Italian National Report – Research and Construction

ASSOCIAZIONE ITALIANA TECNICO ECONOMICA DEL CEMENTO

ANNO / YEAR LXXXAPRILE / APRIL 2010 854

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7637

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l’ industria italiana

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The 3rd International fib CongressWashington DC, May 29 – June 2, 2010

Italian National ReportResearch and Construction

854

SPECIAL ISSUE

RI VI STA DELL’AS SO CIA ZIO NE ITALIANA TEC NI COECO NO MI CA DEL CE MEN TO (AI TEC) l’industria italiana

del Cemento

Contents854 Anno/Year LXXXAprile/April 2010

RESEARCH

CONSTRUCTION

SEISMIC BEHAVIOR

6 Earthquake Engineering of Reinforced Concrete Structures: The ItalianState-of-the-art L. Ascione, E. Cosenza, G. Mancini, G. Manfredi, G. Monti, P. E. Pinto

74 Experimental Research on Seismic Behavior of Precast Structures F. Biondini, G. Toniolo

CONCRETE

80 State-of-the-art on Research on Structural Concrete in Italy M. Collepardi

L’AQUILA EARTHQUAKE

88 Damages of L’Aquila earthquake G. Manfredi

124 Reconstruction between temporary and definitive: the CASE projectG.M. Calvi, V. Spaziante

CIVIL ENGINEERING WORKS

154 Italian High-Speed Network. A special focus on concrete structures• HS railway cable-stayed bridge over Po river• “Piacenza” viaduct• “Modena” system viaducts• “Savena” viaduct• “Caivano” variation structures• Tunnels in the Florence-Bologna stretch of High-Speed Line• New stations for Italian High-Speed Network

188 “Colletta” cable-stayed bridge192 Rio S’Adde viaduct196 Bridge over Vajont creek200 “Cesare Cantù” cable-stayed bridge204 Bridge between La Maddalena and Caprera Islands208 “Don Bosco” bridge. Architecture, white as light212 Viaduct for State Road (SS) 23 216 “Roccaprebalza” viaduct220 “Sandro Pertini” bridge upgrade224 Cable-stayed footbridge over the Frodolfo river228 Bridge over Mazzocco creek 232 Bridge over the Sacco river236 Bridge over the Santa Caterina channel 240 Bridge over the Cimadolmo branch242 “Isola della Scala” bridge244 The “Strada dei Parchi”248 A24 – Completion of the motorway Roma-L’Aquila-Teramo250 “S. Antonio” viaduct254 Viaduct for the Algeria East-West Motorway

3 Foreword (M. Menegotto)Di ret to re re spon sa bi leManaging EditorLaura Negri

CollaboratoriAssistantsMarco Veronesi

Grafica e ImpaginazioneDesign & EditingStudio Mariano - Roma

EditorePublisher

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Tutti i diritti di riproduzione sono riservati. Nessuna parte di questaRivista può essere riprodotta in nessuna forma. La Rivista nonassume la responsabilità delle tesi sostenute dagli Autori e delleattribuzioni relative alla partecipazione nella progettazione edesecuzione delle opere segnalate dagli stessi Autori.All rights of reproduction are reserved. No part of this Magazine may bereproduced in any form whatsoever. The Magazine assumes noresponsibility for the theses put forward by the authors or for the authors'indications of their responsibilities regarding participation in the designand construction of the works.

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Stam pa/Printed by: Grafica Ripoli snc - Via Paterno - Villa Adriana (Tivoli)

In copertinaOn front coverPantheon dome, Rome. Inside view

BUILDINGS

258 MAXXI – Center for the contemporary arts262 “Olympic Palavela”266 New Bocconi University270 Milanofiori 2000 – Corporate Center274 “Acqua minerale San Benedetto” plant278 New Sky Italia Headquarters 282 Light Pavilion286 Agenzia Spaziale Italiana new headquarters290 “Altra Sede” for the Regione Lombardia294 New “Sant’Anna” Hospital 298 “Verdi” Theatre302 Banca Lombarda Center306 Boglietti Palace310 “Cuore immacolato di Maria” parish complex314 “San Giovanni Battista” parish complex318 “Somada” Business Center

Contents

l’industria italiana del Cemento

Edited by A.I.C.A.P. (Associazione Italiana Calcestruzzo Armato Precompres -so) and “l’industria italiana del Cemento – iiC”

The 3rd International fib CongressItalian National Report Research and Construction

The 3rd fib International Congress takes place during (hopefully towards the end of) a global downturn in the economy.

The impact was felt in all fields, including construction industry, and in all countries.

Crises are opportunities for rethinking activities and habits. The need to face high costs of materials and production stimulates

the search for improvements. This parallels the awareness of further needs, not of immediate return, such as saving resources,

building more durable constructions, recover for reuse or recycling at least of materials, or better yet of whole structures or parts

of them.

Design and construction usually take account of safety and economy but more and more they consider too service life

requirements, products’ life cycle assessment, resource saving and other environmental issues, in one word, sustainable

development.

CEB and FIP, now fib, were accompanying and leading the progress of structural concrete design and construction during the

past half century and have been a great partner in cultural interchange for our country.

AICAP, the national association for structural concrete, mirrors fib and takes advantage also of its actions to contribute in its

turn to the dissemination of knowledge and the improvement of practice. Among other initiatives directed toward encouraging

engagement in better design and execution, AICAP launched for the first time a national award for the best concrete structures

for buildings and civil engineering works, that will be given every second year, and edits this National Report, where a

selection of works using structural concrete and completed in Italy in the last four years is illustrated, among which both 2009

award winners. Rome’s Pantheon is the exception, with its record 43.4 m dome shown on the cover, cast altogether 20 centuries

ago, using a lightweight concrete quite similar to the modern one.

Research is playing a decisive role in advancing techniques to fulfil new requirements and has progressed in Italy during these

years, promoted also by the industry, more sensitive to it while under pressure of the crisis. Therefore, this Report also describes

some results related to the national state of the art.

Italy is a highly seismic country, thus most researchers in recent past have worked in the field of seismic engineering, aiming at

better design of new structures as well as at assessing and retrofitting older ones.

Unfortunately, one year ago this science was called to prove its ability after the L’Aquila earthquake. The authors of the articles

in this Report were personally involved also in the field operations. Their contributions helped the surveys in the immediate

post-event days, the directives for the repairs and the implementation of the CASE project. The latter has represented quite a

successful system response, up-to-date and praised worldwide, that provided a large population of homeless with sets of

permanent houses, built in few months on seismically isolated platforms, located in sites purposely selected and equipped in the

city neighbourhoods.

AICAP is then glad to present the 2010 National Report, which comes in to being too thanks to the renowned magazine iiC,

which is publishing this special issue, in spite of the difficult times.

Marco MenegottoHead of the Italian delegation to fib

3iiC•4/2010

Foreword

RESEARCH

6 RESEARCH - Seismic behavior

2–FC: Calibration of Confidence Factors3–IRREG: Assessment of the Nonlinear Behavior of Buildings, with

Emphasis on Irregular Ones4–MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)

Buildings5–TAMP: Influence of Infills on Structural Response6–SCALE: Behavior and Strengthening of Stairs7–NODI: Behavior and Strengthening of Beam-Column Joints8–BIAX: Behavior and Strengthening of Columns under Combined

Axial Load and Biaxial Bending and Shear9–PREFAB: Behavior and Strengthening of Prefabricated Industrial

Structures

For the sake of clarity of exposition, given the large variety of differentsubjects, the structure of the paper, in each of its sections, follows theTask organization above.

1.1 MND: Non-Destructive Methods for the Knowledge of ExistingStructuresTask MND specifically focuses on the knowledge of the constituentmaterial properties of Reinforced Concrete (RC) existing structures.Particularly, Task MND is devoted to the estimation of the in-situ con-crete strength by using destructive and non destructive methods. Dataon material properties from several in-situ and laboratory investigationswere collected and analysed with the major objective of defining reli-able as well as not very expensive procedures and criteria for the esti-mation of the in-situ concrete strength. Further, methods for the treat-ment of the uncertainty that characterizes experimental data obtainedthrough in-situ and laboratory investigations were analysed and sometheoretical simulations of the influence of material properties on theseismic capacity of existing buildings were carried out.

1.2 FC: Calibration of Confidence FactorsA fundamental phase in the assessment of existing reinforced concretebuildings and in their strengthening design is the knowledge processthat one has to follow to acquire the necessary information. This isbased on the collection of different kinds of information regarding: a)the structural system configuration, b) the materials strength, c) thereinforcing steel details, and d) the conditions of the structural ele-ments. The Italian Code (OPCM 3431, 03-05-05, Annex 2) as well as the mostadvanced International Codes (FEMA 356, EC8 Part 3) specifies datacollection procedures about the configuration of the structural system,as well as material strength and condition of the structural elementscomprising the building, and ensuing Confidence Factors (CF) to applyto the mean values materials properties, based on the quantity and

This extended paper summarizes part of the results, those related toreinforced concrete (RC) structures, of the largest research program

on earthquake engineering ever held in Italy. The ReLUIS project fund-ed by the Department of the National Civil Protection granted15.000.000 € and involved more than 600 researchers all over the coun-try between 2005 – 2008. The ten research tasks (lines) ranged from theseismic risk of existing structures to new design paradigms, includinggeotechnical earthquake engineering issues and innovative approaches toseismic risk reduction as earthquake early warning systems as well asemergency management. In the following the ReLUIS research lines regarding: (I) assessment ofexisting RC buildings (Line 2); (II) bridges (Line 3); and (III) retrofit ofRC structures via innovative materials (Line 8), coordinated by theauthors, are described in their development and findings. The three sec-tion of the paper are structured as stand-alone, including their own intro-duction, conclusions, vision and references, for readability purposes. Moredetails, references and products of the project may be found in theresearch section of the ReLUIS website (http://www.reluis.it/). Because of timeliness of the ReLUIS project and the involved scientists,the work described in the following is likely to express the state of the artof earthquake engineering concerning reinforced concrete structures inItaly.

I - ASSESSMENT AND REDUCTION OF THEVULNERABILITY OF EXISTING REINFORCEDCONCRETE BUILDINGS

1. INTRODUCTION

Research Line 2 focuses on the assessment of the seismic performanceof existing reinforced concrete buildings, covering a wide spectrum ofproblems, each one treated within a single Task. These aspects spanfrom those related to the preliminary knowledge phase, to the use ofnonlinear assessment methods, while placing emphasis on peculiarmodeling problems, such as those related to the presence of stairs,infills, beam-column joints, and biaxial behavior of the elements. TheResearch Line is also devoted to the study of mixed-type (masonry/RC)buildings and of prefabricated industrial buildings. The following Task list collects and organizes the entire scientific activ-ity:

1–MND: Non-Destructive Methods for the Knowledge of ExistingStructures

Earthquake Engineering of ReinforcedConcrete Structures: the Italian State-of-the-art

7RESEARCH - Seismic behavior

quality of the information gathered (the so called Knowledge Level). Inthe current approach, the CFs are given through tables.Aim of the Task has been: a) evaluation of CF effects on the assessmentof buildings seismic performances; b) development of a procedure forthe evaluation of concrete and steel strength, to be reliably used inassessing members capacity; c) new definition of CF, evaluated by aclosed-form equation as a function of number, kind and reliability ofeach testing method employed, and of the reliability of prior informa-tion.

1.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular OnesTask IRREG deals with problems related to the definition of plan andelevation irregularity and the effects of irregularity on the structuralbehavior and its prediction though different methods of analysis pro-vided by the current design codes. More specifically, the main focuses of this tasks are:a. Study of the definition and effects of plan irregularity on the responseof RC buildings up to the Ultimate and Collapse limit states;b. Comparison and calibration of different linear and nonlinear meth-ods for reinforced concrete structural members, with emphasis onpushover and nonlinear dynamic analyses;c. Comparison between research-oriented and professional-orientedstructural analysis software, in order to identify analytical tools that sat-isfy both modeling precision and computational speed.

1.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)BuildingsThe work has been developed on the evaluation of the seismic responseof mixed-type buildings behaving as parallel systems, with regard toboth local (interaction between masonry and RC elements) and globalfeatures, by performing a series of non linear numerical analyses. The research activity has been focused on: a. the classification of the main geometrical characteristics of such kindof buildings and the study of their response – behaving as parallel sys-tems – subjected to horizontal forces;b. the problems concerning the modelling of mixed-type buildings andon the distribution of the seismic action between masonry and rein-forced concrete elements by performing a series of numerical analysesobtaining the capacity curves of individuals resistance elements andthe building as a whole.

1.5 TAMP: Influence of Infills on Structural ResponseSeveral theoretical and numerical analyses, and, above all, the damagedistributions on buildings that have suffered an earthquake show that

masonry infills can modify substantially the expected seismic responseof framed structures although special devices connecting the infill pan-els with the surrounding meshes of frame are not applied.In spite of that, most seismic codes (the more recent too) give some pro-visions in order that the resisting elements of frame bear theunfavourable effects of a non-uniform infill distribution in plan or/andin elevation, but they do not suggest any procedure to quantify theseeffects or the favourable lateral stiffness and resistance contributionsthat infills give when they are uniformly located.This gap occurs since the influence of the masonry infills on the seis-mic response of framed buildings is a still open research topic, whereunivocal and general results do have not been achieved. The presentstudy refers to this subject with the following main objectives:- mechanical characterization of the masonry infill kinds that are com-monly utilized in the Italian country by means of experimental tests ontheir components (resisting elements and mortar) and masonry samples;- experimental investigation on infilled meshes of RC frames, with theaim of calibrating a pin-jointed equivalent diagonal strut model.A companion study has also been devoted to verifying the influence ofthe infills on the seismic response of RC framed structures; for this pur-pose, shaking table tests on a 1:2 scaled 3D building and numericalnonlinear analyses of multi-storey frames have been carried out.

1.6 SCALE: Behavior and Strengthening of StairsThe main objectives of this Task are the following: Identification of themain stairs typologies used in the past construction practices;Numerical investigation of the influence of the stair substructure on thestructural seismic response. In particular, both global and local seismicperformance have to be investigated with reference to frame and stairsmembers connections; Construction of building sub-assemblages,including a stair substructure, for experimental tests execution, specif-ically targeted at understanding their seismic performance.

1.7 NODI: Behavior and Strengthening of Beam-Column JointsThis Task aims at investigating the experimental behaviour of RC struc-tural members, particularly beam-column joints without or withstrengthening, thus providing a contribution to a more reliable evalua-tion of the seismic vulnerability of RC existing buildings. In particularof great interest is the understanding and the validation of capacitymodels relevant to the joint panel zone in beam-column sub-assem-blages reported in the literature and in seismic codes. Further, there isa need of knowledge in the field of strengthening and retrofit systemsthat can be used taking into account the actual geometry of joints: e.g.presence of slab and other framing elements that could prevent aneffective arrangement of the retrofitting system. To this purposes, wide

Luigi Ascione1, Edoardo Cosenza2, Giuseppe Mancini3, Gaetano Manfredi2, Giorgio Monti4 and Paolo E. Pinto41 Università degli Studi di Salerno, Fisciano (SA), Italy2 Università degli Studi “Federico II”, Naples, Italy3 Politecnico di Torino, Turin, Italy4 Sapienza Università di Roma, Rome, Italy


bibliographic research on the experimental investigations on beam-col-umn joints and on different repairing/strengthening techniques as wellas experimental researches on different joint specimens have been car-ried out.

1.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearModern approach to safety assessment of existing reinforced concretestructures and design of strengthening interventions, in particular thoseaimed at increasing ductility of columns, are based on enhanced andcomplex methods for structural analysis (seismic demand), but also onthe availability of data concerning performances of members at failure(seismic capacity).On the other hand, common constructions are not necessarily affectedby regular shapes and/or regular distribution of seismic resistant sub-structures, so that seismic actions result in complex deformation pathson columns and in general on compressed resisting members.This is the reason why Task BIAX research activity has been devotedto provide an insight on the response of r.c. members subjected to biax-ial bending and axial load. In particular, some aspects have beenanalysed in detail. In compliance with the overall objectives of theresearch programme as a whole, Task BIAX duties were the definitionof a set of reliable and well documented data and procedures concern-ing: (a) rotation capacity of r.c. members subjected to generalised bend-ing and axial forces; (b) development of simplified methods of analysisfor general r.c. cross sections for design safety checks; (c) developmentof refined methods for assessment of generalized moment-curvaturerelationships of cross sections; (d) extension of results to r.c. membersreinforced with FRP materials.

1.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresThe assessment and reduction of seismic vulnerability of a widespreadcategory of precast structures typically used for industrial buildings isa topic of high importance. The production of these structures startssince from the years ‘50s of last century with elements and constructionsolutions which had a relevant evolution through the subsequent times.It is a social important interest to know the state of this wide buildingheritage with respect to its seismic vulnerability so to address, follow-ing rational criteria, possible interventions of upgrading of inadequatestructures.

2. BACKGROUND AND MOTIVATION

The essential motivation for each Task stems from recognizing some

gaps in the code, related to certain procedural and methodologicalaspects in the seismic assessment of existing buildings.Specifically, for Task MND, it is noted that, in the case of non-destruc-tive testing, a lack of clearness exists about the relative importance ofsuch tests with respect to destructive ones for the evaluation of materi-al properties.The data acquisition modality has immediate consequences on the cal-ibration of Confidence Factors, treated in Task FC, which may assumedifferent values from those given in the code, in case one accepted toinclude results from non-destructive tests in addition to – or even insubstitution of – destructive ones.Moving to the level of analysis methods for seismic safety evaluation,the need of a deeper insight into the usual assessment techniques isrecognized, with particular emphasis to their predictive capacity whendealing with irregular buildings, dealt with in Task IRREG. It would beexpedient to identify, for example, a synthetic parameter capable ofquantifying the level of irregularity and, possibly, an associated applic-ability threshold that helped selecting the most appropriate assessmentmethod, be either of simplified nature, such as pushover analyses, ormore refined, such as nonlinear dynamic analyses.For mixed-type (masonry/RC) buildings, the lack of code provisions,which could guide the designer towards the assessment of the com-pound behavior in a unitary manner – also accounting for interfaceactions between different constructive typologies – , is absolutely strik-ing and appropriate methods and provisions should be identified inTask MIX.A different remark is needed for the influence of infills on the structur-al response, where the motivation for the research carried out by TaskTAMP stems from the absence of code provisions to account for themeaningful interactions that develop between infills and structure, withsignificant effects, both, at the global level (behavior factors), and at thelocal level (collapse mechanisms induced by the presence of localizedforces).The following three Tasks SCALE, NODI and BIAX, related to stairs,beam-column joints, and biaxial behavior of columns, respectively, dealwith three aspects where the need of providing the designers with oper-ational tools is imperative, especially for as regards the assessment ofthe capacity of such elements. In the first case, the motivation is toobtain a deeper insight about the influence of stiffening elements – thestairs – on the structural response. Generally, when modeling the resist-ing system, these elements are either neglected or modeled with unac-ceptable simplifications. In the second case, that relative to beam-col-umn joints, it is necessary to develop more accurate capacity models,accounting for the joint panel behavior, but also for the presence of sec-ondary phenomena significantly modifying the resisting mechanism,


rials strengths.Difference in the knowledge procedure about the single structural para-meters and the actual possibility of propagation to the structure as awhole of information gathered on single members unlikely can beaccounted for by a single CF to be applied to mean materials strengthvalues.Material strength is characterized by, both, an intrinsic spatial variabil-ity and an epistemic uncertainty, caused by either workmanship (forinstance not compliance with the original project, execution of struc-tural elements in different times with different materials strength), orreliability of testing methods, or degradation of material properties withtime, or a combination of the former. On the other hand, amount anddetailing of reinforcement, defective detailing, etc., neglecting theintrinsic uncertainties, are characterized by epistemic uncertaintiesonly, mainly due to lack of the original project and/or not compliancewith it; collected data on one structural element are certain but do notallow to eliminate uncertainties about other elements.Objectives of recent studies (Franchin et al. 2008, Jalayer et al. 2007)have been the evaluation of the effect of CF on the assessment of thestructural reliability and new proposals for calibration of a CF.

2.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular OnesMany of the existing RC structures were built without accounting forseismic actions, thus much attention has been paid in recent years tothe development of reliable methods of analysis and assessment. Linearmethods seem inappropriate in most cases; many current seismic codesand guidelines include provisions for nonlinear analysis (Eurocode 8,2003a, EuroCode 8, 2003b, FEMA 356, 2000, ATC-40, 1996), whichseems to be the natural choice for existing structures subjected to mod-erate and strong design earthquakes. This is obviously a big issue inItaly, a seismically active country where many buildings were erectedin the ‘60s, ‘70s and ‘80s usually accounting for only gravitationalactions. Furthermore, the new seismic zonation classifies areas previ-ously considered non-seismic as seismic, thus new assessment areneeded even on recently built structures. Following the publication of the most recent Italian Seismic Codes, theReLUIS program of the Italian Department of Civil Protection intendsto validate and improve the new code, to propose alternate procedureswhen deemed necessary, and to provide practical examples to practic-ing engineers. These activities are particularly important for newmethodologies, such as nonlinear methods of analysis. Focus of thesestudies is not only the application of the nonlinear methods of analysis,but also the use of the results of the nonlinear analyses to assess theseismic vulnerability of structure.

such as concentrated forces ensuing from hook-bent bars, or bond-slipin rebars. In the third case, that of biaxial behavior of columns, themotivation for research stems from the awareness that the capacityequations currently available in the code are calibrated on the mono-axial behavior, besides, without interaction with shear.Finally, for the prefabricated structures studied in Task PREFAB, theintention is to provide the normative framework with more completeindications than those currently available, with the objective of bridg-ing the current information gap through the proposition of specificguidelines for the seismic assessment and strengthening of such build-ings.The above considerations are expanded in the following sections.

2.1 MND: Non-Destructive Methods for the Knowledge of Existing StructuresModern seismic codes require that a knowledge level (KL) is defined(e.g. 3 KLs in EC8 part 3: limited, normal and full knowledge) in orderto choose the admissible type of analysis and the appropriate confi-dence factor values in the evaluation. Among the factors determiningthe KL, there are the mechanical properties of the structural materials.In RC structures, the compressive strength of concrete has a crucial roleon the seismic performance and is usually difficult and expensive toestimate. Reliable procedures to take into account the factors influenc-ing the estimation of in-situ concrete strength, particularly in case ofpoor quality concrete, are not currently available. According to variouscodes (e.g. in Europe EC8-3, in Italy NTC 2008) estimation of the in-situ strength has to be mainly based on cores drilled from the structure.However, non-destructive tests (NDTs) can effectively supplement cor-ing thus permitting more economical and representative evaluation ofthe concrete properties throughout the whole structure under examina-tion. The critical step is to establish reliable relationships between NDTresults and concrete strength. The approach suggested in most codes(e.g. in EC8-3) is to correlate the results of in-situ NDTs carried out atselected locations with the strength of corresponding cores. Thus, NDTscan strongly reduce the total amount of coring needed to evaluate theconcrete strength in an entire structure.

2.2 FC: Calibration of Confidence FactorsData collected for the assessment of a building are obtained from avail-able drawings, specifications, and other documents for the existing con-struction, and must be supplemented and verified by on-site investiga-tions, including destructive and nondestructive examination and testingof building materials and components. As a function of the completeness of as-built information on buildings(Knowledge Level) the Italian Code specifies different analysis methodsand Confidence Factors (CF) to be applied to the mean values of mate-


2.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)Build ingsFrom the early 20th-century the combined RC-masonry buildings wide-ly spread in European, Mediterranean and Southern America countries.Despite the diffusion of this combined building typology, the interna-tional guidelines have not followed building evolutions; nowadays,international guidelines are not exhaustive to deal with specific prob-lems of this building typology, such as: horizontal loads repartition, con-nections between different technology elements and over strength fac-tor. The Argentinean guideline (NAA-80) points out the fundamental roleperformed by slab, on the base of the own relative stiffness, for sharingseismic action between vertical different technology resistant elements.During the years, the Italian guidelines have provided discordant indi-cations. The Italian guideline (D.M. 1996) suggested to assign the total seismicaction to masonry walls in the case of new buildings, while for existingbuildings, the combined RC-masonry buildings should be consideredas structural elements typology that prevalently supports horizontalloads, generally masonry walls. Regarding masonry buildings, the Italian guideline (O.P.C.M. 3431)allows to employ different technology elements to support gravity loads,only if the seismic action is fully supported by elements with the sametechnology. In the need to consider the collaboration of masonry wallsand different technology systems to sustain the seismic action, a non-linear analysis should be carried out according to O.P.C.M. 3431. Thelatest Italian code (D.M. 2008) confirms the instructions provided byO.P.C.M. 3431 by which the real structural system should be consid-ered with particular attention to, both, stiffness and strength of theslabs, and the connections effectiveness between the structural ele-ments.

2.5 TAMP: Influence of Infills on Structural ResponseThe very numerous papers that concern the behaviour of infilled framesare quite uniformly distributed within the last forty years (Figure 1a).This subject has kept topical mainly because of the following reasons:- different materials that can be utilized for the infill panels; - difficul-ty in modelling the frame-infill interactions; - high number of parame-ters governing the lateral response of an infilled mesh of frame. It follows that the models that have been proposed by differentresearchers are strongly related to the kind of masonry infills that wereexamined and to the experimental tests validating the models them-selves. As regards this, Figure 1b shows how the section of the equiva-lent diagonal strut by different authors is differently related to the samesynthetic parameter lh’, which depends on the geometrical and

mechanical properties of the two sub-systems (frame and infill). In thefigure, w denotes the height of the section and d is the diagonal lengthof the infilled mesh (Refs. 3-7). Further discordant results can be foundconsidering models including the post-elastic hysteretic behaviour.Therefore, the main motivation of the present study lies in the non-availability of an univocal approach, able to define an appropriate infillmodel depending on the properties of the masonry utilized.

2.6 SCALE: Behavior and Strengthening of StairsIn general the presence of a stair creates a discontinuity in a regularreinforced concrete skeleton frame made of beams and columns; in fact,from the geometrical point of view, a stair is composed by inclined ele-ments (beams and slabs) and by short (squat) columns. These elementscontribute to increase the stiffness of the stair due to the elastic behav-iour of inclined elements and of squat columns. For these reasons theelements that constitutes the stair are often characterized by a high seis-mic demand: the squat columns are subjected to high shear force thatcan lead to a premature brittle failure; the inclined beams, differentlyfrom the horizontal beam, are defined by high variation in axial forcesthat can modify the resistance and deformability of all these elements.Although this is well known, no studies have been conducted byresearchers to evidence the role of stairs on the seismic capacity ofexisting RC buildings; the identification of the weakest elements of thestructure and the failure type considering the presence of the stairs areof particular interest. In this way, the knowledge of structural solutionsand design practice of stairs is an important step in order to define theirreal geometric definitions and to understand their seismic perfor-mances.

