify the functionality of CPTu in Finland calibration tests on multiple testing sites were needed. On such sites, CPTu tests should be carried out with high-accuracy equipment and the results should be compared to reliable comparison tests to find correlations.
This research is a part of the project referred to as FINCONE and a parallel research study to evaluate preconsolidation stress and deformation properties of CPTu conducted by Di Buò (2020).
1.2 Aim and objectives of the study
The scope of study is:
� To create a practical and sufficiently accurate solution for interpreting undrained shear strength from CPTu results.
The term practical means:
� Simple enough for daily use in design practice.
� The additional parameters to improve interpretation should be quick and easy to define at low-costs.
The goal for “sufficiently accurate” could be described as:
� The confidence level for the interpretations should be such that the values can be used in low risk design cases without any reference strength tests.
� Highly accurate interpretations for high risk design cases can be done with calibrating the CPTu interpretations by high quality reference test(s).
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In addition, the study aims at evaluating, for the first time, the anisotropic nature of the undrained shear strength of Finnish clays. Correlations from CPTu measurements will be given to assess passive, direct and active undrained shear strengths.
Apart from this doctoral dissertation, the project also includes a popularization of CPTu testing in Finland. In this connection, training has been given to soil investigators and geotechnical engineers.
1.3 Defining the problem framework
The aim is to create interpretation equations for assessing the undrained shear strength of Finnish clays by CPTu sounding. The equations for CPTu interpretation can be seen as logical models where input parameters will be processed by equation to the outcome of interest. The outcome is the undrained shear strength and the input parameters are the CPTu measurements with an additional stress state consideration. The interpretation equation between input and outcome should be simple but sufficiently accurate for the design purpose. The factors in such equations should be linked with simple index properties or parameters that could be estimated with local knowledge.
A theoretical solution for cone penetration considering all the factors influencing this complex problem may not yet be introduced for clayey soils. However, theoretical approaches may offer valuable knowledge about the main influences that need to be considered. These influences can be listed as follows: 1.) soil stiffness or rigidity index; 2.) cone and sleeve roughness; 3.) ratio of horizontal to vertical stress; 4.) strength anisotropy; 5.) soil sensitivity; 6.) degree of pore pressure dissipation during penetration; and 7.) viscous rate effects (Schneider et al. 2008). A CPTu cone including pore pressure measurement enables the use of an effective and/or total stress analysis in the interpretation.
To build a bridge between CPTu measurements and calibrationstrength results, a high demand should be placed on both results. To minimize variation in results, procedures and devices should be carefully considered and kept constant during the whole period of the research. It is recognized when studying soft clays, that there is a demand for high accuracy with small values at the beginning of the loading range of cone resistance (Sandven 2010).
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To calibrate the empirical solutions, some high-quality testing sites should be well characterized so that calibration can be justified. A lack of national testing sites in Finland has resulted in the need to first establish sites with high quality testing for use in calibration.
To have highly reliable data for estimating soil mechanical behaviour, high quality samples are needed (Lunne et al. 1997). Based on previous experiences, piston samplers with relatively small inner diameters could sometimes have problems offering good quality samples. Therefore, the required high-quality sampling first needs to be solved.
Finally, the correlations between the reliable results of the CPTu measurements and calibration tests will be analysed. The use of index properties in the correlation equation as predictors will be evaluated by statistical analysis. Adding predictors should increase the accuracy, indicated by statistical parameters such as the relative standard error (RSE). The practical perspective will be kept in mind so that the evaluation of undrained shear strength by CPTu will be suited to engineering practices. This can be seen as beneficial, as no additional predictor, defined by laboratory tests, will be needed. This will mean that at least a preliminary interpretation can be done immediately. In more demanding design cases, uncertainties related to the CPTu interpretation can be decreased by adjusting the interpretation by site-specific reliable calibration results.
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2 UNDRAINED SHEAR STRENGTH
2.1 Definition of shear strength “The shear strength of a soil mass is the internal resistance per unit area that the
soil mass can offer to resist failure” (Das 2002). The shear strength is dependent on drainage condition. While the drainage of pore water is prevented or the loading is sufficiently fast i.e. the rate of loading is significantly larger than the rate of consolidation, then the fully saturated sample/soil responds undrained and the shear strength is called the undrained shear strength (su). When the volume of soil can change and excess pore pressure can be fully diminished by time, then the shear strength is called the drained shear strength.
The shear stresses exist when unequal stresses are loading the soil mass. The critical difference between the highest and the lowest principal stresses correspond to the strength of the soil and any attempt to increase that stress difference will lead to failure. Fifty percent of the highest difference in the undrained condition is the undrained shear strength of the soil. The undrained shear strength can be expressed by total and effective main stresses as in Equation 2.1, where q is deviatoric stress, σ1 is total major stress, σ´1 is effective major stress, σ3 is total minor stress and σ´3 is effective minor stress.
�� = �� = ���
� = ��� ��� (2.1)
The shear strength is formed by friction acting on the shear plane and by interparticle bonding known as cohesion. The magnitude of the friction is associated with the size of the effective normal stress in the shear plane. The term “effective” illustrates only the load which is carried by the soil skeleton as Terzaghi (1925) concluded. The determination of the friction can be made from the change of shear strength compared to the change of effective normal stress. The cohesion is the resistance caused by the forces tending to bond or hold the soil particles together.
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Mohr-Coulumb failure criterion as shown in Equation 2.2 is the most used failure criterion to present the maximum shear stress by the components of effective cohesion (c´), effective friction angle (φ´) and effective normal stress (σ´n).
� = �� + ��� ∙ ����� (2.2)
2.2 Factors influencing the undrained shear strength
Many factors affect the magnitude of undrained shear strength as well the behaviour of the soil during shearing. This can be expected as su is an emerging strength property, as opposed to the critical state friction angle which is unique for a material at a certain effective stress level regardless of initial void ratio or loading history (at least in theory). Some of these factors are listed and categorized below.
2.2.1 Stress history
Stress history includes the effects caused by constant stresses and stress changes with respect to time after the soil layer has formed. The constant stress period and stress changes especially near the normally consolidated state decreases the void ratio. The decrease of the void ratio makes the soil denser, which increases the interaction of soil particles making the structure at least more stable against loads and usually increasing the strength as well. The void ratio change under a constant load is the result of the creep. The creep is a viscous property and it is later categorized as well under viscous effects. In this study, influences which changes the void ratio is considered to be the stress history effects.
Based on the results of delayed consolidation in one-dimensional compression, Bjerrum (1967), suggested that the relationship between the void ratio and the logarithmic effective vertical stress cannot be described by one curve but requires a system of parallel timelines. This concept is known as the concept of isochrones and the principle is shown in Figure 2.1. Later Leroueil and Marques (1996) showed that Bjerrum’s conclusion concerning the time effect of one-dimensional compression are also valid in triaxial stress space.
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Figure 2.1. Bjerrum’s consolidation curves at corresponding parallel timelines (Bjerrum 1967).
Preconsolidation pressure (σ´c) is the limit pressure between stiff and soft response. In Figure 2.1, pc is used to denote preconsolidation pressure. Often the soil is assumed to behave as elastic below preconsolidation pressure at a corresponding strain rate. Current stress state compared to preconsolidation pressure have an influence on how the soil behaves under undrained loading. Dense structure with a relatively low stress state increases the tendency to dilate in shearing. Accordingly, loose structure close to a normally consolidated state has a tendency to contract.
2.2.2 Anisotropy
Magnitudes of soil properties are direction-dependent properties related to permeability, deformation and strength properties making the soils anisotropic. Anisotropy is composed of inherent anisotropy and stress induced anisotropy. Inherent anisotropy is the result of geological formation and stress-induced anisotropy is due to an anisotropic stress state. The effect of anisotropy in a direct simple shear test with lightly over consolidated samples is presented in Figure 2.2. The samples were cut in different direction and the tests were carried out in the same normal and shear stress state as occurred in the field. Bjerrum (1973) found this to
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be caused by the higher development of pore pressure when the stress rotation is increased compared to the initial state.
Figure 2.2. Variation of undrained shear strength results observed in a direct simple shear test with differently directed shear planes in the stress state as those occurring in the field (Bjerrum 1973).
An example of shear modes on a slip surface is presented in Figure 2.3. On a slip surface, mobilized undrained shear strength tends to vary as the inclination of the failure surface changes. Thus, relevant anisotropic shear strengths can be applied by adapting corresponding shear test results in different parts of a slip surface.
Figure 2.3. Shear modes in different parts of a slip surface under an embankment (Larsson 1980 after Bjerrum 1973).
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The anisotropic shear strength of the soil is not only dependent on minor and major stresses. Intermediate stress also influences the strength of the soil. Often strength anisotropy is studied by undrained triaxial tests where the intermediate stress is kept either equal to major or minor principal stress. For the intermediate stress between the minor and major principal stress values, obtained strength values are somewhere between these extreme cases.
Earlier, anisotropy of Finnish clay has been studied experimentally and by modelling by Karstunen and Koskinen (Karstunen et al. 2005, Karstunen and Koskinen 2008).
2.2.3 Viscous effects
Viscous effects include time-dependent behaviours like creep (volume, deviatoric, rupture), stress relaxation and strain rate effects. Undrained shear strength is well known to be strain rate dependent due to viscoplasticity resulting in a higher strength with a higher strain rate of loading. From the results of varying strain rates in undrained triaxial tests, it can be seen that higher strain rates lead to a decrease in shear induced pore pressure with reconstituted samples (Sheahan et al. 1996) and natural samples (Länsivaara 1999). Previously, Länsivaara has conducted some strain rate tests on Finnish clay (Länsivaara 1996).
In an isotach manner, the unique stress-strain-strain rate effect seems to be valid to be generalized for all loading types (Vaid&Campanella 1977; Leroueil et al. 1985). Vaid and Campanella (1977) stated that constant stress creep results can be used to predict stress, strain and strength response under a constant rate of strain loading. In a very low strain rate, some ageing effects might occur and can cause deviation in the stress-strain curve compared to the higher strain rate tests as noted by Leroueil et al. (1985) in their various constant rate strain oedometer results in Figure 2.4.
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Figure 2.4. Leroueil et al. (1985): constant rate of strain oedometer results on Batiscan clay.
2.2.4 Temperature effect
Due to the viscoplastic nature of the soil, the temperature effect could be included in the viscous effect. The temperature rise causes a decrease in the shear resistance of the soil (Mitchell 1976, Moritz 1995), thus increasing the creep rate (Mitchell et al. 1968). Therefore, the increase of the creep rate causes a rise of the pore pressure in the undrained condition. Baldi and Hueckel (1990) demonstrated the thermal softening of clays with increasing temperature and developed a constitutive rationale for it.
Leroueil et al. (1985) proposed that the isotache concept can be extended to also cover the temperature effects. This is supported by the empirical results of Boudali et al. (1994) and Moritz (1995). Figure 2.5 presents the temperature effects on the oedometer test results at three different temperatures where the preconsolidation pressure decreases with an increase in temperature (Moritz 1995).
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Figure 2.5. The results of three balanced temperatures with oedometer tests (Moritz 1995).
2.2.5 Structure effects
The structure of the soil has a huge effect on the behaviour of a wide range of soils including soft and stiff clays, granular soils and residual soils (Leroueil&Vaughan 1990). In this study, the structural effects are the ageing effects on strength and stiffness which cannot be accounted for by void ratio and viscous effects. This definition omits the creep from the structure effects even though it is usually included in the ageing effects. The ageing effects for the structure can be classified according to Sorensen (2006) as being made by inherent (thixotropy, bonding, cementation) and environmental (weathering, chemical changes in pore water from external sources, heat- and pressure-induced changes to the soil structure) effects. These effects can make the structure more stable or unstable on in situ stress state.
The most common way to assess the structure effects might be to measure the intact strength and disturbed strength in the same void ratio and compare these values together resulting in a sensitivity value. Burland (1990) proposed this method to assess the influence of the soil structure by comparing the intact consolidation curve to the intrinsic sedimentation curve with a reconstituted sample as shown in Figure 2.6. The difference with these curves represents the structure effects of an increase in the stiffness and strength. The increase in stiffness and strength leads to a steeper collapse towards the intrinsic sedimentation curve illustrating a more brittle behaviour. Similar curves in a normal consolidation range might be an indication
34
that the soil is behaving like an ideal soil defined as a soil having a less pronounced effects on the soil structure. Based on Skempton’s (1970) consolidation curve results the intrinsic consolidation curve is a liquid limit dependent property but Burland (1990) suggested preferably defining the curve by the consolidation test.
Figure 2.6. Oedometer results on undisturbed and reconstituted samples of Bothkennar clay (Burland 1990).
2.2.6 Sample disturbance
Sample quality has a crucial effect on laboratory tests done for the determination of the strength and deformation properties of soft sensitive soils (Lunne et al. 1997). The sample may be subjected to change in stress conditions, change in water content and void ratio, disturbance of the soil structure, chemical changes, mixing and segregation of the soil constituents, all of which cause disturbance according to Hvorslev’s basic classification (Hvorslev 1949). Some of disturbance effects are time- dependent like stress-relief due to dissipation of negative pore pressure (Skempton&Sowa 1963), drying and chemical changes due to oxidation (Mitchell 1976).
Even change in water content and void ratio, chemical changes, mixing and segregation can be prevented by carefully sampling, although some stress and strain change will occur in sampling and sample handling. These changes may cause a destruction of the soil structure causing a loss of strength. Partial destruction may decrease the apparently elastic stress range while increasing the plastic deformations during reloading, resulting in smaller undrained shear strength values and
35
deformation modules in an over consolidated stress range (Lunne et al. 1997; 2006, Landon et al. 2007, Löfroth 2012, Holtz et al. 1986, Sandven et al. 2004, Tanaka et al. 1996, Amundsen et al. 2016). According to several researches (Leroueil et al. 1979, Hight et al. 1992, Lunne et al. 2006 and Karlsrud&Hernandez-Martinez 2013), sample quality has a considerable influence on triaxial compression tests while extension tests are not so sensitive as regards sample quality. Lunne et al. (1997) presented the results with different samplers corresponding to the different degrees of disturbance as can be seen in Figure 2.7, where the lower peak shear stress values correspond to lower quality.
Figure 2.7. Anisotropic consolidated undrained compression test results with different initial sample sizes and the sampler types (Lunne et al. 1997).
2.2.7 Softening
With structured soils especially, the post peak behaviour is strongly controlled by strain softening causing a loss of strength. The loss of strength is mainly governed by more localized shear strains causing a greater orientation of soil grains in the shear band. Some statements have been presented about cohesion or friction angle drop after peak (i.e. Skempton 1964, Burland 1990). Recent studies have tended to suggest that the strength reduction is just greater shear induced pore pressure, at least for clays having low OCR (Thakur 2007, Gylland et al. 2014). Some difficulties could
36
occur in pore pressure measurement when the pore pressure is measured at the base of the sample. Figure 2.8 shows the effect of shear induced pore pressure in the idealization of undrained strain softening as Thakur et al. (2014) presented. Thus, the softening can be assumed to follow a failure line down to residual strength values.
The stress state in soil is initially anisotropic. The shear stresses are mobilized differently on soil planes and thus the strain needed to reach peak strength depends on the direction of loading (Bjerrum 1973). Comparing anisotropically consolidated undrained compression (CAUC) and extension (CAUE) tests and direct simple shear test (DSS) with natural clay samples, the post peak reduction is observed to be the smallest in CAUE tests and the highest in CAUC tests and DSS tests somewhere in between (Karlsrud&Hernandez-Martinez 2013).
Figure 2.8. Idealization of undrained strain softening in soft sensitive clays seen at the laboratory strain level up to 20% (Thakur et al. 2014).
Karlsrud and Hernandez-Martinez (2013) listed some trends for post-peak strength reduction and the trends show that strength reduction is increases with increasing sensitivity and decreases with increasing OCR. The data was gathered mainly from Norwegian clay deposits except one, the Bothkennar site in the UK. Lunne et al. (2006) mentioned that in large strains the shear resistance is governed mainly by the water content but Karlsrud and Hernandez-Martinez did not find that kind of trend in strength reduction. The mineralogy and pore water composition have their own influence on large strain shear resistance (Mitchell 1976, Helle 2017) which might explain the difficulty of seeing the connection between water content and strength reduction.
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As mentioned in Chapter 2.2.5, the structure effects might show some extra strength compared to the initial void ratio. This extra strength is sensitive to strains and the extra strength can be easily lost. The extra strength can be seen in the stress-strain curve of triaxial compression test as a sharp peak strength and a steeper drop of strength after the peak undrained strength is reached compared to the behaviour of soil that has a less pronounced effect on the soil structure. Thereafter, the increase in the strains will diminish the effect of the structure and the soil will start to follow more closely the void ratio dependent behaviour at certain strain rates and temperatures. Roughly the behaviour of structured soft fine-grained soil has a significant strain softening behaviour after the undrained peak strength in the triaxial compression whereas soil that has a less pronounced effect on the soil structure and/or overconsolidated fine-grained soil behaves more closely to being perfectly plastic.
2.3 Estimation of undrained shear strength by SHANSEP method
Ladd and Foott (1974) proposed the SHANSEP method (Stress History and Normalized Soil Engineering Properties) to investigate properties of soil in the laboratory. In this method, samples are first consolidated into a new preconsolidation pressure and then the samples are sheared in different OCR states. Thus, the stress history of the samples is more precisely known, and then the undrained shear strength can be estimated by Equation 2.3 including the S parameter for certain shear modes. As the preconsolidation pressure is artificially made, the behaviour of samples represents the behaviour of reconstituted soils. However, the SHANSEP procedure is also applicable for natural clays consolidated under yield stress, especially when the dependency of the S parameter on the index parameters is considered. The variation of the S parameter in Equation 2.3 and 2.4 (Ladd&Foott 1974, Mesri 1975, Jamiolkowski et al. 1985) might have more variation because of the effect of the natural structure of soils and the disturbance level of the samples affecting the oedometer tests and shear tests. Jamiolkowski et al. (1985) suggested the use of Equation 2.4 for most low OCR clays.
�� = � ∙ ���� ∙ ���� (2.3)
�� = � ∙ ��� (2.4)
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In Figure 2.9, the normalized behaviour of Boston blue clay in direct simple shear tests is presented. From the results, the nonlinear effect of OCR can be defined as well the S-parameter corresponding to the value when the OCR is 1.
Figure 2.9. Example of the normalized behaviour of Boston blue clay in direct simple shear tests (Ladd&Foott 1974).
Larsson (1980) compared the strength results of inorganic clays from Sweden and Norway by using the data of natural samples collected by Berre (1972), Berre and Bjerrum (1973) Aas (1976) Karlsrud and Myrvoll (1976) and Larsson (1977). He proposed that undrained triaxial compression strength over preconsolidation pressure can be approximated by 0.33. He proposed that the same ratio for direct simple shear tests and extension tests increases as plasticity increases.
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2.4 Estimation of undrained shear strength using the concept of Critical State Mechanics
Schofield and Wroth presented the concept of Critical State Soil Mechanics (CSSM) in 1968. The concept describes the ideal elastic/plastic behaviour of the material in the original Cam-Clay model. The behaviour of the soil has been defined by saturated reconstituted soil samples in triaxial compression tests where the minor and the intermediate principal stresses are equal. The principle of the concept is that there is a unique combination of deviatoric stress (qf), mean effective (p’f) stress and void ratio (ef) at a critical state where the continuous shearing will eventually lead. This unique stress point of p’f and qf lies on the Critical State Line corresponding to the critical stress ratio (M) of q and p’. Later Roscoe and Burland (1968) modified the yield surface to be an ellipse. The model was named a modified Cam-Clay (MCC) and it is one of the most used soil models because of its simplicity. In both models, CSL intersects the yield surface at the maximum value of deviatoric stress where the failure is attained.
Equations 2.5 and 2.6 together define the critical state for soil, where � is the void ratio under mean effective stress of 1 kPa in normally consolidated state and λ corresponds a coefficient to illustrate a relation between the logarithmic change of mean effective stress to a change in the void ratio along a normal consolidation line. At the top of Figure 2.10 the yield surface of modified Cam Clay is indicated. Failure of the soil is assumed to occur at a stress state where the mean effective stress is half of the isotropic preconsolidation pressure (p´c). This stress state corresponds to the critical void ratio as shown in the lower Figure 2.10.
� = ∙ !� (2.5)
" = � − # ∙ $�(!�) (2.6)
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Figure 2.10. The relation between the void ratio, mean effective stress and maximum deviatoric stress in Critical State Soil Mechanics as presented in the modified Cam Clay model (Carter et al. 1979).
Unloading increases the over consolidation of the soil and leads to expansion of the soil by the elastic strains along the expansion line in Figure 2.10. The relation between the initial mean effective stress (p´0) and the assumed critical mean effective stress (p´f = ½·p´c) is an important aspect affecting soil behaviour in undrained conditions. Depending on the p’f in relation to the initial mean effective stress (p´0) in undrained conditions, the p’0 needs to increase or decrease in order to attain the critical state line. The p’0 change is associated with the plastic volumetric change corresponding to the void ratio change. Volumetric strain in undrained conditions does not occur, so this tendency for plastic volumetric strain should be compensated by pore water pressure. In yielding, where there is a significant tendency for plastic volumetric strain to occur, the development of pore water pressure can be positive or negative as regards failure. This depends on the tendency of the soil particles to move over each other in an existing stress state.
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To define the limit of yielding, where plastic strain starts to occur in two- or three-dimensional space, a yield surface will be needed. In the modified Cam Clay material model, a symmetrical yield surface with respect to the isotropic axis is adopted and the size and the shape of the yield surface is defined by preconsolidation pressure and critical stress ratio (M) (Roscoe&Burland 1968). Equation 2.7 expresses the yield surface for MCC. Later Tavenas and Leroueil (1977), based on their results, proposed that the yield surface differs from theoretical shape implied in the Cam-clay models. They suggested that the shape of the yield surface of a natural clay is an elliptical shape, centered on the K0 line of normally consolidated clay.
% = �� + � ∙ !� ∙ (!�� − !�) = 0 (2.7)
With the yield surface and the Critical State Line, the undrained behaviour of soil can be illustrated in a p´-q coordinate system. Inside the yield surface of the soil it can be assumed to behave similarly to an elastic corresponding vertical effective stress path when p´=(σ´1+σ´2+σ´3)/3. When the stress path reaches the yield surface from the left side (“dry side”) of the Critical State Line this leads to dilative behaviour as shown in Figure 2.11 with the stress path B-D1-D2 in a consolidated-isotropically undrained compression (CIUC) test. Continuation of the loading causes a decrease in the pore water pressure until the stress path again strikes the failure line at a higher mean effective stress level compared to the initial state. (Leroueil et al. 1990) Then the ultimate failure is attained.
After the deviatoric stress has reached the yield surface on the right side (“wet side”) of the Critical State Line, the compressive behaviour occurs as a shearing like stress path A-C1-C2 as illustrated in Figure 2.11. The pore water pressure increases during continuous loading causing a decrease in the mean effective stress. The shear stress is able to increase until the failure line is reached. (Leroueil et al. 1990) This shear hardening during yielding is the property of the soil.
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Figure 2.11. Pore pressures generated during undrained triaxial tests. (Leroueil et al. 1990)
In elastic zone inside the yield surface, the excess pore water pressure only consists of the change of mean total stress (Δumean = Δp = p - p0) illustrated by Hooke’s law. When the yield surface has been reached, the second component related to the yielding also starts to generate (Δushear = Δp’ = p’0 - p’f ). Thus, the excess pore pressure can be expressed by Equation 2.8 where the second term is dependent on the history and nature of the loading (Wood 1990).
∆' = ' − '� = ∆'�*,� + ∆'-.*,/ = (! − !�) − (!� − !�� ) (2.8)
As mentioned before, the plastic volumetric strain is compensated by the pore pressure development in the yielding in an undrained condition. The plastic volumetric strain itself is controlled by two factors: a flow rule and a hardening rule. The flow rule points to the relative direction of the plastic flow at the current stress state where yielding occurs, and the hardening rule defines the size of the plastic strain increments.
The pore pressure response is linked to the compressibility of the clay (Hansen and Gibson 1949). In the Cam Clay material model, this same basic idea is utilized. The plastic volumetric hardening parameter can be defined from the oedometer test results where the slope of the unloading-reloading line defines the elastic volumetric change parameter (κ) and the slope of the normal compression line defines the total
43
volume change parameter (λ). Smaller differences between λ and κ is an indication of the greater potential for hardening.
By using the critical state concept, the undrained shear strength can be assessed with the critical stress ratio (M), σ´v0, OCR and plastic volumetric strain ratio (Λ=(κ-λ)/λ) (Schofield&Wroth 1968). Instead of using the vertical effective stress, use of the mean effective stress would be more appropriate. In practise, the mean effective stress is unlikely to be known, and an estimation of the effective vertical stress can be more easily and accurately done. Thus, the effective vertical stress is adopted as the expression in Equation 2.9. (Wroth 1984) Likewise the OCR is defined by the vertical preconsolidation pressure instead of the isotropic yield stress.
�� = 1� ∙ ���� ∙ 2345
� 67 (2.9)
Another way to assess su is based on the excess pore pressure (Δu) failure, if it is known. Assessing Equation 2.11 is derived from Equation 2.10 where the soil is initially in an isotropic state.
∆' = 8��� − � 9 − 8� � − ���� 9 = :; ∙ � + ���� − ���� ∙ 2345
� 67 (2.10)
�� = ;∙<∆�>?�@A� ∙B2CDEF 6G:HI
� (2.11)
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3 PRINCIPLES OF USED LABORATORY AND FIELD TESTS TO ASSESS THE BEHAVIOR OF SOIL
Undrained shear strength can be measured or assessed directly or indirectly in the laboratory or in the field. For the direct methods in this study, the field vane tests were conducted in the field, and the triaxial and direct simple shear tests in the laboratory. The CPTu test can be categorised into an indirect method to assess shear strength as the cone resistance especially, is known to be highly dependent on shear strength. It is known that between the preconsolidation pressure and the undrained shear strength there is a strong correlation therefore the oedometer test can be used for the indirect method in order to assess the undrained strength e.g. by using the S-parameter from the SHANSEP method.
3.1 Triaxial compression test
The triaxial apparatus has been described in detail i.e. in Bishop and Henkel (1962). As the name triaxial test indicates, three directions of loading are considered. Traditional triaxial test devices are only capable of controlling the cell pressure and/or the axial loading. Thus, the intermediate stress should be equal either to the minor or major stress. This kind of triaxial device will be considered later. Figure 3.1 shows the triaxial chamber with its components and the pressures that are exerted on the soil specimen during the triaxial test.
Triaxial compression tests can be conducted in drained or undrained conditions by controlling the valve which enables the water to flow out from the sample. To measure compression shear strength using the triaxial test, a cylindrical sample is sheared at the desired strain rate by increasing the stress difference between the axial and radial direction keeping the axial stress as the main stress. Before shearing, the sample can be consolidated into either an isotropic or anisotropic stress state.
Using pore pressure measurements or by ensuring that the pore pressure has been dissipated in drained tests, the effective stresses can be determined. For the practical
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reason of saving time, drained tests in soils with low permeability are rarely done. Usually, effective strength parameters of the cohesion and friction angle are estimated by conducting a set of three undrained tests consolidated to three different consolidation states. For example, the relation of effective normal and maximum shear stress in a shear plane can be solved by Mohr’s circles, where the failure line will be placed to touch tangentially on Mohr’s circles. Depending on the nature of the problem, the failure criterion can be associated with e.g. peak deviatoric stress, peak obliquity, peak pore pressure or limiting strain.
Some general issues to consider in triaxial testing are e.g., sample preparation/sealing, area correction, membrane resistance, frictional ends, seating, filter drain resistance, bedding error, saturation, system friction (Baldi et al. 1988, Germaine and Ladd (1988).
Figure 3.1. Triaxial test device with named components on the left side and on the right, the loadings for the sample (Day 2010).
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3.2 Triaxial extension test
The triaxial extension test differs from the compression test in that the major stress in the failure will be in the cell pressure. To become a major stress, the cell pressure can be increased, or the axial stress can be decreased. Perfect contact between the load cell and the top cap is necessary and can be assured e.g. by the suction cap (Jardine et al. 1984).
The reorientation of the principal stress during the test can cause more deformation of the sample in the extension test compared to the sample in the compression test at least when the samples are consolidated anisotropically. Due to the stress reorientation and higher strains, there may be a lack of a clear maximum shear stress in the results. Sometimes with increasing shear strain, the shear stress continues to increase slowly. Then the shear strength may be defined at a certain level of shear strain.