2.7 NODI: Behavior and Strengthening of Beam-Column JointsObservation of the damage caused by strong earthquakes on RC build-ings designed to resist only to gravity loads showed that the main mech-anisms that characterize structural collapses are, beyond the yielding ofprimary elements such as column and beams, slippage of longitudinalbars in columns and beams and joint failures (e.g. Braga et al. 2001). Based on these observations and on the results of extensive experimen-

Fig. 1- a) Main published papers ; b) Ratios w/d proposed by different authors.

a b


tal campaigns, some provisions were inserted in the Italian technicalregulations imposing performance criteria for the design of new RCstructures placed in seismic zones. The capacity design approach pro-vided by current Italian and European codes (NTC 2008, CEN 2004)aims at preventing brittle failure mechanisms in beam, column andjoint members as well as at ensuring a weak beam-strong column glob-al collapse, being more favourable in terms of overall ductility. Forexisting RC structures, designed without anti-seismic criteria, there isthe problem of a reliable assessment of their seismic resistance also inorder to identify the more appropriate strengthening intervention sys-tems. Improving knowledge on capacity models, particularly as for typ-ical Italian building structures, is the principal thrust for the researchactivity of task 7 (NODI) in the framework of DPC-ReLUIS 2005-2008Project.The workflow in terms of literature review, experimental testing andnumerical analysis performed by the RUs involved in the task is, then,finalized to the analysis and validation of the provisions of Italian andEuropean codes in order to improve them and make them more adher-ent to the reality of the Italian existing RC building stock.

2.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearModeling of reinforced concrete members is really a traditional topic ofstructural engineering, but some aspects need further developmentwhen seismic assessment of existing constructions is concerned. In fact,well-established results for modern concrete structures do not cover alarge population of members built with obsolete materials and structur-al details like smooth bars. This is actually a relevant issue, since bondbetween steel bars and the surrounding concrete is poor and anchoringmechanical devices can play a relevant role in the development of plas-tic deformation and therefore of the drift capacity. A number of models characterized by different models of complexitycan be found in the National and International technical literature (fib,2003; Panagiotakos and Fardis, 2001; Park and Paulay, 1975) and pro-vide an estimation of the rotation capacity at yielding and at failure ofcolumns member. However, they generally are able to well representresponse of r.c. members where deformed bars are used. Based on sucha background, the research on columns subjected to biaxial bendingand axial force has been conceived to cover the lack of knowledge at thetime of proposal. In fact, advances in seismic Codes and increasingneed of data for design purposes can be addressed among the primarymotivations of Task BIAX. On the other hand, since tools for the esti-mation of strength and deformation of bare cross sections were not soconsolidated, a specific focus on columns strengthened with FRP mate-rials is certainly of applicative interest. This circumstances confirm the

rational basis of the research and above all the actual usefulness of itsresults.

2.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresPrecast structures passed through the check of weak and strong earth-quakes and have been submitted to a wide specific experimental andtheoretical investigation performed in the main international researchcentres. From these experiences some key aspects turned out to bedeterminant for the good seismic behaviour of precast structures. Thesekey aspects are listed below:- dry friction supports, not suitable to avoid the loss of bearing;- diaphragm action, important to avoid joint distortions;- lateral supports, necessary to avoid the overturning of beams;- 2nd order effects, to be considered to avoid early collapses;A positive condition of the existing buildings of concern is the possiblepresence of a bridge-crane which required a structural design with rel-evant horizontal forces and a proportioning of the columns which couldbe adequate also for seismic action even in the presumption of low duc-tility.The regulation in force for the design of structures in seismic zones atthe time of construction is obviously a conditioning aspect which affectsthe seismic capacity of existing buildings. Actions and rules for designhave been taken from that regulation which may result inadequate onthe base of the today knowledge. The problem concerns the seismiczoning on one hand and the design criteria on the other.

3. RESEARCH STRUCTURE

In the following sections, the objectives pursued in each Task aredescribed.

3.1 MND: Non-Destructive Methods for the Knowledge of Existing StructuresResearch has been mainly focused on the evaluation of the role of themain factors affecting the estimation of the in-situ concrete strengththrough destructive and non-destructive tests, on the determination ofthe design concrete strength, on the evaluation of the possible damageon core specimens due to drilling and, finally, on the load bearingcapacity of structural members subjected to drilling before and afterrestoration interventions. Other activities were mainly devoted to ana-lyze the correlations between the various methods and the possible spa-tial variability of concrete properties throughout the surveyed members.Regarding data collection, a large amount of experimental data fromdestructive and non destructive in-situ investigations on real strucureswere collected and analysed, also through treatment of uncertain vari-


ables with different mathematical nature. In addition to the above top-ics, the seismic behavior assessment of buildings with structure com-posed of unidirectional RC frames was carried out by means of non-destructive in situ tests, with the objective of estimating their horizon-tal load-carrying and dissipative capacity. Finally, another importantobjective was the evaluation of the dispersion of experimental resultsfrom non-destructive measurements based on a critical review of datareported in the literature.

3.2 FC: Calibration of Confidence FactorsThis Task had two objectives. The first one was to propose a methodol-ogy for the calibration of the CF for materials strength, taking intoaccount the uncertainties characterizing existing building and theeffects on the reliability of the assessed structural performance. Amethodology was also sought for the evaluation of material strength bydestructive and non destructive in situ testing methods taking intoaccount the relevant reliability. The procedures were meant to be basedon the application of the Bayesian method. The proposed methodologyand the equation developed for FC have been validated on several sim-ulated cases and on tests made on several buildings. The second objec-tive was to develop a probabilistic methodology for seismic assessmentof existing buildings taking into account explicitly the uncertainties inthe material properties and the structural detailing parameters andimplementing the available test and inspection results. This methodol-ogy may be used for determination of confidence factors.

3.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular OnesThe main objectives are:Validation of available modelling alternatives for RC buildings, mainlylumped-plasticity and distributed plasticity models, both in commercialand research software.Validation of current methods of analysis for the seismic assessment ofexisting RC buildings, with emphasis on nonlinear methods and theirapplicability to plan-irregular buildings.The above validations were carried out through the analysis of severalbuildings selected by the different research units. Three buildings(shown in Figure 2) were selected as common tested structures: one isa doubly symmetric rectangular building, one is an L-shaped building,and the third is a rectangular building with an internal court. Thesebuildings are representative of the structural buildings commonly foundin Italy. Several commercial and research programs were used for thenonlinear analyses, including SAP2000, OpenSees, Midas, etc.The final objective of this task is the compilation of a document thatcontains: an introduction to nonlinear modeling of RC buildings and to

the nonlinear methods of analysis of the European seismic codes: adescription of the main sources of nonlinearities in existing RC build-ings: the application of different modeling techniques to the seismicvulnerability assessment of the three building mentioned above. Thedocument is intended to be a primer for practicing engineers who wantto use nonlinear methods of analysis.

3.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)BuildingsWith reference to the first goal, technical literature and internationalguidelines have been studied in order to define the classification of themain geometrical characteristics of such kind of buildings and the studyof the response of mixed-type buildings – behaving as parallel systems– subjected to horizontal forces. With reference to the second goal, a series of numerical analyses basedon different and progressively refined modelling assumptions have beenperformed in order to investigate the seismic action distribution bet -ween different technology elements, changing the size and then thestiffness of the RC elements, but retaining the geometry of the buildingand comparing the seismic behaviour of the mixed-type building withthe original masonry one. Pushover analysis have been performed byusing a lumped model to evidence critical zones and possible failures.

3.5 TAMP: Influence of Infills on Structural ResponseA first experimental investigation was devoted to determining themechanical properties of three typical kinds of resisting elements, com-monly used for infill masonry, and of the mortar utilized for their assem-

Fig. 2- Tested buildings.


bly. Then, several infill samples were subjected to compressive tests byassuming orthogonal or parallel loading directions with respect to themortar layers. Further results were obtained under diagonal compres-sive loading, to determine shear modulus and resistance. At the end ofthis phase, the experimental values of elastic moduli and resistanceswere compared with the values that Italian M.D. ’87 provides by link-ing the mechanical properties of masonry elements to those of theircomponents.A second phase of the experimental research was devoted to acquiringthe response of square infilled meshes of RC frames subjected to acyclically varying lateral forces. Two 1:2 scaled samples were tested foreach of the three kinds of infill that had been mechanically character-ized previously. The results of these tests have made it possible to cal-ibrate the hysteretic model of pin-jointed diagonal strut proposed inCavaleri et al., 2005.A further experimental investigation was carried out by means of shak-ing table tests on a 1:2 scaled 3D infilled RC frame, reproducing anactual non-infilled building, previously subjected to pseudo-dynamictests at the ELSA-JRC-Ispra. These tests had the following objectives:- to quantify the lateral stiffness and resistance contributions that infillscan provide; - to verify the influence of the infills on the crack distrib-ution and the collapse mechanism. The same objectives were pursuedby nonlinear numerical analyses on multi-storey RC frames subjectedto natural seismic accelerograms. These analyses also showed the neg-ative effects of non-uniform infill distribution along the height. Another experimental campaign on infilled r.c. frames (1:1/2 scale), onmaterials (concrete, steel, blocks and mortar), and on subassemblages(small panels) has been performed. These tests had the objective of cal-ibrating equivalent strut models, through comparison of experimentalresults on bare and infilled frames, in order to evaluate the infill con-tribution as well as its uniaxial force-displacement relationship. Theconstitutive models for masonry infills have been also calibrated inorder to predict the cyclic response of infilled frames.

3.6 SCALE: Behavior and Strengthening of StairsWith reference to the first goal, several available manuals and books atthe time of construction have been studied in order to define the typol-ogy classification and the corresponding evolution of this classificationduring the years with the increasing knowledge on the use of the mate-rials and of computational machines. An analysis of the codes usedfrom 1909 to the 1980ies has been conducted in details with a criticaljudgement based on the actual knowledge. Examples of stairs designedfor only gravitational loads have been studied with reference to differ-ent typologies.With reference to the second point a series of numerical analyses based

on different and progressively refined modelling assumptions and crite-ria has been performed in order to investigate the principal failuremodes. A critical study has been conducted on the different shearstrength formulations present in literature (Biskinis et al.2004; Sezen etal. 2004; Zhu et al., 2007), in order to simulate potential shear failurein squat columns, which can be easily found in most buildings. Thepushover analysis by using a lumped model has been performed to evi-dence critical zones and possible failures. With reference to the third point, a test set-up has been defined in orderto investigate the experimental behaviour of a building sub-assem-blages, including a stair substructure.

3.7 NODI: Behavior and Strengthening of Beam-Column JointsA wide experimental campaign on beam-column joints representativeof typical members present in Italian existing buildings was planned,designed and carried out. In particular, the research activities weredevoted to outline the influence of some parameters on the mechanicalbehaviour and the failure mechanism of the joints, such as axial force,amount of reinforcing steel and earthquake design level. Furthermore,the research focused on the code expressions for the evaluation of theultimate rotation of RC elements in order to highlight possible discrep-ancies between the theoretical and experimental results. Anotherresearch objective was the analytical modelling of beam-column jointsby using DIANA software to analyze the main parameters affecting theirseismic performance and, specifically, the analytical modelling ofexperimental tests conducted on external joints. Other experimentaltests on beam-column joints relevant to existing buildings were per-formed as well, following an experimental program complementary tothe above-mentioned one, that is, in this case specimens reinforced withsmooth bars were tested and, in some cases, after the first test, jointswere retrofitted to evaluate the effectiveness of some retrofit systems.Finally, supplemental activities followed two main branches: on onehand, a series of either reinforced or unreinforced base joints were test-ed thus evaluating the performance of several strengthening systems,and, on the other hand, a wide database of tests on beam-column jointswas built and analyzed.

3.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearThe activities have been planned to cover four main objectives: (1)review of technical literature with specific reference to available exper-imental data; (2) development of refined and simplified models for bareand FRP reinforced members; (3) experimental activity on columnssubjected to cyclic actions; (4) drafting of a technical report summariz-ing the main applicative aspects of the work.


3.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresThe study envisages a preliminary classification of the industrial prefab-ricated building typologies existing in Italy, from which the most frequentcharacteristics of element-to-element connection types will emerge. Thisfirst cataloguing phase is then followed by a purely experimental phase inwhich some connections, identified as more vulnerable (e.g., friction con-nections), are subjected to a series of cyclic tests to simulate seismic con-ditions. Results and information obtained from the experimental tests willserve as a basis to develop practical models for assessing the capacity ofsuch connection zones and to orient towards the definition of criteria andtechniques for strengthening interventions.The main results obtained by each Task are summarized in the follow-ing sections.

3.10 MND: Non-Destructive Methods for the Knowledge of Existing StructuresThe results obtained during the Project are mainly made up by the exe-cution and analysis of experimental investigations either on in situ realstructures or on laboratory specimens, by the implementation of someprocedures to estimate the in situ concrete strength and by the uncer-tainty treatment of the structural characteristics of existing strucures.A wide experimental program was carried out, comprising more than 20RC beam and column members, several hundreds of non destructivetests (NDTs) and more than 50 destructive tests (cores). Analysis ofresults has shown a large scatter of the core concrete strength both in asingle member and among members extracted by the same story of abuilding. Lower scatters have been detected for the NDT results withthe exception of the surface ultrasonic velocity (see Figure 3). As aresult of these findings, the role of some factors influencing the in situconcrete properties has been carefully evaluated, and some criteria tosuitably select locations for sampling have been provided. A procedurefor the evaluation of the concrete strength based on the Sonreb method,using both core and NDT measurements, has been set up and widelyvalidated, clearly showing its higher prediction capacity when com-pared to the relationships currently available in the technical literature.It requires that the relationship between the in situ concrete strengthand the NDT measurements is experimentally derived for the specificconcrete under test. As for the possible damage on core specimens due to drilling, theresults have shown that the strength reduction suffered by cores can besignificantly influenced by the original strength value of the in situ con-crete. Consequently, adopting a constant coefficient to take into accountdrilling damage, as suggested in the technical literature, can determineincorrect results. On the contrary, it appears suitable adopting coeffi-cient values, obtained during the research, which are inversely propor-

tional to the core strength as provided by the compression test. Finally,some important results regarding the effect of core drilling on the struc-tural members, performing tests before and after a possible restoration.Further, some factors influencing the relationship between the “local”strength provided by core specimens and the in situ strength of thestructural member as a whole, have been highlighted.Regarding the variability of concrete mechanical properties throughoutsingle structural members and among different sampling locations,investigations based on a wide series of experimental data gatheredfrom surveys carried out on structures assessed for seismic vulnerabil-ity were carried out. Main results obtained are briefly outlined below:• no general trends have been recognized regarding the spatial vari-ability of the key mechanical properties of concrete throughout columnmembers as a possible result of the combination of the effects of theload pattern and the casting process;• although carried out on members already cracked and damaged, theresults of sonic tests are affected by scatters smaller than those deriv-ing by the compression tests on concrete samples; further, the ratiobetween ultrasonic velocity derived by indirect and direct measures arenormally distributed around the average value of 0.75;• rebound tests have confirmed the substantial impossibility of recog-nizing general trends in the spatial variability of the mechanical prop-erties of concrete and led to values affected by scatters quite similar tothe destructive ones.

Fig. 3- Role of past applied loads on in-situ measurements: (a) qualitative bending moment due to vertical loads (a), and (b)test results along the lower part of the extracted beam (rebound number S, direct velocity V, surface velocity Vs, core strengthfcore).

b

a


CF for a Limited knowledge (CF=1.35).For each value of the CF the demand/capacity ratio has been evaluatedand positioned in the structural performance distribution, for the threeknowledge levels, with reference to the structure at hand. The CF valueconsidered as the exact one has been defined as the value correspond-ing to the demand/capacity ratio value with 5% exceedance probabili-ty. In this way, the CF for material strength is evaluated on the basis ofthe probabilistic evaluation of structural performance taking intoaccount all the intrinsic and epistemic uncertainties, including uncer-tainties on the testing methods.A simplified method is proposed too, based on a limited number ofMonte Carlo simulations, which is able to approximate the probabilitydistribution of the structural parameter. This can be a basis for thedevelopment of simple procedure to use in for evaluation of structuralsafety. A second objective was to develop a procedure for evaluation of mater-ial strength and calibration of CF based on the application of Bayesianmethod, to take into account the number and the reliability of the in-situ tests carried out. The Bayesian method allows to employ destructive and non-destructivetesting results to update a prior probability distribution function.Destructive and non-destructive testing results are separatelyemployed, taking into account individual testing reliability (reliabilitydue to testing errors and errors in regression curve that provides thematerial resistance as a function of the testing parameter). More thanone test method can be employed performing consecutive up-dating ofthe probability distribution function.The statistics reliability of the mean value is improved by applying theconfidence interval for the mean; a 95% lower confidence level is con-sidered, which represents the value for the structural assessment.In order to facilitate its evaluation, a simplified procedure is defined.The material strength value for structural assessment can be obtainedscaling with an appropriate Confidence Factor a weighted mean of thesampling mean values obtained by, both, different testing methods andprior information:

(1)

where:

(2)

where x� DM and x� NDM are the sampling mean of the destructive and nondestructive tests, respectively; nDM and nNDM are the correspondingsampling dimension.Generally, if Mi is the i-th testing method adopted, the material strength

mD = � m̆'''inf,m̆f

mFC

m = [ ]m'f+nDM�xDM+nNDM�xNDM

1+nDM+nNDM

Regarding the treatment of uncertainties in determining the character-istics of materials and more generally of existing building parameters, afuzzy-logic based approach for uncertainty treatment has been set upand a computer code for its implementation has been developed. In addition, the prediction capability of some formulations provided bythe current technical literature was verified based on experimentalinvestigations through non-destructive and destructive tests on existingstructures.Based on a a critical review of available literature, a database was pre-pared that collects literature data on non-destructive tests on concretespecimens for concrete grade assessment. Dispersion of experimentalresults has been estimated and the influence of uncertainties comingfrom the concrete grade estimation on seismic capacity of RC existingbuildings has been investigated.Some results obtained during the project have been reported in paperspublished on journals and in proceedings of Conferences (e.g. Masi andVona, 2008; Marano et al., 2008; Olivito et al., 2008).

3.11 FC: Calibration of Confidence FactorsA first aim of the task has been the evaluation of the confidence levelon structural safety of existing buildings given by seismic structuralassessment carried out according to the indications of the Italian OPCM3431, 03-05-05, Annex 2. Uncertainties in reliability structural analysis are due to material prop-erties, structural details and condition of the structural elements. Theprior distribution of the considered uncertainties takes into accounttheir mechanical effects. The proposed probabilistic models are subse-quently updated by in-situ information. A parameter describing the structural performance is defined as thedemand/capacity ratio and its probability distribution is assessed by aMonte Carlo simulation. Each realization corresponds to an applicationof the capacity spectrum method and needs the execution of a structur-al linear static analysis. The Bayesian up-dating of the structural relia-bility is carried out by a Markov Chain Monte Carlo algorithm. Thestructural performance prior probability distribution function is evalu-ated in two different cases: 1) taking into account the uncertaintiesabout material properties and structural details; 2) updating the struc-tural assessment based on in-situ tests and inspections.The updating process consists of two different levels: in the first,destructive tests and relating errors are taken into account; in the sec-ond, non-destructive tests and relating errors are taken into account.Structural seismic performance has been evaluated in three cases: a) byusing the mean material strength values (CF=1); b) by using the meanmaterial strength values scaled by the CF for a Normal knowledge(CF=1.2); c) by using the mean material strength values scaled by the


for the assessment is:

(3)

where x� Mi is the sampling mean of the i-th testing method and nMi itsdimension.The CF can be expressed in an explicit form as a function of theBayesian coefficient of variation Vm for the median value of the mater-ial strength: FC = a+c�Vm

w (4)The parameter Vm, which estimates the reliability of the available infor-mation, is defined as:

(5)

where s2s,Mi and s2

t,Mi are the sampling variance and the variance of theregression curve of the i-th testing method, respectively.The Eq. (4) has been calibrated for concrete strength by applying theleast squares method. A Monte Carlo method has been used to simulate sampling withdestructive and non destructive testing and the resulting equation forthe CF is:

(6)

The Eq. (6) is effective if samples have been extracted from homoge-neous zones of the structure. If in the structure potential non homoge-neous zones are identified, the t-Student test can be executed on themean values extracted from the two zones. If the t-student test identifynon homogeneous zones these must be separately evaluated consider-ing two different median values for concrete strength with two CF. A method has also been investigated for the evaluation of reliability ofthe correlation function for the assessment of material strength by insitu tests.

3.12 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular OnesThe main results of the project are as follows:• Regarding the validation of the modified pushover procedure pro-posed by Fajfar (2002), it was found that it is conservative for plan-reg-ular framed structures with respect to NTHA, while it is unreliable forshear wall structures. For irregular frames, multi-modal procedureshave to be used in order to improve the accuracy of static non linearanalyses. Furthermore, predictions drawn from non linear static analy-

FC = 0.9 + Vm�

m = [ ]m'f+�inMi�xMi

1+�inMi

nMi�xMi

s2s,Mi

+s2t,Mi

�i

smmm

s2s,Mi

+s2t,Mi��i

nMi

Vm = =

ses were often found un-conservative in terms of interstorey drifts orchord rotations;• A new pushover method that explicitly takes into account the tor-sional behaviour of asymmetric-plan buildings was defined. The effec-tiveness of the proposed procedure was evaluated by comparing theseismic demand of selected case studies with that obtained throughboth nonlinear dynamic analyses and other pushover methods;• The nonlinear analyses on the regular and irregular buildings haveshown the importance of damping in nonlinear dynamic analysis. Morespecifically, as the hysteretic model improves, the damping should bedecreased. Furthermore, viscous damping is hard to assign and little orno indications are given in the published literature to guide the user. Avalue of 2-3% damping for the elastic modes appear reasonable, but nofinal indications were found;• In NTHA, for both natural and generated accelerograms, the applica-tion of the seismic input in the principal directions of the structure mayunderestimate the demand, the structural demand varies considerablyas the seismic input direction changes, more so for natural accelero-grams. However, as for the number of input ground motions to use innonlinear dynamic analyses, enhancements to the EC8 requirementswere proposed;• For irregular buildings, pushovers applied in different directions indi-cated demands and capacities that depend on the direction considered.Partial results on the L-shaped building are shown in Figure 4.• Shear collapse in existing buildings not designed according to thecapacity design rules, is often dominant. If only the flexural capacity ismodelled, most frames are verified both at the Ultimate and at the NearCollapse Limit State. However, additional checks show that the shearcapacity has already been reached on several elements prior to reach-ing the design accelerations. This indicates the need to use models thatconsider shear failure too (this is very rare in the available software).Furthermore, early shear failures point to possible retrofitting actions,

Fig. 4- Pushover analyses on L-building: Influence of seismic input angle on the demand: displacements (left) and rotationratios (right).


Table 1. Literature review on mixed-type buildingsCodes National International Total

7 15 22Damages after earthquake1 National International Total

16 63 79Mixed-type building2 Experimental Analytical studies Total

7 1 8Confined masonry3 Experimental Analytical studies4 Total

27 24 51

1 Reports and other papers enclosing field observations after earthquakes.2 Building in which elements of reinforced concrete and masonry are not bonded.3 Building in which bearing masonry walls are confined by reinforced concrete elements.4 Analytical studies, including also proposal for design and for the modelling of confinedmasonry structures.

The main conclusions inferred by the study can be summarized as fol-lows. In many codes the confined masonry structures are taken intoconsideration, even though usually just empirical dimensional rules aregiven, while methods for assessment and design are lacking and, whenpresent, they differ substantially from a code to another. Concerning themixed structures (unconfined masonry), only a small number of codesgives some indications, which are usually about the behaviour factorand/or the distribution of horizontal forces among different structuralelements. A similar trend was found also when gathering papers on fieldobservations, experimental tests and analytical studies: a lot of datawere found concerning confined masonry structures while just a fewwere gained about other kinds of mixed structures. In general it is pos-sible to state that confined masonry can withstand to seismic actions,given that materials are of good quality and good constructions rules arefollowed. On the contrary, some deficiency, such as small amount oftransverse steel in columns, high thickness of mortar joint, lack of RCelement near the openings, can generate the bed behaviour of this kindof structures. Concerning the mixed structures (other than confinedmasonry), the main problems which arise in the assessment and designare related to the distribution in plan of the horizontal actions, to thedetermination of the behaviour factor and to the study of the connec-tions between masonry and reinforced concrete elements. In literature,indications about these points are few and further studies are needed. With reference to the second goal, the main problems concerning thenon-linear analyses on RC-masonry mixed buildings have been high-lighted. Particular attention has been devoted to the seismic action dis-tribution between different technology elements. The results of theanalysis carried out on 3D RC-masonry mixed building (see Figure 5)with external walls and internal frames, underline the fundamentalmasonry role to withstand horizontal action, while a significant transla-tion increase could be offered by introducing RC frames. This researchhas shown the growing capacity offered by mixed building to bear theseismic action by increasing the RC elements stiffness. The increasing

such as shear strengthening of beams and columns;• Synthetic expressions were developed that express the plan-regular-ity of a building with respect to stiffness and strength distributions.Also, simple corrective coefficients were developed to compute thestructural demand increase as the structural eccentricity increases. Inalternative, a simplified procedure was developed that can help opti-mize the structural design by producing a plan-regular building;• Severe convergence problems were encountered in commercial codesthat use lumped plasticity models. These problems were related to thesoftware limitations rather than the modelling selection. As expected,fibre section models provide a better, more physically-based predictionof the section response. It is however important to extend the model toinclude shear failure, in order to avoid post-processing of the results;• Different Limit State definitions for performance assessment wereconsidered, such as interstorey drift, plastic rotation and chord rotation.It was shown that different measures lead to different, sometimes verydifferent, capacity predictions. The limits usually assumed for inter-storey drift result in larger chord and plastic rotation than limits pro-posed by the European and American Codes. Furthermore, the proce-dure provided by Eurocode 8 to compute the chord rotation capacityyields predictions that are very different from those obtained analyti-cally, that is integrating the member curvature throughout the plastichinge length;• A simplified method for deriving fragility curves and evaluating theprobability of failure was proposed. The method is based on an incre-mental application of the so called N2-Method with natural responsespectra, whose irregularity covers the record-to-record variability of thestructural response without the need for performing non-linear dynam-ic analysis.

3.13 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)Build ingsThe main activities of the first goal have involved both the literaturereview of mixed-type buildings and the modelling aspects related tothe RC substructure of such structures. A report was produced(Decanini et al., 2006) including the main results obtained in theTask.The work involved the study of 160 papers, as indicated in Table 1.Confined masonry is listed separately to underline the difference inthe amount of studies between those constructions and structures inwhich elements of different technologies (masonry/reinforced con-crete) are not bonded. The latter case includes constructions in whichmasonry and reinforced concrete frames are present at the same leveland construction in which reinforced concrete frames are placed inthe upper floors only.