Undrained triaxial extension tests might include more uncertainties than compression tests. Lade and Tsai (1985) concluded with remoulded clay samples with over consolidation ratios of 1, 2, 5 and 15 that this difference is caused by different failure modes; 1.) line failure and; 2.) zone failure. In extension tests, the different failure modes have a more significant influence on the stress-strain behaviour and strength causing lower strength values in line failures. The failure mode which will occur, depends on the boundary conditions, the uniformity of density of the specimen or the clay deposit, and the tendency of the clay to dilate. (Lade&Tsai 1985)
3.3 Direct simple shear
The principle of a standard direct simple shear (DSS) is to shear consolidated sample by increasing the horizontal load and measure the shear stress and normal stress on the horizontal plane during tests. The DSS apparatus is shown in Figure 3.2. Horizontal stiffness for the consolidation is provided by the copper rings or with a wire-reinforced membrane around the samples. During shearing, the copper rings and wire-reinforced membrane can slide in a horizontal direction. The sliding causes some extra friction, especially with the copper rings, which may be significant in soft soils. Thus, consideration should be made for this friction.
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An undrained condition is ensured by keeping the volume constant in shearing. This can be done by maintaining a constant height. Then the vertical stress will decrease during shearing and compensate for the development of pore pressure as shown in the results of Dyvik et al. (1987) in a comparison between standard and truly undrained DSS devices for normally consolidated clay.
Only the average shear and normal stress are measured in a DSS test. These two alone do not provide enough information to define the actual stress state in the sample. The horizontal plane can be either a maximum stress obliquity or a maximum shear stress at failure. Additionally, De Josselin de Jong (1971) proposed that the sliding could occur in conjunction with rotations on vertical planes. In practice, the horizontal plane is assumed to be a maximum stress obliquity providing often a conservative design value.
At large shear strains, the maximum stress ratio (τ/σ´n) may be achieved and the maximum friction angle can be estimated for normally consolidated soils. Thus, the maximum friction angle corresponds to the maximum obliquity stress defined in the triaxial test. (Dyvik et al. 1987)
Figure 3.2. Direct simple shear apparatus with aluminum confining rings (Kjellman 1951).
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3.4 Oedometer test
The oedometer test is a one-dimensional compression test where the vertical stress and vertical strain is measured during compression of the cylindrical sample. It is commonly used in Nordic countries. The sample is confined by a stiff ring to ensure that all deformation occurs in the direction of the loading. From the results, it is possible to define the limit pressure where the soil response changes to a softer response. This limit pressure is called the preconsolidation pressure (p´c).
The loading in a standard oedometer test is performed either with incremental loadings (IL) or a constant rate of strain (CRS). In an IL test, the sample is loaded with load steps keeping the load constant for 24 h. In this study, a CRS oedometer test is used and thus the main interest is on those results.
In a CRS test, pore pressure is measured. In Figure 3.3 a schematic sketch of the oedometer device is shown where the pore pressure is measured by a pressure transducer. With the measurements of the pore pressure, effective vertical total stress can be calculated from the total effective stress. Consolidation can occur at the top of the sample while the bottom is kept closed for drainage and to measure pore pressure. An advantage of conducting the test with a strain rate is that when some excess pore pressure does not fully dissipate, the permeability of the sample can be assessed. In such cases, for effective stress calculations, the distribution of excess pore pressure in the sample should be considered.
Many methods to interpret the preconsolidation pressure from oedometer data have been proposed. The most common way to formulate the interpretation is based on a graphical fitting to a stress-strain curve. Comparison of the interpretation methods of Casagrande (1936), Janbu (1963) Pacheco Siva (1970), Becker et al. (1987) and Karlsrud (1991) with high quality samples did not show significant difference especially with the rather low preconsolidation pressures concluded by Paniagua et al. (2016).
Excessive friction between the soil sample and the confining ring should be minimized by using silicon oil between the ring and the sample.
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Figure 3.3 Schematic sketch of oedometer apparatus with fixed confining rings (Sällfors 1975).
3.5 Field vane test
The field vane test is conducted by inserting and rotating a four-bladed vane at the desired depth and measuring the applied torque and angular rotation (ASTM D2573-1 2007). To ensure an undrained condition, high rotation such as 0.1˚/s is applied. For a successful test, the key issues are: i) insertion of the blade to the desired depth with as little disturbance of the soil as possible, ii) measurement of the actual torque produced by the soil shearing. Measuring the actual torque of the soil shearing may be more challenging when the measurement is conducted above ground level. (Ortigao&Collet 1988); this is due to the rods used to transfer the torque to the vane being more vulnerable for extra friction. New devices have become available where the measurement and rotation are applied close to the vane. The principle of the vane device with rotation and measurements above ground level is shown in Figure 3.4.
The applied torque can be converted to shear stress by considering the vane dimensions and applying several assumptions in the conversion. Those assumptions are 1.) Penetration of the vane causes negligible disturbance, both in terms of changes in effective stress and shear distortion; 2.) No drainage occurs before or during shearing; 3.) The soil is isotropic and homogeneous; 4.) The soil is sheared on a cylindrical shear surface; 5.) The diameter of the shear surface is equal to the
50
width of the vane blades; 6.) At the peak and remoulded strength there is a uniform shear stress distribution across the shear surface; 7.) There is no progressive failure, so that at maximum torque the shear stress at all points on the shear surface is equal to su (Chandler 1988). Despite the fact that the assumptions are not truly correct as several researchers have demonstrated, such as, Cadling&Odenstad 1948, Flaate 1966, La Rochelle et. al. 1973, Roy&Leblanc 1988; Kimura&Saitoh 1983, Kulhawy et al. 1983, Mitchell 1976, Donald et al. 1977, Silvestri&Aubertin 1988, Chandler 1988; Gylland et al. 2012, Wroth 1984, De Alencar 1988, Griffiths&Lane 1990, the general trend of the vane results is often reasonable compared to the reference shear strength values; especially when the vane results are corrected to account for anisotropy and rate effects, as suggested by Bjerrum (1973).
Figure 3.4 A Vane device with casings to minimize friction between the rods and soil (Ortigao&Collet 1988).
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3.6 Piezocone test
The principle of the piezocone test (CPTu) is to push the CPTu probe straight into the soil with a constant rate of penetration (20±5 mm/s). During penetration, the tip resistance (qc), pore pressure (u) and the sleeve friction (fs) are measured, usually at logging intervals of 2 cm. (ISO. 2012) Accessories often included in the CPTu probes are an inclinometer and a temperature meter. The measured values can be used for interpretation of the stratification, classification of the soil type and evaluation of the engineering soil parameters. (ISO. 2012) Usually in the interpretation, the measured values need to be compared with the initial stress condition values such as the effective or total stresses and initial pore water pressure.
A better accuracy for cone resistance and sleeve friction can be implemented by using a more sensitive structure for the CPTu probe or by limiting the calibration to a specific range (Lunne et al. 1997). For a more sensitive structure, the usage range of the CPTu probe should be limited to avoid overloading and/or breakage of CPTu probe. Therefore, a preliminary investigation might be needed to discover a suitable layer for the sensitive cone. The classification of the soil type is based on the understanding and/or finding the trends as regards how the property is affecting the CPTu measurement(s). Often the classification is made by using charts which are divided into areas for different soil types. Figure 3.5 is an example of the identification charts proposed by Robertson thirty years ago (1990). Updates have been proposed for the chart types (e.g. Schneider et al. 2008) and for the calculated terms but the basis of the classification of soil behaviour type (SBT) can be seen very well in this figure. Figure 3.5 includes the placed points from one CPTu sounding (Sandven&Watn 1995). To place the points on the charts, some terms are needed for the calculations. These terms are normalized cone resistance (N more usual denoted as Qt), friction ratio (Rf) and pore pressure ratio (Bq). Equations for these terms are shown on the axis of the charts in Figure 3.5.
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Figure 3.5 Classification of soil type regarding the soil behaviour type charts proposed by Robertson (Sandven&Watn 1995).
3.6.1 Tip resistance
The tip resistance is the axial force divided by the cross-sectional area of the cone which is needed to push the cone into the soil while the cone displaces the soil. The tip is a conical tip with a 60˚ apex angle. A cross-sectional area of 10 cm2 has been used but different areas can be used to increase the accuracy or penetration depth.
Shear strength, soil stiffness, initial stress state and roughness of the cone and friction sleeve have an impact on the magnitude of tip resistance in homogenous soil (e.g. Vesic 1972, Baligh 1985, Teh&Houlsby 1991). Soil strength and stiffness are dependent on many factors. These factors increase the complexity of understanding how the tip resistance is formed. Additionally, some variation in corrected cone resistance can be noticed when comparing the results of the CPTu cones from different manufacturers at the same testing sites. Even with same CPTu cone, some variation can be noticed. In some cases, the reason for the variation was thought to be the result of poor preparation before the testing or a difference in the cross-sectional dimension of the tips (Lunne et al. 2018).
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3.6.2 Pore pressure
The pore pressure is the pressure composed of the initial pore water pressure (u0) and the excess pressure generated during the penetration of the cone into soils with low permeability. The pore pressure can be measured on the face of the cone (u1), right behind the cylindrical part of the cone (u2) or just behind the friction sleeve (u3). These locations are illustrated in Figure 3.6. The most common location for pore pressure measurement is u2 and that is used in this study. Later, only u2 pore water pressure is considered, unless otherwise is stated. It is useful to remember that in soft soils u1 is the highest and the u3 is the smallest measurement.
Pore pressure distribution around the cone is not constant and the distribution is a soil property dependent factor (Powel et al. 1988). The main soil property in fine-grained soils that has an influence on the generated pore pressure is over consolidation. In heavily over consolidated soils, the measured pore pressure at location u2 may even be negative due to dilatation. Decrease of the over consolidation ratio will increase the excess pore pressure due to the shear induced component having more influence on the generation of the excess pore pressure.
Figure 3.6 Three possible locations for pore pressure measurement (Lunne et al. 1997).
The well-prepared saturation of the piezocone is a primary factor in having a good pore pressure response from the pressure sensor. In a comparison between the pore pressure measurements of cones of different manufacturers, the results were similar when the preparation of the saturation was done well (Lunne et al. 2018).
The filter is needed to isolate the effective stress from the total stress so that only the pore pressure is measured. Two types of filters are commonly used, slot and
54
porous filters, which should be saturated with incompressible viscous mediums before testing. Both types are accepted in the ISO (2012) standard.
3.6.3 Sleeve friction
The sleeve friction is the friction acting over the standard 150 cm2 sleeve area. With the u2 pore pressure measurement, the friction sleeve is located right above the u2 filter.
The sleeve friction is formed by the horizontal effective stress and the coefficient of the effective friction between the soil particles and the sleeve (Larsson&Mulabdic 1991). The influence of the cone penetration on the initial stress state changes the effective horizontal stress around the cone shaft.
In soft soils especially, where the measured sleeve friction is low and typically under 10 kPa, the accuracy and variation of sleeve friction measurements have been noticed to be the poorest of the CPTu measurements. The fs measurement is sensitive for diameter changes along the cone body and the pore pressure difference at the ends of the friction sleeve have their own effects on measurements too (Lunne et al. 2018). With a well-designed CPTu cone, the accuracy of the sleeve friction can be improved but also some corrections can be made after testing.
3.6.4 The corrections for the measurements
The pore water pressure is known to affect the measurement of the tip resistance and the sleeve friction, especially in fine-grained soils (Campanella et al. 1981). The structure of the CPTu cone contains filter element(s) and sealed slots where the pore water pressure can act. These slots, where the distributed pore pressure is affected, are presented in Figure 3.7 for u1, u2 and u3 which correspond to the possible location of the pore pressure measurements. In the case of tip resistance, u2 decreases the measured value through the filter element. In soft soils especially, the relative magnitude is high compared to the tip resistance. The correction for the tip resistance is made with Equation 3.1 where a is the correction factor defined in the pressure chamber.
�J = �� + '� ∙ (1 − �) (3.1)
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Influence on the sleeve friction is more complex due to the unknown magnitude of the pore pressure at the level of the top of the sleeve friction corresponding the level of the u3 measurement. This u3 can be estimated by the Equation 3.2 based on the measurements on Swedish soft soil areas (SGI Rapport 42). Equation 3.3 is used for correction of sleeve friction where Asb is the cross sectional area at the bottom of the friction sleeve, Ast is the cross sectional area at the top of the friction sleeve and As is the area of the friction sleeve (ISO. 2012).
'; ≈ 0.7 ∙ ('� − '�) (3.2)
OJ = O- − (�F∙PQR�∙PQS)PQ
(3.3)
Figure 3.7 Correction for the pore pressure effects on tip resistance and sleeve friction (Modified by Mayne 2007, after Jamiolkowski et al. 1985).
The penetration length (l) can be corrected to correspond to the penetration depth, when the inclination (α) of the cone is measured. This effect is more significant at deeper depths. Equation 3.4 can be used for correction.
T = ∫ �V� (X)Y� Z$ (3.4)
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4 ANALYTICAL THEORIES BEHIND THE INTERPRETATION OF UNDRAINED SHEAR STRENGTH FROM CONE PENETRATION
Over the years, many attempts have been made to explain cone resistance by using analytical theories such as bearing capacity theory, cavity expansion theory and strain path method. These theories provide good background knowledge about the factors effecting the empirical fitting between the CPTu measurements and calibration data. Below a summary of these theories is presented together with a discussion of their limits.
4.1 Bearing capacity theory
In the bearing capacity theory, the ultimate load that the soil can sustain is calculated to satisfy the limit equilibrium of rigid incompressible material following the Mohr-Coulumb failure criterion. The resistance against the load is composed of the shear strength along an assumed slip surface (Terzaghi 1943). In the CPTu test, the cone resistance is assumed to correspond to the ultimate load along the failure plane.
The failure is suggested to occur on a predefined slip plane as a function of the friction angle and the roughness of the base of the footing. Many variations have been proposed for the shape of the slip plane as regards shallow and deep foundations.
The load at the footing is resisted by shear stresses at the edges of the shear zones and the overburden pressure. The shear zones are divided to three zones: 1.) The Active Rankine zone, below the foundation 2.) The Prandtl zone, a radial zone pushed by the active zone 3.) The Passive Rankine zone, pushed by the Prandtl zone.
Originally, the bearing capacity solution was proposed for relatively shallow footings. To deal with the cone penetration problem a correction factor to account for the shape effect is needed and should be adopted. Equation 4.1 shows the bearing
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capacity of a strip footing for effective stress analysis and in Equation 4.3 for an undrained analysis as proposed by Terzaghi (1943). In Figure 4.1 the shear zones for a rough foundation in frictional soil proposed by Terzaghi and Meyerhof (1951) are shown. The bearing capacity factors are solved with the assumption that the principle of a superposition is valid. Thus, the effects influencing the bearing capacity factors (Nc, Nq, Nγ) can be considered separately by assuming one bearing capacity factor of the three to be zero. With a deep foundation problem, the effect of the soil weight in the shear zones can be assumed to be negligible compared to the weight of the soil besides the foundation. Thus, Nγ is negligible so Equation 4.1 can be simplified into Equation 4.2. In a total stress analysis, cone factor Nγ is zero when the friction angle is zero.
��Y[ = �� ∙ \� + �]� ∙ \� + 0,5 ∙ `� ∙ a ∙ \b (4.1)
��Y[ = �� ∙ \� + �]� ∙ \� (4.2)
��Y[ = �� ∙ \� + �] (4.3)
One of the most used bearing capacity factors Nc and Nq are presented in Equations 4.4 and 4.5. Meyerhof proposed these factors in 1963 and later the factors were adopted by Hansen (1970) and Vesic (1975) to their bearing capacity formulas. Meyerhof included in his bearing capacity equation separate correction factors for the bearing capacity factors to observe the shape, depth and inclination effects. For example, in the case of a circular footing in a total stress analysis, the cone factor Nc with the shape factor will be 7.41.
\� = "c∙[,�d� ∙ ��� (45° + �� 2⁄ ) � (4.4)
\� = 8\� − 19 ∙ �V��� (4.5)
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The bearing capacity theory is the one of the first methods used in cone penetration analysis owing to its simplicity. To simplify the problem, several assumptions are needed. Specifically ignoring the soil stress-strain behaviour and the change of initial stress condition during continuous penetration have their own disadvantages for interpretation. The assumed shape of the slip surface also has huge effect on the bearing capacity.
Figure 4.1 Meyerhof’s (1951) comparison between his and Terzaghi’s (1943) plastic zones in the case of a rough foundation in frictional soil a.) shallow b.) deep.
4.2 NTH method
The basis of the NTH method is on the bearing capacity theory for strip footing. Instead of assuming that the size of the slip plane continues at a certain assumed level, the size can be controlled by the angle of plastification (β) in the bearing capacity factor as shown in Equation 4.6. Thus, with the limited length of the slip plane, the bearing capacity factor decreases while the length decreases. When the angle of plastification is zero, the bearing capacity factors are the same as Meyerhof (1963) proposed. In silts and fine sands, the angle of plastification is observed to be dependent on porosity (Senneset et al. 1988).
\� = "(c�∙g)∙[,�d� ∙ ��� (45° + �� 2⁄ ) � (4.6)
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The NTH method is an effective stress-based method solving the equilibrium of effective cone resistance by the effective stresses in rigid plastic soil. Thus, the pore pressure needs to be known or measured. To account for the effect of the excess pore pressure on a shear surface, the bearing capacity factor Nu is adopted in the bearing capacity calculation. Some assumptions concerning pore pressure behaviour and distribution on a shear surface are needed. Additionally, the measured value of the pore pressure is dependent on the location of measurement. Senneset et al. (1982) proposed the approximation formula for Nu presented in Equation 4.7.
\� = 6 ∙ ����� ∙ (1 + �����) (4.7)
The interpretation equation for CPTu penetration is expressed in Equation 4.8. The bearing capacity Equation 4.1 is termed as cohesion, but in NTH method attraction (a´) is applied.
��*[ = 8\� − 19 ∙ (���� + ��) − \� ∙ ∆'� (4.8)
The NTH method has some problems for interpretation. The theoretical analysis has been made for a plane strain condition without a suggestion for the shape factor (Lunne et al. 1985). The angle of plastification has been observed to include more uncertainties and variation (Lunne et al. 1997) than Sandven et al. proposed (1988) based on the type of soil. The continuous penetration of the CPTu cone into the soil causes complex loading for the soil element effecting the excess pore water pressure and its magnitude at a certain level and distance compared to the cone (Teh&Houlsby 1991).
However, the empirical comparison even with the simplified solution (β = 0 and a´ = 0) to estimate the effective friction angle in the NTH method seems to give reasonable values (Ouyang&Mayne 2017).
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Figure 4.2 Idealized failure plane and angle of plastification to control the size of the plane (Janbu&Senneset 1974).
4.3 Cavity expansion theory
In the cavity expansion theory, the cylindrical object with the initial radius (Ri) starts to expand to radius (Ru) as Bishop et al. (1945) suggested for the metal indentation problems. Later Vesic (1972) extended the analysis for compressible soils covering spherical and cylindrical expansions. To simplify the axisymmetric problems, Vesic assumed the soil to be weightless and initially in an isotropic stress state. In Figure 4.3, the cavity is expanded from radius Ri to radius Ru turning the surrounding soil into a plastic state till the radial distance of Rp. The plastic zone is surrounded by the unbounded elastic zone. The properties of the elastic zone are defined by elastic modulus (E) and Poisson’s ratio (v). The compressibility of the soil can be taken into account in a drained analysis allowing the plastic volumetric strain in the plastic zone.
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Figure 4.3 Expansion of the cavity (Vesic 1972).
To expand the cavity from the initial state, the pressure inside the cavity must be increased. When the cavity stress is increased, the surrounding soil must move outwards. With symmetrical expansion of the cylinder or the sphere in an unbounded homogenous soil with isotropic initial stresses, the displacement vectors are directed at any point to the center point of the sphere or the axis of the cylinder of the cavities. Based on these assumptions, the spherical and the cylindrical symmetries are valid. The equilibrium equation for the soil element can be written as in Equation 4.9, where k is 1 for the cylindrical expansion and 2 for the spherical expansion.
i�ji/ + k ∙ �j�l
/ = 0 (4.9)
Following Vesic’s analysis of the spherical cavity expansion in an undrained condition, the radial stress at the limit equilibrium (2·su = σr – σθ) can be expressed by Equation 4.10 integrated from Equation 4.9 and solved by using the boundary condition of σr = Pu when r = Ru. Pu is the ultimate pressure in the cavity and r is the radial distance.
�/ = m� − 4 ∙ �� ∙ $� 2 /5n
6 (4.10)
The change of the cavity volume causes the radial displacement of the surrounding soil and the radial displacement ur at a distance of r can be obtained from the Equation 4.11 when the volume change is zero. The expansion of the cavity will cause the radially decreasing displacement of the surrounding soil. The displacement up corresponds to the smallest displacement where the yielding occurs when the
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displacement up causes the radial stress of σp. With Lame’s solution, the stresses and the radial displacement can be associated and the displacement of up can be presented as shown with Equation 4.12, where σ0 is initial isotropic stress.
��; − �o; = �/; − (�/ − '/); (4.11)
'p = :?��∙q ∙ �p ∙ 8�p − ��9 (4.12)
The radial stress σp can be defined with Equation 4.10 when r = Rp and be substituted to Equation 4.12. Thus, the substituting previously defined equation into Equation 4.11 and by neglecting the higher powers of up and assuming the �o; to be small compared to ��;, the ratio of Rp / Ru can be obtained. This radial distance Rp would be the boundary between elastic and plastic zones and the boundary is dependent on the undrained shear strength and the elastic modulus (E). Instead of the elastic modulus, the shear modulus is usually used and the boundary can be found by using the ratio of the shear modulus and the undrained shear strength (G / su) termed the rigidity index (Ir). Equation 4.13 shows that the ratio of Rp / Ru is only dependent on the rigidity index Ir.
5r5n
= s q�∙(:?�)∙-n
= tu/ (4.13)
Thus, the relative distance to the boundary of the elastic and the plastic zone is known. Vesic proposed this formula for the bearing capacity based on the spherical cavity expansion theory as shown by Equation 4.14.
m� = �� ∙ \� + �� = �� ∙ v; ∙ [$�(u/) + 1] + �� (4.14)
Later in 1977, Vesic updated this bearing capacity factor using a more realistic failure pattern according to observations of models and the penetration of full-size piles. This assumed failure pattern under the pile point is shown in Figure 4.4 for zone I.) the conical wedge, zone II.) the radial-shear zone in dense soil, although in relatively loose soil, visible radial slip surfaces have been not observed, and zone III.) the plastic zone as a cavity expansion region. Equation 4.15 presents the bearing capacity factor for a frictionless soil as Vesic (1977) proposed. By using a characteristic value of 100 for the rigidity index in clay (Mayne et al. 2009), the bearing capacity factor will be close to 10.
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\� = v; ∙ [$�(u/) + 1] + c
� + 1 (4.15)
The advantage of the cavity expansion theory is that it takes into account the soil deformation behavior and illustrates fairly simply relationship between the deformation behavior and the cone resistance.
Figure 4.4 Tri-zonal failure mechanism as Vesic (1977) proposed.
4.4 Strain path method
The strain field around an expanded cavity in undrained conditions have been shown to be independent of soil properties, whereas the stress field changes with the soil type (Ladanyi 1963, according Sandven 1990). This observation provided the basis for the Strain path method (Baligh&Levadoux 1980, Baligh 1985).
Baligh (1985) treated the quasi-static penetration problem as a steady state problem keeping the rigid object like cone fixed while the homogenous soil passes the cone as steady state flow in an undrained condition. In Baligh’s solution, the soil behaves as rigid and perfectly plastic under isotropic conditions.
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In the Strain path method, the problem is assumed to be essentially strain-controlled due to the kinematic constraints in deep penetration. The individual soil element goes through the strain path in the soil stream passing the cone. To approximate the strain paths, the velocity fields can be estimated by a potential theory treating the soil as an incompressible, inviscid and irrotational fluid (Whittle&Aubeny 1993).
By integrating the velocity fields, an approximation of the strain field can be done. With the appropriate constitutive law, the deviatoric stress for soil elements may be determined. Thus, the Strain path method enables an estimation of the stress rotations and the magnitude of the loadings experienced by the soil element during cone penetration.
The Strain path method includes an error in the equilibrium of the initial stress state as Teh and Houlsby (1991) mentioned. To correct this initial in equilibrium, they proposed a solution that would balance the unbalanced body forces by equal and opposite forces. They treated the soil as a homogeneous elastic-perfectly plastic material obeying the von Mises yield criterion. Based on the calculations of the finite element method (FEM), they proposed the Nkt factor including the effects of the rigidity index of soil, cone roughness (αf), shaft roughness (αs) and initial stress condition (Δ) for cone resistance. The effects of αf, αs and Δ were solved by varying one factor at a time and proposing the correction factor to include the effect on the cone factor. They especially noticed that the cone resistance is highly influenced by the effective horizontal stresses rather than the vertical stress. The cone factor is presented in Equation 4.16. In typical cases, the Nkt factor will be in the range of 9 to 17 (Teh&Houlsby 1991).
\y[ = v; ∙ (1 + $�u/) ∙ 21.25 + zj
����6 + 2.4 ∙ X − 0.2 ∙ X- − 1.8 ∙ ∆ (4.16)
where:
X = �V�"|V'}ℎ�"��80 ≤ X ≤ 1.09
X- = �ℎ�O�|V'}ℎ�"��(0 ≤ X- ≤ 1.0)
∆= ���� − �.�2 ∙ ��
� = ℎV|�TV���$��|"����Z"�(−1.0 ≤ ∆≤ 1.0)
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The calculated axial and radial strain contours for a cone angle of 60˚ are presented in Figure 4.5 as found in the results of Teh and Houlsby (1991) with a simple calculation. As they noted, below the tip, the strain contours are approximately similar to that predicted by the spherical cavity expansion theory and above the cone tip, the strains are similar to the cylindrical cavity expansion solution.
Figure 4.5 Radial and axial strain contours around a cone (Teh&Houlsby 1991).
4.5 Estimation of undrained shear strength from excess pore pressure
Penetration of the cone into the soil causes development of excess pore pressure (Δumeas) as shown by Equation 4.17. The excess pore pressure is especially significant in clays close to a normally consolidated state. Excess pore pressure consists of two components: the first component is related to the change of mean total stress (Δumean) in loading and the other component is related to shear induced pore pressure (Δushear) corresponding to the change in mean effective stress (Wood 1990).
�'�*,- = ∆'�*,� + ∆'-.*,/ (4.17)
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To estimate undrained shear strength, both components of the excess pore pressure components need somehow to be estimated. For example, by using the Cavity expansion theory, the change of mean total stress near the cone can be estimated by Equation 4.18. in the case of spherical expansion (Vesic 1972). Spherical expansion is thought to correspond better cone resistance, whereas the sleeve friction is more influenced by cylindrical expansion.
��� = ∆'�*,� = v; ∙ �� ∙ $�(u/) (4.18)
In the analytical solution especially, simplifications and assumptions are needed, and one often used simplification is to consider soil behaviour as perfectly plastic. Then, the shear induced component is constant after failure even if the shear strains increase. Thus, the shear induced component may be evaluated by Henkel’s parameter (a) when the relationship between the shear induced pore pressure at failure and maximum deviatoric stress (qmax) are known - as shown in Equation 4.19. This Henkel’s parameter (a) should illustrate the complex loading of cone penetration. (Vesic 1972) Another way is to use the principle of Critical state mechanics to estimate the shear induced component for failure by Equation 4.20. (Mayne 1988) Some assumptions or general parameters are needed for the evaluation.
�'-.*,/ = � ∙ ��,� = � ∙ -n� (4.19)
�'-.*,/ = !�� − ! � = !�� − !�� ∙ 2345� 67 = !�� − �∙-n
1 (4.20)
Some uncertainty is related to the pore pressure measurement location and what is the best single test type to estimate pore pressure behaviour. Likewise, the assumption of a perfectly plastic material does not include consideration of shear induced pore pressure after the soil has been plastified.
4.6 Spherical cavity expansion theory combined with critical state soil mechanics (SCE-CSSM) to assess OCR
Spherical cavity expansion (SCE) theory (Vesic 1972, 1977) combined with Critical state soil mechanics (CSSM) is an application solving the penetration problem with
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effective strength parameters rather than undrained shear strength (Mayne 1991, 1993). Net cone resistance (qnet = qT – σv0) in an undrained condition is usually considered a total stress analysis. Accordingly, the analyses for excess pore water pressure (Δu2 = u2 – u0) and effective cone resistance (qeff = qT – u2) are carried out by using effective strength parameters. Then, it is more practical to solve all three analyses with respect to the over consolidation ratio (OCR). After solving the OCR, the undrained shear strength can be evaluated.
A net cone resistance solution for OCR is utilized for the Spherical cavity expansion theory where the undrained shear strength is substituted with Equation 2.12 from the simple critical state model such as a modified Cam clay. To assess the OCR, the modified Cam clay is also used for the interpretation of the excess pore pressure as shown in the previous paragraph. Combining these to solutions, the OCR can then be assessed from the effective cone resistance as well. The modified Cam clay model is an elasto-plastic material model where yield softening and hardening until failure can be controlled by the parameter Λ. Three Equations 4.21, 4.22 and 4.23 can be found to solve the OCR in ideal soils corresponding closely to elastic perfectly plastic soils (Mayne 1991, Mayne&Bachus 1988, Mayne 1992). The general values like M = 1.2, Ir = 100 and Λ = 1 can be used for first-order approximations (Mayne et al. 2009). The first-order approximations are shown at the end of the Equations 4.21 – 4.23. The initial stress state of the soil is assumed to be isotropic.
��� = 2 ∙ � (� 1⁄ )∙(���@A) �@A�⁄(v ;⁄ )∙(Y�zj?:)?c �⁄ ?:�(: 7⁄ ) = 0.33 ∙ ���@A
�@A� (4.21)
��� = 2 ∙ � 8∆�F �@A�⁄ 9:(� ;⁄ )∙1∙Y�(zj):�(: 7⁄ ) = 0.54 ∙ �F�A
�@A� (4.22)
��� = 2 ∙ � ::.��∙1?: ∙ 2���F
�@A� 6�(: 7⁄ ) = 0.60 ∙ ���F�@A� (4.23)
In ideal soils, all three Equations have been shown to have similar values for OCR in many cases. Some difficulties have been seen in interpretations of highly structured soils with Equations 4.21 - 4.23. Therefore, later updated solutions for structured soils were proposed to consider the incompatibility of strain levels where the maximum values of deviatoric stress and excess pore pressure occurred in triaxial compression tests (Agaiby&Mayne 2017). Agaiby and Mayne (2017) proposed that the incompatibility of strain levels can be considered by two friction angles defined
68
from an undrained stress path corresponding to: a.) peak stress (qmax) and (b) maximum obliquity (σ´1 / σ´3)max. “The measured cone resistance (qt) corresponds to the peak friction angle at φ'qmax while the measured pore water pressure (u2) relates to the value taken at larger strains, corresponding to a maximum obliquity, or φ'MO” as Agaiby and Mayne (2017) stated. By using this hypothesis, they proposed Equations 4.24, 4.25 and 4.26 to interpret OCR, where Mc1 is for φ'qmax and Mc2 for φ'MO.
��� = 2 ∙ � (���@A) �@A�⁄1��∙(� ;⁄ )∙(Y�zj?:)?c v⁄ ?: �⁄ �(: 7⁄ )
(4.24)
��� = 2 ∙ � 8∆�F �@A�⁄ 9:(� ;⁄ )∙1�F∙Y�(zj):�(: 7⁄ )
(4.25)
��� = 2 ∙ �(���@A) �@A�⁄ (1�� 1�F⁄ )∙�8∆�F �@A�⁄ 9:�:.��∙1��?(1�� 1�F⁄ ) �(: 7⁄ )
(4.26)
Combining Equations 4.24 and 4.25, evaluation of the operational rigidity index can be obtained as shown in Equation 4.27 (Agaiby&Mayne 2017).
u/ = "�! �:.�∙��?�.���∙1��∙(�∗:)�∙1�F1��∙(�∗:) �
(4.27)
Where
�J = ��*[ ����⁄
�∗ = ∆'� ����⁄
4.7 Considerations related to evaluations
A practical solution to evaluate the undrained shear strength from the CPTu results should be simple enough that it is straightforward to use. However, the reliability should also be sufficient for the geotechnical design to be able to trust the evaluations. Often this means that interpretations have been tried that will improve the results by additional index parameters, which are easy and cheap to define. These index parameters could have some relationship on mechanical properties that really are resisting the cone to penetrate the soil.
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Agaiby (2018) listed and updated the list of proposed cone factors (Nkt) after the work of Konrad & Law (1987); Lunne et al. (1997); Low (2009) and presented the list in his doctoral thesis (Agaiby 2018). From the list, it was found that the rigidity index is the most relied on parameter to include the stress-strain behaviour of soil. Other often included factors in the cone factors are the roughness factor and the initial stress state factor. The roughness factor is partly controlled by a standardised demand for smoothness (ISO 2012) but of course, it still has its own influence, but this influence is better if limited to a smaller range. Thus, it might be included as a part of a constant numerical value in the interpretation. The effect of the initial stress state is more difficult to evaluate as the undrained shear strength depends on the over consolidation and the initial stress state and the initial stress state and the OCR might not be known. A guess as to the initial stress state might be possible based on e.g. the first-order estimation of OCR or the undrained shear strength if necessary.
Compared to laboratory results, the rigidity index might not be as simple to define for a complex penetration problem. To simplify the penetration problem, the linear behaviour of soil is often adopted. To consider the real behaviour of soil, the nonlinear response of stress-strain with one rigidity index value would be difficult but at least a well-representative value might be possible to find. Additionally, the most predominant failure mode to represent cone penetration is not simple but some recommendations can be found in the literature for representative failure mode as well as for the strain level where the rigidity index should be defined. Broussard (2011) listed some studies where these two elements were considered. To summarise the lists, it seems that for failure mode, the undrained triaxial compression test is the most recommended (Keaveny 1985, Teh&Houlsby 1991, Schnaid et al. 1997, Yu&Mitchell 1998, Yu et al. 2000) and the most recommended shear modulus to define rigidity index is the secant modulus at 50% of the stress to strength (Skempton 1951, Keaveny 1985, Kondrad&Law 1987, Schnaid et al. 1997, Yu&Mitchell 1998, Mayne 2001).
The interpretation of the undrained shear strength from the excess pore pressure or by using an effective stress analysis would need consideration of the location where the pore water pressure is measured as the principal loading direction rotates along the cone towards the shaft. Figure 4.6 presents the main strain types and the predominant failure modes around an advancing cone (Keaveny&Mitchell 1986, Baligh 1984, after Su 2010). The failure mode for the cone resistance seems to be triaxial compression whereas for the pore pressure measurement right behind the cone (u2), the failure mode seems to present more as a pressuremeter test and/or a
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direct simple shear test. Thus, there is more influence on pore pressure development from shearing (Chen&Mayne 1994).
Figure 4.6 Main strain types and predominant failure modes around an advancing cone after Su (2010) combined data of Keaveny&Mitchell 1986 (left) and Baligh 1984 (Right).
The piezocone test is carried out at relatively high strain rate compared to standard laboratory tests. Due to the high strain rate, the undrained condition can be ensured in soils with low permeability as shown in recent extensive studies by Kim et al. (2008) and Lehane et al. (2009). Based on the reviewed data of the triaxial compression tests, Kulhawy and Mayne (1990) agreed with the observations of Ladd and Foott (1974) that each log cycle of strain rate increases the undrained shear strength by 10 %. Accordingly, Chen and Mayne (1994) proposed that the undrained shear strength from the piezocone test should be 53 % higher than su from the triaxial compression test with a strain rate of 1 % per hour. Often, the strain rate effect is not considered in the analytical approaches to the CPTu interpretation even though comparison is made to the undrained shear strength defined with standard laboratory tests.
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5 TEST PROCEDURES USED IN THE STUDY AND QUALITY CLASSIFICATIONS
This chapter includes more specific details about the triaxial tests, the DSS, FV and CRS tests and how they are performed, and some associated details are also mentioned. To attain reliable data, the measuring devices and procedures should be suitable for their purposes. Additionally, the disturbance of the soil caused by performing the test or by the sampling procedure have a significant effect on the results. To estimate the sample quality, several indicators and methods have been proposed. All methods are related to the impacts that are caused by a loss of effective stresses and the interactions between the soil particles. Perhaps the most practical method is based on the void ratio change in consolidation to an in situ effective stress as proposed by Lunne et al. (1997). This method is adopted in this study.
The laboratory tests were conducted in a climatised room with a constant temperature of 20°C. However, the average temperature in soil layers is approximately 6 – 8 degrees in southern Finland. Any temperature or strain rate corrections are not included in strength results, preconsolidation pressure or CPTu data. The effects of a partly higher strain rate that might occur in field conditions and a higher testing temperature compared to in-situ are opposite, i.e. they partly compensate each other.
5.1 Undrained triaxial tests
The details of the triaxial apparatus with the correction procedures used in this study is described in Kolisoja (1990).
Triaxial tests were performed with the cylindrical samples with a height to diameter ratio of 2. The usual specimen diameter was 51 mm. Samples were loaded to failure at a constant axial strain rate of 0.8 %/h in undrained condition. The loading was continued until an 18 % axial deformation was achieved.
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In an undrained condition, the drainage is prevented in shearing and the pore pressure is measured below the sample through the pressure channel. Because of the limitation of the triaxial device used, the backpressure was not used. Before shearing, the sample was consolidated to the desired consolidation stress state. Special attention was given to the de-airing.
Consolidation of the triaxial samples were mostly done isotropically. The cell pressure (σcell) in the isotropically consolidated undrained compression (CIUC) tests and the isotropically consolidated undrained extension (CIUE) tests was chosen as the smallest value between [0.73·σ'1; 0.6·σ'p], where σ'1 is the effective vertical stress and σ'p is the preconsolidation pressure. The basic principles for selecting between these values were as follows:
� to consolidate close to the in situ hydrostatic stress level
� to ensure that the yield surface did not expand during consolidation.
Some anisotropically consolidated tests were performed with same principle as in the isotropic consolidation. To estimate initial deviatoric stress state, a coefficient of earth pressure at rest (K0) should be defined or estimated by friction angle. Due to the special interest in the undrained shear strength, neither of these parameters were defined in advance. Thus, conservative deviatoric stress values were selected.
The consolidation pressure was kept constant for 24 h, while the end of consolidation was verified by the measured volumetric strain. The preconsolidation pressure σ'p was inferred from the CRS oedometer tests on samples from the same tube as the triaxial specimens.
The area correction was made assuming the shape change to be cylindrical but in extension tests a necking of the sample in the upper part occurred as shown in Figure 5.1. Therefore, the tails of the stress paths are biased by too low deviatoric stress values. In addition, some nonlinearity in the membrane correction might have occurred in the high axial deformations. Therefore, the acceptable maximum axial strain in the extension tests was assumed to be at the level of 10 - 12 % strains.
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In extension tests, axial stress was decreased to cause an extension failure. The axel of the load cell was fixed to the top cap via a tiltable joint.
Figure 5.1 Triaxial extension test sample after 18 % of deformation.
5.2 Direct simple shear test
In the DSS tests, the cylindrical samples supported by copper rings were first consolidated to a stress level close to the preconsolidation pressure. They were then unloaded to the in-situ stress state before shearing (Selänpää et al. 2018). During consolidation, an excess pore pressure can dissipate from the top and bottom of the sample. The height of the samples were around 20 mm and the diameters were 50.4 mm corresponding a surface of 2000 mm2.
The DSS device was purchased in the middle of the study. Thus, at some testing sites DSS tests were not done. To adopt the new device and procedure, some verification tests were needed. During verifications, the procedure was improved regarding the consolidation. The rest times after loading and unloading and appropriate stress level compared to preconsolidation pressure were adjusted. The stress levels compared to preconsolidation pressures varied between 70 - 90 % and the rest times 0.5 – 10 hours. A higher stress level corresponds with a shorter rest time. The stress rate in consolidation was 0.1 kPa per minute. This stress rate corresponds to a sufficiently slow loading rate so that significant excess pore pressure should not develop under two-way drainage conditions. In the DSS test, pore
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pressure is not measured. Thus, the stress rate was estimated utilising the oedometer test results taking into account that drainage in oedometer test is one way and the tests have been performed using a constant rate of strain.
During shearing, the copper rings slide in a horizontal direction. The sliding causes friction which has been assessed by the shearing of water balloons. The friction correction is made afterwards for the tests. The sample was sheared approximately at a shear strain rate of 4.8 %/h until a 20% of shearing was attained. The strain rate is higher than in the triaxial test, but the lower strain rate caused unstable friction between the rings. Figure 5.2 shows the sheared DSS sample still inside of the copper rings.
Due to the membrane between the sample and the copper rings, some horizontal deformation might occur before the sample is supported. This might cause a small disturbance in the sample due to the extra deformation.
Figure 5.2 Deformation after a sample shear with supporting copper rings.
5.3 CRS oedometer
The details of the oedometer apparatus used in this study is described in Kolisoja (1990).
The specimen with a diameter of 45 mm and a height of 15 mm was loaded with a constant rate of strain up to a vertical loading of 300 kPa. The strain rate was 0.4 – 0.6 %/h. The higher strain rate was used when a higher silt content was observed by visual observation. After the loading of 300 kPa was attained, the specimen was unloaded to 25 – 50 kPa for an estimation of over consolidation stress-strain behaviour.
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The estimation of the preconsolidation pressure was made by drawing two curves corresponding to over consolidation and normal consolidation regions in the stress-strain graph (Kolisoja 1990). The stress-strain curves were fitted by using two non-dimensional parameters for both regions as Janbu (1963) proposed in his Tangent modulus approach. To find the most representative parameters, an ordinary least squares method was used to verify the curves having the smallest sum of squared errors. Corresponding stress level, where these two curves intersect, is assumed to be the preconsolidation pressure.
5.4 Field vane
In sensitive soft soils, it is reasonable to use a larger vane to minimize the effect of disturbance and guarantee that the maximum torque value is in a favourable range. Then, a height and diameter of 150 mm and 75 mm for the rectangular vane was selected as corresponding to the greatest possible size for this particular vane device. The angular rotation rate was 0.1 ˚/s and the torque was measured after every 0.1˚. To disturb the soil for the measurement of the remoulded vane strength, an angular rotation rate of 6 ˚/s was used, corresponding to the greatest possible angular rotation rate. At least 10 full revolutions were used before measuring the remoulded shear strength with an angular rotation rate of 0.1 ˚/s.
The vane testing was performed by using a so called downhole vane device, where the rotation and the measuring unit are close to the vane. The vane device is shown in Figure 5.3. This configuration may produce a more realistic stress-rotation curve because the torque caused by soil shearing should be less influenced by other friction sources and the reaction torque is transferred to the protective casings which have a large rotational stiffness. The protective casings are fixed to the drilling rig above the ground level during shearing. Distortion of the rotation may occur if the joints of the protective casings are not tightened properly.
Pushing the vane out of the protection shoe to the test depth caused some initial mobilisation of shear stress. Thus, in the results, some deviation from zero may occur. In some comparisons, the shear stress-rotation curve was corrected to begin from zero by assuming a linear line through the first measurements.The measured vane strength should be corrected to correspond to the loading in the design problem. Helenelund’s (1977) correction factors are based on the findings of Bjerrum (1972, 1973) and Aas (1976). According to the analysis of slope failures
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calculated by the measured vane strengths, Bjerrum proposed that the correction factor is a function of the plasticity index (Ip) and Aas (1976) suggested it was a function of ratio of the mean (uncorrected) vane strength (su,measFV) and mean effective stress in the critical slip line. Combining these findings and using the correlation of plasticity index and liquid limit (Hansbo 1957, Ostreman 1960) and by specifying the correction to be valid for liquid limit values over 50% , Helenelund proposed two Equations 5.1. and 5.2. He suggested comparing the correction factors and using the smaller value for the correction. However, it is advisable to only use Equation 5.1 in stability calculations for railway embankments in Finland (Ratahallintokeskus 2005).
�: = :.�:?�� :��⁄ ≤ 1 (5.1)
�� = �.�:?�∙Qn,���Q��
A� (5.2)
Figure 5.3 The vane pushed out of the protective shoe and casings below the drilling rig.
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5.5 Piezocone test
CPTu tests have been executed by following the ISO standard(2012). In every testing site, the usual number of CPTu tests was 4 – 5 soundings in order to attain repeatability. Additional modules of resistivity and seismic measurements were used. Seismic tests caused longer pauses, which might have some effect on the pore pressure measurements immediately after the pushing forward started again. Studies of seismic and resistivity measurements have been dealt by Mäenpää (2016) and Haikola (2018).
To ensure good saturation in soft layers, the dry crust was pre-pushed, and the pre-pushing was extended below the ground water level. The pre-pushing was made by pushing a steel cone with a diameter of 60 mm into soil. The opened hole was filled with water and then used for the temperature balancing of the cone close to ground temperature. Saturation was done by placing the CPTu probe in a vacuum chamber filled partly with silicon oil and then increasing the vacuum pressure to remove the air from the pore pressure measuring system. Figure 5.4 shows the CPTu cone during saturation.
Figure 5.4 CPTu probe in the vacuum chamber.
Due to the main interest in soft layers, a sensitive cone with a measuring range of 0…7.5 MPa for the tip resistance was used. The actualized measuring range based on the results were mainly under 700 kPa, which is under 10 % of the whole range. The tip resistance has a quoted non-linearity of 0.2 % full scale (15 kPa). For sleeve friction, the quoted non-linearity is 0.7 % of full scale (2 kPa) and for pore pressure
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measurement the given value is 0.25 % of full scale (5 kPa). Combining the non-linearity value, and zero shift value in each measurement, maximum error may be estimated. By calculating the maximum error, tests were classified into the application classes in ISO (2012) and the results, which fulfilled the demands of the best class, were accepted for further inspections. The limits for maximum acceptable errors in the best class were 35 kPa or 5 % of the measured value for tip resistance, 10 kPa or 2 % of the measured value for pore pressure and 5 kPa or 10 of the measured value for sleeve friction; these values are rather easy to fulfil with careful preparation of the test. Additionally, an overall view of the results and a comparison between tests at the same site were used to find the most representative CPTu test for interpretation. Particularly, the poorer pore pressure responses were easier to see.
The correction factors to correct the tip resistance and sleeve friction were investigated by the modified triaxial pressure chamber. The pressure chamber had a limited range for pressure. Thus, the maximum pressure of 400 kPa. based on the results at the depths where the samples were taken, was enough. The correction value differed from the value that the manufacturer proposed. The measured value for a was 0.83, and this was used for correction. The sleeve friction was noted to be almost independent of u2 as was proposed by the manufacturer. When it was observed that the sleeve friction measurement had a rather poor resolution of 1 kPa, especially more uncertainty was related to the estimation of excess pore pressure at level u3. Thus, sleeve friction is only corrected by the effect of unequal pore pressures on the ends of the sleeve friction. u3 is estimated by equation 3.2.
The results may be possible to correct afterwards by the values of zero shifts. The CPTu probe used does not have a temperature sensor, which would give valuable information about temperature changes and may be one major effect for zero shifts. With temperature measurements, CPTu measurements may be corrected for temperature changes at depths, where the temperature changes occur. To understand the temperature effect on the sensitive cone used, the effects of the measurements were tested by observing the values at different balanced temperatures. The results are shown in Figure 5.5. Even though there was an attempt to compensate for the temperature changes with the hardware, the temperature effects seemed to be significant for tip resistance and pore pressure measurements. For this reason, the temperature was balanced carefully close to ground temperature before taking the zero values and starting the CPTu test.
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The cone used in this study measures the cone resistance and the sleeve friction from independent load cells. A Wheatstone strain gauge circuit is used for both sleeve and tip measurements. For the pore water measurement, a pressure cell is used. The inclination measurement and temperature compensation are added on the cone. The cone was not equipped with a temperature sensor as mentioned above. Presumably, the accuracy of the cone has improved by the more sensitive structure where the strain gauges of the tip resistance are attached, and the permitted maximum loading has limited range where linearity is better. The more sensitive structure is related to the deformation occurring in the structure under applied load. As the operation of strain gauges is based on the change of the resistance due to its change of cross-sectional area in conductor under applied load, increasing the change of cross-sectional area by increasing the deformation in the body, where strain gauges are attached, make the resistance change more significant. This improves the accuracy of lower values.
Figure 5.5 The temperature effect on the sensitive cone used.
To ensure that the new CPTu probe did not have a systematic error, some comparisons were conducted at three testing sites by comparing the results to results measured by another probe. The Kotka site is not otherwise included in this study. This reference CPTu probe has an equipped pore pressure measurement with a slot filter and the range for tip resistance measurement is 0…10 MPa. The comparison results are presented in Figures 5.6, 5.7 and 5.8 and “Ref” denotes the other CPTu probe. This reference test was performed only once per site so the variation in the
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results of that probe is not known. However, some conclusions can be made: 1.) NP systematic error was found 2.) The highest deviation of corrected tip resistance is 30 kPa in a shallow depth at the Paimio test site. 3.) The measured sleeve friction with the reference cone was generally about 2 kPa lower and this may be due to unequal end areas of the friction sleeve causing an error in the uncorrected values 4.) More variations occurred in corrected tip resistances and pore pressures at shallow depths. For the reference cone, the pore pressure correction value for tip resistance was the same as the manufacturer reported. At the Paimio test site, some doubt arose about the soil in some testing points not being fully natural due to the small distance to an old installed water pipe. In the vane results especially, which are presented later, a very high variation can be noticed.
Based on the results, it is not possible to state which probe measured the values more correctly. In interpretation, it should be remembered that uncertainty is related to all measurements even those values to which the interpretation is targeted. Similar uncertainties and even more variation were found in a more thorough investigation with different probes published by Lunne et al. (2018).
Figure 5.6 Comparison results from the Sipoo testing site.
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Figure 5.7 Comparison results from the Paimio testing site.
Figure 5.8 Comparison results from Kotka testing site.
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5.6 Sample and test quality classification
To classify the sample quality, the method proposed by Lunne et al. (1997), has been adopted for laboratory tests. This method based on the void ratio change occurred during reconsolidation to in-situ stress Lunne et al. (1997). defined the classes based on the NGI (Norwegian Geotechnical Institute) experiences of sampling and the effects of the sample disturbances. The classes are shown in Table 1. The limits for the classes can be defined by the water content at a certain range in the over consolidation ratio instead of the relative change in the void ratio (Δe / e0).
Table 1 The quality classes based on the void ratio change proposed by Lunne et al. (1997).
To classify the tests, the relative change in the void ratio should be defined. In the CRS tests, the void ratio change is estimated from the axial strain and it is selected from the stress level corresponding to the in situ effective vertical stress. The void ratio change in the triaxial test is defined by the water flowing out from the sample, while the sample is reconsolidated to estimate in situ mean effective stress. In the DSS test, the vertical deformation is measured but because of the membrane, the horizontal support needs some compression before acting. Thus, sample quality was estimated from the void ratio change based on the water content measured from the sample before and after the test. The initial water content was measured from the excess soil removed when trimming. Inhomogeneities may have some influence on class classified in the DSS tests and this is also the reason why the two DSS specimens classified as “Good to fair” were accepted for interpretations. Relative void ratio changes at reconsolidation are presented in Figure 5.9. One DSS specimen did not show any void ratio change at reconsolidation, which might also be an indication of natural inhomogeneity in the sample.
The field vane test always included some disturbance caused by the insertion of the vane into soil. The vane test was performed three times at the same depth moving
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each test to a distance of about 2 metres from the previous test. The best of the three results at the same depth was assumed to include the least disturbance.
Figure 5.9 Verified sample qualities at reconsolidations a.) triaxial specimens b.) DSS specimens c.) CRS samples.
CPTu test classification was made according to ISO (2012). The CPTu tests that fulfilled the demands of Class 1, were accepted for further analysis. All CPTu tests were compared to each other to identify random errors. Consequently, the especially poor pore pressure response in shallow depths or the disturbance caused by pre-pushing was easier see in comparison.
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6 SAMPLING AND THE TEST RESULTS
Previous studies have shown that usually larger samples provide a higher quality even when the diameter of the sampler is just enlarged (Lunne et al. 1997, Lanzky&Palmquist 2013). However, the size of the sample does not alone assure high sample quality. In addition, samples should be designed so that the sampling operation, sample handling and the sampler itself cause as little disturbance as possible. It is known that inside clearance, sharpness and the geometry of the cutting edge and the area ratio determine the directions of the loadings and thus also impact the disturbances when a sample is taken (Baligh 1985, Baligh et al. 1987, Siddique 1990, Hopper 1992). For research purposes, two large size samplers have proven their ability to produce high quality samples in various fine-grained soils, namely the Sherbrooke block sampler (Lefebvre&Poulin, 1979) and the Laval tube sampler (Rochelle et al. 1981), whose operation differs fundamentally.
A growing interest in increasing sample quality has also generated some recent research in Finland (Mataić 2016). Undisturbed sampling in Finland is traditionally performed by piston samplers like the Swedish standard piston samplers St I and St II (SGY 1976). These samplers provide samples with a diameter of 50 mm and that size and quality are the basis for Finnish geotechnical designs when samples are needed. Sample disturbance caused by a poorer sampler might partly compensate for higher preconsolidation due to the higher strain rate used in the laboratory; a strain that is greater than that which could occur at a construction site. Recently, Tampere University of Technology (TUT), nowadays Tampere University (TAU), has developed a tube sampler providing samples with a diameter of 132 mm. Some comparison has been carried out to compare the sample disturbance between this large tube sampler and the St I sampler.
6.1 Improvement of sample quality by large tube sampler
The design of the large tube sampler is strongly inspired by the SGI large-diameter sampler with a diameter of 200 mm (Larsson 2011). The SGI sampler is a hybrid between the Laval sampler and a modified NGI 200 mm sample as summarised by
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Löfroth (2012). In addition to the difference in dimensions, the most significant difference in the TUT sampler compared to the SGI sampler is that the sampling tubes are changeable, and the sample can be stored inside the tubes. Different lengths of sampling tubes can be utilized by adjusting the thread bars corresponding to the lengths of the sampling tubes. Some dimensions and pictures of sampler can be found in Di Buò (2020). The sampling phases are listed below:
1. Opening and cleaning the hole with an auger drill
2. Filling the borehole with water or slurry
3. Lowering the sampler on the base of the borehole
4. Slowly pushing the sampling tube into the soil while the top valve is open
5. Increasing adhesion between the sample and sampling tube by waiting for a time
6. Closing the top valve, cutting the sample by pulling the cutting wire, feeding in air/water and retrieving the sampler
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7. Preparing the sample for transportation
Figure 6.1 A sketch of the TUT sampler drawn and designed by the author.
The sample quality comparison between the TUT sampler and the St I sampler was conducted at two testing sites Paimio and Sipoo with CRS oedometer test results. The results are shown in Figure 6.2 and in the publication Di Buò et al. 2018. The results from the comparison indicate that the results are consistent and no significant differences in preconsolidation pressure occurred. Perhaps the sample quality of the TUT sampler provides a more constant sample quality than the St I, but one enormous benefit with the TUT sampler is that it enables the trimming of two or three triaxial samples at exactly the same depth.
In general, the sample quality with the TUT sampler is good to excellent (Di Buò et al. 2018) but sometimes the sampling was challenging, and this caused a loss of the samples during withdrawal or just poor quality in the test results. Equality of the sample qualities between the St I and TUT sampler may prove that the results with the tube samples does not need further consideration for reduction of peak values
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as the Finnish design practice based on the sample quality provided by St I seems to be same. Only a consideration to reduce the values due to the rate effect is valid. The tube sampler and piston sampler cause some extension stress in the samples due to suction when retrieving; this was shown in the results of Schjetne (1971) when monitoring pore pressure during piston sampling. This suction might be one reason for the higher disturbance compared to the block samples taken by the Sherbrooke sampler (Lunne et al. 1997), which does not cause as much total stress change in the sample (Amundsen et al. 2017).
The Perniö test site included some high-quality samples taken by a Sherbrooke-type mini block sampler (Emdal et al. 2016). For two out of eight of these high-quality samples, a considerably higher active undrained shear strength was measured, in comparison with the other test results. These two tests are the only samples that were prepared and trimmed for testing within 24 h of sampling. (Selänpää et al. 2018) It might be that negative pore pressure inside the sample has endured well, thus minimizing the volume change in the samples - keeping the intact structure more unaffected. Later in Figure 7.3, these results can be noticed at a depth of 6.65 m where there are significantly higher triaxial compression results below and above this point.
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Figure 6.2 CRS results with the TUT sampler and the St I from several depths at the Paimio testing site (Di Buò et al. 2018).
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Figure 6.3 CRS results with the TUT sampler and the St I from several depths at the Sipoo testing site (Di Buò et al. 2018).
By using the TUT sampler, some experiences and observations have been gained and some improvements can be made in the future to improve the sampling procedure and the sampler itself:
� The possibility to feed water through the channel into the bottom of the auger drill to avoid suction while retrieving the auger drill.
� Better control of the geometry of the sampling tube.
o The cross section of the tubes was not always perfectly round.