RC elements stiffness is significant for the seismic action distributionboth in linear and non-linear field. From the analyses it has emergedthat, by increasing the stiffness of internal RC frames, the maximumsustainable seismic action of mixed building increases, while the rateof seismic action supported by masonry walls decreases (Nardone et al.,2008). The important slab role in sharing the seismic action betweenthe vertical resistant elements and the importance of slab to performthis function in order to avoid undesirable behaviour of the building hasbeen underlined.In addition to that, concerning the first model, the analyses highlightedthat the presence of the RC frame is generally detrimental for themasonry wall. In fact, the resisting total base shear of the mixed systemis smaller than that of the masonry wall alone when the contribution ofthe frame wall is smaller than about 15-20% of the total base shear.Moreover, when the top connection is non effective, hinges develop atthe top and the bottom of RC columns. Therefore, it seems appropriateto design the RC elements considering seismic action deriving from thepertinent vertical loads. While, if the connection between masonry andframes are effective, it would be expedient to assign the whole of thehorizontal loads to the masonry (Decanini et al., 2008).In general, the analyses performed on the 3D model highlighted the dif-ficulties in modelling the masonry and the connections between differ-ent elements (beam-masonry, floor-masonry) and indicated the stronginfluence on the global structural behaviour of the connections effec-tiveness. The parametric analyses showed the importance of the mason-ry tensile strength among other mechanical parameters on the globalresponse and highlighted that the RC elements remain elastic and givea negligible contribution to the overall performance. Finally, the com-parison between the mixed structures results and those obtained for themasonry structures confirmed how the practice of replacing masonrywalls with RC frames, in the interior of old masonry buildings, can havenegative consequences on the vulnerability of the building themselves.

3.14 TAMP: Influence of Infills on Structural ResponseThe resisting elements that were chosen to construct commonly usedmasonry infills were: calcarenite ashlars, hollow clay brick and hollowlightweight concrete blocks. The results of the compressive tests on the

related masonry samples lead to the following main conclusions: i) cal-carenite masonry exhibits middle compressive and shear strengths, andductile behaviour up to the collapse for each of the three consideredloading directions; it behaves like an orthotropic material in which theratio between the two elastic moduli depends on the mortar properties;ii) clay brick masonry is about twice resisting in compression, but duc-tile behaviour was observed only under diagonal loading, when theshear behaviour of the mortar joints is involved; iii) lightweight concretemasonry has very low resistance capacities along all the loading direc-tions; the results obtained from this kind of masonry were rather scat-tered, because vertical mortar joints are not provided for, due to the ver-tical profile of the resisting elements. The experimental values of elas-tic moduli and resistances showed that the values deduced by the M.D.’87 provisions are not always reliable. In particular, the shear strengthvalues are strongly underestimated by this code for all masonry tested.The cyclic tests on the infilled meshes of RC frames have made it pos-sible to verify that the cross-section of an equivalent strut can be deter-mined by using the procedure and parameters shown in Amato et al.,2008, where a not negligible role is exerted by the transverse strainratio in the diagonal direction and the compression level transmitted tothe columns after the masonry infills have been constructed. The later-al resistance of the infill can be deduced from the masonry shearstrength; this can be translated into a compressive strength to be con-ferred to the equivalent strut, by means of a suitable criterion takinginto account mainly the disconnection arising between frame and infill.Finally, these tests showed that the calibration of the adopted modelleads to sufficiently approximate results (Figure 6).The shaking table tests on the 3D scaled building (Figure 7) have beencarried out by assuming a natural input accelerogram (Herceg-Novi,

Fig. 5- Analyzed models (Nardone et al., 2008).

Fig. 6- Cyclic tests on infilled RC frames: a) test set-up; b) experimental results; d) validation of model.

ab

c


sions giving the desired interstory drift.The results of this Task, synthetically presented here, are discussed indetail in the final Report of Task TAMP. The following remarks can bemade, concerning their use and possible improvement of provisions: -the proposed model of equivalent diagonal strut could be utilized foranalyses of existing structures designed to bear only gravitational loads;- the expressions proposed by M.D. ’87 would be revised, also includ-ing values to be assigned to the transverse strain ratios; - infills wouldbe considered for evaluation of the period of vibration of the structureand definition of structural regularity.

3.15 SCALE: Behavior and Strengthening of StairsAccording to the literature, the existing stairs can be classified into twomain categories depending on the static behaviour of the stair steps: (i)stairs with steps performing as cantilever beam, stair type A (see Figure8a), (ii) stairs with simply supported steps stair, type B (see Figure 8b).Generally the stair type B are used worldwide, in Europe and USA,while the stair type A, with inclined beams, are much more adopted inEurope (Tecnica y Pratica del Hormigon Armado, 1989, Reynolds andSteedman, 2002, Guerrin and Lavaur, 1971). According to the USA’smanuals a great scatter in stair structural solutions can be found (Berry,1999). According to the manual design criteria (Marrullier, 1910; Rosci, 1939;Santarella, 1953, 1957; Pagano, 1963; Migliacci, 1977) stair type Acould be designed considering only gravity loads, any seismic actionscould not be taken into consideration. The permanent and live loads onthe steps generate on the beam element (bs1-bs2-bs3) a torsionalmoment T and a distributed load producing on the beam shear force Vand bending moment M. Each flight step is designed modelling it as acantilever beam subjected to a distributed load.As it is explained in several dated manuals (Santarella, 1953, 1957;Pagano, 1963), the design bending moment into the beam (bs1-bs2-bs3) is evaluated on the basis of different static schemes correspondingto different constraints at the extreme ends of the beam. In particular,two extreme constraint conditions are suggested: full constraint andsimply supported. The torsional moment is considered of relevantimportance, it leads to add transversal reinforcement (stirrups) alongthe length of the beam (bs1-bs2-bs3). The adopted values of the tor-sional moment depend on the hypothesis upon the flexural stiffness ofthe inter-storey slab: flexible and rigid diaphragm. About stair type B, manuals indicate two limit structural schemes: (i)an horizontal beam full constraint at the end, (ii) an horizontal beamsimply supported at the extreme ends. Normally bending moment andshear are the internal forces taken into consideration. In the manuals ofthe construction time the only severe prescription is regarding the

Montenegro 1979), scaled to three levels of PGA. These tests showedthat: - the fundamental period of vibration of the structure is influencedby the presence of the infills (PGA = 0.04 g); - the crack distributiondepends on the location of the infills with opening (PGA = 0.3 g); - theinfilled structure bears a PGA value (0.54 g) that proves to be aboutthree times the value that had been deduced by pseudo-dynamic testson the corresponding non-infilled structure.With regard to the numerical investigation, two series of four-storey andtwelve-storey RC frames have been considered. Frames of the firstseries were designed to bear only gravitational loads; the ones of thesecond series were designed according to the EC8 provisions. Theresults lead to the following remarks: - the presence of infills impliesgreater quantity of input energy for the structure; nevertheless, theinfills can dissipate a lot of this energy so that the total balance isfavourable to the resisting elements of frame with respect to the case ofbare structure; - in the frames of the first series the infills can exert adecisive role towards a seismic event; - in the frame of the second seriesa non-uniform distribution of infill in elevation would be considered indefining the structural regularity. Finally, the studies regarding the calibration of a three-strut modelarrived at the conclusion that this approach is useful for the compre-hension of the local behaviour of columns in case of degradation due toinfill-induced shear effects. For what concerns response in terms of dis-placement, single strut models seem to give satisfactory results. Strutparameters can be directly obtained from the mechanical characteris-tics of infill components, i.e., mortar and blocks. Moreover, a design cri-terion for a protecting dissipative bracing systems has been developed.The proposed approach allows to define the required minimum dimen-

Fig. 7- Results of shaking table test (PGA = 0.3 g): a) 3D building; b) crack distributions.

b

a


design of the steel bars: a reinforcing bar should not bend to form anglesthat favour pull-out of the concrete cover (Pagano, 1963). The Italian stair design practice during the period 1954-1980 has beenanalyzed in order to identify the most common typologies and the effec-tive adopted design criteria. As already remarked, according to thedated technical manuals and codes stairs could be designed for onlygravity load. The predominant stair type in the studied building sampleis type A; the flight steps are cantilever elements constraint to inclinedbeams having one point of discontinuity in the 53% of the cases, twopoints of discontinuity in the 37% of the cases and is directly connect-ed to the column without any discontinuity in the 5% of the cases. Thestair type B is present in the sample with incidence of 3%. The designpractice of the most common type of stairs, composed by flight stepsconstrained into a beam, is herein studied. The static design schemehas a great scatter; beams were designed considering a maximummoment M+=qL2/a in the midspan with a=12 (30% cases) or a=8(30% cases), while the minimum moment at the extremes of the beamis obtained with a=12 in most cases (76% cases). Regarding the influence of stairs on the seismic capacity, this prelimi-nary study on the structural typologies of the building sample has evi-denced the following problems related to the presence of stairs: distri-bution of seismic forces (not considered in the design), different model-ling design of stair structure, material strengths, element detailing.The structural typology of stairs generally introduces discontinuitiesinto the typical regular reinforced concrete skeleton, composed bybeams and columns; in fact, the sub-structure “stair” is an assemblageof inclined elements as slabs or beams. All these elements contribute toincrease the stiffness of the stair due to the elastic behaviour of inclinedelements and of squat columns. For these reasons the elements thatconstitute the stair are often characterized by a high seismic demand:the squat columns are subjected to high shear demand that can lead toa premature brittle failure; the inclined beams, differently from the hor-izontal beam, are defined by high variation in axial forces that can mod-ify the resistance and deformability of all these elements.All these aspects are discussed with a series of analysis on a RC build-

ing representative of the studied sample; non linear static analyses (sta-tic push over analysis) finalized to the evaluation of the role of the stairs,of their elements and modelling is performed. The building without any stair is defined as reference. Two models havebeen considered to study stair type A with inclined beams and stair typeB having reinforced concrete slab. For each structure, different model-ling have been adopted to evidence the influence on the global responseof: biaxial bending modelling in the beams of the substructure “stair”;bending moment-axial force (M-N) interaction into the inclined ele-ments (beam and slabs); bending moment-shear (M-V) interaction intothe inclined elements and columns.In general, the presence of stair brings to an increase of lateral strengthand to a reduction in displacement capacity with respect to the build-ing without stair (Cosenza et al., 2007a). On the contrary, the resultshave confirmed the need to utilize biaxial bending modelling and toaccount for the interaction of the different internal forces (Cosenza etal., 2008) as: bending moment-axial force interaction that characterizesthe inclined elements, and the bending moment-shear interaction thatgoverns the behaviour of squat columns. Shear failure becomes pre-dominant in the squat columns and in the reinforced concrete slabs andprecedes the conventional ductile failure (see Figure 9).

3.16 NODI: Behavior and Strengthening of Beam-Column JointsMain results obtained in this Task are made up by the execution andanalysis of the experimental program on beam-column joints withoutstrengthening, as well as by the execution and analysis of some tests onstrengthened and unstrengthened beam-column joints and on basejoints. Numerical-experimental comparisons have been performed, aswell, regarding some highly representative test campaigns available inthe literature and relevant to some of the tests carried out. A careful lit-erature review was carried out relevant to: (i) experimental programscarried out by other researchers, (ii) joint capacity models, and (iii) codeprovisions on beam-column joints.A wide experimental program was carried out on full scale externalbeam-column joints relevant to typical existing RC buildings having

Fig. 8- Stair typologies: (a) stairs type A with steps as cantilever beams, (b) stairs type B with simply supported steps. Fig. 9. Results of the push-over analysis in terms of base shear versus roof displacement (Cosenza et al., 2008).

a b


A parallel experimental program was devoted to the test (under imposedincreasing cyclic displacements applied to the beam end) of 4 externalreduced scale beam-column joints provided with smooth bars, laterretrofitted after a first series of tests. For 2 of them only the anchorageof the beam bars had been restored by welding threaded bars to theends of the longitudinal beam bars and bolting them to steel platesplaced on the column external surface. For the other 2 joints, besidesrestoring the anchorage, also vertical and horizontal carbon fiber fabricshad been applied on the column, below and above the joint panel.Moreover, other 2 real scale beam-column joints were built and testedunder cyclic loads, either with or without retrofitting, to evaluate theincrease of joint performance. As for the first 4 joints, the restoration ofthe anchorage of the beam bars proved to be very efficient, since it pro-vided increases in strength up to 300% with respect to not retrofittedjoints. The additional carbon fiber reinforcements did not providenoticeable increases in strength, because the cracking of the joint,which would have required the carbon fiber contribution, did not occur. Regarding the other 2 joints, it was observed that the not-retrofitted oneattained low strength values due to the lack of steel reinforcement fornegative moment in the beam. As regards the joint retrofitted beforetesting, the carbon fiber fabrics applied on the beam significantlyincreased both strength and displacement ductility.Additional research activities were mainly focused on experimentaltests on full scale RC columns (base joints not-strengthened andstrengthened) tested under constant axial load and monotonic or cyclicflexure. The strengthening systems consisted in confinement by par-tially wrapping unidirectional carbon or glass layers around the elementat the base. On some specimens, in addition to confinement, steelangles (in some cases, anchored at the foundation) were placed in cor-respondence of the member corners. The specimens were designedaccording to the Italian codes in effect during the ’60s and ‘70s with theaim of reproducing typical dimensions, rebar amount and details com-mon at that time. Main results achieved during the Project are as fol-lows: (i) regardless of the axial load value, the FRP confinement pro-duces significant increases in terms of ductility, especially if a GFRP(glass) jacket is used; (ii) the arrangement of the longitudinal steelangles unconnected to the foundation leads to higher ductility levelsthan those measured for members strengthened by only FRP systems;this ductility gain is lower for columns tested under n=0.40, even if inthese cases the unconnected angles also provide an improvement of theflexural strength; (iii) when the longitudinal angles are anchored to thefoundation the flexural strength of the RC columns significantlyincreases, but a reduction of the available ductility is observed.A numerical-type activity focused on the analysis of the performance ofRC beam-column joints through numerical simulations by using the

different Earthquake Resistant Design (ERD) level. Quasi-static testshave been performed with 3 loading cycles for each drift value gradu-ally increased from 0.25% up to the total failure of the joint. Resultsshowed yielding force equal to about 20 kN for the NE joints (gravityloads only designed), and about 40 kN for both Z2 joints (designed forseismic zone 2, medium seismicity) and Z4 joints (designed for seismiczone 4, very low seismicity), as a consequence of the minimum require-ments on reinforcement amount prescribed by the Italian code. Observed failure mechanism in all the joints showed a wide and heavycracking in the beam, due to the small amount of the longitudinal rein-forcing bars as the beams were not loaded by the floor slabs in the con-sidered building model. Joint failure was generally caused by the ten-sile failure of the reinforcing bars in the beam. A different as well inter-esting behaviour was displayed by some tests on Z2 joints (Masi andSantarsiero, 2008), where a wide cracking also in the joint panel and asoftening mechanical behaviour (Figure 10) were observed, due to thereduced amount of the applied axial force (�=0.15). Drift value (coinci-dent with chord rotation value) at failure in the NE joints is about 3.0%,while in more ductile Z2 and Z4 joints, values equal to about 4.5% havebeen detected. As for the contribution of joint panel own deformations to the total driftof the sub-assemblage, experimental results showed that it is rather low,being always lower than 10% even in the heavily damaged specimens.Further, interesting results have been found by comparing the strengthof the joint panel provided by the tests and that one obtained applyingthe European (CEN, 2004) and the Italian Code (NTC, 2008) expres-sions to evaluate the capacity of beam-column joints. Results show agood estimation ability of code expressions that have been able to pre-dict which of the specimen was subjected to diagonal cracking of thejoint panel. Regarding the ductile capacity, the difference between NEand seismic specimens was lower. In all the tests with high axial load afailure mechanism with extensive cracking of the beam and evidentdeterioration of the concrete at the beam-column interface has beennoted.

Fig. 10- Test results on a Z2 joint (design for medium seismic zone, low axial load): a) damage pattern at drift=7%, and b)force-drift relationship.

a b


F.E. software DIANA, validated by means of experimental test cam-paigns available in literature (e.g. by the Shiohara working group) aswell as some of the tests performed. In particular, many non linearanalyses have been performed on typical existing external beam-col-umn joints as they can be found in real buildings built in the past. Thedefinition of typical deficiencies found in real beam-column joints andthe analysis of the main parameters governing the structural behaviourof such joints allowed to highlight some effective strengthening solu-tions, especially for external joints. Further, such work allowed to vali-date some theoretical models able to predict the behaviour of bothexternal and internal joints, as well as to validate the expressions pro-posed by some codes to predict the failure of the joint panel.Some results obtained during the project have been reported in paperspublished on journals and in proceedings of Conferences (e.g. Masi etal., 2008).

3.17 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearThe presentation of the results follows the research outline that hasbeen above discussed. In particular, methodological and numericalcontributions are presented separately from the main experimentalfindings. This choice is actually related to the nature of the results andtheir impact on applicative aspects of structural seismic design, name-ly codes and design and assessment practice.In particular, the detailed review of technical literature that has carriedout during the early stages of the research pointed out the relevance of thetype of reinforcement used for construction of existing buildings. In fact,it has been demonstrated that experimental response of r.c. members canbe affected by type of reinforcement, smooth or deformed, and by bondinteraction between steel and concrete especially in post-yielding phase.As a consequence, a concrete effort has been devoted to perform a com-parative analysis of inelastic performances of members depending ontype of reinforcement; this means that both numerical and experimentalactivities have been calibrated in order to cover at local -strength andductility of cross sections, bond under static and cyclic loads of rebars-and at global – stiffness and rotation capacity of members-. Results of both theoretical and experimental activities have been pub-lished in the context of National and International conferences andmeetings. Specific attention has been paid also to continuous educationof young engineers and technical updating for practitioners. Diffusionof software packages (free download of executables and tutorials) andresearch results has been supplied on Reluis website.

The first set of results refer to the development of methods for numeri-cal analysis of cross sections subjected to generalised bending and axial

force. Different approaches have been proposed by the different teamsinvolved in the Task, comparative analyses have been carried out andmultiple applicative perspectives have been covered.A specific software has been developed and made available at the pro-ject website (www.reluis.it) for download. It is based on a Fortran lan-guage procedure and takes advantage of a user-friendly Visual Basicinterface and multiple language platforms. The computational engine ofthe software has been also used for the development of an extendedparametric analysis aimed at the evaluation of the influence of the biax-ial actions not only on the ultimate strength but especially on the crosssection ultimate curvature. In particular, such influence was studied onsquare RC cross-section characterized by different values of axial loadand geometrical percentage of reinforcement. The study has beendeveloped in order to define simplified analytical formulation to easilypredict the ultimate curvature reduction in the case of biaxial bendingand axial load with respect to the case of uniaxial bending. Figure 11gives a view of the program interface (top right), an example of typicalresults in terms of generalised relationship between the reduction ofultimate curvature compared with the reference uniaxial value(�=�u,biax/�u,uni), normalised axial load � and the inclination of the

Fig. 11- Effect of biaxial bending on cross section local deformation (Di Ludovico et al. 2008a, 2008b).


At local scale, a number of bond tests on smooth rebars were devoted tothe definition of a constitutive stress-slip law to be used in numericalsimulations.Figure 13 reports sample data concerning specimens (left), typicalcyclic experimental data (center) and finally the idealized constitutivelaw calibrated against the test (right). At large scale, eight experimental tests on r.c. square or rectangular fullscale columns under constant axial load and uniaxial bending were per-formed. Monotonic or cyclic action were applied on specimens. Detailsabout the experimental program are reported in Table 2 and Figure 14.Each test was performed under displacement control.

Table 2. Summary of experimental tests on r.c. columnsUNIAXIAL TESTS - Cross Section BxH (cmxcm)30x30 30x30 50x30 30x50

Longitudinal Reinforcement 812 812 1212 1212 Steel Rebars type Plain Deformed Plain Deformed Normalized Axial Load, � 0.2 0.2 0.1 0.1 Type of action Monotonic Cyclic Cyclic Cyclic Number of tests 2 2 2 2

Primary experimental outcomes clearly indicate that the contribution ofthe base rotation on the global deformation mechanism is noticeablydifferent in case of columns reinforced by plain or deformed rebars,however, the overall member global deformation and energy dissipationcapacity is not strongly affected by the bond performances of the inter-nal rebars. The global deformation capacity in the case of plain rebars,is mainly due to a localized source of deformability at the column foun-

stress plane angle (top and bottom left) and finally a comparisonbetween simplified and refined results (bottom right).An alternative method has been proposed and implemented. It arrivesat defining closed-form equations for performing assessment of existingRC columns with two-way steel reinforcement, under combined biaxialbending and axial load. Starting from the load contour method, an effi-cient procedure for estimating the strength/deformation section capaci-ty has been developed. In addition, simple closed-form equations forcomputing section uniaxial resisting moments and ultimate/yieldingcurvature has been defined. The method has been also extended to cross sections reinforced withFRP, resulting in a very effective guide for fast implementation ofresults in more general software packages and direct application bystructural engineers using easy to manage electronic sheets.Figure 12 shows an example of results that can be obtained accordingto this design tool, both for bare and FRP-strengthened members andits ability to give reliable results with a reduced computational effort. Another relevant achievement is related to the response of the wholemember under axial force and biaxial bending. In particular, the atten-tion has been focused on the development of the actual stiffness of themember depending on the stress level and influence of biaxial bending.A fibre model of the member has been implemented in order to providerecommendations for characterization of equivalent stiffness to be usedin elastic analysis. The study carried on has been limited, at themoment, to the definition of the problems which must be taken intoaccount while passing from a cross section to the whole member, Boscoet al. 2008. In this context, it is worth noting the contribution of UNICHthat carried out nonlinear analyses of reference regular and irregularreinforced concrete buildings and on problems related to the selectionof input ground motion, Canducci et al., 2008.

In compliance with the main issues derived from the theoretical analy-sis of r.c. members and cross sections, experimental activity has beencarried out at different scales.

Fig. 12- Comparison between exact (fibre method) and approximate approaches for bare (left) and FRP reinforced r.c.members (right) (Monti and Alessandri, 2008).

Fig. 13- Fibre model of the cross section (left), force-displacement curve for a column (center), normalised equivalent stiffness(right) (Verderame et al., 2008a, 2008b).

been drafted and published in a specific booklet (Figure 15). Every typehas a short description with sketches and indications about the years ofproduction, the regions of destination and the relative diffusion withrespect to the global precast production. Some notes are added withindication of possible behaviour deficiencies for seismic destinations.Being a key point of precast construction, special attention has beenaddressed to the seismic behaviour of connections. Some tests havebeen performed to quantify their seismic capacity following a standardapproach.Five principal categories of connections have been considered:- floor-to-floor connections between adjacent floor or roof elements;- floor-to beam connections between floor or roof elements and the sup-porting beam;- beam-to-column connections between the beam and the column;- column-to-foundation connections which provide the base support tothe columns;- cladding-to-structure connections for the support of the wall panels.Two level of tests have been performed:- particular tests: referred to the qualification of single connectorsinserted between two overdimensioned blocks and subjected to theprincipal action expected in the structural system (Figure 16, left);


dation interface (fixed end rotation), while a more diffused crack patternalong the column end was observed in columns reinforced by usingdeformed steel rebars. The significant influence of P-� effect on theglobal behavior of specimens has also clearly emerged by the experi-mental tests; if such effect is neglected, the ultimate rotation recordedon columns reinforced by deformed steel rebars is clearly less than thatobserved in columns reinforced by using plain rebars. However, due toP-� effect the ultimate rotations related to the two different columnstypologies is very close. Such result can be explained by consideringtwo main aspects: the higher strength of members reinforced withdeformed rebars and the higher slope of the softening branch of theshear-drift curves, if P-� effect is considered. A calibration of a numer-ical model able to take account of specific aspects related to bond ofsmooth rebars and anchoring end details has been also developed(Verderame et al., 2008c).

3.18 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresThe first activity consisted of a wide survey of the existing buildingsproduced from the ‘50s up to today and of the rendering in a reasonedway of the investigation results. This activity enjoyed the support of theNational Association of prefabrication industries Assobeton whichinvolved ten member companies to provide the design documentationof a number of constructions built in different times. So about 150 pro-ject documentations have been collected covering some decades of pro-duction. Since information about far times were lacking, this survey hasbeen integrated with the historical memory of some experts, exploitingtheir knowledge together with the old bibliography of specific journals.From the examination of the project documentations a syntheticdescription for any building has been recorded in a standard format,summarising its features in a specific form of easy reference. This workled to the printing of the booklet “Precast structures: list of projects ofexisting buildings” which provides a good evidence of some decades ofprecast production.A complete catalogue of the different types of precast structures has

Fig. 14- Experimental tests on columns.

Fig. 15- FrontPage cover of two produced catalogues on precast concrete buildings.

Fig. 16- Details of testing of structural connection.

been examined (Felicetti et al., 2008).In parallel, a testing frame has been set up for push over and cyclic testson beam-to-column connections. The common type with a couple ofbars protruding from the column and passing through the holes of thebeams is examined both in longitudinal and transverse direction.The experimental test results in terms of force displacement allowed tocharacterise the connection roof element to beam, as global ductility,ultimate strength and dissipative capacity. In Figure 18, force-dis-placement curves are illustrated (monotonic test on rigid blocks) com-paring the traditional existing connection with a modified solution of theconnection. While a brittle behaviour is expected for the traditionalexisting connection (concrete crash of the beam edges), a ductile behav-iour is then obtained if the steel plate in the connection is opportunelyreduced. Almost similar conclusions can be obtained for quasi-staticcyclic tests as reported in the research reports and papers. Some other tests have been performed on dry bearing in order to mea-sure the friction factor of neoprene pads over the concrete surface andtheir deformation parameters. Tilting tests and on inclined plane havebeen made with different levels of normal loads (Magliulo et al., 2008).The comparison between tests results and friction coefficient valuesprovided by PCI Handbook (1999), CNR 10018 (1999), and UNI-EN1337:3 (2005), is shown in Figs. 19 and 20. In Figure 19 the neoprenecompressive stress (s) is reported on the horizontal axis, while the shearone (�) on vertical axis; in Figure 20 on this axis the friction coefficientis presented. It is evident that PCI Handbook and CNR 10018 curveswell approximate the experimental data linear regression curve; thisdoes not happen in the case of UNI-EN 1337:3 curve. However CNR10018 provides a bit larger friction strength with respect to the experi-mental results, while PCI Handbook provides larger friction strengthonly for compressive stress lower than 3 N/mm2. Further more, the testsresults confirm the light increment of friction strength, which corre-

- local tests: referred to the qualification of the connection includedbetween two significant portions of the elements, representing the struc-tural arrangement and subjected to the relevant components of theaction (Figure 16, right);Previously a standard protocol for testing has been drafted. It definesthe six parameters:- strength: maximum value of the force which can be transferredbetween the parts;- ductility: ultimate plastic deformation compared to the yielding limit;- dissipation: specific energy dissipated through the load cycles;- deformation: ultimate deformation at failure limit;- decay: strength loss through the load cycles compared to force level;- damage: residual deformation at unloading compared to the maximumdisplacement;to be measured both by:- push over tests following a monotonic increase of displacement:- cyclic tests following an alternate history of displacements (Figure 17).A number of tests on roof-to-beam connections (8 push-over and 11cyclic) have been performed. A complete report is available with theresults of the experimentation. Three types of steel connectors have


Fig. 17- Specific energy and cyclic behaviour of the connection.

Fig. 18- Push-over test: comparison of the two different solution (traditional vs. slightly modified).

Forc

e

Displacement

sponds to a light decrement of the friction coefficient, as the compres-sive stress increases.On the base of experimental results, the following relationships for neo-prene-concrete friction coefficient are proposed:m= 0.49 if s� 0.15N/mm2

if 0.15 < s� 5N/mm2

where s� is the compressive stress and =0.055 N/mm2; s�=5 N/mm2

is the neoprene maximum compressive strength according to CNR10018. These formulations along with the linear regression curve oftests mean results are plotted in Figure 21: the two curves are almostcoincident.

m = 0.1+ s�

4. DISCUSSION

4.1 MND: Non-Destructive Methods for the Knowledge of ExistingStructuresThe research activities scheduled within the Task have been to a greatextent carried out without significant delays or anticipations, and themain objectives have been achieved.

4.2 FC: Calibration of Confidence FactorsThe results obtained agree with those expected, concerning the devel-opment of a Bayesian procedure for material strength evaluation andthe calibration of confidence factors.

4.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEm phasis on Irregular OnesThe results obtained by the Task are in line with the objectives origi-nally established. Several papers by the different research units havebeen published or are in press in international journals or conferences.The only problems encountered by some of the units originated fromconvergence issues in nonlinear codes. This is a well-known problem,but some software failed to converge on a regular basis, thus delayingadvances in the research. However, overall, the task followed the sched-ule of work originally outlined.

4.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)BuildingsThe Task has pursued the proposed objectives. In particular, aspects ofmodelling the behaviour of mix-type buildings through non linear analy-sis have been investigated. These analyses have highlighted the differentsteps in which the resistant elements withstand the seismic action.The analyses performed allowed the identification of the main parame-ters affecting the structural behaviour of mixed building and gave indi-cations on feasible modelling and verification criteria.

4.5 TAMP: Influence of Infills on Structural ResponseThe objectives that have been pursued by this research are consistentwith the expected ones. Nevertheless, it must be observed that theexperimental calibration of the parameters defining the cyclic behav-iour of the proposed equivalent diagonal strut model is affected by thefollowing main limitations: - it has been made considering only squaremeshes of infilled RC frames; - the possible presence of an infill withopening has not been considered.These limitations, due to not sufficient time and resources, did not allowthe model to be validated by its use to reproduce the experimentallydetected response of the 3D infilled building subjected to the shaking


Fig. 19- Comparison between compressive-shear stress curves provided by PCI Handbook, CNR 10018 and UNI-EN 1337:3and tests regression

Fig. 20- Comparison between compressive stress – friction coefficient curves provided by PCI Handbook, CNR 10018 andUNI-EN 1337:3 and tests regression curve.

Fig. 21- Proposed compressive stress - concrete–neoprene friction coefficient relationship.

actually exhaustive, since trial biaxial tests have been designed de -pending on the findings of the theoretical work, but not completed. Thisresults in the need to extend and validate the results obtained in thecontext of the present task and give a direct contribution to the devel-opment of specific design recommendations for members under com-bined axial load and biaxial bending.

4.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresThe research activities scheduled within the Task have been to a greatextent carried out without significant delays or anticipations, and themain objectives have been achieved.

5. VISIONS AND DEVELOPMENTS

5.1 MND: Non-Destructive Methods for the Knowledge of ExistingStructuresA large amount of RC buildings, both private and public, now placedin seismic zones, were originally designed taking into account onlygravity loads and without explicitly provide ductile detailing. Anextraordinary rehabilitation program needs to be implemented onsuch buildings, where an accurate evaluation of the available seismiccapacity is important to set up cost-effective interventions.Investigations have a crucial role to adequately know the structure tobe evaluated. For this reason, there is an increasing need to set upand put at disposal of technicians and other involved stakeholderssufficiently reliable as well as not very expensive methods to estimatein-situ material properties. Number of tests required to suitably applythese methods have to be as low as possible, thus making the totalrequired budget sustainable to building owners and thus furtherencouraging their use. To this purpose, the results obtained in thisTask confirm that a smart combination of NDTs and direct tests (suchas core extraction) gives effective solutions from both the economicaland technical point of view.Future research work should be devoted to the following:• Provide methods more and more capable of achieving effectiveresults in terms of prediction capability of concrete properties takinginto account both intrinsic randomness and epistemic uncertainty. • NDTs currently available on concrete do not provoke damage onstructural members but on some other building components (e.g. par-titions, infills, plaster, etc.) thus determining remarkable repair costs:new methods are necessary to make them really not very expensive. • As for reinforcement, taking into account the heavy damage causedby the extraction of steel bars, non destructive methods to estimate itsmechanical properties need to be set up.

table tests.

4.6 SCALE: Behavior and Strengthening of StairsThe Task has pursued the proposed objectives. A detailed investigationof the main stairs typologies and of the most used design procedureshave been performed; in particular, a report including the main resultswas developed (Cosenza et al., 2007b). Numerical investigations havebeen performed in order to understand the seismic behaviour, and thepossible failure mechanisms of buildings having the most common stairtypologies. The results have confirmed the need to utilize biaxial bend-ing modelling and to account for the interaction of the different internalforces as: bending moment-axial force interaction that characterizes theinclined elements, and the bending moment-shear interaction that gov-erns the behaviour of squat columns. Shear failure becomes predomi-nant in the squat columns and in the reinforced concrete slabs and pre-cedes the conventional ductile failure. An experimental set-up has beendesigned on the basis of some simulations performed by using differentmodelling: dimensions of a single span frame with inclined beam, loadsand resisting-wall have been defined.

4.7 NODI: Behavior and Strengthening of Beam-Column JointsThe research activities scheduled within the Task have been to a greatextent carried out without significant delays or anticipations, and themain objectives have been achieved.

4.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearSummary of results reported in the previous sections leads to recognizethat the development of the work basically complies with the initialschedule. In particular, as numerical analyses and software deliver-ables are considered, a good agreement with the program can beaddressed.In fact, different approaches and numerical procedures have been setand made available to technical community. They cover at different lev-els the need of tools for checks required by modern codes in terms ofstrength and local deformation. This applies both to design and assess-ment of existing un-strengthened concrete structures and to seismicupgrading using FRP materials. Interaction between groups involved inthe study of irregular structures and of use of FRP for seismic upgrad-ing of structures is another positive aspect of the work. When experimental program are concerned, it is worth noting that a rel-evant contribution to the knowledge of bond mechanisms for smoothbars has been given and an approach to the comparative analysis of per-formances in terms of rotation capacity of full scale r.c. members withsmooth and deformed bars has been accomplished. The work is not


5.2 FC: Calibration of Confidence FactorsThe work performed by the Task has focused on problems related to thedefinition of a reliable assessment of structural seismic performance. A probabilistic model for structural performance has been developedand a method for calibration of Confidence Factors has been proposed. Furthermore, a procedure for evaluation of material strength has beendeveloped based on the application of the Bayesian method, taking intoaccount, both, amount and reliability of the in-situ tests performed. In addition to that, with reference to the prior probability distribution ofstructural detailing parameters, a preliminary list of possible structuraldefects was prepared, in which, for each defect, a set of possible valuesand their relative weights are envisaged.To proceed further along this path, the following developments areneeded:• A complete procedure should be developed defining all phases of theknowledge process of an existing building. The procedure shall give aCF calibrated on the basis of the distribution of the assessment resultsconditional on the acquired knowledge.• Further studies are needed regarding the evaluation of materialstrength values from in-situ non-destructive tests, whose reliabilitydepends on the reliability of the regression curves used to transform thetest parameter into a material strength value.• A questionnaire, meant to be addressed to professional engineers asa survey, has been prepared. The results of such survey shall be usefulin creating a thorough database of structural defects and their proba-bility distributions.

5.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular OnesSeveral directions for future work have emerged from the project, mostof them related to seismic code enhancements:• Further studies for assessing the applicability of nonlinear pushoverprocedures to plan irregular buildings;• Further studies on different engineering demand parameters, such asinterstory drift, chord rotation, plastic hinge rotation, as means forassessing the structural response;• The need for the definition of the damping to be used in NTHA,depending on the model level of refinement. Using 5% damping resultsin an un-conservative assessment;• A clear definition of spectrum compatible accelerograms (accordingto EC8) to be used in NTHA;• The need to establish a framework for Performance Based EarthquakeEngineering, based on a fully probabilistic approach;• Assessment of the current modelling capabilities for shear failure pre-diction, with the possible development of simplified approaches that

can provide an engineering estimate without adding complexity to theconvergence procedure;• Possible guidelines on how to consider other sources of nonlineari-ties/failures, such as bond-slip, beam-column failure, etc.

5.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)Build ingsConsidering the obtained results, the development of research aimed atevaluating the seismic behavior of mixed-type buildings can be envis-aged as follows: • Establish through non-linear analyses the influence of RC-masonryconnections on the global behavior of the building. Particular attentionshould be paid at intersections between beams and perimeter masonrywalls.• Evaluation of q-factors. This objective should be pursued through theimplementation of (static and dynamic) non-linear analyses.• Evaluation of the seismic response (numerical and not only) of otherstructural configurations of mixed-type buildings not included in thisstudy.

5.5 TAMP: Influence of Infills on Structural ResponseThis research gave useful results concerning the effects of masonryinfills on the lateral response of RC frames, supported by experimentalvalidations. Possible developments could be aimed to pursue the following furthermain objectives: • to revise the available empirical expressions linking the masonryproperties to those of its components, by carrying out further tests onmasonry samples made of other kinds of resisting elements; • to generalize the proposed model of equivalent diagonal strut, by eval-uating experimentally the role of factors that have not been consideredhere: infilled mesh geometry, different compression levels on the framecolumns after the masonry infills have been constructed, presence andsize of an opening in the infilled mesh of frame.

5.6 SCALE: Behavior and Strengthening of StairsConsidering the obtained results the future research developing wouldbe oriented to the evaluation of the influence of the stair on the seismicresponse analysing the following aspects:• Experimental assessment of the strength capacity and deformabilityof squat columns. In this way, the behaviour of squat columns would bebetter understood considering the large number of models found in lit-erature that are not completely exhaustive for a reliable assessment ofthe seismic behaviour of reinforced concrete structures.• Experimental evaluation of the seismic response of 2-D and 3-D


fruitful approach to the development of tools for the theoretical estima-tion of strength and curvature ductility of members has been carriedout. Flexural mechanisms are clearly identified both at local and glob-al scale and an encouraging capacity of simulation is demonstrated byboth static and cyclic proposed models of members with smooth bars.This means that numerical sensitivity analyses can give a positive con-tribution to an optimized design of an experimental program able toconfirm numerical forecasts, show possible points of weakness of thetheories and/or give an insight on specific aspects of the behavior undersevere cyclic loads. The large variety of existing constructions and thediffusion of smooth bars in many very urbanized areas exposed to seis-mic risk point out the need to continue the investigation on such type ofmembers and even develop customized strengthening techniques usingadvanced materials. Despite such positive feedbacks of the research, it is worth noting thatfurther work is strongly recommended to assess:• the behavior of short columns, and • the flexure-shear interaction in presence of smooth reinforcement. In fact, the observed deformation mechanisms can produce effects onthe strength and ductility of members subjected to complex forces, likethose generated on columns of irregular constructions.

5.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructuresThe experimental campaign and the numerical investigations carriedout have highlighted that the most vulnerable buildings are those withdisarticulated diaphragm behaviour. However, as emphasised inPalermo et al. (2008), an accurate study on the modelling of connec-tions needs to be done in order to correctly predict the overall responseof precast concrete buildings.

6. MAIN REFERENCES

6.1 MND: Non-Destructive Methods for the Knowledge of ExistingStructures- Masi A., Vona M. (2008). “La stima della resistenza del calcestruzzoin situ: impostazione delle indagini ed elaborazione dei risultati”,Progettazione sismica, Anno I, No. 1, IUSS Press, ISSN 1973-7432.- Marano G.C., Morrone E., Mezzana M. (2008), “Approccio ibridofuzzy per l’integrazione e l’interpretazione delle prove non distruttive”,Atti del convegno Valutazione e riduzione della vulnerabilità sismica diedifici esistenti in cemento armato, E. Cosenza, G. Manfredi, G. Montieditors, Polimetrica International Scientific Publisher, Roma, Italy.- Olivito R.S., Spadea G., Carrozzini A., Spadafora A.R. (2008). “Strut -ture esistenti in cemento armato: controlli e verifiche mediante tecniche

frames with inclined elements and squat columns. The comparison ofdifferent test results would allow to evidence the influence of the intro-duced specific elements such as inclined beams and squat columns.• Numerical and experimental evaluation of the influence of the stair inbuildings with different location of infill walls. The numerical analysiswould be performed by using static and dynamic tools. This studieswould allow the evaluation of the interaction between stairs and infillsthat are commonly modelled in different manner.

5.7 NODI: Behavior and Strengthening of Beam-Column JointsThe behaviour of beam-column joints can strongly affect the seismicglobal behaviour of RC building structures. Some mechanisms (e.g.concrete cracking, slippage of longitudinal reinforcing bars, etc.) can beresponsible of additional deformability, on one hand, while, on the otherhand, can alter the capacity design assumptions on the framing struc-tural members (beams and columns) and on the joint member itself. Forthis reason, research activities need to be increasingly devoted to devel-op accurate capacity models of beam-column joints to reliably evaluatethe performances of RC building structures.Research carried out in this Task already provided important resultsrelevant to the role of the key behavioural parameters of RC beam-col-umn joints, thus giving useful suggestions on the reliability of currentcode expressions and on possible improvements. However, many otherissues need to be further studied regarding both as-built and strength-ened joints (already damaged or not damaged), skilfully combining pur-posely designed experimental investigations, review of experimentalcampaigns reported in the literature, and accurate numerical simula-tions. Among others, some future research developments can be recognized asfollows: • design and execution of extensive experimental programs on jointspecimens having different characteristics (e.g. internal or external, bi-or tri-dimensional, shape, beam type, etc.) well targeted on the typesrepresentative of the Italian and European built environment; • experimental and numerical validation of different strengtheningtechniques based on the comparison of the relative performance andapplication limits, particularly dealing with tri-dimensional joints andjoints with embedded beams; • codification of test protocols to make possible and promote experi-mental result exchange.

5.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and ShearThis Task research activity dealt with a specific, but relevant aspect ofseismic design and assessment of r.c. constructions. An effective and


non distruttive”. Atti del convegno Valutazione e riduzione della vulne -rabilità sismica di edifici esistenti in cemento armato, E. Cosenza, G.Man fredi, G. Monti editors, Polimetrica International Scientific Pu -blisher, Roma, Italy.

6.2 FC: Calibration of Confidence Factors- Franchin P., Pinto P. E., Pathmanathan R., (2008). “Assessing theadequacy of a single confidence factor in accounting for epistemicuncertainty”, Convegno RELUIS “Valutazione e riduzione della vulne -rabilità sismica di edifici esistenti in cemento armato”, Roma 29-30maggio 2008.- Jalayer F., Iervolino I., Manfredi G., (2007). “Influenza dei parametridi modellazione e dell’incertezza associata nella valutazione sismica diedifici esistenti in cemento armato”, Proceedings XII Convegno ANI-DIS, Giugno 2007.- Jalayer F., Iervolino I., Manfredi G., (2008). “Structural modellinguncertainties and their influence on seismic assessment of existing RCstructures”, submitted to Structural Safety, 2008.- Monti, G., Alessandri S., Goretti, A., (2007). Livelli di conoscenza efattori di confidenza. Proceedings XII Convegno ANIDIS, Giugno 2007.- Monti G., Alessandri S., (2008). “Confidence factors for concrete andsteel strength”, Convegno RELUIS “Valutazione e riduzione della vul-nerabilità sismica di edifici esistenti in cemento armato”, Roma 29-30maggio 2008.

6.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, withEmphasis on Irregular Ones- EuroCode 8 (2003a). Design of Structures for Earthquake Resistance,- Part 1: General Rules, Seismic Action and Rules for Buildings, Draftno. 6.- EuroCode 8 (2003b). Design of Structures for Earthquake Resistance,- Part 3: Strengthening and repair of buildings, Draft no. 3 (FinalProject Team – Stage 34), 2003.- Fajfar P. (2002). “Capacity Spectrum Method Based on InelasticDemand Spectra”, Proceedings of the 12th European Conference onEarthquake Engineering, Paper 843, London (UK).- FEMA 356 (2000). Prestandard and Commentary for the SeismicRehabilitation of Buildings, Prepared by A.S.C.E., Washington, D.C.(U.S.A.).- ATC 40 (1996). Seismic Evaluation and Retrofit of Concrete Build -ings.

6.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)Build ings- D.M. Infrastrutture, 14 Gennaio 2008. G.U. 4-2-2008, N.29, Norme

tecniche per le costruzioni.- D.M. Lavori Pubblici, 16 Gennaio 1996. G.U. 5-2-1996, N.29, Normetecniche per le costruzioni in zona sismica.- Decanini L.D., Liberatore L., De Sortis A., Benedetti S. (2006).Rapporto sui comportamenti, osservati in laboratorio e dopo terremotiseveri, sulle prescrizioni di diverse normative, nazionali e internazio -nali, a proposito di edifici a struttura mista muratura-c.a.. Rapporto diricerca RELUIS. Novembre 2006.- Decanini L.D., Tocci C., Liberatore L., Benedetti S. (2008). Analisinumeriche non lineari agli elementi finiti su modelli piani e spaziali diedifici a struttura mista muratura-c.a.. Rapporto di ricerca RELUIS.Gennaio 2008.- NAA-80, 1980. Normas Antisismicas Argentinas.- Nardone F., Verderame G.M., Prota A., Manfredi G. (2008). Analisicomparativa su edifici misti C.A.-Muratura. Valutazione e riduzionedel la vulnerabilità sismica di edifici esistenti in c.a. Roma 29-30Maggio 2008. MIX-04.- O.P.C.M. n.3431, 2005. Suppl. Ord. n.85 alla G.U. n.107 del10/05/2005 e s.m.i.

6.5 TAMP: Influence of Infills on Structural Response- Amato G., Cavaleri L., Fossetti M., Papia M. (2008). “Infilled frames:influence of vertical loads on the equivalent diagonal strut model”,Proc.14th WCEE, Beijing, China, CD-ROM, Paper 05-01-0479.- Cavaleri L., Fossetti M., Papia M. (2005). “Infilled frames: develop-ments in the evaluation of the cyclic behaviour under lateral loads”,Structural Engineering and Mechanics, Teckno Press, (21) 469-494.- Decanini L.D., Bertoldi S.H., Gavarini C. (1993). “Telai tamponatisoggetti ad azioni sismiche, un modello semplificato, confronto speri-mentale e numerico”, Atti 6° Convegno ANIDIS, Perugia.- Durrani A.J., Luo Y. H. (1994). “Seismic retrofit of flat-slab buildingswith masonry infills”, Proceedings from the NCEER Workshop onSeismic Response of Masonry Infills-Report NCEER-94-0004, 1-3.- Mainstone R.J. (1971). “On the stiffness and strengths of infilledframes”, Proceedings Institution of Civil Engineers, London,Supplement IV, Paper 7360S, 57-90.- Stafford-Smith B. (1966). “Behavior of square infilled frames”,Journal of the Structural Division, ASCE, 92 (1), 381-403.- Stafford-Smith B., Carter C. (1969). “A method of analysis for infilledframes”, Proceedings Institution of Civil Engineers, London, (44) 31-48.

6.6 SCALE: Behavior and Strengthening of Stairs- Berry R. (1999), The construction of buildings. 5th edition Vol. 2.Blackwell Science.- Biskinis D.E., Roupakias G.K, Fardis M.N., (2004).“Degradation of


CEN Comité Européen de Normalisation (2003). Eurocode 8 – Designof structures for earthquake resistance - Part 3: Assessment and retro-fitting of buildings (draft n. 6)”, prEN 1998-3, Brussels.- Masi A., Santarsiero G. (2008). “Sperimentazione su nodi trave-colon-na in c.a. progettati con diversi livelli di protezione sismica: primi risul-tati”, Atti del convegno Valutazione e riduzione della vulnerabilità sismi-ca di edifici esistenti in cemento armato, E. Cosenza, G. Manfredi, G.Monti editors, Polimetrica International Scientific Publisher, Roma,Italia.- Masi A., Santarsiero G., Moroni C., Nigro D., Dolce M., Russo G.,Pauletta M., Realfonzo R., Faella C., Lignola G.P., Manfredi G., ProtaA., Verderame G.M., (2008). “Behaviour and strengthening of RCbeam-column joints: experimental program and first results of theresearch activity in the framework of DPC-Reluis Project (ResearchLine 2)”, Proc. of the 14th World Conference on Earthquake Engineering,Beijing, China.- NTC 2008, Decreto Ministeriale 14 gennaio 2008, Norme tecniche perle costruzioni, Suppl. or. n.30 alla G.U. n.29 del 4/2/2008.

6.8 BIAX: Behavior and Strengthening of Columns under CombinedAxial Load and Biaxial Bending and Shear- Alessandri S., Monti G. (2008). “Design equations for the assessmentand FRP-strengthening of reinforced rectangular concrete columnsunder combined biaxial bending and axial loads”. Journal Mechanics ofComposite Materials, 44 (3), pp. 197-324.- Di Ludovico M., Verderame G.M., Iovinella I., Cosenza E. (2008a).“Domini di curvatura in pressoflessione deviata di sezioni in c.a. –PARTE I: Analisi a fibre”. Proc. Reluis2Rm08 “Valutazione e riduzionedella vulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 mag-gio, E. Cosenza, G. Manfredi, G. Monti Eds., Polimetrica InternationalScientific Publisher, ISBN 978- 88-7699-125-5, pp. 629-638.- Di Ludovico M., Verderame G.M., Iovinella I., Cosenza E. (2008b).“Domini di curvatura in pressoflessione deviata di sezioni in c.a. –PARTE II: Valutazione semplificata”. Proc. Reluis2Rm08 “Valutazionee riduzione della vulnerabilità sismica di edifici esistenti in c.a.” Roma29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., PolimetricaInternational Scientific Publisher, ISBN 978- 88-7699-125-5, p. 639-648.- fib (2003) “Seismic assessment and retrofit of reinforced concretebuildings”. Bulletin 24, ISBN 978-2-88394-064-2.- Bosco M., Ghersi A., Leanza S. (2008). “Force displacement relation-ship for r/c members in seismic design”. Proceedings of the 14th WorldConference on Earthquake Engineering. October 12-17, Bejing, China. - Canducci G., Camata G., Spacone E. (2008). “Problematiche con-nesse alla descrizione del moto sismico”. Proc. Reluis2Rm08 “Valuta -

shear Strength of Reinforced Concrete Members with Inelastic CyclicDisplacements. ACI Structural Journal title no. 101-S76.- Cosenza E., Mariniello C., Verderame G.M., Zambrano A. (2007), Ilruolo delle scale sulla capacità sismica degli edifici in C.A. Atti Del XIIConvegno Anidis “L’ingegneria Sismica In Italia”, 10-14 Giugno 2007,Pisa, Italia. (in Italian).- Cosenza E., Verderame G.M., Zambrano A. (2007), Rapporto sul-l’analisi tipologica delle scale e criteri di progettazione statica e sismi-ca. Rapporto di ricerca RELUIS. (in Italian).- Cosenza E., Verderame G.M., Zambrano A.(2008), Capacità sismicadegli edifici esistenti in C.A.: L’influenza delle scale, Proc. Reluis2rm08“Valutazione e riduzione della vulnerabilità sismica di edifici esistentiIn C.A.” Roma 29-30 Maggio, E. Cosenza, G. Manfredi, G. Monti Eds.,Polimetrica International Scientific Publisher, Isbn 978-88-7699-125-5. (in Italian).- Marrullier E. (1910), Guida Pratica per la costruzione degli edifici conspeciale riguardo al cemento armato, Editori Torinesi. (in Italian).- Migliacci A. (1977), Progetti di strutture, raccolta delle lezioni tenutepresso il Politecnico di Milano negli anni accademici 1966-67, 1967-68, Parte seconda, Edizioni Masson, Milano. (in Italian).- Pagano M. (1963), Strutture, Liguori Editore. (in Italian).- Perri E. (1966), Ingegneria Sismica, Unione tipografico- EditriceTorinese. (in italian).- Reynolds C. E., Steeedman J.C., (2002). Reinforced concrete design-er’s handbook, Spon Press Taylor & Francis Group London.- Rosci L. (1939), Manuale pratico di volgarizzazione del Calcolo delCemento Armato, G. Lavagnolo, II edizione, Torino. (in Italian).- Santarella L. (1953), Il cemento armato, le applicazioni alle co -struzioni civili ed industriali, Vol. II, 13, Edizione Hoepli. (in Italian).- Santarella L. (1957), Il cemento armato, monografie di costruzioni ital-iane civili ed industriali, Vol. III, 13, Edizione Hoepli. (in Italian).- Sezen H., Moehle J.P., (2004). “Shear Strength Model for LightlyReinforced Concrete Columns, Journal of Structural Engineering,ASCE November 2004, pp.1692-1703.- Tecnica y Pratica del Hormigon Armado. (1989) Ceac ed. Barce -lona.(in Spanish).- Zhu L., Elwood K.J., Haukaas T., (2007). “Classification and seismicsafety evaluation of existing reinforced concrete columns”, Journal ofStructural Engineering, ASCE. 1316-1330.

6.7 NODI: Behavior and Strengthening of Beam-Column Joints- Braga F., De Carlo G., Corrado G.F., Gigliotti R., Laterza M., Nigro D.,(2001). “Meccanismi di risposta di nodi trave-pilastro in c.a. di strut-ture non antisismiche”, Atti del X Congresso Nazionale “L’ingegneriaSismica in Italia”, Potenza-Matera, Italia.


zio ne e riduzione della vulnerabilità sismica di edifici esistenti in c.a.”Roma 29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., Poli -metrica International Scientific Publisher, ISBN 978-88-7699-125-5,pp. 327-334.- Panagiotakos T., Fardis M.N. (2001). “Deformation of R.C. membersat yielding and ultimate”. ACI Structural Journal, 98 (2), pp. 135-148. Park R., Paulay T. (1975). “Reinforced concrete structures”. J. Wileyand Son, Inc., N.York, 769 pp.- Verderame G.M., De Carlo G., Manfredi G., Fabbrocino G.(2008a).“L’aderenza ciclica in campo elastico delle barre lisce. Parte I: la spe -rimentazione”, Proc. Reluis2Rm08 “Valutazione e riduzione della vul-nerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E.Cosenza, G. Manfredi, G. Monti Eds., Polimetrica InternationalScientific Publisher, ISBN 978-88-7699-125-5- Verderame G.M., Ricci P., Manfredi G., Fabbrocino G. (2008b).“L’ade renza ciclica in campo elastico delle barre lisce. Parte II: la mo -del lazione”. Proc. Reluis2Rm08 “Valutazione e riduzione della vulnera-bilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E. Co -senza, G. Manfredi, G. Monti Eds., Polimetrica International ScientificPublisher, ISBN 978-88-7699-125-5 - Verderame G.M., Ricci P., Manfredi G., Cosenza E. (2008c). “Lacapacità deformativa di elementi in c.a. con barre lisce: modellazionemonotona e ciclica”. Proc. Reluis2Rm08 “Valutazione e riduzione dellavulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E.Cosenza, G. Manfredi, G. Monti Eds., Polimetrica InternationalScientific Publisher, ISBN 978-88-7699-125-5, p. 617-628.