� Consideration of developing a sample catcher for tube sampler thus avoiding the need for top suction.
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� Developing a better cutting procedure for the sample by improving the feeding of water/air into the shear zone or by a different cutting mechanism.
o Clay is sticky material and could stick on the shear zone requiring force to separate the sample from the soil below.
� Disposing the inside clearance.
o It was subsequently noticed that the inner diameter of the cutting shoe was 1 mm smaller than in the sampling tubes even though the author had designed and dimensioned it to be same.
o For success with samplings having a smaller diameter in the cutting shoe, slurry is needed in the empty space between the sample and the tube to ensure top suction.
6.2 The cutting of the tube sample
The principle of the layouts of the specimens in the tube sample is shown in Figure 6.4. Some differences in layouts can be found such as in the case where the DSS was not available and sometimes it was necessary to repeat the previous test because of homogeneity or difficulties in trimming. The upper part of the sample tube will always be partly empty because of the sampler design.
The testing was begun by pushing the deepest part of the sample out and then trimming at least one oedometer specimen and a specimen to define the sensitivity and liquid limit of the fall cone (FC). If there was a day or days between the testing at the next level, usually 10 mm of the height was removed. CRS1 provided the first result of the preconsolidation pressure enabling the consolidation pressure for triaxial tests and DSS to be defined. A constant OCR was used to calculate preconsolidation pressure for different depths in the tube sample. Preconsolidation pressure was needed to define the consolidation pressure for the triaxial and DSS tests. The oedometer test was repeated after the strength tests. From the results, it was possible to assess the sample quality and storage effect. When the sample quality was verified to be similar compared to first CRS test, it enabled some extra tests like the friction angle test series with a sample diameter of 35.8 mm to be conducted.
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Figure 6.4 The principle of the layout for specimens.
6.3 Influences on relative void ratio changes at reconsolidation
In Figure 6.5, the relative void ratio changes at reconsolidation in the triaxial compression tests using high quality samples taken with the TUT sampler are compared to several parameters in order to detect if the TUT sampler is inappropriate for a specific soil type or if the greater relative void ratio change was caused because of the sampling procedure or the consolidation stress state. Any significant effect was not found between the change of relative void ratio and sensitivity, liquid limit, organic content and water content. Correlation with the organic content showed a very small trend but it might also have been influenced by one sample having an organic content of 1.8 %. Originally, the sample quality criteria proposed by Lunne et al. (1997) was developed for inorganic clays so the influence of an increasing organic content with respect to volume change at reconsolidation may need its own limits for classification.
Sampling depth may have some influence on sample quality. Greater effects might be caused at deeper depths where the sampling needs more time for all the sampling phases e.g. more time for stress release at the bottom of the borehole and more time for extension stress to have an influence due to suction while the sampler is retrieved. The greatest influence was found between OCR and the void ratio
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change. This is a natural result of being further away from yield surface in the reconsolidation stress state as the OCR is greater and thus, the creep should have less influence. An attempt to assess this OCR influence is made in the consolidation criteria presented in Chapter 5.1.
Excluding the poorer quality samples in the comparison will result in a better suitability of the sampler but in this comparison, the purpose was to check if the TUT sampler caused some notable distortion in certain soil. Thus, the behaviour of the soil may be more distorted in those soils where the sampler has more difficult to obtain high quality results and then this could lead to incorrect connections in correlations proposed later.
The storage time effect is considered in Figure 6.6 for samples taken with the TUT sampler. Partly because of the poor capacity of the test devices and human resources, storage times are in general too long. Previously, the effect of long-term storage was studied using the CRS results which had over half a year between the CRS tests (Di Buò et al. 2018). The results were from the same sample tube but not from exactly the same depth. No significant effect was found when the measured water content was the same, which indicated that the soil type was the same. For this reason, the sample quality is also assumed to be valid for triaxial tests. Short-term storage effects have not been studied even though it is known that stress relief effects are the most significant immediately after sampling, this is because of the lower undrained shear strength (Skempton&Sowa 1963). For comparison, in Figure 6.6a, only isotropically consolidated samples with the TUT sampler and mini Sherbrooke sampler were accepted. Figure 6.5b also includes anisotropically consolidated samples. Later some speculation is presented about the effect of the consolidation type. Storage times have been defined by the sampling date to the day when the shear phase in triaxial testing was started. Storage time between the sampling date and the day when the sample was installed in the triaxial cell under some cell pressure would be a more appropriate interval. In addition, the normalized shear strengths in Figure 6.6b could be compared against the time between the CRS and triaxial testing. From the results presented in Figure 6.6, it can be concluded that the results are not clear, however, it can perhaps be stated that some effect might occur as the storage time increases. Storage time is not the only factor which may have an influence on sample disturbance. Other uncertainties also increase the variation in the results.
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In Figure 6.7, axial strains at failure are compared to relative void ratio changes at reconsolidation. There is a minor indication that higher change in void ratio during reconsolidation indicates a higher axial strain at failure. It is though difficult to estimate the influence on undrained shear strength. For that, comparison tests for samples at the same depth with a different level of disturbance would be needed.
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Figure 6.5 Relative void ratio change compared to a.) Sensitivity b.) Liquid limit c.) Organic content d.) Water content e.) OCR f.) Sampling depth.
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Figure 6.6 Influence of storage days to a.) Relative void ratio change at reconsolidation b.) Normalized undrained shear strength by preconsolidation pressure.
Figure 6.7 Comparison between axial strains at failures and relative void ratio change at reconsolidations from undrained compression tests
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7 TESTING SITES
For the calibration of the CPTu interpretation, data was collected from several test sites in Finland. Soft clay areas in Finland are mostly located close to the coast of the Baltic sea. Therefore, the testing sites are mainly located in those areas as can be seen in Figure 7.1. In Finland, there is an interest in increasing the axle loads of cargo trains. The test sites at Masku, Paimio, Sipoo and Lempäälä were chosen in connection to this and based on some preliminary stability calculations. (Andersson-Berlin 2012). The goal was to also utilise the results of this study for more accurate stability evaluations, in order to reduce the need for stability improvements. Perniö is also located close to the railroad but this area has been reinforced earlier. This site has been used for failure test performed by loading the freight wagons until the old sidetrack failed (Mansikkamäki 2015, Lehtonen et al. 2015). In a parallel study conducted by Di Buò (2020) preconsolidation stress and deformation properties were evaluated by CPTu. The detailed CRS results and all conducted CPTu results can be found in Di Buò’s dissertation while here only one CPTu test per site used for interpretation is shown.
All the samples have been taken and tested for this study and the Di Buò’s study. Only a little preliminary data was found about the testing sites. Perniö is an exception. The Perniö site has been investigated in previous studies (Mansikkamäki 2015, Lehtonen 2015, Mataić 2016, D´Ignazio 2016), but the results needed to correspond to the same sample quality by using the same large tube sampler at all the sites. In Perniö, sampling had previously been performed twice in 2015 and 2016. In 2016, the mini Sherbrooke sampler was tested with guidance of NTNU doctoral student Helene Amundsen.
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Figure 7.1 The location of the testing sites. [https://asiointi.maanmittauslaitos.fi/karttapaikka/]
The laboratory tests were started with oedometer tests to assess the preconsolidation pressure. From the preconsolidation pressure, the consolidation pressures for the triaxial tests which define the undrained shear strength were defined. An oedometer test takes about 4 days to complete so usually the first set of triaxial tests were done on samples stored for at least one week before trimming. The short-time effects of sample storage concerning sample quality and the effects on soil behaviour have not been studied. For some soils, it has been observed that in-situ response is more realistic immediately after sampling - at least with high quality block samples (Amundsen et al. 2017).
The sampling was conducted between depths of 2…9 meters even if the soft homogenous layer continued deeper. The reasons for this are that the stability problems are more dependent on the upper layers, the sampling is more time-and-cost-consuming at deeper depths and the sample quality may be poorer partly due
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to the sampling method. The planned testing points for all sites were inside an area of 4 metres times 4 metres where the distances to another test point was 2 m. CPTu and vane testings were done around the sampling point. The ground surface was relative flat at all sites.
The depths for the samplings were chosen based on the CPTu data, in an endeavour to take the samples for comparison from the most homogenous layers. The tip resistance in the CPTu test is dependent on the properties of the soil close to the tip but also the surrounding layers located outside of the shear zones have impacts on the tip resistance. Thus, the tip resistance value is a complex average value weighted by the distances to the tip. To compare the calibration test results to values defined by CPTu test, it is easier to have the samples from the homogenous layer that have a sufficient distance from another layers. The homogeneity level might be difficult to establish due to the inherent variations of structure and composition in the microscale. One problem caused by inhomogeneity is also related to calibration tests. The DSS test may be the most sensitive for small weak layer(s) in the sample. The vane test and the compression test might average these small differences in the horizontal planes due to the different inclination of the shear plane.
An inclusive set of index tests were done. These index tests are not always exactly from the same depths as the strength test. The closest values have been used.
The usual layering in natural soft soil areas in Finland consists of soft layers with low variation in OCR below the top layer. The top layer is either dry crust or peat. Over consolidation changes gradually from a heavily consolidated state to a nearly normally consolidated state below the dry crust. The layers with a constant OCR over 3 and over 2 m thickness are relatively rare. At least the sites that where proposed for investigations did not have these properties. It may be that for sites with clay which has a higher OCR, stability is not a problem.
The sampling does not include the samples from the top layers. For an effective stress calculation, some assumptions for unit weights were used. The unit weight of 17 kN/m3 was adopted for the dry crust and for the organic layer, the unit weights of 12.5…13.5 kN/m3 were used.
A high capacity cone was used to define suitable depths for a sensitive cone to avoid overloading and to define the starting level as being below ground water level where there is less the risk of losing saturation. The high capacity cone was in some sites pushed through the dry crust which caused a lack of saturation and poor pore
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pressure measurements. Only one CPTu with a sensitive cone and classified as the best was used for further evaluation and interpretation.
In the following chapters, the results of undrained shear strengths from different laboratory tests and from the field vane tests are presented. The graphs include trend lines for every test type. The right edge of the lines is drawn through the highest strength values. The trend lines presented do not accurately represent the strength profiles and the layers of the soils are not considered. The thickness of the line is about 2.5 kPa without any fundamental reason. The intention is to illustrate the sample quality effect in sensitive soils. With a high quality sample, the maximum shear strength points are likely to be located on the right side of the thick line. Correspondingly poor-quality samples are likely to be located on left side of the line. The sample quality is illustrated with different colours for the strength and oedometer tests: green corresponds to the quality of “Very good”, yellow is “Good” and red means “Poor”
In the CPTu results, the presented sleeve friction value is an averaged value over ±4 cm at a corresponding depth because of the relatively poor resolution of the sleeve friction measurement. The resolution was 1 kPa, which is a little poor for very soft soils. The charts of the CPTu results include the normalised friction ratio (Fr) and pore pressure parameter values (Bq). These parameters are defined in Equations 7.1 and 7.2:
a� = �F�A���@A
(7.1)
%/ = ����@A
∙ 100% (7.2)
For all sites, the CPTu data is presented also as SBT charts similar to those shown in Figure 3.5. Some conclusions in the site descriptions utilise the trends shown in Figure 3.5 for increasing OCR and sensitivity. The datapoints in the SBT charts would be better specified according to the depths. When comparing the Bq and Fr results in the figures where the CPTu results and the SBT results are shown, some conclusion can be made as regards the depths. The equation for normalised cone resistance (Q) is shown in Equation 7.3:
�[ = ���@A�@A� ∙ 100% (7.3)
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7.1 Perniö, Salo
At the Perniö site, the thickness of the dry crust layer was about 1 m and below the dry crust, there was a two meter thick layer of fine-grained soil where the OCR (defined as σ´c/ σ´v0) decreased from highly over consolidated state to an OCR of just above 1.3 until a depth of 3.5 m was reached. In this transition layer the top part of it had more organic content and the organic content decreased to close to zero at the bottom of the layer, as shown in Figure 7.2. Slight over consolidation continued until a depth of 7.5 meters. Initial water content (w) was in all the measured depths higher than the liquid limit (wL) defined by the fall cone. The difference between the liquid limit and the plastic limit (wP) corresponding to the plasticity index (IP) was mainly between 20 and 40. The sensitivity (St) varied between 40 and 70 defined by the fall cone. The clay content seemed to increase as the depth increased. The data points were categorized by the sampling year either 2015 (15) or 2016 (16). The elevation from sea level is about +7.4 m in the N60 vertical coordinate reference system. The estimated initial water level was 0.8 meter below the ground surface.
The strength profiles are shown in Figure 7.3. In soft layers the undrained triaxial compression strengths show the highest values and triaxial extension strengths correspond to the lowest values while the measured field vane strengths and DSS strengths are between these minimum and maximum values. The magnitudes of the strengths are mainly between 7 – 24 kPa.
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Figure 7.2 The results of the oedometer tests, the index tests, the estimated initial pore water pressure and the effective stress curves at the Perniö testing site.
The CPTu data profiles are presented in Figure 7.4. Based on the CPTu result measurements, the cone resistance is mainly under 400 kPa and the measured pore pressure is under 300 kPa. The cone resistance follows well the preconsolidation and the compression strength profiles. In the pore pressure measurements, the response of the higher over consolidation can be seen in the shallow depth while the pore pressure measurements are significantly lower. Lower pore pressure values compared to cone resistance are caused by the natural shear behaviour of stiffer soil but some dilatation could also have occurred. Additionally, some poorer response in the measured values may have occurred at shallow depths. The profiles of the normalised friction ratio (Fr) and pore pressure ratio (Bq) also indicate a higher OCR in shallow depths where the Bq is lower but the Fr is higher. Additionally, higher values of Fr can be influenced by higher organic content. From Figure 3.5, where the SBT charts are presented with arrow trends, the previous conclusions can be justified. While the accuracy of the sleeve friction measurement might be poor, relative changes in the Fr profile are more useful rather than single values. Figure 7.5 presents the SBT charts from Perniö.
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Figure 7.3 Measured strengths at the Perniö testing site.
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Figure 7.4 CPTu test results from Perniö and the calculated values of normalized friction ratios and pore pressure ratios.
Figure 7.5 Normalised soil behaviour type (SBT) charts from Perniö a.) Qt - Fr b.) Bq - Qt.
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7.2 Masku
The elevation from the sea level is about +9.15 m in the N2000 vertical coordinate reference system. The ground water level is estimated to be 1.2 m below ground surface. The ground level is slightly inclined towards the Maskunjoki river. The layers at the Masku site, starting from the top layer was a dry crust continuing to about 1.2 m below the ground surface, the soft layer started almost immediately after the dry crust and continues on to a depth of over 16 m.
The water content was higher than the liquid limits, but the relative differences were smaller than at the Perniö site, as shown in Figure 7.6. The sensitivity was lower compared to the values at Perniö. The plasticity indexes were between 40 and 60 and the clay content was over 60 at all depths. At a depth of 5 m, the organic content was a little bit higher compared to other depths and that may cause higher plasticity index values and a higher water content. The OCR was around 1.6 at the depths where the oedometer tests were conducted.
The strength results in Figure 7.7 are between 8 and 25 kPa at the levels where the samples were taken. The vane tests were conducted at deeper depths than the those where the sampling was performed. At deeper depths, the vane strength is at its highest, almost 50 kPa.
The CPTu results show constant values for Fr and Bq in Figure 7.8. In the case of Bq especially, this may indicate that the OCR is constant. The same conclusion can be made from the oedometer results and the SBT charts in Figure 7.9, where the data points are close to each other. A small gradual change in qc and u2 can be seen at level 7.2 m in Figure 7.8.
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Figure 7.6 The results of the oedometer tests, the index tests, the estimated initial pore water pressure and the effective stress curves at the Masku testing site.
Figure 7.7 Measured strengths at the Masku testing site.
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Figure 7.8 CPTu test results from Masku and the calculated values of normalised friction ratios and pore pressure ratios.
Figure 7.9 Normalised soil behaviour type (SBT) charts from Masku a.) Qt - Fr b.) Bq - Qt.
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7.3 Sipoo
The ground level is around +22.3 m in the N2000 vertical coordinate reference system, and the estimated initial ground water level is one meter below ground surface. The thickness of the dry crust is about 1 m. Below the dry crust the soil softened quickly at first until the 1.5m level and then the softening became slower based on the CPTu results in Figure 7.12 where Bq indicates changes towards lower OCR values with increasing depth. Variation of the OCR can also be seen in the SBT chart presented in Figure 7.13. At about the 4 m, the soil reached the level where the behaviour of the soil seems to be at its weakest. From that level, no strength tests were conducted. Figure 7.10 shows that the highest water content is almost 120 at the 5 m level while at the 3 and 8 m levels the water content is around 90. The higher water content at level 5 m may be due to the somewhat higher organic content. In soft structured soils, high water content often correlates with high clay content but is not only dependent on that. The clay content was over 60 % at all depths where it has been measured. The OCR stayed under 2 while the lowest value was around 1.4. Based on the result at the 6 m level, some variation in preconsolidation results can be found.
Figure 7.10 The results of the oedometer tests, the index tests, the estimated initial pore water pressure and the effective stress curves at the Sipoo testing site.
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In Figure 7.11, where the strength values are presented, the field vane results at the two first levels show high variation. It might be that all the values do not represent the most appropriate value. Comparison with the DSS and extension test results indicate that the value measured with the vane was too low. One explanation could be that the insertion of the vane at this testing depth has caused a greater disturbance in the soil causing a loss of strength. During maintenance after the vane tests, some dry crust was observed sticking to the vane blades. This extra soil on the vane blades may be the reason for the greater disturbance as the intruded vane with the dry crust stuck on the blades may have caused greater strain on the surrounding soil at that test depth. The DSS and triaxial compression results might be the most realistic values. At the 3 m level, the extension result is low compared to other tests. In the extension tests, the shear bands in the samples are more horizontally directed than in the compression test and thus thin weak layers or inhomogeneities will have more influence in the extension test.
Figure 7.11 Measured strengths at the Sipoo testing site.
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Figure 7.12 CPTu test results from Sipoo and the calculated values of normalised friction ratios and pore pressure ratios.
Figure 7.13 Normalised soil behaviour type (SBT) charts from Sipoo a.) Qt - Fr b.) Bq - Qt.
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7.4 Paimio
The top layer at Paimio was a dry crust with thickness of 1.2 meter. From depth of 1.2 m to 2.2 m, the soil got softer. At a depth of 3.2 m, the OCR was 1.8 and at a depth of 6.3 m the OCR dropped to 1.3. Below 6.3 m, the OCR stayed approximately constant based on the CRS results in Figure 7.14 and the Bq results in Figure 7.16 The water content was lower between 3 and 5 m compared to the results from a depth of 6 to 9 m. The liquid limit followed the same trend. The highest water content corresponding to about 110 was at the same level at which the highest organic content was measured. The sensitivity was slightly higher at depths between 3 and 5 m compared to lower depths even though the water content was lower. The explanation can be found by calculating the liquidity index which is accordingly higher at shallower depths. The liquidity index and the sensitivity has been shown to correlate well in fine grained soils (Wood 1990). The sensitivity was higher than 60 at all depths. The same trend as noted before between the water, clay and organic contents were found from the results. Higher organic and clay contents can cause a higher water content and liquid limits, at least in natural soils. The laboratory results are presented in Figure 7.14. The ground level is at an elevation of +22.25 m in the N2000 vertical coordinate reference system and the ground water level is estimated to be 1.2 m below the ground surface.
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Figure 7.14 The results of the oedometer tests, the index tests, the estimated initial pore water pressure and the effective stress curves at the Paimio testing site.
The strength results in Figure 7.15 show that measured undrained shear strength values are under 25 kPa at all depths. The vane results again showed some variation at shallow depths and the variation in the extension test results seems to occur randomly. It might be that some of the tests were conducted too close to an old water pipe causing variations, especially in vane results at shallow depths. Possibly the results at depths of 4.3 and 7.3 meters do not represent the extension strengths correctly. The strength profile of the DSS results nicely follows the same profile as the preconsolidation pressures (Fig. 7.14). The triaxial compression test result at the shallowest depth seems to differ from the general trend. Some extra disturbance may have occurred in this case, even though sample quality corresponds to the best class. Based on the results in Figure 7.17, the soil can be classified as a sensitive soil by many CPTu measurements.
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Figure 7.15 Measured strengths at the Paimio testing site.
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Figure 7.16 CPTu test results from Paimio and calculated values of normalised friction ratios and pore pressure ratios.
Figure 7.17 Normalised soil behaviour type (SBT) charts from Paimio a.) Qt - Fr b.) Bq - Qt.
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7.5 Lempäälä
The topmost layer at Lempäälä was gyttja/peat with a thickness of 3 metres. Through visual observation during the drilling of the hole for tube sampling with the auger, it seemed as if the organic content decreased downward. Below the first layer, there was an organic clay layer between the depths of 3 to 4.5 metres from ground level. Below the elevation of 4.5 m, inorganic homogenous clay was observed, and it continued to an elevation of 7 metres. Below 7 m, the soft layer continued but with more variation in the CPTu results as shown in Figure 7.20. Using the N2000 vertical coordinate reference system, the measured elevation was +83.15 and the ground water level was estimated to be 0.4 m below the ground surface.
Lempäälä site was the first site where the new tube sampler was tested. This might be the reason for the variation in the sample quality due to the unfamiliar sampling procedure. In another site with similar index properties, the sample quality was better.
Figure 7.18 The results of the oedometer tests, the index tests, the estimated initial pore water pressure and the effective stress curves at the Lempäälä testing site.
Based on the results in Figure 7.18 at a depth of 3.7 m, a higher variation in the index results can be noticed. The OCR stayed at around 1.4 at all depths where the CRS
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tests were performed. The measured water content was higher than the liquid limit indicating structuration of the soil. The measured strengths at depths below 4 metres are the lowest of all of the testing sites presented in Figure 7.19. This could be caused by the low overburden pressure while the upper layers are organic layers with low unit weight and a high ground water level. The corrected cone resistance also shows the lowest values at the same depths in Figure 7.20. Based on the results in Figure 7.21, which presents the SBT charts, the upper layers were not classified as organic soils.
Figure 7.19 Measured strengths at the Lempäälä testing site.
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Figure 7.20 CPTu test results from Lempäälä and the calculated values of normalised friction ratios and pore pressure ratios.
Figure 7.21 Normalised soil behaviour type (SBT) charts from Lempäälä a.) Qt - Fr b.) Bq - Qt.
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8 MEASURED RESULTS
In this chapter, the dataset is first considered using the index properties to consider the relationships of one item to another. Later, the influences of the index properties on the strengths and normalised strengths are observed. Partly, this attempt to find inner correlations of the dataset could explain later why some index properties both show some influence on the parameter under study. Additionally, the strain behaviours are presented for each strength tests.
8.1 Some correlations between the index test results for Finnish clay
In Figure 8.1 a) the theoretical relation between the water content and the unit weight for fully saturated soils with a specific gravity of 2750 kg/m3 is presented together with the measured values. In Figure 8.1 b) the corresponding organic content is presented for the measured values. It is known that the organic content has an influence on this relationship, but with the typically quite low values in Finnish clays, the influence is minor. Therefore, the water content can be used to calculate the unit weight and also the total stresses for low organic soils below the ground water level.
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Figure 8.1 Correlation of unit weight to natural water and organic content in the dataset.
The sensitivity of soil and liquidity index is observed to correlate well together as can be seen in Figure 8.2 where sensitivity of the soil in the dataset is considered. The liquidity index is defined in Equation 8.1. The dataset includes all index test results without consideration of the sample quality which does have an influence on the sensitivity value. Possibly the sensitivity is more pronounced in the small remoulded strength value rather than the slightly underestimated undisturbed value. Nevertheless, the dataset shows that low remoulded strength values correlate with the sensitivities and even better with the liquidity index (Figures 8.2 b and c). However, the curve is so steep when the liquidity index is lower than 1.5 that an accurate estimation of the remoulded strength may be difficult. Naturally, the ratio of the liquid limit and water content compared to the remoulded strength correlate well. The liquid limit can be defined with the fall cone test as well the remoulded strength. The liquid limit defined by the fall cone corresponds approximately to the strength state where remoulded strength is 1 kPa. Thus, a higher water content compared to a liquid limit will result in a lower remoulded strength and vice versa.
¢u = ��£���£
(8.1)
Sensitivity is also been compared to the liquid limit, plasticity index, clay content and ratio of water content and liquid limit in Figure 8.3. As was mentioned earlier, sensitivity is more pronounced in the remoulded strength rather than in the sensitive
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peak value for sample disturbance. Due to having samples with the qualities of fair to excellent, a clear correlation can be seen between the sensitivity and ratio of water content and liquid limit. When the sensitivity is compared to the liquid limit a significantly poorer correlation is found. Using the ratio of water content and liquid limit to assess sensitivity would be useful as these properties are not as dependent on the sample quality as sensitivity.
Figure 8.2 Comparison between a.) Sensitivity and Liquidity index b.) Sensitivity and Remoulded strength c.) Liquidity index and Remoulded strength d.) Ratio of liquid limit and water content and remoulded strength.
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Figure 8.3 Sensitivity compared to a.) Liquid limit b.) Plasticity index c.) Clay content d.) Ratio of water content and liquid limit.
The relationship between water content and organic content, sensitivity, clay content and liquidity index are presented in Figure 8.4. The best correlation can be found for the clay content (8.4c). Increasing organic content causes an increase in water content and perhaps this could cause a greater variation in results. Water content alone does not define the sensitivity of soil as sown in Figures 8.4b and 8.4d where clear trends are not found as to the sensitivity and liquidity index. In a remoulded state, electrochemical forces can have significant influence on the remoulded strength (Rankka et al. 2004).
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Figure 8.4 Water content compared to a.) Organic content b.) Sensitivity c.) Clay content d.) Liquidity index.
In Figure 8.5, more focus has been put on the Atterberg limits and especially as regards the liquid limit. Correlation between the liquid limit and plasticity index is shown in Figure 8.5a. A significant difference would not occur if the plasticity index was estimated by only using the liquid limit value. The water content and liquid limit tend to increase when one increases in relation to each other (8.5b). The same kind of trend is found between the water content and clay content (8.5c) but more variation occurs at the top part of the range. Increasing the organic content could to a certain extent offer some explanation for the variations in Figures 8.5b and 8.5c.
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Additionally, for example, the pore water chemistry, mineralogy and grain size distribution can have their own influence on the liquid limit value (Mitchell 1976).
Figure 8.5 Liquid limit compared to a.) Plastic limit b.) Water content c.) Clay content d.) Organic content.
Increasing the organic content in clay causes a rise in the rate effects, su/σ´c ratio and deformation to failure as summarised in the SGI report 38, which includes results from Finland and Sweden. In these results, the organic content varied in a much larger range compared to the results of this study. However, the influence of the organic content to plastic limit, liquid limit, remoulded strength and sensitivity is shown in Figure 8.6. The organic content seems to have the most significant effect on the plastic limit (8.6a). For the liquid limit, a poorer correlation was found (8.6b)
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and the remoulded strength and sensitivity does not show any correlation with the organic content (8.6c,d).
Figure 8.6 Correlation of organic content to a.) plastic limit, b.) liquid limit, c.) remoulded strength, d.) sensitivity.
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8.2 Field vane comparison between uphole and downhole devices
Field vane (FV) tests have been carried out using a new commercial vane device with rotation and measuring immediately above the vane (referred to as a downhole device) (Selänpää et al. 2017). To check the results of the downhole device with respect to the traditional way i.e. with the measuring and rotation above the ground level (referred to as an uphole device), a comparison between the results with downhole and uphole devices was made at three test sites. Murro and Kotka are additional test sites that were omitted from any further analysis. In Murro, the soil layers are more organic to a depth of 15 m as they have higher values than the specified limit of 2 % for this study (Koskinen et al. 2002). In Kotka, the higher organic content occurs at shallow depths and deeper depths; the sample quality was classified as being insufficient for CPTu calibration. The uphole device was specified to include casings to isolate the rotation rods from the surrounding soil and the measuring device was a Nilcon device modified by electric rotation instead of a hand crank. In a Nilcon device, the data is recorded on a paper graph and the results are measured by hand from the paper graph. FV tests were usually performed 2-3 times at each depth with both devices to observe the variability of the maximum vane strength. The results are presented in Figure 8.8 and 8.9 using circle symbols for the uphole device results and crosses for the downhole device results. It seems that with the downhole device the results are slightly higher, which may be the result of better accuracy or just smaller disturbance occurring with the vane with a smaller blade thickness (2mm vs. 3mm). This better accuracy at low values was clearer in remoulded strength results (Fig. 8.9) where the relative difference is higher. Measuring very small values by hand from a paper graph could also be challenging while taking care of the non-linearity of the torque spring at low values. Additionally, some extra friction can occur between the vane and the measuring device causing apparent increase in the soil strength even though the critical parts are lubricated.