6.9 PREFAB: Behavior and Strengthening of Prefabricated IndustrialStructures- Felicetti R., Lamperti M., Toniolo G., Zenti C., (2008). “Analisi speri -mentale del comportamento sismico di connessioni tegolo-trave distrutture prefabbricate”. XVII Congresso C.T.E., Roma, 5-8 Novembre,Italy, pp. 867-874.- Magliulo G., Capozzi V., Fabbrocino G., Manfredi G., (2008). “Deter -minazione sperimentale del coefficiente di attrito neoprene-calcestruz-zo per la valutazione della vulnerabilità sismica delle strutture prefab-bricate esistenti”. Proc. Reluis2Rm08 “Valutazione e riduzione dellavulnerabilità sismica di edifici esistenti in c.a.”, Roma, 29-30 maggio, E.Co senza, G. Manfredi, G. Monti Eds., Polimetrica International Scien -ti fic Publisher, ISBN 978-88-7699-129-5, p. 717-724.- Palermo A., Camnasio E. and Poretti M., (2008). “Role of DissipativeConnections on the Seismic Response of One-Storey Industrial Build -ings”, Proc. of the 14th World Conference on Earthquake Engineering,Beijing, China.

II - SEISMIC ASSESSMENT AND RETROFIT OF EXISTINGBRIDGES

1. INTRODUCTION

The perception of the risk associated to the seismic vulnerability ofthe transportation infrastructure, and in particular to that of bridge

structures, on the part of both the relevant authorities and the profes-sion is a quite recent acquisition in Italy. This is possibly due to the factthat in the last two major events that have struck the Country in the sec-ond half of the 20th century (Friuli 1976 and Irpinia 1980) the trans-portation infrastructure has not suffered significant distress. In particu-lar, in Friuli the construction of highways was just at the beginning. Inthe Apennine crossing of the A16 highway the bridges did undergosome damage, mainly due to the inadequacy of the bearing devices, butthis was promptly remedied by the owner through the systematic adop-tion of the then innovative technique of seismic isolation.

On the other hand, it can be observed that this delay in the apprecia-tion of the risk is not exclusive to Italy. For example, it is enough tomention that it took twelve years after the spectacular failures of quitemodern bridges (Figure 1, left) during the San Fernando (1971) earth-quake, for the Federal Highway Administration (FHWA) to publish afirst document titled “Retrofitting guidelines for Highway Bridges”(FHWA-ATC, 1983). Still, in 1989, despite of the large retrofit programset up (later proved to be fully inadequate), the Loma Prieta earthquakeexposed substantial deficiencies in bridges in California (Figure 1,right).


Fig. 1- Damage to bridges during the San Fernando, 1971 (left) and Loma Prieta 1989 (right) events.

The situation as briefly outlined above is sufficient to understand thatthe state of the art on seismic assessment and retrofit of bridges stillneeds to be advanced in several areas. The research undertaken withinthis Line of the DPC-Reluis Project aimed at providing a contributionin this direction. The areas considered to be of prioritary interest wereassessment methods, retrofit criteria and techniques, abutments andfoundations, with the final goal of producing a comprehensive docu-ment with guidelines and example applications. This result, which hasbeen achieved, represents the first European document on the topic and

text on the seismic assessment and retrofit of existing bridges.


The program was articulated into five main tasks:1) Identification of bridge typologies:Bridge typologies characterising the Italian road- and railway systemswere identified through contacts with the main national and regionaladministrations, as well as with major contractors. In particular, exist-ing contacts were exploited and new ones were established with ANAS,Autostrade, Italferr, RFI, Ferrovie della Calabria in order to acquiredetailed documentation on a number of representative structures.2) Assessment methods:Existing methods developed for the assessment of buildings wereextended to the deal with bridge structures, and methods specificallydevised for bridges were further developed. The goal was to fine-tuneseveral methods of increasing level of accuracy and required effort to beused according to the importance and size/regularity of the bridge.These include displacement-based linear and non linear static meth-ods, as well as simplified behaviour-factor-based methods. Specialattention was also devoted to modelling for non-linear analysis.3) Retrofit criteria:This task focussed on the specific aspects of the application of tradi-tional and innovative retrofit techniques to structural elements typicalof bridges. The program of activity included the execution of an exper-imental campaign aimed at establishing the effectiveness of alternativeretrofit techniques. This task also included the development of seismicisolation solutions to be applied in the retrofit of bridges that were notinitially designed to be isolated, as well as bridge-specific seismic iso-lation design criteria.4) Assessment of abutments, earth-retaining structures and founda-tions:Abutments and foundations are often weak elements in existingbridges. The goal of this task was to advance the state of the art in theseismic analysis and assessment of these components, an area stillcharacterised by the widespread use of mostly empirical or convention-al approaches.5) Model applications to bridges of different typologies:It was planned that under this task at least one bridge for each of themain typologies identified under task 1 was to be subjected to a detailedassessment and retrofit design, and then documented in an applicationmanual to complement the pre-normative document.The guidelines and the companion manual represent the mainoutcome of the project.The five broad tasks outlined above were split into a number of sub-

could be envisaged to form the basis for a future addition to theEurocodes system.


Starting from the year 1992 on funding from the FHWA, a vast researchprogram has been undertaken in the US to clarify several aspects relat-ed to the seismic assessment and retrofit of bridges.The first product of the above research appeared in 1995 in the form ofthe “Seismic Retrofit Manual for Highway Bridges” (FHWA, 1995): itsfurther development has led to the “Seismic Retrofitting Manual forHighway Structures: Part 1 Bridges” and the “Seismic RetrofittingManual for Highway Structures: Part 2 Retaining structures, Slopes,Tunnels, Culverts and Roadways” (FHWA- MCEER, 2005).In Europe the Eurocodes system includes a normative document for theseismic design of new bridges, which is at least partially based on therecent concepts of performance-based design: Eurocode 8 Part 2 (CEN,2005a). This document, however, is not matched by a companion onecovering existing bridges, differently with the situation of buildings, forwhich such a document is available in the form of Eurocode 8 Part 3(CEN, 2005b). In the year 2003 a firm change of direction towards the harmonisationwith Eurocode 8 has occurred in the Italian normative framework forseismic design. In that occasion the priority was given to the drafting ofdocuments for the design of new structures, both buildings and bridges.A document for existing structures was also introduced, but again lim-ited to buildings. These documents served later as the basis for the pro-duction of the seismic chapter of the current Eurocodes-aligned nation-al design code produced by the “Ministero delle Infrastrutture”(DM2008).The need for a document dealing explicitly with the problem of assess-ing and retrofitting bridges in seismic areas dates back actually to 2003,when the update of the seismic design code was accompanied by theobligation of assessing, within the time limit of five years, all the strate-gic structures and infrastructures in the Country. Adhering to this oblig-ation and with reference to bridges, with funding from the CivilProtection Department, the National Agency for Roads and Highways(ANAS) has launched a program for the assessment of all its bridgestructures. Further, the theme is of pressing interest due to the wide-spread activity currently ongoing on the Italian highway network toincrease its traffic capacity.

Within the above context the DPC-Reluis research project, with its Line3 is intended to respond to the outlined needs, and in particular to thatof producing a document to be used as a proposal of a future normative


tasks according to the table below. The table also shows the progress ofactivity over the whole project duration of three years. The progressachieved during the second year is briefly summarised in the followingTable 1.

4. MAIN RESULTS

The main results of the research activity are summarised in the followingaccording to the research structure described in the previous section.

4.1 Task 1: Identification of bridge typologiesThe main bridge typologies on several seismic-prone portions of theItalian railway and road/highway networks have been identified duringthe first year of activity. In summary, the data collected from varioussources (mainly national/regional administrations) pertain to the Torino-Bardonecchia-Frejus (TBF) and the Parma-La Spezia (PLS) highways(Politecnico di Torino), the Firenze-Bologna (A1FiBo) portion of the A1Milano-Napoli and the Apennine portion of the A16 highways (Univer sitàdi Roma “La Sapienza” and Università di Roma Tre), the Adriatic por-tions of the A14 Bologna-Canosa highway and of the SS16 Adriaticastate-road (Università di Chieti-Pescara), the Roma-Viterbo (RMVT) andRoma-Sulmona (RMSu) railways (Università di Roma Tre), the regionalrailway and roadway networks of Calabria (Università di Cosenza).

Structural typologies characterizing the TBF and the PLS highways arequite different. The first highway, built in between 1983 and 1992,includes rather uniform typologies: a) about 300.000 m2 of precast seg-mental box girder bridges with pier heights up to 90m and span lengthsbetween 40 and 100m, b) about 200.000 m2 of girder bridges in con-crete and in composite steel-concrete with pier heights between 5 and30m and span lengths between 20 and 80m. Representative bridgesare: the Borgone viaduct (20+26×40+20 m), the Ramat viaduct

(50+9×100+50 m and tall piers), Bardonecchia bridge (7×42 m) and theMillaures bridge (6×80 m, composite steel-concrete). The second high-way, built between 1965 and 1975 shows a greater typological variabil-ity, which can be reduced, however, to a few homogeneous sets.Representative bridges are: the Borgotaro viaduct (slab bridge with sev-eral interconnections), the Narboreto bridge (4×30 m), the Rio Verdeviaduct (2?65+6×95+76 m and very tall piers, h=150 m) and theRoccaprebalza South viaduct (13×45 m and tall piers), Rio Barcalesa(7×43m).

As it regards the infrastructure in the Abruzzi region typologies andconditions were monitored along A14 and SS16. Data sheets, consist-ing of 7 sections that provide location, type, category, geometrical andenvironmental characteristics, condition, and photographic description,were compiled to catalogue all 52 bridges of the latter road. Bridgeswere classified according to structural type, material and geometry. Thebridge conditions, the piers, and the cracking and vegetative maps wereconsidered.

Bridge structures on the A1FiBo and A16 were scrutinised, and aselected number of bridges either representative of the most frequenttypologies or significant for their outstanding design was identified. Afurther screening of the set including these structures and those identi-fied on the TBF and PLS has led to the definition of the final case-stud-ies for the detailed applications and the development and calibration ofassessment methods.

In the roadway and railway system of Calabria the most common typolo-gies are single-stem or frame piers with full, single-, or multi-cellularhollow-core cross-section; simply supported decks, made up of rein-forced or prestressed concrete girders and a cast-in-place RC deckslab. Two study cases were selected for the study of non-conventionalprotection/retrofit techniques. The first one is the Follone viaduct on theA3 Salerno-Reggio Calabria highway, where the spans have been con-nected with by means of longitudinal devices, while the second case isthe Val di Leto viaduct, on a provincial road, which was recently retro-fitted using oleodynamic devices.

Finally, the data collected during the survey of the two railway linesRMVT and RMSu, have allowed selection of four masonry arch bridges,two per line, to be used as case-studies for the calibration of analysismethods for masonry bridges.

4.2 Task 2: Assessment methodsIn summary, Task 1 has shown that the relatively many important


Table 1. Subdivision of the research activity into tasks and sub-tasks.

m = 0.3mpila + mpulv + mimp (1)

(2)

(3)

�u = �y + (�u – �y) � lp(H–lp / 2) (4)

(5)

(6)

The guidelines and application manual give a detailed description ofthe method, of the safety verifications of bearings, piers and founda-tions, and a complete worked-out example.

4.2.2 VERIFICATION OF APPLICABILITY OF MPA METHOD TO BRIDGE STRUC-TURES

The MPA method by Chopra and Goel (2002) has been devised for theanalysis of tall buildings. Its applicability as an alternative to inelasticdynamic analysis and adaptive pushover methods for the assessment of

H � (mpulv + 0.3mpila)Hp + mimpHimpm

�y = �yH2 / 31�

�T = 2� mi / k = 2� mi�y / Vy��max= SDe(T)

q�max= SDe(T) 1+(q–1) T < Tc

TcT[ ]

T � Tc o q 1

bridges crossing wide valleys in the mountain tracts of the Central andSouthern Apennine (A1FiBo, PLS, A3) represent a negligible percent-age of the total bridge stock, made up essentially of bridges with sim-ply-supported decks (prestressed or reinforced concrete girders plusslab) with single stem of frame piers. For this reason a special effort hasbeen devoted to devising a simplified non-linear method suitable for theanalysis of bridges with simply-supported decks (Università di RomaLa Sapienza). As it regards statically indeterminate bridges (continuousdecks) a distinction was made between those with special configuration,for which inelastic dynamic analysis is in most cases the method ofchoice, and simpler bridges, for which two methods have been thor-oughly explored: the Modal Pushover Analysis (MPA) method(Università di Roma La Sapienza) and the Secant Mode Superposition(SMS) method (Università di Pavia). Finally research has also focussedon two more issues, namely the always debated problem of directionalcombination rules (Università di Chieti-Pescara) and the non linearmodelling of seismic protection devices (Università di Cosenza).

4.2.1 SIMPLIFIED NON LINEAR METHOD FOR BRIDGES WITH SIMPLY-SUPPORT-ED DECKS

For these bridges it is possible to set up an ad hoc assessment proce-dure which represents a convenient trade-off between simplicity andaccuracy. The reference model is that of a vertical cantilever with a con-tinuous distribution of mass, on top of which rest the pier cap and thedeck masses. As long as the pier height is not such as to make highermode contributions significant, in the transversal direction each pierrepresents a single-degree of freedom oscillator (see Figure 2a). In thelongitudinal direction the entire bridge can also be represented as aSDOF system if seismic restrainers are provided that minimise the rel-ative movements of adjacent decks on top of the pier caps (see Figure2d). In this case the system has mass equal to the sum of the tributarymasses of the piers and resisting force sum of the resisting forces of thepiers (assuming that maximum displacements are permitted by theabutments joints).

The method consists of a simplified non linear static analysis in whichthe force-displacement laws are constructed based on the results ofmoment-curvature analysis of the pier bases (see Figure 2b). The fol-lowing equations give the tributary mass, Eq.(1), the effective height inthe transversal direction, Eq.(2), the yield and ultimate displacement(see Figure 2b and c), Eq.s (3) and (4), the period, Eq.(5), and the cor-responding demand displacement, Eq.(6), as for single-mode conven-tional pushover analysis. The effective height equals the pier’s height inthe longitudinal direction. Displacement capacity follows from ultimatedisplacement with appropriate safety factors.


Fig. 2- Simplified non linear method for the assessment of bridges with simply-supported decks.

a

b

c

d

bridge structures has been investigated through its application to the RioTorto viaduct (see Figures 3 and 4), one of the case studies selected forthe project. The structure, built at the end of 50’s, is characterized bythirteen-span twin decks realized with two girders and top slab. Thetwelve supports consist of a pair of framed piers, one under each deck.Each pier is a multi-storey reinforced concrete frame with variableheight, realized with two circular columns of diameter D =120÷160 cm.The results from inelastic dynamic analysis have been taken as bench-mark for the purpose. The response has been compared for severalintensity levels (to assess the influence on accuracy of the level of non-linearity in the response) and in terms of different response quantities,both local and global (section curvatures and element displacements).

The comparison provided the following indications: • The location of maximum modal displacement is the best choice asthe reference degree of freedom (DOF) for estimating the demand on thestructure. Each significant mode is therefore characterized by its ownreference DOF.

• The variation of the lateral load distribution, from one based on themodal (elastic) displacement shape to another based on the (plastic)displacement shape at failure, does not affect appreciably the results.• The best estimate of the displacements by the MPA (i.e. the onederived taking the optimal reference DOFs) matches reasonably wellthat from TH. It is worth noting that a comparable amount of approxi-mation on the response of the structure is obtained both in the elasticand in the plastic response regimes. This observation, together with theprevious one, indicates that the main approximation of the method, i.e.being based on the initial elastic modal vector, may not represent amajor limitation.• Differences between the nodal displacements estimated by the MPAwith respect to those by the TH are found to be in the order of 15%,independently of the intensity level of the ground motion. Analogousresults are observed also for the curvatures at members end-sections,resulting in almost coincident patterns of plastic hinge location and pre-dictions of members failures.

For the considered case, the application of the MPA method has shownto lead to fully acceptable results. Such a favourable conclusion stillawaits substantiation from a larger number of applications. Theseresults have led to the introduction of the method among the allowedmethods in the draft guidelines.

4.2.3 SMS METHOD

The Secant Mode Superposition method consists essentially of an iter-ative multi-modal response spectrum analysis on a structural modelwith secant stiffness properties and equivalent viscous damping. Theprocedure can be summarised in the following steps:• Step 0: A starting displacement profile and stiffness distribution areassumed;• Step 1: The stiffness matrix of the equivalent linear structure isassembled;• Step 2: Modal analysis is carried out;• Step 3: Displacement in each vibration mode are obtained either froman over-damped elastic or from an inelastic displacement spectrum;• Step 4: Modal contributions are combined to yield displacement pro-


Fig. 3- Longitudinal profile of the Rio Torto viaduct (A1FiBo).

Fig. 4- Two piers of the Rio Torto viaduct (A1FiBo).

• Pier height of 6 m (aspect ratio equals 4); •Longitudinal reinforcement: 80�10 (rL = 1.05%) with an overlappinglength equal to 20 diameters (200mm) at the base of the pier;•Transversal reinforcement: stirrups �6/150mm (rV = 0.38%);• Axial load equal to 1000kN (� = 4.3%) or 2000kN (� = 8.6%);• Concrete Rck400;• Steel FeB44K.

The retrofit intervention aimed to restore the tensile stress path from thepier section to the foundation, avoiding at the same time any plasticiza-tion of the overlapping region. The new stress path created using longi-tudinal FRP strips applied to the overlapping region is expected tocause the plastic hinge shift upwards where the longitudinal steel iswell anchored allowing for an efficient energy dissipation.During the design phase different possible solutions have been consid-ered concerning the retrofit materials (carbon, aramid or glass FRP), theretrofit geometry (width and length of the region to be retrofitted), thetechniques to be used for the anchoring of the FRP strips to the foun-dation. This was possible employing a numerical FE model developedto predict/reproduce the tests results.Regarding the materials, the final choice was to use carbon FRP (C-FRP): the analyses indicated that this material is the only one able tosustain the acting tension forces. Too many FRP layers would have beenneeded to carry the same force using glass or aramid fibres, affectingthe effectiveness of the retrofit intervention.

For what concern the geometry of the retrofit intervention, the finalsolution was to apply longitudinally two C-FRP layers on the four sidesof the specimen. As far as the exploitation of the material strength is

file and moment distribution (different combinations rules were exam-ined);• Step 5: Two response indices are computed, that evaluate conver-gence on displacement profiles and force distributions, while checkingthat the structural capacity is not violated.• Step 6: A final response index is obtained as an average of the firsttwo and checking the convergence of the proposed iterative procedure.

The method has been thoroughly tested on the six “typological” bridges,with regular and irregular configurations, and different number of spansand span length. Verification of the method is versus non-linear time-history analysis in terms of maximum deck displacement, and maxi-mum pier shear forces has been carried out.

4.3 Task 3: Retrofit measuresThe experimental part of the research activity of Line 3 has been car-ried out at the University of Pavia and of Roma Tre. The two experi-mental campaigns have focussed with different goals on the testing ofpiers. The tests performed in Pavia were aimed at ascertain the effec-tiveness of FRP retrofit measures in order to confine hollow-core pierswith insufficient lap-splices, while those performed in Roma Tre wereaimed, using large-scale specimens, at the characterisation of theresponse of frame piers built in the ‘60s.

Finally, as it regards masonry bridges, a comprehensive survey of theexisting retrofit techniques for this type of structures has been carriedout by the University of Genova.

4.3.1 EXPERIMENTAL ACTIVITY ON FRP STRENGTHENING FOR INSUFFICIENT

LAPS PLICE

Four 1:2 scaled bridge piers were designed with an insufficient over-lapping length of the longitudinal bars across the critical zone thatshould lead to an early loss of the lateral strength due to bar slippage.The built specimens (see Figure 5 left and middle) have the followingcharacteristics:• Hollow-core rectangular cross-section (see Figure 5, right) with ex -ternal dimensions 800x1500mm and wall thickness of 150mm;


Fig. 5- Pier section and specimens built at the University of Pavia.

Table 2. Considered retrofit materialsMaterial Description fu[MPa] e[MPa] eult[%] Layer thickness [mm]SRP 3x2 High Density 1167 77773 1.50 1.1938SRP 12x High Density 948 64811 1.46 1.1938CFRP High Modulus 3000 390000 0.77 0.165GFRP Alkali Resistant 1700 65000 2.62 0.23AFRP High Modulus 2800 105000 2.67 0.214

concerned this choice appears to be questionable since the fibresapplied to the pier sides parallel to the imposed motion will not havethe same stress as those on the other two sides, but the adopted solutionseemed to be the only possibility to assure the maximum stress diffu-sion across the pier section. It is worth mentioning that even thoughanchoring 1500kN force to the foundation of the scaled specimen wouldhave been probably feasible using a steel collar fixed to the foundationwith some high-strength steel bars taking advantage of the deep foun-dation of the specimen, moving back to real structures the anchoring tothe foundation of the tensile force induced in the FRP by a seismicexcitation would have been much more difficult, if not unfeasible.Spreading the tensile force on the four sides of the pier, the anchoringis clearly easier. Between different possibilities initially considered toanchor such force, final choice was to use an anchoring system realisedwith FRP too. The idea was to employ aramid connectors, normallyused to transfer shear stresses. If this solution will be found to be effec-tive, as it seems from its design, multiple advantages will arise both onthe economic and technologic sides.

Due to external constraints only two piers have been tested within theduration of the project, those without the FRP retrofit in the lap-spliceregion. The tests confirmed that, as expected, lap-splice with an over-lapping length equal to 20 times the diameter of the spliced bars isinsufficient to assure the anchoring of the bars (see Figure 6a). The testsalso underlined that the effectiveness of the lap-splice decreases whilethe axial load increases: that is because of the higher stresses and dam-ages (such as partial concrete spalling) in the overlapping region. Figure 7 shows the base shear-top displacement diagrams derived from

the performed tests: here the lateral load carrying capacity drops quitequickly because of the bars sliding. The red curves are the result of thenumerical model.

4.3.2 FINITE ELEMENT MODELLING CALIBRATED TO THE EXPERIMENTAL TESTS

RESULTS

Given the large number of seismically under-designed bridges, thatneed to be assessed and potentially retrofitted due to insufficient lap-splicing, the development of an efficient analytical model to simulatethe response of FRP-retrofitted elements was deemed critical. A quitesimple though effective finite element model was developed usingSeismostruct (Seismosoft, 2006). Figure 8 shows the adopted numericmodel. The longitudinal FRP layers have been represented like an ele-ment itself. Rigid links have been used to place the FRP at the rightdistance from the longitudinal axis of the retrofitted member in order forthem to be able to give the right contribution to the flexural strength ofthe pier. Furthermore, each FRP element has pinned connection at bothends in order to be subjected to pure axial load. On the other hand, theFRP wrapping can be modelled in approximation without adding ele-ments to the model, since its main effect is the increased concrete con-finement that can be represented by the confinement factor already pre-sent in adopted concrete stress-strain representation (Mander, 1988;Martinez-Ruenda and Elnashai, 1997). To tests the effectiveness of the


Fig. 6- Test Set-up (a) and open crack at the pier base during the test with 2000kN axial load (b).Fig. 8- FRP retrofitted pier model.

Fig. 7- Base shear-top displacement diagrams of the two “as-built” pier (a) N = 1000kN (b) N = 2000kN.

adopted finite element model, the behaviour of 1:4 scaled square hol-low section piers from previous experimental campaigns (Calvi et al.,2005 and Pavese et al., 2004) has been reproduced through push-overanalysis.

4.3.3 EXPERIMENTAL ACTIVITY ON LARGE-SCALE SPECIMENS OF FRAME PIERS

Large-scale tests on framed piers have been undertaken at theUniversità of Roma Tre. This typology, characteristic of many oldviaducts of the Italian highway system, has been chosen for its high

beam, (see Figure 11a), a suitable grid of displacement transducers hasbeen placed on this beam, in order to measure the cracks amplitude, inthe other two specimens. In the second two mock-up’s a different fail-ure mechanism occurred on the same transverse beam. In particular,during the second test both beam-column joints collapsed (see Figure11b-c), while in the third one, only the left end of the beam failed inshear (Figure 11d), with a simultaneous failure of the right beam-col-umn joint (Figure 11e). This outcome was a nice experimental verifica-tion of the effect of material fluctuation on determining which amongstsimilarly resistant failure mechanisms actually occurs in reality.

seismic vulnerability. Among the representative bridges, a framed pierfrom the “Rio Torto” viaduct has been chosen (see Figures 3, 4 and 9).For the experimental program three mock-up’s of pier 12 without retro-fit have been realized and tested, with the goal of characterizing itscyclic response and relative collapse mechanism (see Figure 10).Subsequently, one or more reinforcing systems were meant to beapplied to the tested piers, for repeated testing to check the efficiencyand the reliability of the proposed reinforcing solutions.A ductile flexural failure was predicted with formation of plastic hingesfor this pier while, on the contrary, all three specimens failed in shear,either in the transverse beam or the joints, suggesting that the formulaemployed for the evaluation of the shear strength tends to overestimatethe ultimate shear.

Since the first test has shown a premature shear failure of the transverse


Fig. 11- Failure mechanisms of the transverse beam in the three tests.

Fig. 9- The viaduct “Rio Torto”.

Fig. 10- The viaduct “Rio Torto”.

Fig. 12- Experimental force-displacement cycles.

a b c d e

The differences between the failure mechanisms of the three piers, how-ever, have a little influence on the global behaviour, as shown in Figure12, which compares the global force-displacement cycles of the threespecimens.The experimental results have been compared with the results of anumerical model, which was set up using the non-linear code“OpenSEES”. Shear failure has been introduced using a shear force-deformation relationship with a tri-linear backbone and an appropriatedegradation law, included in a fiber non-linear element, using the sec-tion aggregator command. The yield-penetration at the base of the col-umn effect is particularly relevant due to the presence of plain steelbars. This phenomenon, if neglected, can induce an overestimation ofthe structural stiffness. This effect has been taken into account using azero-length element placed at the column base with a properly modifystress-strain law of the steel bars. Finally, the buckling phenomenonhas been taken into account using a corrected constitutive law of steel.The FE model used was able to reproduce accurately both the global aswell as the local behaviour, as shown in Figure 13.

4.4 Task 4: Abutments and foundationsThe activity under this task has been carried out at University of RomeLa Sapienza, and has dealt with two distinct problems: a) the develop-ment of an efficient non linear method for the analysis of diaphragm-type abutments, free standing and retrofitted with tie-backs; b) thereview of the literature on soil-foundation-structure interaction with thegoal of providing detailed indications for practitioners to be includedinto the assessment guidelines.

4.4.1 A SIMPLIFIED NON LINEAR DYNAMIC MODEL FOR THE ANALYSIS OF

ABUTMENTS

A simplified model for the dynamic analysis of diaphragm walls retain-ing dry cohesion-less soils with horizontal back-slope subjected to seis-mic excitation has been developed (Franchin et al, 2007a). The modelis based on the well-known one-dimensional Winkler approximationand on the non-linear shear-beam model for the ground layers on bothsides of the wall (see Figure 14). The model can include anchor-ties andcan account for non-linearity in all of its elements (retained soil,anchors and wall). According to preliminary numerical applications,which include validation of the proposed model results versus those ofa refined plane-strain nonlinear finite-element analysis carried out witha commercial code, the model appears to yield quite accurate predic-tions of static and dynamic bending moment distributions and perma-nent wall displacements.Next the developed model has been applied for the analysis of theresponse of a diaphragm abutment prior and after upgrading interven-tion with change of the support conditions and insertion of tie-backs(Franchin et al 2007b). The analysed structure is represented in Figure15.The application of the model has shown its versatility in assessing thesystem response in its existing state and in progressive states of upgrad-


Fig. 13- left: Comparison between theoretical and numerical force-displacement curves; right: base column rotations.

Fig. 15- Diaphragm abutment retrofitted with anchor ties.

Fig. 14- The developed model for diaphragm type. Fig. 16- Results of the abutment analysis: left, moment diagrams; right, top displacement time-histories.

soil-foundation system, consisting of the stiffness and damping func-tions (of the frequency) to be assigned at the pier base. This impedanceincludes the evaluation of the frequency dependent “dynamic” groupeffect, i.e. the modification of the impedance obtained as a simple sum-mation of the individual pile impedances to account for the interactionof the wave-fields produced by each pile.– Evaluation of the response in the frequency domain. This has beendone both with a purpose-made code and with a commercial finite ele-ment software that implements frequency-domain analysis (Sap2000).The resulting power-spectrum of the displacement components can beintegrated to yield the root-mean-square (RMS) or standard deviation ofresponse, from which maxima to be used in verification are readily

ing, in terms of both forces (Figure 16 left) and dynamic displacements(Figure 16 right). To the extent that it has been validated at present, themodel represents a very efficient tool for realistic design and assess-ment purposes.

4.4.2 CRITICAL REVIEW AND RECOMMENDATIONS ON METHODS FOR THE

ANALYSIS OF SOIL-FOUNDATION-STRUCTURE INTERACTION

A comprehensive review of the literature on the treatment of theresponse of deep foundations has been carried out. This has led to iden-tifying the available methods and their pros/cons. After the survey aselection has been made of those procedures considered suitable forpractical application and some numerical applications have been car-ried out to assess the relevance of the phenomenon (input motion mod-ification by kinematic interaction and foundation flexibility), in terms ofthe response of the superstructure.One example is the bridge structure shown in Figure 17. It is a simply-supported prestressed concrete deck of span length 30.0m typical of theItalian highway construction practice of the ’50s-‘70s with piers con-sisting of a single-stem with hollow-core circular section. The dimen-sions are in the figure. The foundation consists of a mat on 5 piles of


Fig. 17- Simply-supported deck on single-stem hollow-core pier founded on piles. Fig. 18- Complex impedance at the pier base: stiffness (top), camping (bottom), translation (left) and rotation (right).

1.5m diameter. Pile length is 20m. The bridge has 6 spans and a pierof height 20m has been considered. Soil can be classified based on theavailable information as type D. The structure has been modelled asshown in Figure 17e, i.e. as a three-degree of freedom system (includ-ing horizontal and rocking component of the base).The analysis has been carried out in the frequency domain using thesubstructuring approach. The steps of the analysis include:– Evaluation of the modification of the surface free-field motion (sup-plied as an acceleration response spectrum) due to the kinematic inter-action between soil and pile group. This step provides the power spec-trum of input displacement at the pier base.– Evaluation of the complex frequency-dependent impedance of the

obtained by multiplication for the peak factors.

Figure 18 shows the real (stiffness, top) and imaginary (damping, bot-tom) parts of the complex impedance at the base of the pier, for thetranslation (left) and rocking (right) displacement components. Theseare reported for two different values of the shear wave velocity Vs, bothcompatible with the soil type D. The figure also reports the stiff -ness/damping obtained by simple summation of the single pile contri-butions. Comparing the latter with those of the group allows to appreci-ate the frequency-dependent effect of the pile-to-pile interaction. Thiseffect reduces, by more than 50% in this case, the total stiffness.Finally, Figure 19 shows the power spectral densities of the response in

TRANSLATION ROTATION

CA

MPI

NG

CA

MPI

NG

Cq

STIF

FNES

S

STIF

FNES

S K

q

terms, on the left, of total displacement (relative to input motion, i.e.sum of the foundation translation, the structure deflection and the trans-lation due to rigid foundation rotation), and on the right of the structur-al deflection only. Results are reported for the two Vs values and, forreference, for the fixed-base response. As it can be seen, as expected,the fundamental period of the system elongates considerably due to theintroduction of the foundation flexibility: it starts at T=0.83 s in thefixed base case, and reaches about 1.5s and 1.75s for Vs=200m/s and100m/s, respectively. This increases the total displacements. The drifts,however, are considerably reduced as shown in Figure 19b.The above application, as well as the others carried out, allowed tointroduce in the guidelines quantitative indications on the need forinclusion of SSI into the modelling.

4.5 Task 5: Numerical application to case-studiesAll research units have contributed in producing a vast amount of case-studies that have been of considerable usefulness in checking consis-tency and practicality of the indications that now form the guidelines for

assessment. In this section only a limited overview of the applicationsis given to illustrate the work done. A more detailed description can befound in the final report for the Line 3. Table 3 reports all the analysedbridges.

4.6 Guidelines and Application manualThe activity only briefly summarised in the previous sections has rep-resented a necessary support for undertaking the task of writing whatwas the final product expected from Research Line 3: a proposal for aguidance document on the seismic assessment of existing bridges, anda companion set of example applications. The task, carried out byUniversity of Rome La Sapienza, has gone through several rounds ofscrutiny by all the units. In its final version it represents the firstEuropean document on the topic and could be envisaged to form thebasis for a future addition to the Eurocodes system. Indeed, the docu-ment is fully in line with Eurocodes and reflects to some extent theexperience on the seismic assessment of existing structures gained withthe use of Eurocode 8 Part 3 on buildings. It is also in line with the rel-evant chapters of the DM2008, related to seismic design of bridges, andincorporates its most recent developments on the definition of seismicaction.

The document produced consists of four chapters and two appendices:• Chapter 1: gives an introduction to the problem of seismic assessmentof existing bridges;• Chapter 2: contains the guidelines;• Chapter 3: is an overview of the most common retrofit measures andcriteria employed, without entering into the specifics of their design,making reference for this purpose to specialised texts on the topic;


Fig. 19- Results of SSI analysis on a bridge pier: power spectral densities of the response in terms of total displacement (left)and structural deflection (right).

Table 3. List of case-studies analysed according to the assessment guidelinesUnit Case-study Description AnalysisTorino Narbareto (PLS) Elastic RS analysis + q-factorTorino Rio Barcalesa (PLS) Elastic RS analysis + q-factorTorino Borgotaro (PLS) Elastic RS analysis + q-factorTorino Rio Verde (PLS) Elastic RS analysis + q-factorTorino Ramat (TBF) Elastic RS analysisChieti Vasto Marina (SS16) Elastic RS analysis + q-factorChieti Della Valle (A25) Elastic RS analysis + q-factorRoma Tre Rio Torto (A1FiBo) Inelastic time-history analysisLa Sapienza Rio Torto (A1FiBo) Modal pushover analysisLa Sapienza Standard viaduct (E45) Simplified non-linear method, pushover,

linear dynamicCosenza Follone (A3) Inelastic time-history analysisCosenza Val di Leto Inelastic time-history analysis

4 simply supp. spans, circular hollow-core piers7 simply supp. spans, polygonal bi-cell. piersHollow-core slab deck, highly irregular plan, Figure 219 spans, steel deck with hollow-core RC piers5 spans, box-section, steel pier15 spans, frame piers10 spans, box-section, single-stem hollow-core piers13 spans, inelastic time-history analysis13 spans5 spans, simply-supported, and continuous after sectionwidening4 simply supp. spans, retrofitted with “link system”5 simply supp. spans, retrofitted with oledynamic devices

TOTA

L D

ISPL

ACEM

ENT

Us

(m)

STRU

CTU

RAL

DEF

LEC

TIO

N (m

)

5 spans, simply-supported, and continuous after sectionwidening

Simplified non-linear method, pushover, linear dynamic


Fig. 20- The Narbareto viaduct.

• Chapter 4: contains the numerical examples that illustrate the appli-cation of the methods presented in the guidelines. There are four appli-cations covering:– assessment, by means of the simplified non linear method, of a typi-cal simply-supported bridge with single-stem cantilever piers in its pre-sent state;– assessment, by means of linear and pushover analyses, of the previ-ous bridge in two different configurations, with a new continuous, wider,

composite steel-concrete deck, with and without seismic isolation;– assessment of the Rio Torto viaduct by means of inelastic time-histo-ry analysis.• Appendix A: presents the fundamentals of the response to multiple-support excitations and reviews a number of methods that can beemployed to analyse bridge structures for this effect;• Appendix B: presents the fundamentals of the soil-foundation-struc-ture interaction phenomenon and reviews a number of methods that can

Fig. 21- The Borgotaro viaduct.

North carriageway South carriageway


Fig. 22- The Della Valle viaduct.

Fig. 23- The Follone viaduct.

STATIC SCHEME

RETROFIT VIA STEEL BARS AND NEOPRENE PADS

the possibility of defining the concept of “residual” reference life.Though it is admitted that in our Country it seldom occurs that the deci-sion to demolish a bridge can be taken several years in advance, it mayhappen that, due to planned substantial modification of the trafficcapacity of the link, it will be economically more convenient at a futuredate to replace the bridge. In this case, if seismic upgrade must beundertaken, the concept of residual reference life may be invoked toassign to VR a more realistic reduced value. This possibility is not cur-rently included in the guidelines, though it is regarded as being in linewith the possibility allowed for existing structures to derogate fromstandard safety levels dictated for new structures.

4.6.2 CHAPTER 2, GUIDELINES: METHODS OF ANALYSIS

With respect to classification of methods in static and dynamic, linearand non linear, now common to all modern seismic design codes andgiving rise to the usual four alternatives, the guidelines restrict some-what the field of applicability of linear analysis. This is not unexpected.For new well-designed structures the role of analysis is a relativelyminor one, due to the many constraints (arising mainly from global andlocal capacity design) that guide the design. On the other hand, whenassessing an existing structure, the accuracy in the analysis may havea major economic impact on the retrofit, possibly avoiding it altogether.

The guidelines admit linear analysis of two types only: modal analysiswith unreduced elastic spectrum and verifications in terms of deforma-tion/forces (subject to stringent conditions on the response regularity),and modal analysis with a spectrum reduced by a limited value of thebehaviour factor of q=1.5.The main methods put forward by the guidelines are non linear staticand dynamic analyses. As already anticipated in § 4.2.1, a simplifiednon linear static method is proposed for the very frequent case ofbridges with simply supported decks. For continuous irregular bridgesthe use of more recent pushover variants (adaptive and/or multi-mode)is introduced as an alternative to full-fledged inelastic time-historyanalysis. The allowance for more than single-mode invariant pushoverrepresents a small step forward with respect to Eurocode 8 Part 2,which builds upon the results of recent wide-ranging studies on the per-formance of such methods in the analysis of bridges [see for ex.(Casarotti 2005), (Kappos et al, 2005), (Isakovic and Fischinger, 2005),(Lupoi et al, 2007), (fib, 2007), as well as the draft document “Inelasticmethods for seismic design and assessment of bridges” by Task Group11 of the European Association of Earthquake Engineering].

4.6.3 CHAPTER 2, GUIDELINES: SAFETY VERIFICATIONS

The guidelines introduce a format for bi-directional verification for both

be employed to analyse bridge structures for this effect.

The main body of the manual is represented by chapters 2 and 4, aswell as by the appendices. In the following the most significant or prob-lematic aspects are briefly reviewed and commented.

4.6.1 CHAPTER 2, GUIDELINES: DEFINITION OF THE SEISMIC ACTION

The seismic action is defined, in line with DM2008, by means of anelastic acceleration or displacement response spectrum characterizedby an average return period specified as a function of the limit state ofinterest.

The return period TR is obtained from the probability of exceedancePVR over the reference life VR. The latter is given in DM2008 as theproduct of two factors, the nominal life VN and the “use factor” CU. Theminima for PVR for each limit state are given in DM2008.

In the tentative applications of the guidelines it was raised the problemof the value to be attributed to VN and CU, especially with reference tothe first one. The uncertainty may arise in the choice between 50 and100 years for VN, when considering bridges over highways. TheDM2008 indicates 50 years for bridges of ordinary dimensions, typolo-gy and importance, and 100 years for bridges of large dimensions and“strategic” importance. One would then be probably directed towards100 years, in consideration of the importance of the bridge (it is on ahighway). The next choice is that of CU which leads unambiguously to2.0, since highways are roads of type A according to the Italian classi-fication of roads (i.e. considering, again, the functional importance ofthe road on which the bridge is located). The above choices would implya reference life of 200 years and, for the life-safety limit state, a TR ofabout 2000 years. It is observed that this conclusion would not to be inline with the safety criteria contained in Eurocode 8 Part 2 (Bridges)which indicates for highway bridges an importance factor �I=1.3 to beapplied to the action with TR = 475 years. This multiplication leads inmost of Italy to an action with a return period of about a 1000 years.This latter in turn is consistent with a reference life of about a 100years, which is also the design life specified in the Eurocodes for otheractions (e.g. corrosion).

An official response to the mentioned problem, whose relevance needsnot to be underlined, cannot but come from the competent authorities,which are in charge of choosing the safety levels.

Within the framework of the definition of the reference life one aspectthat deserves particular consideration in the case of existing bridges is


deformations and forces. In particular the format reads:

(7)

where Dx and Dy denote the demand quantities along the two orthogo-nal axes x and y, with Cx and Cy denoting the corresponding capacities.This format becomes, in terms of chord rotations and shear forces:

(8)

(9)

In the above equations the demand terms are understood as the com-bined effect of both orthogonal components of the seismic action. Forexample, with reference to chord rotation, for the case of multi-modalnon linear static analysis one has:

(10)

where the directional combination is of the SRSS type and the summa-tion is over the modes.

4.6.4 APPENDICES

The matter covered in these two appendices, i.e. the response of bridgestructures to different motions at the piers’ bases and the effect of thesoil-foundation system deformability in modifying the input motion aswell as the response of the structure, has been always mentioned incodes without, however, neither precise quantitative indications on theinstances in which these phenomena have to be accounted for, nor ofphysically sensible yet practically applicable methods to do it. The rea-son for this resides clearly in the insufficient advancement on basicresearch. In drafting the guidelines, however, it was considered appro-priate to provide a presentation of selected state-of-the-art approacheswhich are susceptible of practical application.

For what concerns the effect of multiple-support excitation, the guide-lines indicate that the phenomenon should be accounted for wheneversoil conditions along the bridge belong to different soil categories. Theguidelines also present:• A stochastic model of the motion at the supports (Der Kiureghian,1996) that can be used either to generate samples of correlated motionsto be used in time-history response analysis or in linear random vibra-tion analysis;• The multiple response spectrum method (Der Kiureghian andNeuenhofer, 1992), which provides a solution for the random vibrationsproblem of a system subjected to multiple inputs based on the use of thecorresponding input displacement response spectra;• A simplified proposal for time-history analysis employing indepen-

(�x/�u,x)2 + (�y/�u,y)

2 �1�(Vx/Vu,x)

2 + (Vy/Vu,y)2 �1�

�x=�xG± �[(�xEx,i–�G)2

+ (�xEy,i–�G)2]� N

i=1

+ � 1�2 2Dy

Cy) )((DxCx

dent motions at the supports representative of the local soil conditions,which can be applied using currently available commercial finite ele-ment codes (Monti and Pinto, 1998);

For what concerns soil-foundation-structure interaction the guidelinesgive a classification of the approaches and present with some detail thesubstructuring method, in its application to pile (Novak 1974, Makrisand Gazetas, 1991 and 1992) and caisson foundations (Gerolymos andGazetas, 2006a,b). In this method the structure and the soil-foundationsystem are separated and studied accordingly. The study of the soil-foundation system consists of the solution of so-called kinematic inter-action and inertial interaction problems, leading to the modified inputmotion for the structure and to (complex) impedance to be put at thestructure base, respectively. Then the structure is analysed, with a flex-ible support condition, under the previously determined modifiedmotion. All the formulas necessary to perform this procedure are pre-sent in the Appendix.

5. DISCUSSION

The main objective of the project, which was the production of the draftguidelines and their application manual, has been met. In this respectthe Research Line was successful, since the product has been deliveredand its quality is believed to be high.Though it wasn’t explicitly included into the remit for the Line, it mustbe noted that the research group initially intended to cover in the guide-lines both structural concrete and masonry bridges. In spite of theresearch carried out, however, this more ambitious goal could not beachieved.Research on this front was essentially under the responsibility of theunit of Genova. This unit has produced during the three years of theproject a considerable amount of high-quality research that has beenregularly documented in the annual as well as the final reports, and itis also available in research reports from the unit uploaded on the pro-ject website. Quoting from the final report the issues dealt with by theunit cover the following: “i) statistical characterization of the Italianbridge population; ii) mechanical models for solid clay brickwork,needed for detailed and simplified structural models; iii) in field testingof masonry bridges, aiming at the identification of the main mechanicalproperties of the materials and of the bridge as a whole; iv) laboratorytesting of brickwork prisms; v) reduced scale testing aiming at identify-ing the load carrying capacity and the collapse mechanisms of shallowand deep arches taking into account the fundamental collaboration ofthe so called “non structural elements”; vi) reduced scale testing aim-ing at identifying the dynamic properties of shallow and deep arches


tion of expected loss related to any given bridge. Looking now at theproblem of bridge protection from an higher perspective, the attentionshould be directed at the bridges as components of road links forminga transportation infrastructure. The seismic performance of the singlebridge would then be put in relation with the performance of all otherbridges to be able to estimate the overall decrease in functionality of thewhole infrastructure. In this respect the very challenging problem ofdetermining the loss in traffic capacity of a damaged bridge representsan essential element.

7. MAIN REFERENCES

- Casarotti C. (2005), “Adaptive pushover-based methods for seismicassessment and design of bridge structures” PhD thesis ROSE School,Pavia, Italy.- CEN (2005), “Eurocode 8 Part 2: Seismic design of bridges”European Committee for Standardization, Brussels, Belgium.- CEN (2005) “Eurocode 8 Part 3: Assessment and retrofitting of exist-ing structures” European Committee for Standardization, Brussels,Belgium.- Chopra A.K., Goel R.K. (2002), “A modal pushover analysis proce-dure for estimating seismic demands for buildings” EarthquakeEngineering & Structural Dynamics Vol 31(3), pp. 561-582.- Der Kiureghian, A. (1996), “A coherency model for spatially varyingground motions” Earthq. Eng. & Struct. Dyn. Vol. 25, pp. 99-111.Der Kiureghian, A., Neuenhofer, A. (1992), “Response spectrummethod for multi-support seismic excitations” Earthq. Eng. & Struct.Dyn., 21: 713-740- DM2008 (2008), “Nuove norme tecniche per le costruzioni” DecretoMinisteriale del Ministero delle Infrastrutture 14/1/2008.- FHWA (1995), Seismic Retrofitting Manual for Highway Bridges,Publ. FHWA-RD-94 052, Federal Highway Administration.- FHWA-ATC (1983), “Retrofitting guidelines for Highway Bridges”Report ATC-06-2, Applied Technology Council, Redwood City,California.- FHWA-MCEER (2006), Seismic retrofitting manual for HighwayStructures. Part 1- Bridges.- fib (2007) “Seismic bridge design and retrofit – structural solutions”Bulletin 39, International Federation for Structural Concrete.- Franchin P., Pinto P.E., Noto F (2007), “A nonlinear dynamic modelfor seismic analysis of earth-retaining diaphragm-walls” Proc 4th Int.Conf. Earthquake Geotech. Engng, Thessaloniki, Greece. - Franchin P., Pinto P.E. (2007), “Analysis Of Diaphragm-Type BridgeAbutments Before And After Seismic Upgrading” Proc 1st US-Italyworkshop on seismic design and assessment of bridges, Pavia, Italy.

taking into account the collaboration of the so called “non structuralelements”; vii) Limit Analysis procedures for the analysis of masonrybridges taking into account the contribution of all the bridge elements;viii) retrofitting techniques for the bridge and its components.” As itmay be seen all of the investigated topics are of clear scientific interest,though not specifically relevant to seismic assessment of bridges. Thisis simply the unavoidable consequence of the international lack of fun-damental knowledge on the seismic behaviour of masonry bridges.


The research carried out within Research Line 3 has included the state-of-the-art into a document usable for assessing the protection level ofbridges against a number of limit-states.There are certainly several areas where improvement is possible anddesirable, and in particular these are:• The non-linear static analysis for bridges of complex geometry;• The ultimate strength and deformation capacity of structural memberssuch as those encountered in bridge structures (e.g. polygonal multi-cell. hollow-core cross-sections)• The generation of ground motions for multiple-support excitation.While generated motions are being progressively replaced with record-ed ones for the analysis of buildings, their use appears unavoidable forthe analysis of bridges whenever different motions must be consideredat the supports. Currently available procedures are in need of consider-able improvement.• The vast literature on SSI needs to be acquired and digested by struc-tural engineers to become a practical tool. This a crucial aspect in viewof the displacement-based framework of the guidelines and the corre-sponding need for more accurate evaluations of deformations.

The guidelines do not cover the seismic isolation technique. The reasonfor this choice is that the design of seismic isolation does not varybetween new and existing bridges. Seismic isolation, however, will cer-tainly see much further diffusion in the coming years, for new as well asexisting bridges, while isolation device technology continues to evolverapidly with the ensuing need of developing appropriate analysis anddesign techniques. In this respect this can be regarded as an ongoingresearch topic.

To the extent that solutions to the problem of assessing the protection ofa bridge against its ultimate state can be considered to be sufficientlymature, the next important passage is that of being able of estimatingstructural and monetary damage as a continuous discrete function ofseismic intensity. Achievement of this goal would allow for the estima-


- Gerolymos N., Gazetas G. (2006), “Winkler model for lateral responseof rigid caisson foundations in linear soil” Soil Dyn. & Earthq. EngngVol.26, pp. 347-361.- Gerolymos N., Gazetas G. (2006), “Development of Winkler model forstatic and dynamic response of caisson foundations with soil and inter-face nonlinearities” Soil Dyn. & Earthq. Engng Vol.26: 363-376.- Isakovic T., Fischinger M. (2005), “Higher modes in simplified inelas-tic seismic analysis of single-column bent viaducts”, StructuralEngineering International.- Kappos A.S., Paraskeva T.S., Sextos A.G. (2005), “Modal pushoveranalysis as a means for the seismic assessment of bridge structures”Proc. 4th European workshop on the Seismic behaviour of irregular andcomplex structures, Thessaloniki, Greece (Paper 49).- Lupoi A., Franchin P., Pinto P.E. (2007), “Further probing of the suit-ability of push-over analysis for the seismic assessment of bridge struc-tures” Proc. of COMPDYN’07, Crete, Greece- Makris N., Gazetas G. (1991), “Dynamic pile-soil-pile interaction.Part I: Analysis of Axial Vibration” Earthq. Eng. & Struct. Dyn. Vol.20,pp115-132.- Makris N., Gazetas G. (1992), “Dynamic pile-soil-pile interaction.Part II: lateral and seismic response” Earthq. Eng. & Struct. Dyn.Vol.21, pp 145-162.- Monti G., Pinto P.E. (1998), “Effects of multi-support excitation onisolated bridges” Tech. Rep. MCEER 98/0015 pp. 225-247.- Novak M. (1974), “Dynamic stiffness and damping of piles” CanadianGeotech. Jnl Vol.11: 574-598.

III - INNOVATIVE MATERIALS FOR THE VULNERABILITYMITIGATION OF EXISTING STRUCTURES

1. INTRODUCTION

The use of fiber reinforced polymer (FRP) materials for the strengtheningof masonry and concrete structures, represents a valid alternative to tra-ditional techniques. Indeed, many advantages are provided by usingFRPs: lightweight, good mechanical properties, corrosion-resistant, etc. In Italy, the use of FRP materials for reducing the seismic vulnerabilityof existing structures has been allowed for the first time through O.P.C.M.3274 and more recently by the D.M. 14.01.2008, that refer to the ItalianNational Research Council Design Guidelines (CNR-DT 200/2004) forthe external strengthening of existing structures with FRP materials.These guidelines provide, within the framework of the Italian regula-tions, a document for the design and construction of externally bondedFRP systems for the strengthening of existing structures. In particular,several issues concerning the seismic rehabilitation of ReinforcedConcrete (RC) and masonry buildings have already been dealt but a fur-ther investigation is still required.Within this context, the main aim of this research line has been theexperimental validation of design indications provided by the CNR-DT200/2004 guidelines. The main topics investigated in this research task can be summarizedas follows:- the mechanical behaviour of FRP materials;- the cyclic behaviour of RC elements strengthened by means of FRP;- the mechanical and chemical anchorage devices for FRP systems;- the ductility increasing of RC columns confined with FRP;- the RC joint strengthening with FRP;- the masonry strengthening with FRP;- the historical structure strengthening with FRP;- the quality control and monitoring of FRP applications;- the innovative fibers (steel fabrics, natural fibers, FRP grids, etc.) and

matrices (organic and inorganic);- the mechanical behaviour of concrete structures reinforced with fiber

reinforced polymer bars;- innovative strengthening techniques (near-surface mounted (NSM)

technique, FRP prestressed systems).


The research activity has been performed through experimental testsand theoretical studies mainly devoted to the development of simplemethods of analysis and design rules in order to improve the indications


interaction force between FRP materials and the masonry substrate,experimental tests are needed. In this research line, semi-destructiveand non-destructive techniques have been also investigated for thequality control and monitoring of FRP applications to masonry struc-tures, according to CNR DT 200/2004 Guidelines.


In order to guarantee an optimal organization of the research, theResearch Units have been grouped into the following ten Tasks, eachone with a specific topic:- Task 8.1: the mechanical characterization of FRP systems at fixed

environmental conditions under cyclic actions;- Task 8.2: the delamination under cyclic actions and design of anchor-

age mechanical devices for FRP systems;- Task 8.3: the confinement of RC and masonry columns subject to com-

bined flexure;- Task 8.4: the strengthening in flexure and in shear of RC structural

elements with FRP fabrics and near surface mounted (NSM) rods;- Task 8.5: the beam-column and beam-foundation joint reinforcement

with FRP;- Task 8.6: the design criteria for the seismic retrofit of RC and RC-

masonry composite structures with FRP;- Task 8.7: the design criteria for the seismic retrofit of masonry struc-

tures with FRP;- Task 8.8: the strengthening of masonry structural elements with FRP

systems;- Task 8.9: the strengthening of masonry vaulted elements with FRP

systems;- Task 8.10: the quality control and monitoring of FRP applications to

existing masonry and RC structures.

4. MAIN RESULTS

4.1 Task 8.1: the mechanical characterization of FRP systems at fixedenvironmental conditions under cyclic actionsThe aim of the sub-task was to mechanically characterize FRP systems.The study has been focused on durability and mechanical behaviour ofstructural adhesives and FRPs, the mechanical characterization of FRPbars and strips, the values of the safety factors proposed in CNR DT200/2004 and the effects of elevated temperatures and freeze-thawcycling on FRP.