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Figure 8.7 The combination used for the uphole device: the vane, casings and partial protection shoe.
In Figure 8.10 a comparison is shown of the remoulded strengths defined with the vane and fall cone tests from a publication by Selänpää et al. (2017). The publication discusses the reasons why the remoulded strength values measured with a downhole device are higher than those measured with the fall cone and the fact that the difference is not constant. The publication gives several explanations for the higher remoulded strength values with the downhole vane such as 1.) unstable vane position during rotation, 2.) some extra friction caused by vane shaft, 3.) accuracy, 4.) dissipation of excess pore pressure 5.) and just the difference of two methods with different boundary conditions. Due to these reasons, only the fall cone sensitivity values are used in this study.
Figure 8.8 Comparison of the Vane peak strength between the uphole and downhole devices at three sites.
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Figure 8.9 Remoulded vane strength comparison between uphole and downhole devices at three sites.
Figure 8.10 Remoulded strength values of Perniö clay as defined by the field vane and fall cone (Selänpää et al. 2017).
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8.3 Comparison between anisotropic and isotropic consolidation in triaxial tests
Consolidation of the triaxial samples was mostly done isotropically. To investigate the difference between isotropically and anisotropically consolidated samples, some of the undrained compression samples were consolidated anisotropically. Anisotropically consolidated samples were consolidated to the estimated in situ stress and isotropically consolidated to an isotropic stress corresponding to the anisotropic in situ mean effective stress (p´=(σ´1+2·σ´3)/3). The result of such a comparison from the Perniö clay is shown in Figure 8.11. As can be seen, there are no differences in the undrained shear strength values. However, as is known, the peak values are reached at a smaller axial strain in compression for the anisotropically consolidated samples but there is no great difference in the stress path after the isotropic stress path past the anisotropic consolidation stress state. To generalise the results to the extension triaxial tests the same kind of comparison would be needed.
Figure 8.11 Comparison of undrained compression tests with anisotropically and isotropically consolidated samples from Perniö.
To cause a failure for the isotropically consolidated sample, the change of deviatoric stress must be greater compared to the anisotropically consolidated sample. Thus, excess pore pressure to compensate for the tendency to change volume before
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yielding is also higher, as can be seen in Figure 8.12 for the sample from Perniö at a depth of 6.3 m. This difference in excess pore pressure is a natural cause of the greater increase in deviatoric stress change. Pore pressure-based interpretation of CPTu measurements biases the interpretation if the initially anisotropic stress state is not applied.
Both means of consolidation appear to offer reasonable values for undrained shear strengths. Along with biased strain behaviour and excess pore pressure, one more disadvantage was found with the isotropic consolidation compared to the anisotropic consolidation. The isotropic test consolidated to the same mean effective stress as anisotropic consolidated test causes a slightly greater relative void ratio change at reconsolidation, as shown in Figure 8.13. Relative void ratio changes are compared to normalised undrained shear strength (8.13a) and to normalised mean effective consolidation pressure (8.13b). Figure 8.13a shows that the greater void ratio change has not caused higher undrained shear strength values as can be presumed with naturally structured samples. Due to the manually controlled regulator for cell pressure, small differences occurred between comparison tests. Therefore, the consolidated mean effective pressures normalised by the preconsolidation pressures do not give the same values between the comparison tests in Figure 8.13b. Two of the four test pairs have the same normalised value and still the relative void ratio change is higher with the isotropically consolidated samples. This normalisation means the consolidation pressure gives some estimation of how close the stress state has approached to the yield surface on the isotropic axis. Relative distance to the yield surface at different stress states is not known. One can assume that in an anisotropic stress state, this relative distance is more advantageous with an inclined yield surface compared to the isotropic stress state. In some cases, this might change the classification of the sample quality for isotropically consolidated sample to a lower class.
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Figure 8.12 Difference in excess pore pressures in CIUC and CAUC tests at depth 6.3 m.
Figure 8.13 Relative void ration change at reconsolidation compared to a.) Normalised undrained shear strength by preconsolidation pressure, b.) Normalised mean effective consolidation pressure by preconsolidation pressure.
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8.4 Normalized undrained shear strength and its relationships to the index properties
To compare undrained shear strengths at different depths, the normalisation can be used to consider different stress states either by normalising the undrained shear strength with an effective vertical stress or a preconsolidation pressure. Normalized undrained shear strength is compared to index properties to observe influences.
Only the samples identified to have “very good to excellent” quality were included in the comparison. Both samples in the strength test and CRS test at a relatively close level should fulfil this demand. Some doubt arose about extension results as compared to the other tests in Figure 7.9 and 7.12 when the results were compared to each other at depths such as 4.35 m and 7.3 m in Paimio (Fig. 7.12) and at a depth of 2.9 m in Sipoo (Fig. 7.9). These results were excluded from the comparison. These results are shown in Fig. 8.14 and marked as crossed-out.
Figure 8.14 Triaxial tests and DSS relations between a.) undrained shear strength and preconsolidation pressure, b.) normalised undrained shear strengths by vertical effective stress and over consolidation ratio.
In Figure 8.14a, undrained shear strengths defined by laboratory tests are compared with preconsolidation pressures. The triaxial compression tests have been divided into two groups based on the sampler used i.e. the TUT sampler or the mini Sherbrooke sampler. A correlation factor of the linear regression analysis corresponding to the R-squared values are shown in figures. Later, the correlation
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factor refers to the R-squared value (R2). Using R2, a value is given a proportional value depending on how well the two variables explain each other. The best correlation (R2 = 0.971) was shown by the compression test results sampled by the mini Sherbrooke at the Perniö site, which was the only site where it was used. Multiplying the preconsolidation pressure by 0.393, the undrained shear strength can be evaluated. The same values were found at all the test sites, but the samples taken with the TUT sampler were 0.359 with an R2 = 0.759. DSS test results at the Perniö, Paimio and Sipoo sites showed good agreement between shear strength and preconsolidation pressure with a value of 0.227. For the correlation, the value used was 0.854. Whereas in the extension results, the corresponding values used were 0.176 and R2 = 0.368.
SHANSEP parameters m and S determined for different laboratory tests are presented in Figure 8.14b where exponent for x corresponds to the m parameter and the factor multiplying OCR corresponds to the S parameter. For example, these parameters are dependent on the soil type, the mode of shearing, shearing type (strain rate) and the variation in the sample quality in the dataset. Based on the results in Figure 8.14b, the mComp and SComp are 0.359 and 1.318 for the compression strength corresponding sample quality of the mini Sherbrooke at the Perniö site for Perniö clay; 0.360 and 0.93 for the compression strength at all site sampled by TUT sampler; 0.243 and 0.809 for the DSS strength, and 0.206 and 0.787 for the extension strength. The SHANSEP parameters SComp and mComp can be converted to present apparent values of M and Λ. Thus, the undrained compression shear strength can be estimated when M is 1.36 and Λ is 0.93 when vertical effective stress is used. These values can be used later, when the OCR is estimated by using the su,comp value interpreted from the CPTu measurements. Apparent values can be estimated for all SHANSEP parameters if necessary. Due to the limitations of the dataset, there is some uncertainty related to the m and S values. In the dataset, the highest OCR values correspond to the samples that also have the highest plasticity index and water content value. Thus, especially the m value is highly influenced by this odd occurrence. With a larger dataset, it might be possible for the SHANSEP parameters to be defined for clay samples with the same plasticity index or water content.
For Norwegian clays, Karlsrud and Hernandez-Martinez (2013) proposed the averaged values for SComp and mComp to be 0.35 and 0.75 in the case of compression strength. From their data, they also concluded that SComp and mComp could be water content dependent. Increase in the water content could lead to higher values of SComp and mComp, which could explain why those parameters are higher in Finnish clay with
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a higher water content. Some caution is needed due to the narrow OCR data. The averaged values of SDSS and mDSS for DSS in Norwegian clays were 0.22 and 0.8, which are relatively close to the Finnish dataset.
To find a reasonable SExt and mExt parameter for the extension test results, three results needed to be omitted, as mentioned earlier. Including those results in the comparison, would make the SHANSEP parameters SExt and mExt 0.234 and 0.08 respectively. The value of m does not acquire any support from the proposed values in the literature. Therefore, these results were not used.
In Sweden, values for the SHANSEP parameters have also been proposed. For parameter m, a general value of 0.8 is recommended but it can vary, usually between 0.7 and 0.9 (Larsson et al. 2007). The magnitude of the S parameter has been proposed to be more dependent on the anisotropy strength, and as having a constant value of 0.33 in an undrained compression test. Based on empirical results, Swedish practice relies on a liquid limit dependent S parameter to estimate the DSS and extension strengths. A liquid limit change from 20 to 100, causes the SDSS to change from a value of 0.16 to 0.3. Accordingly, the change in SExt is from 0.10 to 0.290. The SHANSEP parameters from Sweden and Norway offer a guidance of where SHANSEP parameters should be for Finnish clays.
In Figure 8.15, normalised undrained shear strengths by preconsolidation pressure are compared to the index properties of the plasticity index, liquid limit, organic content, sensitivity and water content. No clear correlation can be found but a minor influence of water and organic contents on the normalised strengths of the DSS and triaxial tests can be seen. The extension test shows more correlation compared to the other tests as regards all the index properties. Due to the small number of samples and observed variation, the dataset should be updated and recalibrated in the future to increase the confidence level of the parameters.
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Figure 8.15 Relation of normalised triaxial and DSS strengths by preconsolidation to a.) the Plasticity index, b.) Liquid limit, c.) Organic content, d.) Sensitivity, e.) Water content.
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Figure 8.16 Relation of the SHANSEP S-parameter to a.) the Plasticity index, b.) Liquid limit, c.) Organic content, d.) Sensitivity, e.) Water content.
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In Figure 8.16, a similar comparison is made as in Figure 8.15, but in this figure the relationship of the SHANSEP S-parameter is considered for the case of the triaxial compression and DSS tests. Only those results worth mentioning are presented in Figure 8.16c and 8.16e where the increase of organic and water content show an increasing trend in the SComp. Some influence of the water content can also be found for SDSS.
Previously, the SHANSEP S-parameter for vane testing was defined based on old data on Finnish clays (D’Ignazio et al. 2016). The results were that ScorrFV corresponding to the corrected vane strength was 0.244 for Finnish clays when mcorrFV was 0.763 and when Equation 5.1 was used for correction. Preconsolidation stresses have been defined with both constant-rate-strain (CRS) and incremental loading (IL) oedometer tests. For normalising, IL tests were transformed to correspond to CRS preconsolidation by increasing the IL preconsolidation by 27 % for 162 data points out of 216. Twenty-seven percent is the average value defined by Länsivaara (1999) according to Leroueil’s data of ratios between IL and CRS preconsolidation stresses. ScorrFV was not found to have any significant dependence on the plasticity index (Ip), liquid limit (wL), water content (w)or liquidity index (LI).
Similar kinds of comparisons were repeated with the collected field vane data in this study, including comparisons with the index properties and normalised strengths, as shown in Figure 8.17. Comparisons were made for measured and corrected values. The correction factor was obtained by Eq. 5.1 based on the liquid limit. The correction factor limitation was a maximum value of 1. Vane strengths were selected to include only the best of three test results at the same depth and at depths where a high-quality CRS test was available. Additionally, it was assumed that the ratio of the measured vane strength and preconsolidation pressure was over 0.2. Lower values have been omitted from Figure 8.17. The SHANSEP correlation gives reasonable values compared to those values proposed by D´Ignazio et al. 2018. As this previous data has a larger range of OCR in its dataset, it is likely to be less biased by variation in the data and thus have a higher confidence level. In Figure 8.18 a comparison between the estimation using D´Ignazio et al. 2018 and the corrected field vane results are shown. The prediction gives a good approximation of the values even though a small overestimation might be detected due to higher SHANSEP parameters.
Some influences of the index properties can be seen in Figure 8.17 regarding normalised vane strengths. The plasticity index especially, shows an influence on the
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measured vane strength. As the plasticity index essentially correlates with the other index properties in the dataset, the trends were found in the other index properties too. Correcting the measured vane strength by using the function of the liquid limit, the correlation to liquid limit and the plasticity index gave poorer results. Mesri (1975, 1989) proposed that the ratio of the corrected vane strength and preconsolidation pressure is approximately 0.22 when the OCR is lower than 2. In Figure 8.17c and 8.17d, this ratio is lower when e.g. the liquid limit is over 50. Due to variation in the vane test, it might be that is just coincidence. When updating the dataset in the future, this point of view would be good to consider if some overcorrection is necessary.
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Figure 8.17 Relation of the normalised measures and corrected field vane strengths by preconsolidation pressure to a.) OCR, b.) Organic content, c.) Plasticity index, d.) Liquid limit, e.) Water content, f.) Sensitivity.
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Figure 8.18 Interpreted corrected vane strength by SHANSEP equation with previously defined parameters compared to defined data in this study.
8.5 Anisotropy on Finnish soft soils and relationship to index properties
Instead of conducting a series of extension and compression tests for every practical design case, it was considered desirable to try to find a correlation that would assess anisotropy strength e.g. via the index properties. For decades the anisotropy ratio has been proposed as a ratio correlating with the plasticity index (Bjerrum 1973, Ladd 1977, Jamiolkowski et al. 1985) or liquid limit (Larsson 1980). The global correlations seem not to be valid (Won 2013). Won (2013) collected a wide dataset of natural samples of well-defined anisotropic consolidated triaxial test results and did not find any clear trend. In some regions, like in Scandinavia, local trends have been proposed. Karlsrud and Hernandez-Martinez (2013) compared Norwegian data from Scandinavia which also included Bjerrum’s results for leached sensitive clays. Their conclusion was that the correlation did not work well especially when the data includes non-leached clay samples with normal sensitivity. A better correlation was found for water content and OCR.
In general, the anisotropy ratio has been defined by using two methods: comparing peak strengths together or selecting the extension strength at the same
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shear deformation level where the compression peak strength occurred. The second method leads to smaller extension strength causing higher anisotropy. The first method is used in this study because of the isotropic consolidation biased initial deformation behaviour compared to the anisotropic stress state occurring in the ground.
An anisotropy ratio based on the ratio of the peak values of the extension and compression tests was calculated and compared to the index properties. To summarise the results shown in Figure 8.19, some influence could be found for the plasticity index due to the possible greater influence of the plasticity index on the extension test results. It might be that the extension test is more pronounced for shear hardening, when the yield surface is more rounded, or the results are just too influenced by the two results having a plasticity index close to 60. The average anisotropy ratio for Finnish soft clays is 0.615 based on the result with peak strength values. The anisotropy ratios varied between 0.52 – 0.70.
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Figure 8.19 Anisotropy ratio between maximum undrained extension and compression strengths compared to a.) Plasticity index, b) Sensitivity, c.) Water content, d.) OCR, e.) Organic content.
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Figure 8.20 Ratio between maximum undrained DSS and compression strengths compared to a.) the Plasticity index, b) Sensitivity, c.) Water content, d.) OCR, e.) Organic content.
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The same kind of comparison was conducted in Figure 8.20 for the ratio between the peak values of the DSS and compression tests. The averaged value for the dataset was 0.64 while the ratios varied between 0.56 – 0.72. While the anisotropy ratio from the triaxial tests showed the highest influence with the plasticity index, the water content seemed to have the greatest influence on the ratio DSS and compression tests. Some influence was also found with the plasticity index and organic content.
For example, in Sweden, the same anisotropic ratio values for the same range in the plasticity index (20…60) are 0.45…0.75 for suE/suC and 0.6…0.8 for suD/suC based on the empirical results (Larsson 1980, Larsson et al. 2007). A higher dependency on the plasticity index has been proposed for Swedish inorganic clays. In Norway, the anisotropic ratios have been proposed to correlate with the water content (Karlsrud and Hernandez-Martinez 2013). In the range between 70…120 % the corresponding water content in the Finnish dataset, would be values of 0.48…0.63 for suE/suC and 0.77…0.99 for suD/suC with the empirical Norwegian equations. At corresponding liquid limit and water content ranges with the Finnish dataset, the averaged anisotropic ratios for Swedish clays would be 0.6 for suE/suC and 0.7 for suD/suC and for Norwegian clays 0.55 for suE/suC and 0.88 for suD/suC. suD/suC ; these ratios are higher compared to the Finnish dataset. A more consistent ratio is found with Swedish clay. The Swedish dataset has more similarities regarding the index properties than Norwegian clays which may partly explain the closer result. The average values for the suE/suC ratio are very close to those of Swedish and Finnish clays (0.6 vs. 0.615). Compared to Norwegian clays, a lower anisotropy ratio is found for suE/suC , whereas the ratio for suD/suC gives a reverse outcome.
8.6 Effective strength failure criteria
To define the effective friction angle, it is recommended that a set of three triaxial tests are performed. All these tests should be consolidated beyond the initial yield surface to different stress states and then sheared simultaneously recording the pore pressure to failure in an undrained condition Alternatively, drained triaxial tests may be performed; these take a much longer time but are often considered more reliable. For Swedish clays, Larsson et al. (2010) mentioned that the typical effective friction angle is 30°. According to Larsson (1981), for normally consolidated clays with a brittle structure only a part of this shear strength can be mobilised (Larsson 1981). This partial mobilisation of the shear strength may be explained by the instability of
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soil structure causing a reduction in the shear resistance due to induced pore pressure.
The instability of soil has been studied in clays and sands by many authors e.g. Tavenas and Leroueil (1977) and Lade (1992). According to the model of Tavenas and Leroueil (1977) which is based on the behaviour of Champlain clays, there is a stable yield surface where the soil stays in a stable state even when the deviatoric stress is above the instability line. In a stress state beyond this stable yield surface and above the K0 line, instability behaviour can occur. Lade (1992) explained the behaviour of instability by non-associated plasticity. He performed a series of conventional drained and undrained triaxial tests to define the regions of stable and instable behaviour. He found that in the undrained tests instability can be induced anywhere between the (σ1 – σ3)max and (σ´1 / σ´3)max as shown in Figure 8.21 where the instability line is drawn through (σ1 – σ3)max values. In drained tests, these two conditions are reached simultaneously. As the triaxial tests done for this study were monotonic loading to failure, the triggering mechanism to cause failure differed from Lade’s tests conducted under static loading. Thus, some caution should be kept in mind as similar tests were not conducted.
Lade (1992) wrote, about these two failure criteria, that they are both commonly used as criteria for failures: 1.) (σ1 – σ3)max 2.) (σ´1 / σ´3)max where (σ´1 / σ´3)max describes the true failure surface. Herein, (σ´1 / σ´3)max is used as the ultimate failure criteria. It should be noted that in a stability calculation, it may be more appropriate to use a failure criteria of (σ1 – σ3)max to define the shear strength if material model cannot consider the behaviour of an unstable microstructure where in some cases the instability of the structure is reached before the ultimate failure state. To reach an ultimate failure state, post-peak reduction should occur.
For CPTu interpretation, there is also an application, proposed by Agaiby and Mayne (2017) where the friction angles are defined by both of these failure criteria. This method was mentioned in Chapter 4.6 and there these friction angles were symbolled as φ'qmax for (σ1 – σ3)max and φ'MO for (σ´1 / σ´3)max. The same symbols are now used here. Use of the two friction angles for interpretation is explained by the fact that the method by which the incompatibility of the strain levels where the maximum values of deviatoric stress and excess pore pressure occur in the triaxial compression tests (Agaiby&Mayne 2017).
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Figure 8.21 Schematic diagram of the location of the instability line in the p´- q diagram (Lade 2002).
In this study, the focus is on the determination of undrained shear strength. Therefore, most of the triaxial tests were consolidated to a stress state close to in situ, situated inside the initial yield surface. For the clays with the highest OCR in this study, failure may be reached in the vicinity of the top of the initial yield surface. Then, even the maximum obliquity friction angle may be overestimated. Only some test series with samples consolidated beyond the initial surface were conducted. From these results, both failure criteria can be assessed. Some uncertainty may arise due to the higher strains needed to reach the (σ´1 / σ´3)max state, as the correct area correction is then more important. All the results are presented below and both failure criteria are drawn when possible. (σ´1 / σ´3)max is found from the tail of the stress path. From the DSS tests, the failure lines were also matched but instead of using principal stresses, shear stress and vertical stress were used for estimating φ'MO = (τ / σ´v)max and φ'qmax = (τmax). The horizontal plane was assumed to correspond to a plane of maximum stress obliquity in the case of φ'MO and a plane of maximum shear for φ'qmax. Thus, the friction angles were determined by equations tan(φ'MO) =τ / σ´v and sin(φ'qmax) =τmax / σ´v.
The OC triaxial tests, DSS tests and OC tests together with the NC tests close to the same level are shown in Figures 8.22 – 8.26. Additionally, in Figure 8.22c a triaxial data with maximum obliquity and shear stress lines utilising data from Lehtonen´s thesis (2015) is shown. The test results presented in Lehtonen’s thesis have been sampled at a distance of 10 m from the data presented in here. According to the OC test results, the maximum stress obliquity lines show friction angle values in a range from 32.3° to 35.5° while the cohesion varies between 2.5 and 3.4 kPa. Use of cohesion may not be consistent with the critical state concept but adding it to the failure criteria seems to decrease the variation in the failure criteria. In the NC tests,
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the friction angle based on σ´1 / σ´3max lines were found to be about 2 – 4 degrees lower whereas the friction angles defined from the maximum shear stresses in the NC tests give friction angles values between 22° – 25°. The cohesion values were kept the same in all failure criterions.
For the DSS results, the maximum stress obliquity and maximum shear stress lines were also defined and presented in Figures 8.22 – 8.26. Thus, the maximum shear stress lines with zero cohesion give friction angle values between 24° – 25° while the maximum obliquity values are between 27.5° – 30°.
It should be noted that in the figures only one failure line for each criterion is presented. A better estimation of failure might be done using different criteria for each soil layer. Due to the low number of tests, these general lines are presented.
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Figure 8.22 From the Perniö site, a.) φ'MO line from OC triaxial stress paths, b.) φ'MO and φ'qmax lines from DSS tests c.) φ'MO and φ'qmax lines from OC and NC triaxial stress paths based on Lehtonen’s data (2015).
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Figure 8.23 From the Sipoo site, a.) φ'MO line from OC triaxial stress paths, b.) φ'MO and φ'qmax lines from DSS tests c.) φ'MO and φ'qmax lines from NC triaxial stress paths presented together with OC test.
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Figure 8.24 From the Paimio site, a.) φ'MO line from OC triaxial stress paths, b.) φ'MO and φ'qmax lines from DSS tests c.) φ'MO and φ'qmax lines from NC triaxial stress paths presented together with OC tests
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Figure 8.25 From the Masku site, a.) φ'MO line from OC triaxial stress paths, b.) φ'MO and φ'qmax lines from NC triaxial stress paths presented together with OC test.
Figure 8.26 From the Lempäälä site, φ'MO line from OC triaxial stress paths.
8.7 Stress-strain behaviour
In the penetration of the CPTu cone into the soil, large strains in the surrounding soil will occur. Thus, some of the measured parameters such as pore pressure and sleeve friction are highly dependent on the strength behaviour after peak strength in sensitive soils. Deformation before and after peak strength are illustrated next by stress-strain curves with and without normalisation.
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8.7.1 Field vane stress-rotation curves
In Figure 8.27, measured values using the field vane are presented. In Figure 8.28, normalisations using peak shear stress values are shown. During insertion of the vane at the rotation level, some shear stress was mobilized while the vane rotation during insertion was fixed. In Figures 8.27 and 8.28, the curves have been approximated to zero by linear lines from the first measured values. With normalisation, the relative behaviours of the tests were much closer to each other.
In all cases, failure was reached after a rotation of 3 to 6 degrees when the curves were approximated to begin from zero. Rotation to failure increases slightly as the field vane strength increases. Only the most reliable results were taken for the comparison. Thus, with the vane tests showing a greater disturbance level, the highest peaks were not achieved, and this may have increased the rotation to failure. At a rotation of 45 degrees, a stress reduction of 45 % - 55 % has occurred.
Figure 8.27 Curves of rotation versus measured shear stress using the field vane
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Figure 8.28 Curves of rotation versus normalised measured shear stress according to the maximum value in the vane results.
8.7.2 DSS Stress-strain curves
A similar comparison was conducted for the DSS tests as for the vane tests. The results are presented in Figures 8.29 and 8.30. With the vane test, the estimation of shear strain is impossible as the thickness of the shear rupture is unknown. In the DSS test, an estimation of the shear strain can be done more accurately by assuming that the horizontal movement causes constant shear strain on the whole height of the sample. Some uncertainty is related to the correction used for correcting the influence of friction between the copper rings. It is not possible to consider the real interaction of the soil sample and rings during testing and its influence on ring friction without real measurements concerning the ring frictions. However, the DSS results show that failures occurred at between 3 % and 4 % of shear strain in all the samples. Shear stresses after 10 % shear strain are about 65 % to 90 % of the maximum.
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Figure 8.29 Curves of shear strain versus shear stress in the DSS results.
Figure 8.30 Curves of shear strain versus normalised shear stress according to the peak value in the DSS results.
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8.7.3 Triaxial compression stress-strain curves
In the previous chapters, the triaxial compression test results are presented in Figures 8.31 and 8.32 and the isotropically and anisotropically consolidated tests are presented in the same figures. Partly because of this a higher variation in failure axial strains can be noticed as the failure strains are between 0.7 % and 2.5 %. The shear stress values after 6.7 % of the axial strain corresponding close to 10 % of the shear strain are between 20 % and 50 % lower than the peak values. The consolidation procedure has some influence on this comparison but also, a trend can be seen showing that when the failure strain is higher the strength reduction is smaller. This might be due to sample disturbance or due to the natural behaviour of soil. In soft soils with high sensitivity, two failure modes can occur depending on the sample quality. For high quality samples, the shear deformation can localise on a thinner shear zone whereas the failure in poorer quality samples can occur on thicker zone. Thus, more localised shear deformation will cause more influence on the reorientation of the soil structure increasing excess pore pressure (Thakur 2007). The failure mode might be dependent on small inhomogeneities in the sample. In Figure 8.33 three dried samples after triaxial compression tests with high quality samples from Perniö are presented. The sample in the middle and right are closer to the same shape as after the shear tests, this clearly shows where the failure zones are localised. The shape of the sample on the left does not show an outward clear shear zone, as some barrelling has occurred. An assumption about the cylindrical shape change during shearing is used. With localised behaviour, where the top part of the sample keeps its shape, the real shear surface will be smaller compared to the shear surface formed in the fully cylindrically shaped sample. Thus, some underestimation of shear stress may have occurred as the area correction is based on the assumption of a perfectly cylindrical shape.
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Figure 8.31 Curves of axial strain versus deviatoric stress in the CIUC and CAUC results.
Figure 8.32 Curves of axial strain versus normalised deviatoric stress according to the peak value in the CIUC and CAUC results.
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Figure 8.33 Slightly different formed triaxial compression test samples of Perniö clay after drying.
8.7.4 Triaxial extension stress-strain curves
Triaxial extension stress-strain curves are presented in Figures 8.34 and 8.35. In the extension tests, all the tests followed a similar pattern until the peak strength was reached. Afterwards the behaviour between different tests started to deviate more from each other. This might have been caused by biased membrane corrections due to the membrane sliding. Thus, the assumed extension stress caused by the membrane may have decreased and then overcorrection has occurred.
Strains to failure are close to the values with the compression test when consolidation is made isotropically as can be seen in Figures 8.34 and 8.35.
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Figure 8.34 Curves of axial strain versus normalized deviatoric stress in CIUE results.
Figure 8.35 Curves of axial strain versus normalised deviatoric stress in the CIUE results.
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8.8 Stress paths of triaxial tests
All OC stress paths for the triaxial tests are presented in Figure 8.36 including estimations of the effective strength parameters based on the OC tests. The compression strengths can be estimated by varying the cohesion term between 2.5…5 kPa. In the extension tests, most of the extension results seem to follow the failure line without cohesion. Due to uncertainties, the tails of the extension tests are not reliable which may have an influence on the visibility of the cohesion.
Figure 8.36 Stress paths for all the triaxial tests.
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Figure 8.37 Normalised stress paths according to the preconsolidation pressure for all the triaxial tests.
In Figure 8.37, normalisation of the stress paths is made by preconsolidation pressure to illustrate how the consolidation stress states relate to the yield surface. Smaller values for the normalised mean effective the stresses are the result of the states where the distances to the yield surface are greater. Similarities in the stress paths can be seen. Some variations in the normalized undrained shear strengths are mainly due to influence by OCR. As the OCR increases, the normalised deviatoric stress ratio decreases.
-0,6
-0,4
-0,2
0
0,2
0,4
0,6
0,8
1
0 0,1 0,2 0,3 0,4 0,5 0,6 0,7q /
´ c, [
-]
p' / σ'c , [-]
Perniö [3.05 m]Perniö [3.05 m]Perniö [3.25 m]Perniö [3.25 m]Perniö [3.25 m]Perniö [3.25 m]Perniö [4.0 m]Perniö [4.0 m]Perniö [4.7 m]Perniö [4.7 m]Perniö [6.3 m]Perniö [6.3 m]Perniö [6.3 m]Perniö [6.3 m]Perniö [6.7 m]Perniö [6.7 m]Perniö [7.55 m]Perniö [7.55 m]Masku [3.05 m]Masku [3.05 m]Masku [5.05 m]Masku [5.05 m]Paimio [3.35 m]Paimio [3.35 m]Paimio [4.35 m]Paimio [6.35 m]Paimio [6.35 m]Paimio [7.3 m]Paimio [8.6 m]Paimio [8.6 m]Sipoo [2.9 m]Sipoo [2.9 m]Sipoo [5.0 m]Sipoo [5.0 m]Sipoo [6.1 m]Sipoo [8.6 m]Sipoo [8.6 m]Lempäälä [7.1 m]Lempäälä [7.1 m]
φ′ = 32°�� = 0.025 … 0.1 · !��
φ′ = 32°�� = 0 … 0.05 ¦ !��
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8.9 Oedometer results
In Figure 8.38 some typical oedometer curves from the high-quality samples are presented. Oedometer curves normalised by preconsolidation pressure are shown in Figure 8.39. Some trends were found from the figures. Clays having the highest plasticity indexes are the curves with the highest initial void ratios. Those clays having the greatest compressibility can be seen in the change of void ratios to a 300 kPa loading; fewer plastic clays were found with lower initial void ratios.
The relationship between the plasticity index and compressibility was found by Skempton and Jones (1944) with remoulded samples. They proposed a relationship between the compression index and liquid limit where the compression index increases as the liquid limit increases. For Finnish muds and clays, Helenelund (1951) proposed an equation to estimate the compression index according to the water content. Accordingly, a higher water content indicated a higher compression index. Both these conclusions can be found in this study, and the water content and plasticity index also correlate. Even after a loading of 300 kPa, the samples with higher plasticity indexes have a higher void ratio compared to the low plastic samples. Thus, higher plasticity gives extra resistance against volume change by means of its structure as it can be in a higher stress state with a higher void ratio. Possibly the curves will begin to approach each other if the tests are continued in an extreme loading of over 300 kPa.
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Figure 8.38 Oedometer curves vertical stress on the x-axis and the void ratio on the y-axis.
Figure 8.39 Oedometer curves normalised by means of preconsolidation pressure.
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9 INTERPRETATION OF THE CPTU MEASUREMENTS
In the previous chapter, Chapter 8, soil behaviour in laboratory tests were discussed, as well as how the results correlated with the index properties. Now, the previous tests results including the OCR and strength results are compared to the CPTu results. Index properties are used to check whether the correlations are dependent on these. Correlations are evaluated by using the statistical values of relative standard error (RSE) and the square of the correlation (R2).
Undrained failure showed strain softening behaviour after peak shear stress and the undrained strength is anisotropic by nature. Strain softening behaviour and anisotropic strength might cause some problems concerning theoretical interpretations owing to the initial assumption that soil perfectly follows isotropic elastic behaviour.
Possibly, the influence of the simplifications can be illustrated by the SCE-CSSM method originally proposed for ideal soil. A problem may arise when the effective stress-based theory is adopted where pore pressure distribution and softening after the peak strength is not considered. The initial anisotropy and anisotropic strength have also been simplified. Pore pressure distribution is highly affected by the location where the pore pressure is measured due to location dependent loadings and strains for the surrounded soil. The SCE-CSSM method can be used with the general values (φ´=30 , Ir = 100, Λ = 1) for first-order approximations as Mayne et al. (2009) have proposed (equations 4.21 – 4.23). When ignoring stress rotation especially, the anisotropy and strains in soil with a high strain softening behaviour, the interpretation of OCR with net cone resistance, the excess pore pressure and effective cone resistance will lead to the order shown in Figures 9.1 and 9.2. These figures show that the excess pore pressure-based interpretation gives the highest OCR. In Figures 9.1 and 9.2, all the OCR results are included without consideration of the sample qualities. Including other influences together with strain softening such as the cone roughness, the anisotropic initial stress state and anisotropic strength requires too many necessary initial assumptions or other test results to obtain
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reasonable results. Using two of three variables of the friction angle (φ´), the rigidity index (Ir) and the plastic volumetric strain ratio (Λ) in SCE-CSSM as operational values, interpretations can be made at least for the net cone resistance. Then the real meaning of these parameter could be lost. A simple way is to define empirical correlation for an estimation in a similar condition and finally, some discussion could be considered as to it being the result of sensitive soil behaviour.
Figure 9.1 Simplified interpretation using ideal soil parameters at the Perniö, Sipoo and Lempäälä sites.
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Figure 9.2 Simplified interpretation using ideal soil parameters at the Masku and Paimio sites.
9.1 Estimation of the over consolidation ratio by means of the index properties
To keep the calibration test results as comparable as possible, only high-quality samples taken with the TUT sampler were adopted. A uniform dataset is easier to adjust later if necessary.
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It is known that accurate measurement of properties for thin layers surrounded by significantly different deformation and strength properties can be difficult for CPTu as the surrounding layers influence the values by averaging the measured resistances. Because of this, an attempt was made to have the samples in thick enough homogenous layers so that the CPTu measurements could be assumed to present true resistances for individual layers. Due to layering at a depth of 7.1 m at the Lempäälä site, the demand for a homogenous soil condition cannot be assumed to be valid and thus, this depth has been left out of the comparison. In Figure 7.16, this small layer causes the peak values for the cone resistance and the drops in pore pressure. The accepted quantity of samples was then limited to 17.
The simplest correlation for OCR can be found by comparing the measured OCR values with the normalised net cone resistance, excess pore pressure and effective cone resistance. At this point, normalisation was done by using effective vertical stress. In Equations 9.1 – 9.3 and in Figure 9.3, the relationships are demonstrated. Based on the dataset, the mean values for the cone factors, in the OCR interpretation, are 0.29 for Nkt,OCR, 0.37 for NΔu,OCR and 0.65 for Nke,OCR. Note that the mean values are not same as the constants in Equations 9.1 – 9.3, which are the trendlines based on the data. Relative standard errors (RSE) may be utilised assuming that the oedometer results present true values. Thus, relative standard errors for Equations 9.1 – 9.3 are 6.8 % (9.1), 6.3 % (9.2) and 8.3 % (9.3). RSE values are also shown in Figure 9.4 and Table 4. These results will be the basis for later comparisons when improvements in the predictions are considered.
��� = 0.173 ∙ ���@A�@A� + 0.533 (9.1)
��� = 0.370 ∙ �F�A�@A� + 0.013 (9.2)
��� = 0.271 ∙ ���F�@A� + 0.779 (9.3)
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Figure 9.3 Relationship between OCR and normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
The most accurate estimation can be done with Equation 9.2 in similar soil conditions as in the dataset. A similar accuracy can be achieved with a net cone resistance-based interpretation whereas the interpretation based on the effective cone resistance shows more variation.
To investigate possibilities of improving the interpretation with the index properties, the cone factors have been solved by using Equations 9.4 – 9.6. Due to the narrow range of OCR in the dataset, the plastic volumetric strain ratio (Λ) is assumed to be 1. Possible error in the adopted value of Λ may have an influence on the cone factors too.
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\y[,345 = k = 345¨�© @A
@A� (9.4)
\∆�,345 = 345�nF©nA
@ª� � (9.5)
\y*,345 = 345¨�©nF
@A� (9.6)
Figure 9.4 Trends and relative standard errors (RSE) of the measured OCR values. Equations a.) 9.1, b) 9.2, c.) 9.3 have been used to interpret the OCR.
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Cone factors solved with Equations 9.4 -9.6 were compared to the plasticity index, water content, organic content and sensitivity. Comparisons and equations for the cone factors as a function of the index property can be found in Figure 9.5. Additionally, relative standard errors are presented in Table 4. These RSE values show that prediction of the OCR will not improve while the index property is coupled with cone factor. Some improvement was shown in the excess pore pressure-based interpretations when the plasticity index, organic content or sensitivity were added to the cone factor. However, the improvements are not significant.
Figure 9.5 Relation of the cone factors for OCR based on the net cone resistance, excess pore pressure and effective cone resistance compared to a.) the Plasticity index b.) Water content c.) Organic content d.) Sensitivity.
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9.1.1 Improvement of correlation using other variables
Lunne et al. (1985) proposed that effective cone interpretation to define undrained shear strength correlates with the pore pressure parameter, Bq. This observation was also found to be true for the Finnish dataset in the case of effective cone resistance (Selänpää et al. 2018). The correlations between Bq and all the cone factors are shown in Figure 9.6. The RSE values were 6.3 % for the net cone resistance, 13.6 % for excess pore pressure and 7.4 % for effective cone resistance. An improvement in the effective cone resistance-based interpretation can be noticed. The disadvantage of using effective cone resistance for the interpretation is that the interpretation is very dependent on the accuracy of both the measurement of the tip resistance and excess pore pressure. The presented correlation for the effective cone factor given in Figure 9.6, is valid for Bq ranging from 0.65 to 0.9.
Figure 9.6 Dependency of Bq on the cone factors in the OCR interpretations.
9.1.2 Consideration of the influence of the rigidity index on the cone factor of the net cone resistance
In the cavity expansion theory, the rigidity index is a key parameter together with the undrained shear strength in the analysis of the penetration problem. The rigidity index is dependent on which stress level secant shear modulus is defined. As
169
mentioned in Chapter 4.7. some uncertainty is related to the level at which it should be defined for the CPTu analysis. Thus, it may be more suitable to present the rigidity index in a similar way as the modulus reduction curves are often presented.
To define the rigidity index, the shear modulus should be measured or assessed. To measure shear modulus by triaxial device, extra accuracy is needed in small strain measurements. Normal triaxial devices were used for triaxial tests so the rigidity indexes are now assessed by some assumptions and correlations.
Langø (1991) proposed for Norwegian clays that Gmax can be evaluated by using equation 9.7 where normalized shear modulus gmax is mainly dependent of water content. In Equation 9.7, n is 1 for clays. A later Langø’s dataset was updated which had similar findings (Long and Donohue 2010). Similar results from Sweden were found by Touiti and Van Impe (2017) when they reanalysed the results of Larsson and Mulabdic (1991). Additionally, a good correlation was found by Mäenpää (2016) for Finnish clay by using gmax =20 000 % / w % as Länsivaara (2005) has suggested.
«� = }�,� ∙ 100:� ∙ (!� − �)� (9.7)
By Equation 9.7, the maximum shear modulus can be estimated as gmax is defined by the correlation of 20 000 % / w %. Due to lack of information about friction angles needed for the estimation of the mean effective stress together with the OCR , the friction angles for K0,OC estimations are done by two ways: 1.) using constant friction angle value of 30 , 2.) by correlation ����´ = 0.35 − 0.1 ∙ $�up as Wood (1990) has proposed. The friction angles together with measured OCR values are assigned to equation ¬�,34 = 81 − ����´9 ∙ ���-o�d´ proposed by Mayne and Kulhawy (1982). As the key in CPTu interpretation is the rigidity index, estimated Gmax values are divided by the measured undrained compression shear strength, which results in a maximum rigidity index for the compression loading. At first, all maximum rigidity indexes are estimated based on the data of water contents and OCR values. For friction angles, the constant value of 30 is used. These maximum rigidity index results are compared to the plasticity indexes and the water contents as shown in Figure 9.7; this naturally shows some correlation for both index properties as the water content was used to estimate the maximum shear modulus.
The most recommended shear modulus to define rigidity index is the secant modulus at 50% of the stress to strength as mentioned in Chapter 4.7. Often modulus reduction curves are approximated with the hyperbolic stress-strain
170
relationship proposed by Kondner (1963). This relationship is shown in Equation 9.8. The reference shear strain can be estimated by using Equation 9.9 (Hardin&Drnevich 1972b) and/or by Equation 9.10 (Länsivaara 1999). Instead of using τmax, a more appropriate stress value may be the shear stress change from the initial shear stress state to failure in Equation 9.9 as Länsivaara (1999) proposed based on Kondner’s study (1963). Länsivaara’s suggestion (Equation 9.10) is used later. For τmax, the values measured by undrained compression triaxial test are used. To define an initial shear stress state, an initial deviatoric stress (q0) state should be estimated.
Some test data has been selected for the further analysis to consider the rigidity index reduction curves. These selected data are from the Masku results at a depth of 5.0 m, from Sipoo at a depth of 2.9 m, from Paimio at depth of 3.3 m and 6.3 m and from Perniö at a depth of 3.2 m. The results have been selected to have a high range of water content and plasticity indexes. Water content varies between 67 and 117 and accordingly the plasticity index between 15 and 59. Summary of the data are shown in Tables 2 and 3. Table 2 is for constant friction angle of 30 and Table 3 for friction angles estimated by the correlation to plasticity indexes as proposed by Wood (1990). By the friction angles and OCR’s, initial stress states are estimated.
In Figures 9.8a and b, the rigidity index reduction curves are shown. In Figure 9.8a, the constant friction angle is used and in Figure 9.8b, the friction angles are estimated by plasticity indexes. The difference of the rigidity indexes between samples is the highest at low strain levels. The level of the strain used for defining the secant modulus is not clear in the penetration problem. However, the results are shown that when the strain increases the differences in rigidity indexes decreases. At higher shear strain levels, some curves have intersected.
In Equation 4.21, the increase of the rigidity index is decreasing the cone factor of OCR as the cone factor is consisting of M, Ir and partly of Λ. This influence of the rigidity index on the cone factor is illustrated by that same Equation 4.21 using 30˚ for the friction angle and 1 for the plastic volumetric strain ratio (Λ). The rigidity indexes are obtained from Figures 9a and b. The results are shown in Figures 9.8c and d, where the assessed cone factors are compared with the normalized shear stress values. As some curves in Figure 9.8a and 9.8b were intersected, same intersections are difficult to see in Figures 9.8c and 9.8d.
In the data, the range of Nkt,OCR varies between 0.25 and 0.34 (Fig. 9.5), showing a difference of 0.09. Value of 0.25 was for clay owing higher plasticity, and value of
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0.34 for clay owing lower plasticity. Same range of Nkt,OCR is found from Figure 9.8c and 9.8d when the normalized shear stress value is under 0.6. Additionally, that corresponds closely the shear stress level, which is the most recommended level for the selection of the rigidity index. Unfortunately, the results are in different order as they should be. In Figures 9.8c and d, the clays owing higher plasticity are getting higher Nkt,OCR values as in the dataset, the lowest Nkt,OCR values are for clays owing the highest plasticity.
The validity of curves is not certain – a lot of uncertainty due to many assumptions. Some shaping of rigidity index curves can be done by increasing curvatures according to plasticity, but that behaviour should be first verified by tests. Also, the curves may be dependent on many factors e.g. viscous effects and the stress level for selecting rigidity index might not be the same for all soils.
Table 2 Summary of selected and estimated values from five test sites using constant friction angle Ip,
[%] w, [%]
σ'm, [kPa]
σ'v0, [kPa]
OCR, [-]
φ´, [-]
K0,OC, [-]
Gmax, [kPa]
τmax, [kPa]
q0 [kPa]
Δτmax/Gmax, [-]
Masku [5m] 59 117 28,9 39,4 1,45 30 0,60 4948 20,6 15,7 0,0026
Sipoo [2,9m] 43 90 21,3 27,3 1,8 30 0,67 4753 17,8 9,0 0,0028
Paimio [6,3m] 36 108 34,4 49,8 1,15 30 0,54 6347 22,3 23,1 0,0017
Perniö [3,2m] 30 94 19,9 28,0 1,28 30 0,57 4255 13,7 12,2 0,0018
Paimio [3,3m] 15 67 25,7 34,7 1,5 30 0,61 7659 16,8 13,4 0,0013
Table 3 Summary of selected and estimated values from five test sites using correlation proposed by Wood (1990) for friction angle estimation
Ip, [%]
w, [%]
σ'm, [kPa]
σ'v0, [kPa]
OCR, [-]
φ´, [-]
K0,OC, [-]
Gmax, [kPa]
τmax, [kPa]
q0 [kPa]
Δτmax/Gmax,
[-]
Masku [5m] 59 117 31,4 39,4 1,45 23,8 0,69 5360 20,6 12,1 0,0027
Sipoo 43 90 22,3 27,3 1,8 25,7 0,73 4994 17,8 7,4 0,0028
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[2,9m]
Paimio [6,3m] 36 108 36,0 49,8 1,15 26,9 0,58 6637 22,3 20,7 0,0018
Perniö [3,2m] 30 94 20,4 28,0 1,28 28,1 0,59 4371 13,7 11,3 0,0018
Paimio [3,3m] 15 67 24,8 34,7 1,5 32,7 0,57 7387 16,8 14,8 0,0013
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Figure 9.7 Maximum rigidity index defined with an estimated Gmax and the measured undrained compression strength compared to a.) the Plasticity index, b.) Water content.
Q��®
= ::? ¯
¯j�> (9.8)
`/* = ±��®��®
(9.9)
`/* = ²±��®��® (9.10)
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Figure 9.8 Illustration of the rigidity index reduction curves when the reference strain in Kondner’s hyperbolic stress-strain relationship is estimated from ratio of Δτmax/Gmax a.) where a constant friction angle is used b.) where friction angles are estimated via plasticity indexes c.) Illustration of the influence of the rigidity index reductions for cone factors in Equation 4.21 when using rigidity index from Figure a d.) Illustration of the influence of the rigidity index reductions for cone factors in Equation 4.21 when using rigidity index from Figure b.
9.1.3 Consideration of influence of stress exponent
The initial stress state increases with depth and the influence of this increasing stress state on certain parameter could be linear or non-linear. Non-linearity can be considered according to the stress exponent and addressed directly to the stress state or to the function of the stress state e.g. the net cone resistance. In the case of a linear relationship, the stress exponent would be 1.
To consider the stress state influence on OCR, the stress exponent of the normalised net cone resistance may be considered when normalisation is done by
175
means of vertical effective stress in clays. Based on the large database of clay data, Kulhawy and Mayne (1990) presented a linear correlation between the normalised cone resistance (Qt) and an OCR with a constant multiplier (k) of 0.32 for Qt. Later, Chen and Mayne (1996) proposed similar results where the exponent is 1 for Qt in intact clays.
Figure 9.9 General relationship of the net cone resistance to yield stress in various soils (Agaiby 2018, modified in accordance with Mayne 2013).
For a larger variation of soils, the stress exponent is presented in Figure 9.9 based on Mayne’s data (2013) later modified by Agaiby (2018). The stress exponent (m´) varies between 0.7 – 1.1 from clean sand to fissured clays while the constant factor of 0.33 (k) is used for yield stress estimation. For sensitive clays, stress exponent differs from value proposed for intact clays. For sensitive clays, a stress exponent of
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0.9 is proposed while intact clays have a value of 1. The general trend for the stress exponent (m´) is that it increases with the fines content and decreases with the mean grain size.
Multiplier (k) assumed to be 0.33 is shown to vary between 0.1 – 0.5, however, 0.33 is close to the average for clays. The multiplier seems to decrease as both the plasticity index and void ratio increase (Mayne et al. 1998). The factor k depends on both the soil and CPTu cone properties. In order to remember the influences, the most important factors as listed by Schneider et al. (2008) are: 1.) soil stiffness or rigidity index; 2.) cone and sleeve roughness; 3.) ratio of horizontal to vertical stress; 4.) strength anisotropy; 5.) soil sensitivity; 6.) degree of pore pressure dissipation during penetration; and 7.) viscous rate effects. However, some benefits can be found when the stress exponent alone is kept as a variable. Thus, classification can be based on one variable and much work has been done on identifying the stress exponent from the SBT charts. From the SBT charts, the material index (Ic) value can be obtained and with that, the stress exponent can be assessed (Mayne 2017). In the case of sensitive soils, sensitivity itself is a key factor estimated by CPTu measurements. Unfortunately, classification of soil behaviour is highly dependent on sleeve friction which is the most unreliable measurement. Even so, accuracy of the CPTu measurements has been improved during its existence and development is still in progress. At some point inaccuracy may not be a problem.
For Finnish clay, 0.968 for m´ gives the best correlation when the multiplier k is 0.33. Due to sensitivity, the stress exponent may be smaller than for intact clays as m´ seems to have a smaller value with higher sensitivity values. To consider this stress exponent dependency as a means of interpretation would need a larger variation of stresses in the dataset. At this point, it would be reasonable to assume m´ to be 1 and keep the multiplier as the variable.
9.2 Estimation of undrained shear strength
The same kind of analysis can be done for undrained shear strength as was done for OCR. A directly defined correlation between the results of the calibration tests and the CPTu data is compared with the correlations including the index parameters. Similarly, the undrained shear strength can be assessed using the SHANSEP equations defined in Chapter 8.4, while the over consolidation ratio can be estimated with fairly good confidence at least in similar soil conditions. In the following, the
177
direct correlation is defined without the SHANSEP relation, by adopting the general Equations 9.11 – 9.13.
�� = ���@A³µS
= �¶�S³µS
(9.11)
�� = �F�A³∆n
= ∆��³∆n
(9.12)
�� = ���F³µ�
= ��>>³µ�
(9.13)
9.2.1 Undrained compression strength
Undrained compression strength compared to net cone resistance and excess pore pressure gives a slightly better correlation than was found for OCR. The poorest correlation is shown with the effective net cone resistance similar to the OCR interpretation. The correlations are shown in Figure 9.10.
The average values for the cone factors are 9.1 for Nkt,Su,Comp with a standard deviation (SD) of 0.81, 7.7 for NΔu,Su,Comp with an SD of 0.6 and 4.5 for Nke,Su,Comp with an SD of 0.56. Based on these values, cone factors show little absolute variation. Relatively, the smallest variation can be found from the excess pore pressure-based interpretation.
In Figure 9.11, the index properties and pore pressure ratio are compared to the cone factors. When su,Comp is interpreted with the cone factors given in the equations in Figure 9.11, the RSE values shown in Table 4 are obtained. Compared to direct correlation, it seems that some improvement can be achieved when the sensitivity or pore pressure ratio is associated with Nkt,su,Comp or Nke,su,Comp. The accuracy, especially of the effective cone-based interpretation will be improved significantly with the pore pressure dependent cone factor.
In Figure 6.7 it was shown that an increase of change in void ratio during reconsolidation most likely increases the axial strain at failure in the triaxial tests. In Figure 9.12, the relative void ratio change at reconsolidation is compared to resulting Nkt,su,Comp values. As can be seen, the influence is minor, but also here there is an indication of influence. Using the trendline shown in the figure results in a 2.5% difference in the Nkt,su,Comp values when the relative void varies between 1.5 % and 4 %.
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Figure 9.10 Relation between the undrained compression strength and the normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
179
Figure 9.11 Relation of the cone factors for su,Comp based on net cone resistance, excess pore pressure and effective cone resistance when compared to a.) the Plasticity index b.) Water content c.) Organic content d.) Sensitivity e.) Pore pressure ratio.
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Figure 9.12 Influence of relative void ratio change at reconsolidation to Nkt,su,Comp
As the dataset mainly consists of normally consolidated or slightly consolidated samples, it may be useful to present a general equation for a larger range of soils having greater OCR values, even of this verification cannot be done yet. Combining the Equation 2.9 and 9.14 to eliminate OCR, the Equation 9.15 can be derived. While the SHANSEP parameters SComp and m have been defined, they can be converted to present the M and Λ parameters. Thus, M is 1.36 and Λ is 0.93. Adopting the average value for k (=0.285=Nkt,OCR), then the undrained compression shear strength can be solved. A comparison between the measured and estimated values is shown in Figure 9.13. Additionally, the RSE value is 7.3 % and this is fairly close to the best value presented in Table 4 for the net cone resistance-based interpretation.
��� = k ∙ �[ (9.14)
�� = 1∙�¶�SG∙yG∙�@A� �©G��¸G (9.15)
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Figure 9.13 Comparison between the estimated su,Comp values with the general equation 9.15 and the measured values.
9.2.2 Undrained extension strength
Similarly, the same correlation can be found between all type interpretations and extension strengths, as can be seen in Figure 9.14. In addition, the direct correlation gives approximately the same RSE value of 19 % for all cone factors. Greater RSE values compared to other calibration tests might be due to the greater uncertainties related to the extension test results. The averaged value of Nkt,Ext is 18.5 with an SD of 2.95, NΔu,Ext is 14.1 with an SD of 2.34 and Nke,Ext is 8.3 with an SD of 1.32.
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Instead of using direct correlation to estimate extension strength, the accuracy may be improved by adding the index property or pore pressure ratio into the cone factors. This kind of comparison is shown in Figure 9.15 and the corresponding RSE values are shown in Table 4. The RSE values are then changed to be around 15 %. The greatest improvement can be obtained by adding the water content to the cone factor to obtain net cone resistance and excess pore pressure-based interpretations. Correspondingly, the effective cone resistance-based interpretation has the most influence when the pore pressure ratio is the estimator for the cone factor.
Figure 9.14 Relation between the undrained extension strength and the normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
In Chapter 8.5, where anisotropy of Finnish soft clay was considered, the plasticity index gave the highest correlation for the anisotropy ratio. That correlation was not clear owing to the R-squared value being 0.18. From that perspective, a preference
183
for using the plasticity index instead of the water content as a predictor may not be justified.
184
Figure 9.15 Relation of cone factors for su,Ext based on net cone resistance, excess pore pressure and effective cone resistance compared to a.) Plasticity index b.) Water content c.) Organic content d.) Sensitivity e.) Pore pressure ratio.
185
9.2.3 DSS strength
The clearest relationship between CPTu measurements and DSS strength was found for the excess pore pressure with an R-squared value of 0.709. Correlation of the CPTu measurements and DSS strengths can be found in Figure 9.16. the effective cone resistance and DSS strength have a poor correlation, however, a slightly better correlation is found with the net cone resistance.
The average value for the cone factors when assessing the DSS strength from the CPTu results was 15.5 for Nkt,DSS with an SD of 1.71, 11.5 for NΔu,DSS with an SD of 1.07 and 7.2 for Nke,DSS with an SD of 1.18.
RSE values for direct correlation varied between 9.7 – 15.5 % as shown in Table 4. The net and effective cone resistance-based interpretations will be improved when the cone factor for qnet is given as a function of the sensitivity and for qeff when it is given with the pore pressure ratio. The accuracy of interpretation is then improved by 3 %. Figure 9.17 presents all the equations for cone factors being the function of the compared index property or the pore pressure ratio.
186
Figure 9.16 Relation between the DSS strengths and normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
187
Figure 9.17 Relation of cone factors for su,DSS based on the net cone resistance, excess pore pressure and effective cone resistance compared to a.) Plasticity index b.) Water content c.) Organic content d.) Sensitivity e.) Pore pressure ratio.
188
9.2.4 Vane strength
With the vane test, the measured and corrected vane strengths can be compared to the CPTu measurements. First, the relation between the measured vane strengths and CPTu measurements are considered in Figure 9.18. Good correlations were found between the measured vane strengths and the CPTu measurements. The vane results have been limited by using the same principle as mentioned in Chapter 8.4. The determined R-squared values were in a range from 0.82 to 0.90, the excess pore pressure had the highest value while the effective cone resistance had the lowest.
An attempt to improve the cone factors by means of the index property or pore pressure ratio produced the results shown in Figure 9.19 and Table 4. In the case of the excess pore pressure-based interpretation, a direct correlation should be used. The pore pressure ratio seemed to improve the accuracy when used as a predictor of the cone factors for the net and effective cone resistances. The RSE values have then been decreased by 2.1 % for qnet and 5.5 % for qeff.
Figure 9.18 Relation between the measured vane strength and the normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
189
Figure 9.19 Relation between the cone factors for su,FVmeas based on the net cone resistance, excess pore pressure and effective cone resistance compared to a.) the Plasticity index b.) Water content c.) Organic content d.) Sensitivity e.) Pore pressure ratio.
190
When the vane strength was corrected with the factor defined in Equation 5.1, poorer correlations were found for the CPTu measurements and corrected vane strengths as can noticed by comparing the R-squared values in Figures 9.18 to those in 9.20.