Durability and mechanical behaviour of structural adhesives and FRPs

Several tests to determine the mechanical properties of composite mate-

provided by CNR-DT200-2004. FRP are ideal products for structural retrofitting and seismic upgrading.Nonetheless the small knowledge on the durability of the system is oneof the main drawbacks to the use of FRP reinforcement in CivilEngineering. In particular, structural adhesives usually represent theweakest point of the reinforced system and their mechanical behaviourand durability performance need to be investigated. The first problemin using composite materials for structural reinforcement is the deter-mination of their mechanical properties. The bond between FRP andconcrete is a very important issue because the debonding is a very brit-tle failure mechanism and must be avoided. According to performance-based design or seismic evaluation of RCbuildings, it is crucial to provide a correct evaluation of the strength andductility capacity of the RC columns and beams as well as of beam-col-umn joint. Experimental and direct observation of damages occurredduring recent earthquakes strongly highlighted this need. The effectiveness of FRP systems for seismic vulnerability mitigation ofmasonry structures is still in debate, despite it has moved a huge inter-est, becoming the outstanding system in the market for this type ofapplications.Indeed CNR DT 200/2004 has been the first guideline to providedesign criteria for the FRP seismic strengthening of masonry buildings.However, the retrofit design of masonry structures is still not a com-pletely solved problem. This is due to the fact that the masonry struc-ture is load dependent and thus the FRP could be placed in an inactivearea of the resistant mechanism. Furthermore, masonry can activate alarge number of local mechanisms which interact with global behaviourof masonry buildings. The non linear seismic assessment of FRP rein-forced masonry structures is included also in the D.M. 2008 rule. Thenon linear analysis requires the knowledge of the constitutive law of themasonry material both in the unreinforced or strengthened situations. In the recent years, the scientific research has been focused on the safe-guard of historical buildings. Accordingly, CNR DT 200/2004 has beenpublished in order to provide design criteria for the use of FRP systemsfor strengthening existing structures and to avoid their incorrect appli-cation. The Guidelines deal with different types of FRP applications tomasonry and reinforced concrete structures and take into account theimportant phases of quality control and monitoring that should follow astrengthening application. Several aspects affect the effectiveness ofFRP systems such as the surface preparation and FRP installation.Moreover, once FRP strengthening intervention has been carried out,monitoring by non-destructive or semi-destructive tests should be per-formed to ensure the quality and effectiveness of the strengthening sys-tem. It is worth noting that due to the increased number of compositematerial applications and in order to get a better understanding of the


rials and structural adhesives have been performed. Conforming to theASTM requirements, the glass transition temperature (ASTM D3418),porosimetry and the coefficient of thermal expansion (ASTM D360)were determined. Adhesives were also tested under tensile (ASTMD360), compressive (ASTM D695) and flexural loading (ASTM D790).Adhesive shear strength was determined by punch tool tests (ASTMD732). Finally adhesive cylinder specimens were tested under pure tor-sion load.

Adhesive dumb-bell specimens were prepared for tensile testing andthen artificially aged in an environmental chamber in order to analyzepossible detrimental effects on the adhesive mechanical properties.Exposition to deicing salts, freeze-thaw cycles and moisture may in factdeteriorate the mechanical properties with consequences on the dura-bility performances of strengthened structures. Tensile tests were per-formed conforming to the requirements of ASTM D360. In all the con-ditioning treatments, significant losses in adhesive stiffness and tensilestrength were measured. The stiffness and tensile strength reductionsafter exposure to salt spray fog solution may be approximated bystraight parallel lines as described in the Arrhenius life-temperaturerelationship. Fatigue tests on adhesive dumb-bell specimens werefinally performed to attain the fatigue failure curves for the adhesivejoint.

Mechanical characterization of FRP bars and strips

Tensile and relaxation tests were performed on FRP bars with particu-lar attention to the gripping system. Then, experimental tests andnumerical simulations were performed to develop simple, economicaland effective systems for the characterization of composite materialsand adhesives. In particular, an anchor system for tension testing of uni-directional fiber reinforced plastic (FRP) bars of large diameter wasdeveloped. In the system suggested each end of the bar is embedded ina conical polymeric head that fits a conical hole inside the anchoringdevice. In the anchor system, the anchor body shape came from expe-

riences for testing steel ropes and prestressing steel tendons and theshape of the resin head from test investigation. Numerical analyseswere also performed to investigate the effects of anchor parameters suchas cone slope angle, thickness of resin head and friction coefficientbetween the anchor body and the resin head. Pull-out and beam testswere also executed.

Experimental studies and numerical analyses were developed to definepractical tests for the characterization of FRPs and adhesives mechan-ical properties. The main aim of this action was to provide the“Composites Kit Test - COKIT”; a practical tool for professionals andengineers operating in the field of FRPs applications and dealing withFRP materials for structural retrofitting and rehabilitation. The techni-cal document “Istruzioni per la caratterizzazione ed il controllo diaccettazione di materiali fibrorinforzati per il rinforzo strutturale –COKIT ” was thus published and could represent an annex of the CNRDT 200/2004 Recommendations.

Refinement of the safety factors proposed in Design Recommendations

The environmental conversion factors provided in the guidelines of theItalian National Research Council (DT200) were analyzed on the basisof the results of artificially aged adhesive specimens tested under ten-sion. Exposition to deicing salts, freeze-thaw cycles and moisture leads


Fig. 1- a) Execution of the punch-tool test and specimen after collapse, b) execution of the torsion test and specimen aftercollapse.

Fig. 2- a) Anchor system for large diameter GFRP bars, b) numerical analysis.

Fig. 3- Stiffness and tensile strength retention for the structural adhesive subject to freeze-thaw cycles of five hours eachbetween –18° and +4°C for a total duration of about 2 months (FT), to salt spray fog for one month or three months (SF)and to one month humidity (HU).

Effects of elevated temperatures and freeze-thaw cycling on FRP lami-nates behavior

The performances at elevated temperatures and/or at freeze-thawcycling exposure of structural members strengthened by using exter-nally bonded FRP laminates are mainly related to two aspects: the bondbehaviour between FRP and the member substrate; the mechanicalproperties of laminates themselves. The latter aspect has been very lim-ited experimentally investigated; only few tests have been performed toevaluate the residual tension strength of FRP coupons after exposure toelevated temperatures or freeze-thaw cycling. Thus, experimental ten-sion tests on carbon FRP (CFRP) laminates both under controlled tem-perature and relative humidity conditions or after freeze-thaw cyclesexposure have been carried out. In particular, due to reduced capacitythat commercially available resins have to transfer loads over fibresaround glass transition temperature, Tg, two new systems based onepoxy resin have been formulated and characterized by dynamicmechanical analysis (DMA). The main goal of the new formulated sys-tems was to increase Tg, the elastic modulus in the rubber region of theresin and to improve their performances under freeze-thaw cycles. Twodifferent approaches were investigated. First a new epoxy system(namely neat epoxy) was formulated and cured at 60°C after an hour atroom temperature. Secondly, in order to improve the mechanical prop-erties of epoxy matrix by curing at room temperature, a nanocompositesystem was obtained by direct dispersion of preformed nanodimen-sioned silica particles to the neat epoxy resin.

to the deterioration of the mechanical properties of composite materialsand in particular structural adhesives. On the basis of the experimentalresults, the safety factors suggested in the CNR DT 200/2004 recom-mendations may be considered as appropriate, but in aggressive envi-ronments the use of a slightly lower conversion factor seems to be moresuitable.

Tests were performed to refine the safety factor of FRP-steel systems:the fatigue behaviour of steel structures retrofitted by using FRP mate-rials was investigated, S–N curves were defined and the fatigue resis-tance of the steel-CFRP bond was compared to the one of welded detailcategories described in the Eurocode 3.

After performing pull-pull delamination tests on FRP-concrete speci-mens, cylinders were obtained from each concrete prism. Based onEurocode 2 compressive and splitting tests were carried out to deter-mine the conditioning effects on concrete degradation. As a conse-quence of the environmental conditioning, concrete characteristicstrength is assumed to increase by 16% for salt spray fog conditionedspecimens and to decrease by 3% for specimens subject to freeze-thawcycles.


Fig. 4- a) Steel-CFRP specimen b) Reduction in stiffness of retrofitted specimens during fatigue tests; c) S-N curve andcomparison between the fatigue resistance of the steel-CFRP bond for a stiffness reduction of 5% (blue circles) and of 15%(red squares) and that of EC3 welded detail categories.

Fig. 6- Specimen geometry; Temperature and relative humidity exposure profiles; test setup.

Fig. 5- Statistical distribution of the coefficient kG for specimens subject to a) salt spray fog and b) freeze-thaw cycles.

The experimental results point out that the developed formulations ofepoxy resins provide a significant increase of ultimate strength andstrain of CFRP coupons both at room and elevated temperatures withrespect to commercial systems, without significant change of the elasticmodulus. Negligible influence of a low number of freeze-thaw cycleswas observed on the mechanical properties of coupons independently ofmatrices. Experimental outcomes strongly confirmed that the use ofmatrices characterized by higher values of Tg and elastic modulus in therubber region with respect to those traditionally available on the mar-

ket, could allow to overcome one of the main limit of FRP laminatesrelated to their poor performances under elevated temperatures.

4.2 Task 8.2: the delamination under cyclic actions and design ofanchorage mechanical devices for FRP systemsDifferent experimental set-ups can be found in the scientific literaturedealing with FRP-concrete bond tests and it has been observed that dif-ferent test methodologies may give different values of the debondingforce. This task research intended to define a standard FRP-concretebond test to be used to evaluate the maximum transmissible force by anFRP anchorage, to be included in the new version of the Italian code fordesign of strengthening interventions with FRP.

Experimental Round Robin test on FRP concrete bonding

An extensive experimental campaign on FRP-concrete debonding hasbeen carried out by five different Italian Laboratories (University ofBologna, University of Naples Federico II, University of Sannio,Polytechnic of Milan and University of Calabria). The tests were devot-ed to the definition of a standard test procedure for the bond strengthevaluation. According to the Round Robin procedure, 50 concreteprisms (same batch) strengthened with CFRP plates and sheets havebeen prepared by the same operator and subject to bonds test in fivedifferent Laboratories. The sets of homogeneous specimens have thenbeen subject to bond test by five laboratories of the University partnersusing different test set-ups (Figure 7).

Twelve specimens (6 strengthened with sheets, 6 strengthened withplates), with two different bonded lengths (100 mm and 400 mm), havebeen tested by each laboratory, repeating three times the same type oftest. As for the test set-ups (Figure 7), all the Laboratories adopted a

single shear push-pull test. All the tests have been performed underdisplacement control of the FRP free end. In order to evaluate the vari-ability of the results when different set-ups are adopted, the coefficientof variation (COV) for each set of homogeneous experimental tests hasbeen calculated. The scatter of the results is in general small (COVabout 10%), lower than that of the tension strength of the concrete, usu-ally equal to 20-30%. For the plates, the scatter of the results is simi-lar for the different Labs, whilst for the sheets the dispersion is usuallyhigher. The results obtained by Lab 3 are very stable in both cases andclose to the mean values. This study allowed to define a set of rules forthe standardization of bond tests to be used to evaluate the maximumtransmissible force by an FRP – concrete anchorage.

Cyclic tests of FRP-concrete debonding under cyclic loadings

Many strengthened structures are subjected to fatigue loads (i.e. roadsand railways bridges) or to shorter but more intense cyclic actions asseism: in this cases, the FRP-concrete interface is subject to cyclicstress regimes which can lead to premature debonding of the FRP lam-inate from the concrete substrate and cause the FRP failure in mostcases, unless appropriate local measures are taken to prevent it. In order to develop a more economical design for FRP-strengthenedstructures, research line investigated the debonding phenomenon of theFRP reinforcements (both plates and sheets) from the concrete sub-strate under cyclic actions. In particular, tests on little prismatic speci-mens have been performed by applying FRP plates and sheets on con-crete prisms and testing them under both monotonic and cyclic actionswithout inversion of sign.The experimental results of Single Shear Test (SST) performed on CFRPreinforcement applied on little prismatic concrete specimens and char-acterized by high bond length values (400mm) showed that:- the influence of load-unload cycles up to 70% of Pmax,M was negligi-

ble for CFRP sheets and plates; - a low number of load-unload cycles (40) up to 90% of Pmax,M reduced

the debonding load of about 10% in the case of CFRP plates but didnot affect particularly the bonding behaviour of CFRP sheets;

- by increasing the number of load-unload cycles (up to 300) between70% and 90% of Pmax,M, the debonding load of concrete specimensreinforced with CFRP sheets decreased by a percentage factor equalto 10%;

- the transfer of shear stresses at the FRP-to-concrete interface due tothe CFRP reinforcement bond length larger than effective one,allowed to mitigate noticeably the effect of cyclic actions imposed upto 90% of Pmax,M;

- a degradation of interface behaviour has been recorded after the onset


Fig. 7- Experimental set-ups adopted by the five different Laboratories.

LAB4

ratio; applying different FRP reinforcement ratios and checking theconfinement sensitivity to the number of plies.

Tested specimens represent real scale building columns designedaccording to dated codes for gravity loads only. The design concretestrength is 23.1 MPa to simulate concrete mixes used in past decades.Concrete cylinder specimens per each casting have been prepared inorder to characterize the concrete with standard procedures. The usedsteel is characterized by a yield strength of 414 MPa and a modulus ofelasticity of 200 GPa. Particular care has been devoted to constructiondetails, namely: hooks, longitudinal and transverse steel reinforcementratios and concrete cover specifications. Special care has been taken toavoid local failure at the top and the bottom ends of the columns plac-ing steel ties with reduced spacing. Internal steel (bars and ties) rein-forcement ratio is the same for each group of specimens, designed perminimum code requirements. The minimum specimen dimensions are360 x 510 mm2. The load has been applied concentrically under a dis-placement control rate. The load has been conducted in five cycles inincrements of one fifth of the expected capacity for each specimen.Each loading-unloading cycle has been repeated once. Strain gages,potentiometers and LVDTs are used for strain and displacement dataacquisition. In particular, strain gages have been applied on columnsurface and internal bars whereas LVDTs have been placed in order toobtain vertical and horizontal column displacement. Strain data acqui-sition have been obtained by strain gages applied on FRP sheets too.

of debonding, with reduction of maximum shear stress;- the effects of cyclic actions were more significant on plates rather than

sheets and its influence increased with number of cycles;Moreover similar SST tests performed on CFRP reinforcements charac-terized by lower bond length values (50-250mm) allowed to observethat:- design relationships provided by Teng et Al. and by main internation-

al codes for evaluating the effective bond length values are conserva-tive for both sheet and plate reinforcements;

- referring to plates the effective bond length values, experimentallyevaluated by means of the monotonic tests, were noticeably lower thanpredicted;

- even if reinforcement bond length values were significantly low, thecyclic tests outcomes confirmed that the influence of load-unloadcycles up to 70% of Pmax,M was negligible for CFRP sheets and platesdue to elastic behaviour characterizing FRP-concrete interface up tosuch load level;

- Also a further low number of load-unload cycles (10) up to 90% ofPmax,M did not affect particularly the bonding capacity of CFRP rein-forcement due to the transfer of shear stresses at the FRP-to-concreteinterface. Such transfer was more significant on plates rather thansheets: nevertheless, bond lengths particularly conservative for platesallowed to better mitigate cyclic action effects;

- in order to better predict design bond length values, using two differ-ent relationships for sheets and plates, respectively, could be worth-while.

4.3 Task 8.3: the confinement of RC and masonry columns subject tocombined flexureThe main goal of this research task has been to validate the designequations provided by CNR DT200 for the confinement of RC andmasonry members. In particular, experimental tests have been carriedout on both real scale and scaled columns wrapped by using tradition-al FRP (CFRP and GFRP) or an innovative typology of FRP systemmade of basalt material fiber.

Confinement of real scale RC columns subject to axial load

An experimental campaign has been carried out on full scale reinforcedconcrete (RC) columns concentrically loaded and confined by means ofFRP (Glass FRP and Basalt FRP). Five series of tests were planned, foreach series a reference unconfined column was tested and used asbenchmark. The test matrix has been designed to assess the confine-ment effectiveness: applying the same reinforcement ratio and check-ing the effect of the shape, the side aspect ratio and the area aspect


Fig. 8- Debonding force for sheets with anchorage length (a) L=400 mm, (b) L=100 mm, and for plates with anchoragelength (c) L=400 mm, (d) L=100 mm.

Fig. 9- Coefficient of variation (COV) of the debonding force for (a) sheets and (b) plates.

The data have been elaborated in order to investigate on volumetricstrain and Poisson ratio as a function of load level. The main results ofthe experimental campaign can be summarized as follows: a significantincreasing in the axial displacement and a little increase in the ultimateload of the FRP strengthened columns compared to their benchmark.The hollow columns have shown a failure mode characterized bybulging, followed in such case by the rupture of the fibers. A more evi-dent failure mode has been shown by the remaining columns for whichthe fiber rupture has always accompanied the concrete spalling.

Confinement of RC cylindrical specimens strengthened by means ofbasalt fibers and inorganic matrix

The effectiveness of such system as a confinement technique has beenanalyzed by means of an experimental campaign on concrete cylindri-cal specimens. The effectiveness of the proposed technique is assessedby comparing different confinement schemes: 1) uniaxial Glass FibreReinforced Polymer laminates; 2) alkali-resistant fibreglass grid bond-ed with a cement based mortar; 3) bidirectional basalt laminates pre-


Fig. 10- Test Matrix.

Fig. 11- Columns strengthened by means of glass (a) and basalt (b) FRP.

Fig. 12- Load versus vertical and horizontal strains.

(a) Column S-1-5GA (b) Column R-1-8H

schemes were experimentally analyzed in order to evaluate and com-pare the effectiveness of the proposed strengthening techniques: 1) uni-axial glass FRP laminates (GFRP) wrapping; 2) uniaxial carbon FRP(CFRP) laminates wrapping; and 3) uniaxial basalt FRP (BFRP) lami-nates wrapping. In particular 9 tests, were performed on square tuffmasonry (external tuff blocks and inner core filled with tuff chips andmortar) scaled columns (mass density equal to about 1530 kg/m3): sideaverage dimension equal to 220mm; and average height of about 500mm corresponding to 8 courses of tuff bricks (height-width ratio equalto 2.27). Masonry was made by scaled yellow Neapolitan tuff bricks(50x50x100mm) and a pozzolan (local volcanic ash) based mortar(thickness of 12mm). Further 9 tests were performed on square claybrick masonry scaled columns (mass density equal to about 1700kg/m3): side average dimension equal to 260 mm, and average height ofabout 560 mm corresponding to 8 courses of clay bricks (height-widthratio of 2.20). Masonry was made by clay bricks (55x115x255 mm) anda pozzolan (local volcanic ash) based mortar (thickness of 13 mm).

impregnated with epoxy resin or latex and then bonded with a cementbased mortar; 4) cement based mortar jacket. The main objectives of theexperimental program were: a) to investigate on the effectiveness ofconfinement based on basalt fibres pre-impregnated in epoxy resin orlatex and then bonded with a cement based mortar (BRM); and b) tocompare the performance (in terms of peak strength and ultimate axialstrain gains) of different confinement techniques using advanced mate-rials with respect to GFRP laminates jacketing. The investigation was carried out on 23 concrete cylindrical specimenswith a diameter of D = 150 mm and a height of H = 300 mm.


Fig. 13- Typical failure modes.

Fig. 15- Failure modes.

Fig. 16- Specimen details (dimensions in mm): (a) tuff masonry; (b) clay brick masonry.

Fig. 14- BRM wrapping installation procedure.

Experimental outcomes showed that:• BRM confining system could provide a substantial gain both in com-pressive strength and ductility of concrete members inducing a failuremode less brittle than that achieved in the GFRP wrapped members;• lower performance were observed by concrete confinement providedby a primed glass fiber grid bonded with cement based mortar withrespect to BRM and almost no influence was generated by the jacket-ing with mortar only.• maximum ultimate axial strain increases were provided by GFRPlaminates wrapping.

Confinement of rectangular masonry columns subject to axial load

An experimental campaign dealing with 18 square cross-section bothlisted faced tuff and clay brick masonry scaled columns subjected touniaxial compression load. In particular, three different confinement

Masonry columns were tested through monotonically applied axial com-pressive loading under displacements control mode with a rate of 0.005mm/s.

The experimental outcomes showed that:• GFRP and CFRP jackets led to similar compressive strength gainson tuff masonry columns under axial loads. • GFRP and BFRP confining system led to similar compressivestrength gains of brick masonry columns under axial loads. BFRP wrap-ping was more effective in terms of global ductility increase (i.e. ulti-

mate strain gain equal to 413% and 259% for BFRP and GFRP wrap-ping, respectively) even if the mechanical external reinforcement ratioof FRP laminates was lower then GFRP ones; such result could beexplained by the higher values of ratios efl/efu recorded on BFRP lam-inates.• The use of high values of laminates unit height may significant reducethe effectiveness of FRP wrapping systems since it could be detrimen-tal to the quality of confinement execution.• The presence of voids and protrusions on masonry members reducesthe ultimate transverse strain on FRP reinforcement with respect to thattypically achieved on concrete members.

Another experimental campaign has been carried out in order to showthe behavior of columns built with clay or with calcareous blocks, com-monly found in southern Italy, especially in historical buildings.Rectangular masonry columns were tested for a total of 33 specimens;uniaxial compression tests were conducted on columns taking intoaccount the influence of several variables: different strengtheningschemes (internal and/or external confinement), curvature radius of thecorners, amount of fiber-reinforced polymer (FRP) reinforcement,cross-section aspect ratio and material of masonry blocks. Materialscharacterization was preliminarily carried out including a mechanicaltest on plain masonry. For all cases the experimental results evidenced

a significant increase in load carrying capacity and ductility after FRPstrengthening, which identified the columns as ductile elements despitethe brittle nature of the unconfined masonry. Differences in mechanicalbehavior, due to the geometry of the columns, to the nature of differentmaterials, to different strengthening schemes, and to the amount of rein-forcement, have been taken into account. The calibration of designequations recently developed by Italian National Research Council,CNR was conducted to compare analytical prediction and experimentalresults.


Fig. 17- Stress-axial strain relationships and specimens’ failure mode: (a),(c) and (e) tuff masonry; (b),(d) and (f) clay brickmasonry.

Fig. 18- Limestone and clay brick masonry specimens.

The results obtained from the experimental campaign confirmed thatinnovative strengthening techniques, using FRP sheets and bars, areeffective when confinement of masonry compressed elements is need-ed. Two types of masonry were investigated: the first made with claybricks, the second made with limestone blocks. Even if the propertiesof the constituent materials were different, in both cases a significantincrease was measured in terms of peak load and ultimate axial defor-mation. Two construction schemes were considered: full core and hol-low-core columns; the last type reproduces the patterns often found inhistorical buildings. External and internal FRP confinement were test-ed, separately and combined. The proposed techniques are stronglyrecommended when a seismic retrofit is needed, since the external con-finement introduces a plastic behavior of the compressed masonrywhich indicates a large capacity in storing elastic energy which is takenby the fibers placed in the transverse direction. The presence of inter-

nal bars used as an internal confinement system is recommended inaddition to external FRP layers if ductility constitutes a main issue,since in columns strengthened only with bars the ultimate load wasincreased but brittle behavior of unconfined masonry remained.Columns with hollow core also showed a significant increase ofmechanical properties when confinement was applied, especially in thecases of GFRP external sheets combined with internal bars.

Confinement of circular masonry columns subject to axial load

An extended experimental investigation has been performed in order toshow the mechanical behavior of circular masonry columns built withcalcareous blocks that may be commonly found in Italy and all overEurope in historical buildings. Different stacking schemes were used tobuild the columns, aiming to simulate the most common situations inexisting masonry structures. Carbon FRP sheets were applied as exter-nal reinforcement; different amounts and different schemes of confiningreinforcement were studied. The experimental program included a newreinforcement technique made by using injected FRP bars through thecolumns cross section. The structural behavior of masonry columnsdamaged under different levels of load and strengthened by using FRPreinforcements has been also investigated.


• Displacement capacity resulted increased in all cases; strengthenedcolumns tested showed an extended postpeak plastic branch in the loadversus displacement curves;• Columns confined with three 100 mm wide sheets showed highermechanical properties with respect to the same columns confined withtwo 150 mm wide sheets;• Damage caused by overloads applied in the precracking stage beforestrengthening did not reduce the mechanical properties of FRP-con-fined columns;• Presence of internal FRP rebars acted as an effective confining sys-tem for cross sections composed by four blocks;• Application of design equations by Italian CNR furnished conserva-tive results for complete FRP wrapping, whereas prediction of strengthfor masonry confined with CFRP strips showed a reduced scatter withrespect to experimental results.

4.4 Task 8.4: the strengthening in flexure and in shear of RC structuralelements with FRP fabrics and near surface mounted (NSM) rods.A new technique for the shear and flexural strengthening of RC struc-tural elements has been investigated in this research task. In particular,the use of Near Surface Mounted rods (NSM) for structural upgradinghas been deeply analysed by both analytical and experimental investi-gations. Further, the effectiveness of FRP laminates, traditionally used tostrengthen RC or masonry members, has been investigated with refer-ence to full scale prestressed concrete (PC) girders.

NSM bars shear contribution: a calculation procedure

A calculation procedure suitable for practitioners has been developedby simplifying a more sophisticated predictive model recently devel-oped (Bianco 2008; Bianco et al. 2009a-b). That procedure briefly con-sists of: a) evaluating the average structural system composed of the

Fig. 19- Geometry and dimensions (mm) of columns.

Fig. 20- Specimens after failure.

Important remarks follow:• High increase in ultimate strength and strain were evident afterstrengthening;• Complete FRP jacketing was much more effective than discontinuouswraps;

average-available-bond-length NSM strip confined to the correspond-ing concrete prism whose transversal dimensions are limited by thespacing between adjacent strips and the beam cross section width(Figure 20); b) determining the comprehensive constitutive law of theaverage system above (Figure 21); c) determining the maximum effec-tive capacity that the average system can attain during the loadingprocess of the strengthened RC beam by imposing a kinematic mecha-nism and d) determining the NSM shear strength contribution by sum-ming the contribution provided by each strip. The constitutive law(Figure 22) and in turn the equations to determine the maximum effec-tive capacity assume different features depending on the main phe-nomenon characterizing the ultimate behaviour of the average structur-al system of the specific case at hand. Hereinafter, for the sake of brevi-ty, the main features of that computational procedure are shown only forthe case of shallow concrete fracture (u = 4) and a resulting resistingbond length whose value is equal to the effective bond length (Figure23). Further details can be found elsewhere (Bianco 2008).The predictions obtained by that calculation procedure were alsoappraised on the basis of experimental results (e.g. Dias et al. 2007).

The maximum effective capacity for the case of shallow concrete frac-ture and a resulting resisting bond length whose value is equal (u = 4)to the effective bond length can be evaluated by:

(1)

where:

(2)

(3)

Actual Vf and design value Vfd of the NSM shear strength contributioncan be obtained as follows:

(4)

where �Rd is the partial safety factor divisor of the capacity that can beassumed equal to 1.1-1.2 according to the indeterminateness of theinput parameters.