The cone factor equations for su,FVcorr with different predictors are shown in Figure 9.21. The liquid limit dependent correction factor is probably the reason why the plasticity index has improved interpretations, as can be seen in Table 4. However, considerable change in the RSE value was also found in the effective cone resistance-based interpretation when the pore pressure ratio is the predictor. In that case, the RSE values decrease from a value of 14.4 to 8.8 % resulting in the most accurate way to estimate the corrected vane strength.
The average cone factors in the dataset for su,FVmeas and su,FVcorr are 14.7 and 15.9 with an SD of 2.57 and 1.86 for net cone resistance, 11.5 and 12.4 with an SD of 1.08 and 1.46 for excess pore pressure, and 6.4 and 6.9 with an SD of 0.98 and 1.35 for effective cone resistance.
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Figure 9.20 Relation between the corrected vane strength and the normalised a.) net cone resistance, b.) excess pore pressure, c.) effective cone resistance.
192
Figure 9.21 Relationship between the cone factors for su,FVcorr based on the net cone resistance, excess pore pressure and effective cone resistance compared to a.) the Plasticity index b.) Water content c.) Organic content d.) Sensitivity e.) Pore pressure ratio.
193
9.3 Selections of the interpretation equation for the OCR and undrained shear strength
Good interpretation results can be achieved with direct correlations and cone factors with additional parameters. However, some recommendations should be given for interpretations.
First, in Tables 4 - 6, the RSE and R2 values are presented. The RSE values are utilised to assess the improvement in accuracy when the additional parameter and R2 values present the correlation between the cone factor and the additional parameters. In some cases, these two static parameters give different indications as regards improvement. The RSE value gives a clear indication of whether the variation of interpretation has been changed, thus the R2 value is something that could indicate if the cone factor is dependent on the additional parameters. In Table 4, all the RSE values improved by 1 percentage point are bolded and in Table 6 all R2 values 0.40 or higher are bolded. Values lower than 0.4 cause doubt as to whether any correlation can be found between the cone factor and the index property.
Table 4 RSE values for interpreted su and OCR values with direct correlation and by using cone factors as a function of the index property and pore pressure ratio
194
195
Table 5 R2 values between CPTu measurements and calibration test results
Table 6 R2 values between cone factors and additional parameters plasticity index, water content, organic content, sensitivity and pore pressure ratio.
196
197
Below is presented the selected equations to assess OCR and strength by means of CPTu based on the results shown in Tables 4 - 6. These Equations 9.16 – 9.33 are used for interpretations to present the continuous profiles of predicted value versus depth. In the same figures, all the calibration results are presented.
The equations have now been calibrated and these equations should work in a similar condition as in the dataset. Basically, any reasonable equation can be used if the predictor is known. The following are some practical reason why the equations have been included in the selection:
1. The cone factor predictor should have a clear influence on the accuracy compared to the direct correlation. In the study, a decrease of the RSE value by one is selected to be a clear indication.
2. Use of sensitivity as a cone factor predictor might be questionable as its estimation needs high quality samples. The sensitivity value as defined by different strength test methods gives different values so here only sensitivity values defined with the fall cone test could be used.
3. Only in effective cone factor, a pore pressure ratio is selected for the interpretation as it is anyway dependent on accuracy of the cone resistance and pore pressure measurements.
In the case of corrected field vane strength, a correlation with the plasticity index may be needed to include the plasticity index-based correction in the equation even if the RSE is not improved in the data by over 1 percentage point.
Equations to estimate undrained shear strengths can be derived - this includes the OCR interpretations with Equations 9.14 – 9.16 combined with the SHANSEP parameters for certain test type.
The confidence level may be increased if Equations 9.16 – 9.18 all show similar OCR values. This might be an indication of the successful execution of the test and that the soil condition corresponds to this study.
198
��� = 0.173 ∙ ���@A�@A� + 0.533 (9.16)
��� = 0.370 ∙ �F�A�@A� + 0.013 (9.17)
��� = 81.607 ∙ a� − 0.5829 ∙ ���F�@A� (9.18)
��,4¹�p = 0.084 ∙ (�J − ���) + 2.913 (9.19)
��,4¹�p = 0.096 ∙ ('� − '�) + 4.798 (9.20)
��,4¹�p = ���F�.��∙º¨?:�.�� (9.21)
��,J»*�[ = ���@A�.��:∙�?�;.��v (9.22)
��,J»*�[ = �F�A�.��∙�?:¼.�v: (9.23)
��,J»*�[ = ���F::.;:�∙º¨?:�.�� (9.24)
��,½¾¾ = ���@A�.�v�∙¿z?:;.¼�� (9.25)
��,½¾¾ = 0.0641 ∙ ('� − '�) + 3.794 (9.26)
��,½¾¾ = ���F::.:;;∙º¨?:v.v� (9.27)
��,ÀÁ�*,- = 0.058 ∙ (�J − ���) + 1.896 (9.28)
��,ÀÁ�*,- = 0.081 ∙ ('� − '�) + 0.909 (9.29)
��,ÀÁ�*,- = ���F:�.v��∙º¨?:¼.v¼: (9.30)
��,ÀÁ�¹// = ���@A�.:��∙¿z?::.��; (9.31)
��,ÀÁ�¹// = �F�A�.��v∙¿z?:�.��� (9.32)
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��,ÀÁ�¹// = ���F�:.��v∙º¨?�;.�:: (9.33)
Figures 9.22 – 9.26 present comparisons where results from the calibration tests regarding the oedometer and strength tests are shown, and these values are compared to evaluated values from the CPTu results using Equations 9.16 – 9.33. The measured field vane strength interpretation is not presented as it may not offer anything extra. Additionally, it is good to observe that the CPTu based interpretation for triaxial strengths do not deviate much from each other except for shallow depths.
Figure 9.22 Interpretation of a.) OCR with Equations 9.16 – 9.18 b.) undrained compression and extension shear strengths with Equations 9.19 – 9.24 c.) DSS strength with Equations 9.25 – 9.27 and corrected vane strength with Equations 9.31 – 9.33 at the Lempäälä site.
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Figure 9.23 Interpretation of a.) OCR with Equations 9.16 – 9.18 b.) undrained compression and extension shear strengths with Equations 9.19 – 9.24 c.) DSS strength with Equations 9.25 – 9.27 and corrected vane strength with Equations 9.31 – 9.33 at the Perniö site.
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Figure 9.24 Interpretation of a.) OCR with Equations 9.16 – 9.18 b.) undrained compression and extension shear strengths with Equations 9.19 – 9.24 c.) DSS strength with Equations 9.25 – 9.27 and corrected vane strength with Equations 9.31 – 9.33 at the Masku site.
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Figure 9.25 Interpretation of a.) OCR with Equations 9.16 – 9.18 b.) undrained compression and extension shear strengths with Equations 9.19 – 9.24 c.) DSS strength with Equations 9.25 – 9.27 and corrected vane strength with Equations 9.31 – 9.33 at the Paimio site.
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Figure 9.26 Interpretation of a.) OCR with Equations 9.16 – 9.18 b.) undrained compression and extension shear strengths with Equations 9.19 – 9.24 c.) DSS strength with Equations 9.25 – 9.27 and corrected vane strength with Equations 9.31 – 9.33 at the Sipoo site.
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10 DISCUSSION
In the previous chapters, correlation for OCR and strength have been defined. Direct correlation between OCR and the normalised net cone resistance (Q) was found to work well for the interpretation. Based on Figure 10.1a, it can be seen that this correlation is not directed to a zero point. If all the data are on the same line pointing to the origin, the data would have perfect normalisation. However, some property or properties might make the line more horizontal. A similar result was found when the net cone resistance was compared to the undrained compression strengths. Due to this distorted correlation, some underestimation could occur. For example, in the case of the estimation of OCR with Q, underestimation can occur when the value of Q is higher than 7. Even though, estimation of OCR with the equation 9.16 gives a good approximation. Some reasons for the distorted correlation could be listed:
� greater influence of the sample disturbance with certain soil types influencing the reference test results
� greater influence of the strain rate with clays that have a greater plasticity index and lower sensitivity
� a greater initial anisotropic stress state and strength anisotropy than could be excepted based on similar friction angles
� greater variation in the effective friction angle having greater influence on anisotropic behaviour as well inaccuracies in the measurements
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Figure 10.1 a.) Relationship of OCR to Q b.) relationship of Nkt,OCR to plasticity index.
Endeavours were made to minimize the effects of sample disturbance and measurement inaccuracies by following best practices and using accurate measuring devices. Thus, soil behaviour during CPTu penetration should explain the distorted trend. Often, it was found that the plasticity index produced some influence on the observed parameter as well, as can be seen in Figure 10.1b where the plasticity indexes are compared to Nkt,OCR. For some reason, a lower Nkt,OCR value should be used as the plasticity index increases. In a comparison between sensitivity and Nkt,OCR, an even higher correlation was found. As sensitivity increases, the Nkt,OCR also increases as shown in Figure 9.5. Some cross correlation could also occur as the sensitivity and plasticity index have some correlation too, as can be seen in Figure 8.3b.
Cone penetration causes a complex failure for the surrounding soil as shown in Figure 4.6. Thus, trying to find a correlation with one specific test representing only one failure mode at a time might cause a problem if the anisotropic strength together with the friction angle are not constant; even if no significant influences occur due the rigidity index and strain rate effect. It was found that the maximum obliquity line in the undrained compression tests closely follow the same line but the anisotropy ratios su,DSS / su,Comp and su,Ext / su,Comp are not constant because they have a little trend whereby increasing the plasticity index decreases the anisotropy. For isotropic strength, the failure mode does not need as much consideration. For greater anisotropy, greater compensation is needed. Compensation is made in the case of
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the OCR estimation by increasing Nkt,OCR and in the case of the estimation of undrained compression strength, Nkt,Comp is seen to decrease.
Anisotropic strength alone may not distort the correlations. It is known that cone resistance is not only dependent on the vertical stress. For practical reasons, the vertical stress is used when its estimation can be done with reasonable accuracy (Wroth 1984). Based on data from Kenney 1959 and Bjerrum and Simons 1960, Ladd (1977) proposed that friction angles decrease as the plasticity index increases. For an estimation of the initial stress state, the friction angle can be used. In normally consolidated soils especially, Jaky’s well-known equation for estimating K0 gives a reasonable solution. For lower friction angle values, the initial stress state is more isotropic and this means it has a lesser influence on vertical stress normalisation. Normalisation of the CPTu measurement by means of vertical stress causes a problem for soil behaviour classification (Jamiolkowski and Robertson 1988) and for undrained shear strength interpretation (Teh&Houlsby 1991). The greater initial anisotropic stress state is compensated for by increasing Nkt,OCR and decreasing Nkt,Comp. As instability may occur before reaching the ultimate failure state in structured soil, the friction angle for K0 estimation using Jaky’s equation may be better based on (σ1 – σ3)max values as instability is reached at a lower mobilised friction angle.
The strain rate effect on undrained shear strength may be greater in highly plastic clays (Berre and Bjerrum 1973, Vaid et al. 1979). This observation has later questioned (Graham et al. 1983). The strain rate effect has not studied in this research so its influence for CPTu based interpretations cannot be considered in the dataset. However, a higher strain rate seems to enable the (σ1 – σ3)max line to increase and approach the (σ1 / σ3)max line in the triaxial tests with sensitive samples (Berre 1973, Jostad et al. 2006). It could be that the relatively fast cone penetration is more dependent on this (σ1 / σ3)max criteria when strength is considered.
In general, it can be said that the greater strain rate effect needs a greater reduction in su and OCR interpretations through increasing Nkt,OCR and decreasing Nkt,Comp. The strain rate effect together with the anisotropy of the soil may be considered by using the vane results. Vane strength is mainly dependent on stresses on the vertical planes and friction acting on shear planes, especially when the failure zone is cylindrical, and no progressive failure occurs in the vertical planes. In Figure 8.16, the normalised measured vane strength results seem to decrease with increasing sensitivity and increase with an increasing plasticity index. Normalising the measured
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vane strengths using the horizontal effective stresses can be done - as shown in Figure 10.2. Horizontal stresses were estimated with the measured OCR and a constant effective friction angle (30 ), and some relation was found, especially for the plasticity index. The greater disturbance caused by the inserting the vane into the soil might explain this trend or it is possibly due to the increased anisotropy in more plastic clays and/or just due to the rate effects.
The anisotropy of undrained shear strength has been considered in this study as well as the stiffness anisotropy influences on the cone resistance. To study this, anisotropically consolidated triaxial compression and extension tests would be needed. Unfortunately, only isotropically consolidated extension tests were available. An increase in stiffness anisotropy decreases the cone resistance as the same displacement would cause lower pressure.
As the laboratory results and the vane test results did not show a significant difference in strain softening, some distortion might have occurred. In the compression test results, a little influence may have occurred as the initial stress state was unknown and the consolidation stress state was based on the general equation and thus, it was not possible to apply an accurate estimation of the initial mean effective stress. Possibly, a higher mean effective stress state for more plastic clays could lead to a less visible peak strength and strength softening. The vane tests have been conducted in in-situ stress state. However, the vane test has some uncertainty related to the shape of the shear zone in the beginning of rotation. Additionally, during continuous cone penetration the soil goes through a complex strain path where strains are not localised in a certain shear plane as seen in triaxial compression tests and vane tests. Due to this, the strain softening behaviour could be different.
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Figure 10.2 Measured vane strength normalised by means of estimating the effective horizontal stress compared to a.) Sensitivity b.) the Plasticity index.
A better understanding of the influence of the sensitivity and plasticity index via anisotropy as regards cone resistance may need more consideration in future. In the SBT charts, a trend showing increasing sensitivity has been proposed because it is recognised that some properties derived from the CPTu measurements are dependent on sensitivity, such as the friction ratio and pore pressure ratio. In the SBT charts shown in Figure 10.3, sensitive soil zones are marked as zone 1 and the sensitivity increases as the points are more located inside that zone. In general, the measurement of the sleeve friction lacks accuracy, and, for this reason, it should be less trusted. In future, more effort should also be put on the use of the SBT charts to estimate cone factors.
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Figure 10.3 SBT charts, with the data used, together with the calibration results of this study.
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11 CONCLUSION
The aim of study was to create a practical and sufficiently accurate solution for interpreting undrained shear strength from CPTu results. To prove that the solution was sufficiently accurate, it was clear that high demands should put on the calibration tests too. Much effort was put into improving the sample quality and executing the calibration tests with great care.
Five testing sites were investigated using CPTu, field vane tests and taking samples to the laboratory. In the laboratory, oedometer tests, triaxial compression and extension and DSS tests were performed together with determination of index properties. To calibrate CPTu measurements to undrained compression strength and OCR, 17 calibration data points were used. Accordingly, the data points in undrained extension strength calibration were 14, 14 for DSS and 15 for field vane tests.
The performance of the (TUT) tube sampler for Finnish clay conditions was found to be very good when the sample quality was classified based on the void ratio change in consolidation to an in situ effective stress as proposed by Lunne et al. (1997). Suitability of that sample disturbance classification for Finnish clay was not studied.
Some undesired dimensional difference between the tubes and the cutting shoe was noticed as having an influence on the inside clearance, and this may have had some influence on the sample quality; however, at least it had an influence on how well the sample stayed inside the sampler during uplifting.
The Mini Sherbrooke sampler can also offer excellent samples for Finnish clay. The normalisation of undrained triaxial compression strength with a preconsolidation pressure from the same sample showed the highest ratio. The time frame between sampling and testing were in this case the shortest and this may partly explain the better quality.
The Downhole field vane device provides more realistic stress-rotation curves compared to the uphole device, as the rotation and torque are measured close to the vane. Though with the correct configurations and performance, the uphole can offer
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similar peak values. The proportional remoulded strength values between these two devices are high while the remoulded strength values themselves are very low. The differences between measurements could be due to the better accuracy of the downhole device and the lower level of extra frictions.
The anisotropically and isotropically consolidated samples gave the good agreement in undrained compression shear strengths when both samples were consolidated to the same mean effective stress inside the yield surface. Naturally, the isotropic consolidation causes an increase in strain before reaching failure. Isotropic consolidation conducted to the same mean effective stress causes a slightly greater change in the void ratio during consolidation.
The influence of the index properties on the normalised strength values were considered and the SHANSEP parameters defined. The SHANSEP parameters defined from the samples taken with the TUT sampler were: S=0.36 and m=0.93 for the undrained compression strength, S=0.21and m=0.79 for the undrained extension strength and S=0.24 and m=0.8 for the DSS. For the corrected vane strength, the parameters were: S=0.24 and m=0.74 defined by the downhole device. The SHANSEP parameters for the corrected vane strength are close to those values previously proposed by D’Ignazio et al. (2018) as this study had a larger collected database from Finland as regards results measured by uphole devices. For the measured vane strength, the SHANSEP parameters were: S=0.26 and m=0.69.
The plasticity index, liquid limit, organic content, sensitivity and water content have been compared to normalised strength values. The water content shows the greatest influence on the normalised values of compression, extension and DSS strengths. Therefore, the R-squared values are correspondingly 0.16, 0.2 and 0.25. The organic content was given the same R-squared value for normalised compression strength values as the water content. In Figure 8.17, the measured vane strengths normalised by means of preconsolidation pressure had similar dependencies to the plasticity index, liquid limit and sensitivity as the R-squared values were correspondingly 0.4, 0.38 and 0.37. The measured vane strength was corrected by means of the function of a liquid limit. This correction removed the dependency of the sensitivity on normalised corrected vane strengths. However, some dependency was left as regards the plasticity index and liquid limit.
The anisotropy ratios were assessed by comparing the peak undrained strengths as defined by the DSS, extension and compression triaxial tests. For Finnish clay,
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the anisotropy ratio for su,DSS / su,Comp is 0.64 and for su,Ext / su,Comp was 0.615 based on conducted tests.
By normalising the stress strain curves with the peak values, the measured curves became quite similar for all the vane tests, DSS tests and triaxial compression tests. Vane curves showed the greatest strain softening due to the greater shearing. A rotation of 45 degrees caused a reduction of 45 - 55 % in the peak strength. The triaxial compression tests had a reduction of 20 – 50 % compared to the peak value after 10 % of shear strain. The triaxial compression and vane tests had the most visible peaks in the curves. The DSS curves were smoother and had less reduction. Reduction after 10 % of shear strain was 10 – 35 %. The extension test had more variation in the results. After 10 % shear strain, the reduction was between 5 – 45 %. Increasing strains could cause difficulties for corrections e.g. shape and membrane corrections.
Based on the experience gained, executing the extension and vane tests seems to have caused the greatest uncertainty in the results, while the compression tests have the lowest with the DSS somewhere in between. Uncertainty concerning the vane results was controlled by executing two to three vane tests at each site for comparison and thus, reliable results were achieved.
Interpretation of the CPTu measurements was conducted by using net cone resistance (qnet), excess pore pressure (Δu2) and effective cone resistance (qeff) to assess undrained compression, extension, DSS and vane strengths. Normalised values of net cone resistance, excess pore pressure and effective cone resistance were used for OCR interpretation. According to the test data, direct correlations worked well if the correlation was not forced to its origin. In some cases, it was not justified to put any index property in the interpretation as the improvement was not clear. As regards the index properties, the plasticity index, water content, organic content, sensitivity and pore pressure ratio were used. Relative standard error (RSE) was calculated for direct correlation and interpretation where the index property was added into the cone factor. These values were compared, and a 1 percentage point decrease in the RSE value was considered a remarkable improvement. Additionally, the use of sensitivity in the cone factor could be questionable due to the need for at least good quality samples for the strength test. With good quality samples, the strength result itself may be utilised for site specific calibration.
Interpretation of the undrained compression shear strength and OCR using the CPTu measurements (qnet, Δu2, qeff) can be evaluated with the highest accuracy. In
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similar soil conditions with a corresponding sample quality as in the dataset, the relative error in interpretation would be under 7 % for OCR and su,Comp. For su,Comp and OCR evaluations, the net cone resistance-based and excess pore pressure-based interpretations do not need any index property to adjust the cone factors. Similar results were found for su,DSS and su,FVmeas evaluated from Δu2 and for su,FVmeas evaluated from qnet.
Normalised effect cone resistance-based interpretations improved when the pore pressure ratio (Bq) was adjusted for the cone factors. This result was found in evaluations of OCR, su,Comp, su,Ext, su,DSS, su,FVmeas and su,FVcorr. The remainder of the interpretations of su,Ext, su,DSS, su,FVcorr and su,FVcorr seemed to have some benefit if the cone factor was defined as a function of the index property. The more accurate cone factors Nkt,Ext and NΔu2,Ext can be evaluated with the water content, and the plasticity index worked for Nkt,DSS, Nkt,FVcorr and NΔu2,FVcorr. It should be noted that the influence of the plasticity index for Nkt,FVcorr and NΔu2,FVcorr was not clear. In general, the assumption was used that dependency could be found if the vane results were corrected using the liquid limit. The liquid limit and plasticity index had a clear correlation in the dataset.
The relative standard error for the interpretation of the extension strength from qnet, Δu2 and qeff was found to be approximately 15 %. For the DSS strength, the RSE values were between 9 -12.2 % whereas the su,FVmeas interpretation had relative errors of 7 – 9,3 % and correspondingly the su,FVcorr interpretation had 8.8 -10.5 % errors.
If all three means of interpretation (qnet, Δu2, qeff) with the Equations 9.16 – 9.33 had given similar results for strength or OCR, the confidence level for the interpretations would be higher. Additionally, this might also be an indication that the test was conducted in a similar soil condition as in the dataset. The quality and accuracy of the CPTu measurements should also correspond to the quality of this study.
Overall, it can be concluded that the CPTu proved to be well suited as a field investigation method for soft Finnish clays and can provide reliable continuous data on both OCR and undrained shear strength.
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12 FURTHER RESEARCH
In this study, together with the parallel study by Di Buò (2020) a large, high quality database was created on soft Finnish clays. In the future, this database can be used for several further studies. Unfortunately, the testing sites do not have the status of national testing sites and it is unsure how long they all will be available for testing. It would be important to guarantee access for research to these sites in the future.
The database should be extended to cover a larger variation of soil. More interest should be put on clays with higher OCR as well, and silt and organic contents should be included in the dataset. This would further aid in the goal of promoting the use of CPTu as a major soil investigation method in Finland. Additionally, the database should be updated with effective failure parameters and possibly with strain rate effects. Moreover, instability behaviour causing a failure with partial mobilisation of the shear strength could be studied.
To extend the database, comparisons and combining can be done with other databases. Other database should have data with similar soil behaviour as in this study, at least if CPTu correlations for strength are considered. It is also important that the quality of data should be known.
Full flow penetrometers could be something to test in future. Full flow penetrometers are less sensitive to changes in temperature during testing and for pore pressure effect. Additionally, its interpretation has a better theoretical base (Stewart&Randolph 1994, DeJong et al. 2010).
Creating SBT charts with the contours of the cone factors is also something that should be considered in future.
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REFERENCES
Aas, G. (1976) Stabilitetsberegning for fylling på kvikkleire, Ellingsrud, Oslo. Norske Sivilingeni rers Forening. Kurs i Jordartsegenskaper - bestammelse og anvendelse. Gol. 1976. Totalspänningsanalyser, prinsipp, grunnlag. Norske Sivilingei rers Forening. Kurs i Jordartsegenskaper -bestämmelse og anvendelse. Gol.
Amundsen, H. A., Thakur, V., & Emdal, A. (2016). Sample disturbance in block samples of low plastic soft clays. In 17th Nordic Geotechnical Meeting, Island.
Amundsen, H.A., Jønland, J., Emdal, A., & Thakur, V. (2017) An attempt to monitor pore pressure changes in a block sample during and after sampling. Géotechnique Letters 7(2)
Andersson-Berlin, K. (2012). Ratojen luokitus: maanvaraisten pehmeikköjen stabiliteetin selvitys EU-luokituksen mukaisilla kuormilla. Liikenneviraston tutkimuksia ja selvityksiä: 43/2012.
ASTM D2573-01 (2007), Standard Test Method for Field Vane Shear Test in Cohesive Soil, ASTM International, West Conshohocken, PA, 2007.
Baldi, G., Hight, D. W. & Thomas, G. E. (1988) STP977-EB A Re-evaluation of Conventional Triaxial Test Methods, ASTM International, 219.
Baligh, M.M. (1984). The simple-pile approach to pile installation in clay. In Proceedings of the Symposium on Analysis and Design of Pile Foundations, San Francisco, Calif., 1–5 October 1984. pp. 310–330.
Baligh, M.M. (1985). Strain path method, J. Geotech. Div., Am. Soc. Civ. Engrs., 111 (GT9), pp. 1108—1136.
Baligh, M. M., Azzouz, A. S. & Chin, C.T. (1987) Disturbance due to ideal tube sampling, J. Geotech. Engng. Div., Am. Soc. Civ. Engrs, 113 (GT7), pp. 739-757.
216
Baligh, M. M. & Levedoax, J. N. (1980). Pore pressure dissipation after cone penetration. Massachusetts Institute of Technology. Department of Civil Engineering, Cambridge, Mass, pp. 80-111
Becker, D.R., Crooks, J.H.A, Been, K. & Jefferies, M.G. (1987). Work as criterion for determining in-situ & yield stresses clays. Can. Geotech. J. 24: pp. 549-564
Benz, T. (2007). Small-strain stiffness of soils and its numerical consequences (Vol. 5). Stuttgart: Univ. Stuttgart, Inst. f. Geotechnik.
Berre, T. (1972). Sammenheng mellom tid, deformasjoner og spenniger for normal konsolidierte marine leirer. Proc — Nordic Conf. on Soil Mechanics, Trondheim, pp. 23-36.
Berre, T. (1973). Effect of rate of strain on the stress-strain relationship for undrained triaxial tests on plastic Drammen clay. NGI report, (50301-4).
Berre, T. & Bjerrum L. (1973). Shear strength of normally consolidated clays. Proc. 8th ICSMFE, Moscow 1, pp. 39-49.
Bishop, A.W. & Henkel, D.J. (1962) The Measurement of Soil Properties in the Triaxial Test, 2nd edn, Edward Arnold, London.
Bjerrum, L. & Simons, N.E. (1960): Comparison of shear strength characteristics of normally consolidated clays, Proc., Research Conference on Shear Strength of Cohesive Soils, ASCE, Boulder, Colorado, pp.711-726.
Bjerrum, L. (1967) Engineering geology of Norwegian normally consolidated marine clays as related to settlements of buildings. Geotechnique 17(2) pp. 81-118
Bjerrum, L. (1972). Embankments on soft ground, proceedings, ASCE specialty Conference on performance on earth and earth supported structures, Purdue university, Lafayette, vol. II, pp. 1-54.
Bjerrum, L. (1973). Proceedings of the 8th International Conference on Soil Mechanics and Foundation Engineering, Moscow, 3, pp. 111-159.
217
Burland, J. B. (1990) On the compressibility and shear strength of natural clays, Geotechnique, 40(3) pp. 329-378.
Broussard, N. S. (2011). Proper selection of rigidity index (Ir=G/Su) relating to CPT correlations. ECI 284 Term Project. University of California, Davis.
Cadling, L., & Odenstad, S. (1948) The Vane Borer. Proceedings of the Royal Swedish Geotechnical Institute, 2, pp. 1-87
Campanella, R. G., Gillespie, D. G., & Robertson, P. K. (1981). Pore pressures during cone penetration testing. Department of Civil Engineering, University of British Columbia.
Carter, J. P., Randolph, M. F., & Wroth, C. P. (1979). Stress and pore pressure changes in clay during and after the expansion of a cylindrical cavity. International Journal for Numerical and Analytical Methods in Geomechanics, 3(4), pp. 305-322.
Casagrande, A. (1936). The determination of the preconsolidation load and its practical significance. Proc. First Intern. Conf. on Soil Mech. & Found. Eng., Cambridge, pp. 60-64.
Chandler, R. J. (1988) The in-situ measurement of the undrained shear strength of clays using the field vane. Vane shear strength testing in soils: field and laboratory studies, pp. 13-44
Chen, B. S. Y., & Mayne, P. W. (1994). Profiling the overconsolidation ratio of clays by piezocone tests. Georgia Tech Research Corp. Report Nº GIT-CEEGEO-94.
Chen, B. S., & Mayne, P. W. (1996). Statistical relationships between piezocone measurements and stress history of clays. Canadian Geotechnical Journal, 33(3), pp. 488-498.
Das, B.M. (2002). Principles of Geotechnical Engineering (5th ed.): St. Paul, MN (Brooks Cole).
Day, R. W. (2010). Foundation engineering handbook. McGraw Hill.
218
De Alencar, J. A., Chan D. H., & Morgenstern, N. R. (1988) Progressive Failure in the Vane Test, Vane Shear Strength Testing in Soils: Field and Laboratory Studies, ASTM STP 1014, AF
De Josselin de Jong, G. (1971). Discussion Session 2. Stress-strain behavior of soils. In Proceedings, Roscoe Memorial Symposium. Edited by R. H. G. Parry, G. T. Foulis, and Co., Henley-on Thames, United Kingdom, pp. 258-261.
DeJong, J., Yafrate, N., DeGroot, D., Low, H. E., & Randolph, M. (2010). Recommended practice for full-flow penetrometer testing and analysis. Geotechnical Testing Journal, 33(2), 137-149.
D’Ignazio, M. (2016). Undrained shear strength of Finnish clays for stability analyses of embankments. Doctoral thesis. Tampere University of Technology. Publication 1412.
D'Ignazio, M., Phoon, K. K., Tan, S.A., & Länsivaara, T. (2016). Correlations for undrained shear strength of Finnish soft clays. Canadian Geotechnical Journal 53(10): pp. 1628-1645.
Di Buò, B. (2020). Evaluation of the Preconsolidation Stress and Deformation Characteristics of Finnish Clays based on Piezocone Testing. Doctoral thesis. Tampere University Dissertations 222.
Di Buò, B., Selänpää, J., Länsivaara, T. T., & D’Ignazio, M. (2018). Evaluation of sample quality from different sampling methods in Finnish soft sensitive clays. Canadian Geotechnical Journal, 56(8), pp. 1154-1168.
Donald, I. B., Jordan, D. O., Parker, R. J., & Toh, C. T. (1977) The vane test-a critical appraisal. In Proceedings of the 9th Intern. Conference on Soil Mechanics and Foundation Engineering, 1, pp. 81-88
Dyvik, R., Berre, T., Lacasse, S., & Raadim, B. (1987). Comparison of truly undrained and constant volume direct simple shear tests. Geotechnique, 37(1), pp. 3-10.
Flaate, K. (1966) Factors influencing the results of vane tests. Canadian Geotechnical Journal, 3: pp. 18-31
219
Emdal, A., Gylland, A., Amundsen, H. A., Kåsin, K., & Long, M. (2016). Mini-block sampler. Canadian Geotechnical Journal, 53(8), pp. 1235-1245.
Germaine, J. T. & Ladd, C. C. (1988) State-of-the-art paper: Triaxial testing of saturated cohesive soils. In Advanced triaxial testing of soil and rock. ASTM International.
Graham, J., Crooks, J. H. A., & Bell, A. L. (1983). Time effects on the stress-strain behaviour of natural soft clays. Géotechnique, 33(3), pp. 327-340.
Griffiths, D. V., & Lane, P. A. (1990) Finite element analysis of the shear vane test. Computers & Structures, 37(6): pp. 1105-1116
Gylland A. S., Jostad H. P., & Nordal S. N. (2012) Failure geometry around the shear vane in soft sensitive clay. Proceedings of the 16th Nordic Geotechnical Meeting (NGM), Copenhagen, pp. 103-110
Gylland, A.S., Jostad, H.P., & Nordal, S. (2014) Experimental study of strain localization in sensitive clays. Acta Geotechnica, 9(2), pp. 227-240.
Haikola, M. (2018) Water Content Determination of Soft Finnish Clays Using Electrical Conductivity Measurements. Master thesis at Tampere University of Technology (TUT)
Hansen, J. B., & Gibson, R. E. (1949). Undrained shear strengths of anisotropically consolidated clays. Geotechnique, 1(3), pp. 189-200.
Hansen, J.B. (1970) A revised and extended formula for bearing capacity. Geoteknisk Inst., Bulletin 28, pp. 5-11.
Hardin, B. O., & Drnevich, V. P. (1972a). Shear modulus and damping in soils: measurement and parameter effects. Journal of Soil Mechanics & Foundations Div, 98(sm6).
Hardin, B. O., & Drnevich, V. P. (1972b). Shear modulus and damping in soils: design equations and curves. Journal of Soil Mechanics & Foundations Div, 98(sm7).
220
Helenelund, K.V. (1951). Om konsolidering och sättning av belastade marklager: On consolidation and settlement of loaded soil layers. Diss. Tekniska högskolan, Helsingfors.
Helenelund, K. V. (1977). Methods for reducing undrained shear strength of soft clay. In Swedish Geotechnical Institute, Proceedings (No. Report No. 3 Proceeding).
Helle, T. E. (2017). Quick-clay landslide mitigation using potassium chloride, Doctoral theses at NTNU, Trondheim.
Hight, D. W., Böese, R., Butcher, A. P., Clayton, C. R. I., & Smith, P. R. (1992). Disturbance of the Bothkennar clay prior to laboratory testing. Géotechnique, 42(2), pp. 199-217.
Holtz, R. D., Jamiolkowski, M. B., & Lancellotta, R. (1986). Lessons from oedometer tests on high quality samples. Journal of geotechnical engineering, 112(8), pp. 768-776.
Hopper, R. J. (1992) The effects and implications of sampling clay soils, PhD thesis, University of Surrey.
Hueckel, T. & Baldi, G. (1990) Thermoplasticity of saturated clays: experimental constitutive study. Journal of geotechnical engineering, 116(12):1778–1796, 1990.
Hvorslev, M. J. 1949. Subsurface exploration and sampling of soils for civil engineering purposes, Report on soil sampling, U.S. waterways experiment station, Vicksburg.
ISO (2012). Geotechnical investigation and testing – Field testing – Part 1: Electrical cone and piezocone penetration tests, International Standard ISO 22476-1:2012
Jamiolkowski, M, Ladd, C.C., Germaine, J.T. & Lancellotta, R. (1985). New developments in field and laboratory testing of soils. International conference on soil mechanics and foundation engineering, 11, Proceedings, Vol. 1, pp. 57-153. San Francisco.
221
Jamiolkowski, M., & Robertson, P. K. (1989). Future trends for penetration testing. In Penetration testing in the UK: Proceedings of the geotechnology conference organized by the Institution of Civil Engineers and held in Birmingham on 6–8 July 1988 (pp. 321-342). Thomas Telford Publishing.Janbu, N. (1963) Soil compressibility as determined by oedometer and triaxial tests. Proc. 4th European Conference on Soil Mechanics and Foundation Engineering, Wiesbaden.
Janbu, N., & Senneset, K. (1974). Effective stress interpretation of in situ static penetration tests. In Proceedings of the 1st European symposium on penetration testing (Vol. 2, pp. 181-93).
Jardine, R. J., Symes, M. J., & Burland, J. B. (1984). The measurement of soil stiffness in the triaxial apparatus. Geotechnique, 34(3), 323-340.
Jefferies, M.G., and Been, K. (2006). Soil liquefaction – a critical state approach. Taylor & Francis, ISBN 0-419-16170-8.
Jostad, H. P., Andresen, L., & Thakur, V. (2006). Calculation of shear band thickness in sensitive clays. In Proceedings of the sixth European conference on numerical methods in geotechnical engineering, Graz. pp. 27-32.
Karlsrud, K. (1991). Sammenstilling av noen erfaringer med prøvetaking og effekt av prøveforstyrrelse i norske marine leire. NGI report 521500-6.
Karlsrud, K., & Myrvoll, F. (1976). Performance of a strutted excavation in quick clay. Proceedings, 6th European Conference on Soil Mechanics and Foundation Engineering, Vienna, Vol. 1(1), pp. 157-164.
Karlsrud, K., & Hernandez-Martinez, F. G. (2013). Strength and deformation properties of Norwegian clays from laboratory tests on high-quality block samples 1. Canadian Geotechnical Journal, 50(12), pp. 1273-1293.
Karstunen, M., Krenn, H., Wheeler, S. J., Koskinen, M. & Zentar, R. (2005). Effect of anisotropy and destructuration on the behavior of Murro test embankment. International Journal of Geomechanics, 5(2):87–97.
Karstunen, M., & Koskinen, M. (2008) Plastic anisotropy of soft reconstituted clays. Canadian Geotechnical Journal, 45(3):314–328.Keaveny, J. (1985). In-situ
222
determination of drained and undrained soil strength using the cone penetration test, Ph.D. Dissertation, University of California, Berkeley.
Keaveny, J.M., & Mitchell, J.K. (1986). Strength of fine grained soils using the piezocone. In Proceedings of the ASCE Specialty Conference In Situ’86: Use of In Situ Tests in Geotechnical Engineering, Blacksburg, Va., 23–25 June 1986. Geotechnical Special Publication No. 6. Edited by S.P. Clemence. American Society of Civil Engineers (ASCE), New York. pp. 668–685.
Kim, K., Prezzi, M., Salgado, R., & Lee, W. (2008). Effect of penetration rate on cone penetration resistance in saturated clayey soils. Journal of geotechnical and geoenvironmental engineering, 134(8), 1142-1153.
Kimura, T., Saitoh K. (1983) Effect of disturbance due to insertion on vane shear strength of normally consolidated cohesive soils. Soils and Foundations 23, pp. 113-124
Kjellman W. (1951), Testing the shear strength of clay in Sweden, Géotechnique, Vol.2, No. 3, pp. 225-235
Kolisoja, P. (1990). Maan mekaanisten ominaisuuksien automatisoitu määrityslaitteisto. Lisensiaatintyö. Tampereen teknillinen korkeakoulu. Geotekniikan laitos. Rakennustekniikan osasto.
Konrad, J. M. & Law, K. (1987). Undrained shear strength from piezocone. Canadian Geotechnical J. 24, pp. 392-405.
Koskinen, M., Vepsäläinen, P. & Lojander, M. (2002). Modelling of anisotropic behaviour of clays –Test embankment in Murro, seinäjoki, Finland, Finnra Report No. 16/2002. Helsinki, Finland: Finnish Road Administration
Kulhawy, F. H., & Mayne, P. W. (1990). Manual on estimating soil properties for foundation design (No.EPRI-EL-6800). Electric Power Research Inst., Palo Alto, CA (USA); Cornell Univ., Ithaca, NY (USA). Geotechnical Engineering Group.
Kulhawy, F.H., O'Rourke, T. D., Stewart, J. P., & Beech, J.F. (1983) Transmission Line Structure Foundations for Uplift-compression Loading, Load Test Summaries: Appendix to EPRI Final Report EL-2870. Electric Power Research Institute.
223
La Rochelle, P., Roy, & M., Tavenas F (1973) Field measurements of cohesion in Champlain clays. Proceedings of the 8th International Conference on Soil Mechanics and Foundation Engineering, Moscow, 1, pp. 229-236
Landon, M. M., DeGroot, D. J., & Sheahan, T. C. (2007). Nondestructive sample quality assessment of a soft clay using shear wave velocity. Journal of geotechnical and geoenvironmental engineering, 133(4), 424-432.
Ladd, C.C., & Foott, R. (1974). A new design procedure for stability of soft clays. Journal of the Geotechnical Engineering Division. ASCE, 100 (7), pp. 763-786.
Ladd C.C. (1977). Stress-deformation and strength characteristics. Proc. 9th. Int. conf. soil mech. and found. Eng. pp. 421-494.
Lade, P. V. (1992). Static instability and liquefaction of loose fine sandy slopes. Journal of Geotechnical Engineering, 118(1), pp. 51-71.
Lade, P. V. & Tsai, J. (1985). Effects of localization in triaxial tests on clays. Proc. list ICSMFE, San Franscisco, 1985, pp. 549-552.
Langø, H. (1991). Cyclic shear modulus of natural intact clays (Doctoral dissertation, PhD thesis NTH, Trondheim, August).
Lanzky, R., & Palmquist, D. (2013) Sample quality and disturbance in soft marine clay. Master of Science Thesis in the Master’s Programme Infrastructure and Environmental Engineering, Chalmers University of Technology, Göteborg, Sweden
Larsson, R. (1977). Basic behaviour of Scandinavian soft clays. Swedish Geotechnical Institute, Linköping, Report No. 4.
Larsson, R. (1980). Undrained shear strength in stability calculation of embankments and foundations on soft clays. Canadian Geotechnical Journal, 17(4), pp. 591-602.
Larsson, R. (2011). Metodbeskrivning för SGI:s 200 mm diameter 'blockprovtagare' - Ostörd provtagning i finkornig jord (In Swedish) Swedish Geotechnical Institute, SGI. Göta River Commission. GÄU. Subreport 33. Linköping.
224
Larsson, R. & Mulabdic, M. (1991). Piezocone Tests in Clay. Swedish Geotechnical Institute, SGI, Linköping, Report 42.
Larsson, R., Sällfors, G., Bengtsson, P. E., Alén, C., Berghdahl, U., & Eriksson, L. (2007). Skjuvhållfasthet-utvärdering i kohesionsjord.’
Larsson, R., Åhnberg, H., & Löfroth, H. (2010). CPT data in relation to a general model for the undrained behaviour of homogeneous Swedish clays. In 2nd International Symposium on Cone Penetration Testing, CPT (Vol. 10, pp. 145-152).
Lefebvre, G., & Poulin, C. (1979). A new method of sampling in sensitive clay. Canadian Geotechnical Journal, 16(1), pp. 226-233.
Lehane, B. M., O'loughlin, C. D., Gaudin, C., & Randolph, M. F. (2009). Rate effects on penetrometer resistance in kaolin. Géotechnique, 59(1), 41-52.
Lehtonen, V. (2015). Modelling undrained shear strength and pore pressure based on an effective stress soil model in limit equilibrium method. Doctoral thesis. Tampere University of Technology. Publication 1337.
Lehtonen, V. J., Meehan, C. L., Länsivaara, T. T., & Mansikkamäki, J. N. (2015). Full-scale embankment failure test under simulated train loading. Géotechnique, 65(12), pp. 961-974.
Leroueil, S., Tavenas, F., Brucy, F., La Rochelle, P., & ROY, M. (1979). Behavior of destructured natural clays. Journal of Geotechnical and Geoenvironmental Engineering, 105 (ASCE 14657 Proceeding).
Leroueil, S., Kabbaj, M., Tavenas, F. & Bouchard, R. (1985). Stress-strain-strain rate relation for the compressibility of sensitive natural clays. Géotechnique 35, No. 2, pp. 159-180.
Leroueil, S. & Vaughan, P. R. (1990). The general and congruent effects of structure in natural soils and weak rocks. Géotechnique 40, No. 3, pp. 467-488.
Leroueil, S. & Marques, M. E. S. (1996). State of art: importance of strain rate and temperature effects in geotechnical engineering, measuring & modelling time dependent behaviour of soils. ASCE Geotechnical Special Publication 61, pp. 1–60.
225
Leroueil, S., Locat, J., Vaunat, J., Picarelli, L., & Lee, H. (1996). Geotechnical characterization of slope movements In Landslides. pp. 53-74.
Long, M., and Donohue, S., (2010), Characterisation of Norwegian marine clays with combined shear wave velocity and CPTU data Canadian Geotechnical Journal, v. 47, no. July, pp. 709 - 718.
Lunne, T., Christoffersen, H., & Tjelta, T., (1985) Engineering use of Piezocone Data in North Sea Clays, in: proc. ICSMFE-11; San Francisco, 2, pp. 907- 912.
Lunne, T., Berre, T., & Strandvik, S. (1997). Sample disturbance effects in soft low plastic Norwegian clay. In Symposium on Recent Developments in Soil and Pavement Mechanics.
Lunne, T., Berre, T., Andersen, K. H., Strandvik, S., & Sjursen, M. (2006). Effects of sample disturbance and consolidation procedures on measured shear strength of soft marine Norwegian clays. Canadian Geotechnical Journal, 43(7), pp. 726-750.
Lunne, T., Strandvik, S. O., Kåsin, K., L'Heureux, J. S., Haugen, E., Uruci, E., Veldhuijzen, A., Carlson, M. & Kassner, M. (2018). Effect of cone penetrometer type on CPTU results at a soft clay test site in Norway. In Cone Penetration Testing IV: Proceedings of the 4th International Symposium on Cone Penetration Testing (CPT 2018), June 21-22, 2018, Delft, The Netherlands.
Länsivaara, T. (1996), Maan mekaanisen käyttäytymisen mallintaminen. Lisensiaatintutkimus, Geotekniikan laboratorio, julkaisu 34, TTKK
Länsivaara, T. (1999), A study of the mechanical behavior of soft clay, Norwegian University of Science and Technology. NTNU. Doktor ingeniöravhandling, vol. 85.
Länsivaara, T. (2005) Gmax Wizard, Dynaamisten parametrien määritys (Determination of dynamic parameters), Ramboll, 9 s.
Löfroth, H. (2012). Sampling in normal and high sensitive clay–a comparison of results from specimens taken with the SGI large-diameter sampler and the standard piston sampler St II. SGI Varia, 637, 26.
226
Mansikkamäki, J. (2015). Effective stress finite element stability analysis of an old railway embankment on soft clay. Doctoral thesis. Tampere University of Technology. Publication; 1287.
Mataić I. (2016). On structure and rate dependence of Perniö clay. Aalto University School of engineering. Doctoral Thesis.
Mayne, P. W., & Kulhawy, F. H. (1982). K0 - OCR Relationships in Soil. Journal of the Soil Mechanics and Foundations Division, 108(6), pp. 851-872.
Mayne, P. W., & Bachus, R. C. (1988). Profiling OCR in clays by piezocone soundings. In Proc., International Symposium on Penetration Testing. Vol. 2, pp. 857-864. Rotterdam, Netherlands.
Mayne, P. W., & Kulhawy, F. H. (1990). Direct and indirect determinations of in situ K0 in clays. Transportation Research Record, (1278).
Mayne, P. W. (1991). Determination of OCR in clay by piezocone tests using cavity expansion and critical state concepts. Soils and foundations, 31(2), pp. 65-76.
Mayne, P. W. (1992). In situ determination of clay stress history by piezocone model. In Predictive soil mechanics: Proceedings of the Wroth Memorial Symposium held at St Catherine's College, Oxford, 27-29 July 1992, pp. 483-495. Thomas Telford Publishing.
Mayne, P. W., Robertson, P. K., and Lunne, T. (1998). Clay stress history evaluated from seismic piezocone tests. Geotechnical site characterization, Balkema, Rotterdam, The Netherlands, pp. 1113–1118.
Mayne, P.W. (2001). Stress-strain-strength-flow parameters from enhanced in-situ tests, Proceedings, International Conference on In-Situ Measurement of Soil Properties & Case Histories (In-Situ 2001), Bali, Indonesia, pp. 47-69.
Mayne, P.W. (2007) Cone Penetration testing state of practice, NCHRP Project 20-05.
Mayne, P.W., Coop, M.R., Springman, S., Huang, A-B., and Zornberg, J. (2009). State-of-the-Art Paper (SOA-1): Geomaterial Behavior and Testing. Proc. 17th Intl.
227
Conf. Soil Mechanics & Geotechnical Engineering, Vol. 4 (ICSMGE, Alexandria, Egypt), Millpress/IOS Press Rotterdam: pp. 2777-2872.
Mayne, P. W. (2013). Evaluating yield stress of soils from laboratory consolidation and in-situ cone penetration tests. In Sound Geotechnical Research to Practice: Honoring Robert D. Holtz II pp. 405-419.
Mayne, P. W. (2017). Stress history of soils from cone penetration tests. 34th Manual Rocha Lecture, Soils and Rocks, 40, pp. 203-218.
Mitchell, J.K., Campanella, R. G., & Singh, A. (1968). Soil creep as a rate process. Proc. of the American Society of Civil Engineers, Journal of the Soil Mechanics and Foundations Division, 94, pp. 231-254.
Mitchell, J. K. (1976). Fundamentals of Soil Behavior, John Wiley and Sons, Inc., New York.
Mäenpää, J. (2016). Seismisen CPTU-mittauksen käyttö leikkausaallon nopeuden määrittämiseen (Determining Shear Wave Velocity by Seismic CPTU) Diplomityö. Tampereen Teknillinen Yliopisto. Master thesis at Tampere University of Technology (TUT).
Ortigao J.A.R. & Collet H.B. (1988) Errors caused by friction in field vane tests. In Vane Shear Strength Testing in Soils: Field and Laboratory Studies. ASTM International.
Ouyang, Z., & Mayne, P. W. (2017). Effective Friction Angle of Clays and Silts from Piezocone Penetration Tests. Canadian Geotechnical Journal, 2018, 55(9), pp. 1230-1247
Paniagua, P., L'Heureux, J. S., Yang, S. Y. & Lunne, T. L. (2016). Study on the practices for preconsolidation stress evaluation from oedometer tests. In Proceedings of the 17th Nordic Geotechnical Meeting.
Pacheco Silva, F. (1970). A new graphical construction for determination of the pre-consolidation stress of a soil sample. Proceedings 4th Brazilian Conference Soil Mechanincs and Foundation Engineering, Rio de Janeiro, pp. 225-232
228
Rankka, K., Andersson-Sköld, Y., Hultén, C., Larsson, R., Leroux, V., & Dahlin, T. (2004). Quick clay in Sweden. Swedish Geotechnical Institute Report, 65(1).
Ratahallintokeskus 2005 (RHK julkaisuja), B15, Ratahallintokeskus, Helsinki. (In Finnish. In English: Guidelines for embankments stability calculation by the Finnish Transport Agency, publication B15).
Rochelle, P. L., Sarrailh, J., Tavenas, F., Roy, M., & Leroueil, S. (1981). Causes of sampling disturbance and design of a new sampler for sensitive soils. Canadian Geotechnical Journal, 18(1), pp. 52-66.
Robertson, P.K., (1990). Soil classification using the cone penetration test. Canadian Geotechnical Journal, 27(1), pp. 151-158.
Roscoe, K. H. and Burland, J. B. (1968). On the generalized stress–strain behaviour of ‘wet clay’. Engineering plasticity (eds J. Heyman and F. A.Leckie), pp. 535–609. Cambridge University Press.
Roy, M., & Leblanc, A. (1988) Factors affecting the measurements and interpretation of the vane strength in soft sensitive clays. In Vane Shear Strength Testing in Soils: Field and Laboratory Studies. ASTM International.
Sandven, R., Senneset, K., & Janbu, N. (1988). Interpretation of piezocone tests in cohesive soils. In International Symposium on penetration testing; ISOPT-1. 1, pp. 939-953.
Sandven, R., & Watn, A. (1995). Soil classification and parameter evaluation from piezocone tests: Results from the major site investigations at Oslo Main Airport, Gardermoen. Proceedings of CPT’95, 3, pp. 35-55.
Sandven, R., Ørbech, T., & Lunne, T. (2004), September. Sample disturbance in highly sensitive clay. In Proceedings 2nd International Conference on Geotechnical Site Characterisation, ISC.Vol. 2, pp. 1861-1868.
Sandven, R. (2010). Influence of test equipment and procedures on obtained accuracy in CPTu, Proc. CPT'10, Huntington Beach, CA.
229
Schnaid, F., Sills, G.C., Soares, J.M., & Nyirenda, Z. (1997). Predictions of the coefficient of consolidation from piezocone tests. Canadian Geotechnical J. 34 (2), pp. 315-327.
Schofield, A.N. and Wroth, C.P. (1968). Critical state soil mechanics. McGraw-Hill, London, UK.
Schjetne, K. (1971). The measurement of pore pressure duringsampling. In Proceedings of the 4th regional Asian conference, pp. 12–16. Bangkok: ISSMFE.
Schneider, J. A., Randolph, M. F., Mayne, P. W., & Ramsey, N. R. (2008). Analysis of factors influencing soil classification using normalized piezocone tip resistance and pore pressure parameters. Journal of geotechnical and geoenvironmental engineering, 134(11), 1569-1586.
Senneset, K., Janbu, N., & Svano, G. (1982). Strength and deformation parameters from cone penetration tests. Proceedings of the European Symposium on Penetration Testing. ESOPT.
Senneset, K., Sandven, R., Lunne, T., By, T., & Amundsen, T. (1988). Piezocone tests in silty soils. ISOPT-1, Orlando, FL, pp. 955-966.
SGY (1976), Kairausopas III (Soil sampling) Suomen geoteknillinen yhdistys r.y. (Finnish Geotechnical Society) Helsinki, 36 pp., app. ISBN 951-95093-7-2 (in Finnish).
Sheahan, T. C., Ladd, C. C., & Germaine, J. T. (1996). Rate-dependent undrained shear behavior of saturated clay. Journal of Geotechnical Engineering, 122(2), pp. 99-108.
Siddique, A. (1990) A numerical and experimental study of sampling disturbance, PhD thesis, University of Surrey.
Skempton, A.W. and O.T. Jones. (1944). Notes on the compressibility of clays. Quarterly Journal of the Geological Society 100.1-4, pp. 119-135.
Skempton, A. W. (1951). The bearing capacity of clays. In Proc. of Building Research Congress. Vol. 1, pp. 180-189. ICE.
230
Skempton, A. W., & Sowa, V. A. (1963). The behaviour of saturated clays during sampling and testing. Geotechnique, 13(4), pp. 269-290.
Skempton, A. W. (1964). Long-term stability of clay slopes. Géotechnique, (14), pp. 77-102.
Sorensen, K. K. (2006). Influence of viscosity and ageing on the behaviour of clays (Doctoral dissertation, University of London).
Stewart, D. P., & Randolph, M. F. (1994). T-bar penetration testing in soft clay. Journal of Geotechnical Engineering, 120(12), 2230-2235.
Su, S. F. (2010). Undrained shear strengths of clay around an advancing cone. Canadian Geotechnical Journal, 47(10), pp. 1149-1158.
Sällfors G. (1975). Preconsolidation pressure of soft, high-plastic clays. Doctoral thesis, Chalmers University of Technology.
Tanaka, H., SHARMA, P., Tsuchida, T., & Tanaka, M. (1996). Comparative study on sample quality using several types of samplers. Soils and foundations, 36(2), pp. 57-68.
Tavenas, F. and Leroueil, S. (1977). Effects of stresses and time on yielding of clays. In: Proc. of the 9th ICSMFE, vol. 1, pp. 319–326.
Tavenas, F., Flon, P., Leroueil, S., & Lebuis, J. (1983). Remolding energy and risk of slide retrogression in sensitive clays. In Proceedings of the Symposium on Slopes on Soft Clays, Linköping, Sweden. SGI Report. No. 17, pp. 423-454.
Teh, C. I., and Houlsby, G. T. (1991). Analytical study of the cone penetration test in clay. Geotechnique, 41(1), pp. 17-34.
Terzaghi K (1925) Erdbaumechanik auf Bodenphysikalischer Grundlage. Franz Deuticke, Liepzig-Vienna
Terzaghi, K. (1943), Theoretical Soil Mechanics, John Wiley & Sons, New York.
231
Thakur, V. (2007) Strain localization in sensitive soft clays, Doctoral theses at NTNU, Trondheim.
Thakur, V., Jostad, H. P., Kornbrekke, H. A., & Degago, S. A. (2014). How well do we understand the undrained strain softening response in soft sensitive clays?. In Landslides in Sensitive Clays. pp. 291-303. Springer, Dordrecht.
Thakur, V., Degago, S. A., Selänpää, J., & Länsivaara, T. (2017). Determination of remoulding energy of sensitive clays. In Landslides in Sensitive Clays. pp. 97-107. Springer, Cham.
Touiti, L., & Van Impe, W. (2017). Considerations on the Stiffness of Sensitive Soft Soils. In International Congress and Exhibition" Sustainable Civil Infrastructures: Innovative Infrastructure Geotechnology" pp. 20-37. Springer, Cham.
Vaid, Y. P. & Campanella, R. G. (1977). Time-dependent behaviour of undisturbed clay. J.Geotech.Div., ASCE 103, No. GT 7, pp. 693-709.
Vaid, Y. P., Robertson, P. K., and Campanella, R. G. (1979). “Strain rate behavior of Saint-Jean-Vianney clay.” Can. Geotech. J., 16(1), pp. 34–42
Vardanega, P. J., & Bolton, M. D. (2013). Stiffness of clays and silts: Normalizing shear modulus and shear strain. Journal of Geotechnical and Geo environmental Engineering, 139(9), pp. 1575-1589.
Whittle, A. J., & Aubeny, C. P. (1993). The effects of installation disturbance on interpretation of in situ tests in clay. In predictive soil mechanics. Proceedings of the Wroth Memorial Symposium, 27-29 July 1992, St Catherine’s College, Oxford.
Wood, D. M. (1990). Soil behaviour and critical state soil mechanics. Cambridge university press.
Wroth, C. P. (1984). Interpretation of in situ soil tests. Geotechnique, 34(4), pp. 449-489.
Yu, H. S. and Mitchell, J. K. (1998). Analysis of cone resistance: review of methods. Journal of Geotechnical and Geo environmental Engineering, 124(2), pp. 140-149.
232
Yu, H.S., Herrmann, L.R., & Boulanger, R.W. (2000). “Analysis of steady cone penetration in clay.” J. Geotech. And Geoenvir. Engrg., ASCE, 126(7), pp. 594-605.