Bond between NSM bars and surrounding concrete: experimental andanalytical investigation

Pull-out test were carried out to investigate both the qualitative andquantitative influence of some of the involved parameters on the bondperformance (De Lorenzis and Galati 2006, Galati and De Lorenzis2006). Those parameters encompass: ratio between depth and width ofthe slit, kind of epoxy-based adhesive used as binding agent, distanceof the NSM bar from the edge of the concrete prism, distance betweenadjacent bars and employment of external FRP strips used to confinethe joint. Tests were carried out by means of a tangential-pull device toapply the load, LVDT transducers to measure the slip at both the loadedand unloaded extremity and strain-gauges throughout the adheredlength of the bar to measure the deformations along the joint. The mea-sured quantities were processed to obtain the local bond stress-sliprelationship for the different values of the test parameters. Cyclic testswere also carried out subjecting the joint at a limited number of cycleswhose maximum load was assumed equal to different percentages of thepeak static load. The cyclic tests were useful to evaluate the joint resid-ual strength such as the one following a seismic action.An analytical investigation has followed the experimental programabove (Rizzo and De Lorenzis 2007-2009b). In fact, the local bond

maxl1�Rd

1�Rd

Vfd= �Vf= �(2�Nf,int�Vfi,eff �sin)

2��1

Ld�sin(q+)�f,max=�1=

Vfi,eff = � A1�C1 Ld ��f,max+ � arcsin(1–A3��f,max�Ld)+

�

max sfsf

21Ld

{}

[A2�C2

2�A3��f,max

+(1–A3��f,max�Ld)� 1–(1–A3��f,max�Ld)2 – ]�2

sf sf

Lp�J3�l3�sin(q+)4��0�J1

�0�J1

l2

�0�J1

l2

l2�sin(q+)2��0�J1

A1= ; A2=Lp�J3�l; A3= ;

C1=�1– ; C2 =


Fig. 21- Main features of the calculation procedure: a) average-length NSM strip and concrete prism of influence, b) adoptedlocal bond stress slip relationship, c) NSM strip confined to the corresponding concrete prism of influence and semi-pyramidalfracture surface, d) sections of the concrete prism.

Fig. 22- Possible comprehensive constitutive law of an NSM CFRP strip confined within a prism of concrete: (a) concretethat reaches the free extremity (u=1) or strip tensile rupture (u=2), (b) superficial and/or absent concrete fracture andultimate resisting bond length smaller (u=3), equal (u=4) or larger (u=5) than the effective bond length and (c) deep concretefracture (u=6).

Fig. 23- Maximum effective capacity along the CDC for the case u=4: a) comprehensive constitutive law; b) capacityVfi,CDC(�;�); and c) imposed end slip �Li,CDC(�;�); distribution along the CDC for different values of the CDC opening angle �and d) effective capacity as function of the angle �.

contribution to a RC beam.

Experimental investigation on full-scale prestressed concrete beamsstrengthened by means of CFRP

Every year, several prestressed concrete (PC) bridge girders are acci-dentally damaged by over-height vehicles or construction equipmentimpact. Although complete replacement is sometimes deemed neces-sary, repair and rehabilitation can be far more economical, especiallywhen the time and the social cost of the method are drastically reduced. The numerous advantages provided by the use of FRP laminates areleading in a sharp increase on their use for bridge constructionstrengthening. Experimental investigations were conducted in order tovalidate such strengthening technique on PC damaged members andaccurately assess the upper limit of damage amount beyond which FRPlaminates are no longer adoptable as repair solution.Starting from such purposes, an experimental campaign was conductedon five full-scale (13.0m long, 1.05m high) PC double T-beams with areinforced concrete slab, designed according to ANAS (ItalianTransportation Institute) standard specifications. One beam was used ascontrol, and the other four were intentionally damaged in order to sim-ulate a vehicle impact by removing the concrete cover and by cutting adifferent percentage of tendons (17% on two specimens and 33% on theremaining two). The repair, by using externally bonded carbon FRP(CFRP) laminates, aimed at restoring the ultimate flexural capacity ofthe member, taking particular attention to the laminates anchoring sys-tem. In particular, one test was performed on the control beam (refer-enced as S1), two tests were carried out on intentionally pre-damaged,to simulate an over-height vehicle collision, beams (named S2 and S3,respectively) and the remaining two on pre-damaged specimensupgraded by using two and three plies of CFRP laminates anchored byusing U-wraps (named S4 and S5, respectively).In Figure 24 and Figure 25 the test setup and experimental load deflec-tion curves are reported.

The experimental study has shown that: 1) a loss of strands equal to17% and 33% caused a flexural capacity decrease equal to 20% and26%, respectively; 2) to restore the ultimate flexural capacity of theundamaged PC specimen by using CFRP laminates it is necessary toprevent fibers debonding; 3) U-wraps (width wf= 100mm spaced atpf=150mm) were able to significantly delay debonding but if damagedexisting concrete is patched by cementitious mortar, a perfect bond hasto be guaranteed during the cross section restoration to prevent local-ized debonding of longitudinal reinforcement and thus fully exploit thepotential effective FRP strain increase; 4) CFRP laminates increased

stress-slip relationship obtained in the pull-out tests has been modelledby suitable analytical functions whose unknowns were calibrated for thedifferent values of the test parameters. The local bond stress-slip rela-tionship obtained by the cyclic tests was also modelled by analyticalfunctions. Then, the numerical solution of the governing differentialequation has allowed the peak pull-out load be determined as functionof the available bond length. The pull-out tests were also simulated by a FE model, both in the Linearand Non Linear range. The Linear FE model was adopted to evaluatethe bond-induced stresses on a plane transversal to the bar, evaluatingthe maximum stresses for different values of the geometrical andmechanical parameters of the joint and estimating so, local tangentialstress inducing the first-cracking in both resin and concrete. After that,a Non Linear model was developed by modelling: a) the several mate-rials according to the fracture mechanics and b) concrete/adhesive andadhesive/FRP interfaces by employing interface elements.

Experimental and analytical investigation on the shear strengtheningcontribution provided by NSM FRP bars on RC beams

Four points bending tests were carried out on RC beams strengthenedin shear by NSM FRP bars (De Lorenzis and Rizzo 2006, Rizzo and DeLorenzis 2006-2009a). Those beams were designed in such a way thatthe theoretical failure mode, for both the strengthened and un-strength-ened beams was due to shear-tension. Parameters investigated were:spacing, type and inclination of the NSM bars and the shear-span-to-depth ratio. Some beams strengthened by NSM strips were also testedin order to assess the relative effectiveness of the two techniques. Thesystem of FRPs was extensively equipped to measure the deformationsin the bars crossing the CDC. Tests have highlighted the possibility ofa global failure modes consisting in the detachment of the strengthenedcover from the underlying beam core. Such mechanisms had not beenpointed out by previous investigations.Two models were developed to predict the NSM shear strength contri-bution: a) one more simplified and b) a more sophisticated one. The for-mer was based on the Mörsch truss and the employment of a perfectlyplastic local bond stress-slip relationship. The latter takes into accounta more realistic local bond stress-slip relationship and the interactionbetween existing steel stirrups and NSM bars. The different local bondstress-slip relationships obtained in the former phase of the investiga-tion were employed to carry out some comparison. From those compar-isons it was possible to point out the great importance of the fractureenergy as opposed to the shape of the local bond stress-slip relation-ship. This phase of the investigation has led to the development of use-ful formulae for the evaluation of the NSM FRP shear strengthening


both stiffness and flexural moment capacity of PC damaged beams(maximum moment recover equal to about 12% and 20% for specimenswith 17% and 33% of strands loss, respectively; 5) the strengtheningintervention led to weak failure mode with a global ductility loss. The experimental outcomes qualify the application of FRP technique,already adopted in several cases of impacted PC bridges, as an effec-tive tool to restore the flexural capacity of PC girders; however the cal-ibration of theoretical expressions for the computation of the designFRP strain level considering the benefits provided by anchoring sys-tems is strictly necessary.

4.5 Task 8.5: the beam-column and beam-foundation joint reinforcementwith FRPAccording to performance-based design or seismic evaluation of RCbuildings, it is crucial to provide a reliable evaluation of the strengthand ductility capacity of the beam column joints: experimental anddirect observation of damages occurred during recent earthquakeshighlighted this.This task focused on some aspects: namely, cracking of the joint panel,longitudinal reinforcement bars slipping are deformability sources andthey could alter the capacity and interaction of beam and column mem-bers and the joint itself. F.E.M. modeling was adopted as an assessmenttool. The finite element code TNO DIANA 9.1 was adopted to simulateand to analyze numerically some real beam column joint sub assem-blages, characterized by nonlinear mechanical properties and geomet-rical detailing, smooth bars, structural deficiencies, as commonly found

in existing buildings. Such deficiencies were analyzed by means ofparametric analyses to evaluate the possibility to apply externalstrengthening on members characterized by poor concrete quality, lowtransverse reinforcement ratios, inadequate confinement due to lackingstirrups (especially in external joints), low bond performance of smoothand ribbed longitudinal reinforcement in columns and beams.Numerical analyses evidenced typical failure modes, crack patterns,influence of mechanical and geometrical properties on the behavior ofjoints. Some numerical/experimental comparisons were made based onsignificative tests available in scientific literature (for instance per-formed by Prof. Shiohara working group) or tested, during the RELUISProject, by UNIBAS R.U., allowing the numerical F.E.M. model to bevalidated. The behavior of the joints controls the global seismic behav-ior of an entire structure and a building in particular, so that its assess-ment is a crucial task in the strengthening design. Based on such analy-ses, the validity of analytical models for unstrengthened joints, avail-able in scientific literature, was checked. Such check was supported bylocal information provided by the detailed, refined, F.E.M. analyses.This numerical tool was a primary tool to understand the damage evo-lution and to assess the reliability of the main assumptions, equationsand procedures according to the “Quadruple flexural resistance in rein-forced concrete beam-column joints” (Shiohara, 2001) proposed byProf. Shiohara working group. To provide a direct, practical tool, ori-ented to the profession more than a nonlinear refined F.E.M. analysis,it was evaluated the opportunity to extend such consolidated model tothe case of externally bonded FRP strengthening of beam column joints.This model is based on the solution of a system of equilibrium equationsreferred to the four rigid bodies in which the joint can be ideally divid-ed, sometimes neglecting compatibility. This model is able to accountfor different failure modes and for the contribution of externally bond-ed FRP strengthening. The main assumptions can be recalled:• Diagonal cracks in the joint form an angle of about 45°• Normal concrete stresses acting on the main cracks can be reducedto an equivalent force• Longitudinal reinforcement provides only axial forces, so that anydowel action is neglected• There is a global symmetry both on the horizontal and vertical planeThe joint shear can be evaluated according to the sketch in Figure 26.

Vf= T + C's + C'c – Vc (1)Vf= T + T' – Vc (2)

where T and T'are the tractions in the longitudinal bars at the joint sec-tion, respectively; C'c is the compression force in concrete, while C's isthe compression force in the bars. Vj is the joint shear; and Vc is the col-


Fig. 24-- Test setup.

Fig. 25- Experimental load deflection curves.

umn length; and lb is the beam length.

Horizontal and vertical equilibriums of forces follow:

�Fx=0, F1–F2+C1sinq–C2sinq+Vc=0 (6)�Fy=0, –F3–F4+C1sinq+C2sinq–Nc–F7=0 (7)

Where Nc is the vertical force in the column; F7 is the traction in theFRP reinforcement in the y direction and reduced to an equivalentforce.To evaluate the column shear capacity of an unstrengthened joint, theF1, F2 and F5 forces can be assumed equal to the yielding forces of thecorresponding steel reinforcement while F6 and F7 are equal to zero.The five unknowns are Vc , F3, F4, C1 and C2 and can be evaluated solv-ing the system (3)-(7).To evaluate the column shear capacity of a strengthened joint, the F6

and F7 forces can be equal to the strength capacity of the external FRPreinforcement both in the x and y direction, respectively. To account forboth the tensile failure of the strengthening or for a possible FRPdebonding, the FRP capacity is given by the minimum between tensilestrength and debonding force evaluated, for instance, according to CNRDT200.

4.6 Task 8.6: the design criteria for the seismic retrofit of RC and RC-masonry composite structures with FRP.The applications of Carbon FRP (CFRP) and Glass FRP (GFRP) mate-rials have grown during last years; at present, seismic applications havebecome comparable if not more frequent than those related to lacks dueto gravity loads. The Italian guidelines for FRP interventions (CNR-DT200, 2004), deal with the use of composite materials to seismicallyupgrade under-designed RC and masonry structures. From seismicstandpoint, FRP strengthening is regarded as a selective interventiontechnique. Based on the main deficiencies of the existing structure, thedriving principles of the intervention are based on two main strategies:

umn shear. In Figure 27 the joint division is shown: there are four rigidand interacting bodies. Each body can be associated to three equilibri-um equations. The symmetry of the joints allows the system to bereduced to six equations. Moreover, Figure 28 shows the relation bet -ween the column shear, Vc, and joint shear, Vb, based on the equationVb = m � Vc, where m = lc / lb.


Fig. 26- Details of the forces due to the elements converging in the joint panel. Fig. 28- The joint system and the external forces.

Fig. 27- Internal forces.

(a) Forces in reinforcement and FRP (b) Forces in concrete

The independent equilibrium equations are five. Horizontal and verti-cal forces equilibrium and moment equilibrium based on point o (themiddle of the joint panel) can be expressed as follows:

�Fx=0, –F1–F2–F3+C1sinq+C2sinq–Nb–F6=0 (3)�Fy=0, –F3–F4+C1sinq–C2sinq+mVc=0 (4)

(5)

Where Nb is the horizontal force in the beam; F1, F2 are tractions in thelongitudinal bars of the beams at the joint section; F3, F4 are tractionsin the longitudinal bars of the columns at the joint section; F5 is thetraction in the stirrups spread along the height of the joint and reducedto an equivalent force; F6 is the traction in the FRP reinforcement inthe x direction and reduced to an equivalent force; C1, C2 are compres-sions as shown in Figure 27b; jb, jc are the internal lever arms in thebeam and column respectively; q is the inclination angle of the maincracks and assumed equal to 45°; Vc is the column shear; lc is the col-

�Mo=0, mVc+ jb(F1–F2)+ jc(F3–F4)– C2 =0lb2

lj2

12

12

1) preventing potential brittle failure mechanisms (i.e. shear failure, lapsplice failure, buckling of longitudinal reinforcement in compression)and “soft story” collapse mechanism, and 2) increasing global defor-mation capacity of the structure, either by enhancing the ductility ofplastic hinges without their relocalization or establishing a correct hier-archy of strength by relocalizing the plastic hinges according to capac-ity design criteria. The retrofit strategy is obtained by combining theabove principles; the definition of the retrofit scheme depends on thegap between actual and target performance of the specific structure,costs, functional characteristics and importance of the structure. Experimental studies, aimed at validating the effectiveness of FRP toachieve the above goals, are reported in the following.

Seismic strengthening of an under-designed RC structure with FRP

The outlined seismic strengthening strategy effectiveness was experi-mentally investigated within the European research project SPEAR(Seismic PErformance Assessment and Rehabilitation). The structureunder examination was designed and built with the aim of creating astructural prototype featuring all the main problems normally affectingexisting structures: plan irregularity, dimensions of structural elementsand reinforcement designed by considering only gravity loads, smoothreinforcement bars, poor local detailing, insufficient confinement in thestructural elements and weak beam column joints (see Figure 29). Thestructure was subjected to pseudo-dynamic tests, both in its originalconfiguration and retrofitted by using GFRP.

tively. The design of the rehabilitation was based on deficiencies under-lined by both the test on the ‘as-built’ structure and the theoreticalresults provided by the post-test assessment. They indicate that a retro-fit intervention was necessary in order to increase the structural seismiccapacity; in particular, the theoretical results showed that the targetdesign PGA level of 0.30g could have been sustained by the structureif its displacement capacity was increased by a factor of 48%. In orderto pursue this objective, the retrofit design strategy focused on two mainaspects. First it was decided to increase the global deformation capac-ity of the structure and thus its dissipating global performance; suchobjective was pursued by confining column ends with two plies ofGFRP laminates. Moreover, the second design key aspect was to allowthe structure to fully exploit the increased deformation capacity byavoiding brittle collapse modes. To achieve this goal corner beam col-umn joint panels were strengthened by using two plies of quadri-axialGFRP laminates as well as a wall-type column for its entire length withtwo plies of the same quadri-axial GFRP laminates used for the abovejoints (see Figure 30).


Fig. 29- (a) Plan view and (b) 3D view of the SPEAR structure.

Fig. 30- Column confinement and shear strength of corner joints (a); shear strength of wall-type column and retrofittedstructure overview (b).

Fig. 31- Base shear – top displacement curves for ‘as-built’ and FRP retrofitted structure.The structure in its original configuration was subjected to experimen-tal tests with maximum peak ground acceleration (PGA) of 0.20 g. Sinceboth theoretical and experimental results showed that the ‘as-built’structure was unable to withstand a larger seismic action, a retrofitintervention by using FRP laminates was designed. Once the design ofthe GFRP retrofit was provided, the structure was subjected to a newseries of two tests with the same input accelerogram selected for the ‘asbuilt’ specimen but scaled to a PGA value of 0.20g and 0.30g, respec-

The assessment of structural global performance, before and after thestrengthening intervention, was performed by nonlinear static pushoveranalysis in longitudinal direction (positive and negative X-direction,PX and NX, respectively) and in transverse direction (positive and neg-ative Y-direction, PY and NY).

In Figure 31, the theoretical base shear-top displacement curves for the

has been investigated with reference to a PC bridge, named “TorrenteCasale” which is part of the Salerno-Reggio Calabria motorway. Thebridge, built in the seventies, has been recently enlarged (2001) inorder to satisfy the new traffic demands. Due to the recent issuing of anew seismic Italian code, it was decided to assess the bridge capacity,both for gravity and seismic loads, in relation with the new design pro-visions.The bridge existing documentation has been investigated and bothdestructive and non-destructive tests have been performed in order todetermine concrete and steel reinforcement mechanical properties.Once the bridge geometry and the material mechanical properties weredetermined, a theoretical analysis was performed showing that thebridge piers (circular cross-section, diameter D=1m, and total heightH=6m) were not adequate to sustain the seismic actions. Thus use ofSRPs spikes as columns’ flexural reinforcement combined with CFRPlaminates wrapping of columns ends has been investigated to increaseboth member strength and ductility; the structural upgrade was com-pleted by increasing the shear capacity of column cap through CFRPU-wraps (Figure 33). The effectiveness of such technique with respectto a traditional one, based on RC jacketing, has been assessed. Themain construction phases of the rehabilitation intervention are reportedin Figure 33.

4.7 Task 8.7: the design criteria for the seismic retrofit of masonry struc-tures with FRP.The main design problems concerning the seismic vulnerability mitiga-tion of masonry structures are the strengthening of masonry panels forin plane and out of plane actions, the improvement of FRP action bymeans of clamping and connection mechanical devices and proceduresfor the definition of the reinforcement layout in seismic analyses.In order to test the design methods proposed by CNR DT200/2004, sev-

‘as built’ and FRP retrofitted structure are depicted with reference todirection NX (where the maximum capacity-demand gap was recordedfor the ‘as-built’ structure at the significant damage limit state LSSD).Figure 31, clearly shows that the FRP retrofit is able to greatly increasethe global deformation capacity of the structure, slightly affecting itsstrength. The comparison between the seismic structural capacity andboth elastic and inelastic demand is reported in Figure 32 (for directionNX) by using the Capacity Spectrum Approach (CSA) (Fajfar, 2000).Figure 31 clearly shows that column confinement provides the structurewith significantly enhanced ductility, allowing it to achieve the theoret-ical inelastic demand by only modifying the plastic branch of the capac-ity curve.

After that columns and joints were wrapped with GFRP, the retrofittedstructure was able to withstand the higher (0.30g PGA) level of excita-tion without exhibiting significant damage. After tests, FRP wasremoved and it was shown that the RC core was neither cracked nordamaged. The comparison between the experimental results providedby the structure in the ‘as built’ and GFRP retrofitted configurationshighlighted the effectiveness of the FRP technique in improving globalperformance of under-designed RC structures in terms of ductility andenergy dissipation capacity.

Seismic rehabilitation of PC bridges by using FRP and SRP materials

An innovative strengthening technique based on the combined use ofFiber Reinforced Polymer (FRPs) and Steel Reinforced Polymer (SRPs)


Fig. 32- – Theoretical seismic performance comparison at 0.3g PGA between ‘as-built’ and FRP retrofitted structure.

Fig. 33- Details of the designed rehabilitation system.

eral seismic studies of complex monumental masonry buildings havebeen completed by carrying out detailed linear dynamic seismic analy-ses and FRP retrofit design. In the design several combinations ofstrengthening techniques were explored in order to point out the realapplicability of the cited guidelines.The obtained results cover the following areas:a) Verification of the feasibility of a FRP design through simple modifi-cations of the normal design activity;b) Definition of the feasibility bounds for this retrofit technique in caseof monumental buildings;c) Detail types and level of detailing required in real applications;d) Open FRP design problems not covered by the guidelines or lackingof necessary information;e) Comparison of different modelling techniques making use of plate,shell, beam, truss elements to represent the masonry structure;f) Comparison of different material constitutive assumptions for the nonlinear analysis of strengthened walls and buildings.The main results of the completed research are summarized in the fol-lowing:a) The FRP reinforcement net design is easily implemented by a sim-ple modifications of the normal design activity. The only needs are: inte-gration of the stress distributions in order to obtain the stress resultantsand preparation of a verification sheet including the design rules forbending and shear of a FRP reinforced masonry panel. Actually manycompanies have implemented DT 200/2004 rules in freeware softwarewhich can be helpful in this activity.b) The feasibility bounds for this retrofit technique checked for threemonumental buildings analysed, are very different if stuck or mechan-ically fastened reinforcement is used. In fact, only slight diffuseincrease of resistance is obtained by using externally bounded FRPnets. Owing to increase significantly the safety of the building, mechan-ical devices are mandatory. In this last case true building retrofit is pos-sible.c) The set of the detail types needed for practical applications is verylarge and many connection types are not yet fully investigated althoughpractically employed. Fibre ropes, bars, fasteners, metallic inserts aremixed in a way almost never covered by existing experiments andguidelines.d) Open FRP design problems include cement matrices, thermalcycling, delamination in compression, real bond of the regularizationprimer to rough masonry surfaces, behaviour of the mechanically fas-tened FRP elements after internal delamination. e) Several modelling techniques have been employed in order to carryout the seismic studies. Plate elements in linear dynamic analysis allowfor both in plane and out of plane evaluation, but non linear static

analysis with this type of discretisation is not a viable solution. Studieson discrete representation of masonry walls by using a refined trussstructure to represent the compressive load paths inside the masonrypanels showed that this type of discretisation is very simple, allows forcomplicated constitutive laws, allows easily equilibrium checks, andproduces very reliable load – displacement curves.f) The basic choice of associate Mohr – Coulomb or Drucker – Pragerelastic plastic constitutive laws can be effective only in limited cases,where dilatancy is not dominant. More effective analyses require a nonassociate zero dilatancy rule, which is however not common in the pro-fessional engineering software. More refined damage rules are actuallyunder way, but not for normal design activity. Truss elements however,allow to introduce complex behaviour by means of geometric discreti-sation, and by this way, crack tracking can be pursued, even if in a veryrough representation.

4.8 Task 8.8: the strengthening of masonry structural elements with FRPsystems.The research activity has been performed through experimental testsand theoretical study. In particular, the experimental program has beencarried out with reference to masonry panels strengthened with FRPand with SRG and subjected to in plane loads, while the theoreticalstudy concerns the modelling of masonry walls strengthened with FRP.

Masonry panels and building strengthened with FRP

Several experimental tests have been performed with reference tomasonry panels strengthened with different FRP materials (carbon andglass) arranged according to various configurations.In-plane shear–compression tests have been performed on full-scaletuff masonry panels consisted of two-layered walls with the inner partfilled with mortar and chips from yellow tuff blocks, considering differ-ent FRP materials and strengthening configurations. In particular, twosets of panels have been strengthened with grid pattern carbon fibreunidirectional strips (CFRP), made with three horizontal and verticalstrips on each face and two sets of panels have been strengthened withthe same layout, but doubling the number of plies. Similarly, four setsof panels have been symmetrically strengthened with a grid pattern onboth sides of the panels, but with glass bidirectional fibre strips(GFRP). A further set of panels has been made selecting a different geo-metric configuration by arranging FRP laminates along the diagonals ofboth sides of the panels and considering both a CFRP and a GFRPcross layout with either one or two FRP plies.Important experimental evidences have emerged from the performedtests underlying the role of the configuration of the FRP strengthening


damages; in the second phase the prototype has been repaired byselecting GFRP strips arranged according to the Italian Code(Ordinanza 3431/2005, CNR DT 200/2004); in the third phase, dynam-ic tests have been performed on the strengthened prototype.Experimental evidences have shown that the applied GFRP strips allowto perform fast repair interventions in order to make operational mason-ry structures severely damaged by earthquake and at risk of after-shocks. In fact, the repaired prototype has showed reduced openings atthe horizontal joints (about one-third of the maximum opening observedin the case of un-strengthened prototype) and an increasing of the lat-eral strength (about +34%). A further aspect emerged from this testshas concerned the importance to provide effective anchorages for FRPstrengthening elements in order to avoid the delamination phenomenonwhich is particularly influenced by cyclic actions.

Masonry panels strengthened with SRG

Beside the “traditional” Fiber Reinforced Polymers (FRP), the researchinvestigated the possibility of application of innovative composite mate-rials, called Steel Reinforced Grout (SRG), based on high strength steelwires (Ultra High Tensile Strength Steel) forming that are assembledinto a fabric and embedded within a cementitious grout. This applica-tion in fact could combine, to the traditional advantages proper of FRP,the performances of this new material, reducing installation and mate-rial costs, and inducing an increase of ductility. Both composites, FRPand SRG, can be used with perforated or solid brick to form a newstrengthening system, called LATLAM ring – beam, that can be used toeffectively construct the roof ring beams of a masonry structure. Thisnew system, can be also subjected to a pretension force. The followingconclusions may be drawn from the developed research:- the analytical model developed to determine the mechanical behav-iour of LATLAM ring – beams has shown good agreement with experi-mental results and can be incorporated in design provisions;- the experimental tests, performed on full – scale prototypes of LAT-LAM ring – beams, demonstrated good results in terms of load carryingcapacity; - LATLAM ring – beams proved to be a good substitute, either under atechnical and economical perspective, of “traditional” reinforced con-crete ring – beams.

Modelling and analysis of masonry walls strengthened with FRP

A further subject of the task research activity has concerned the mod-elling and the analysis of masonry elements strengthened with FRP.Indeed, recent codes extended the use of displacement-based design

system on the failure mechanism of the tested panels. In fact, while insome cases it has been observed that the debonding of the FRP stripshas been the main responsible of the panels failure, in few cases thetensile rupture of the FRP strips has occurred. The different observedfailure mechanisms have particularly influenced the behaviour of thestrengthened panels both in terms of strength increase and post-peakbehaviour (fracture energy).Further tests have concerned square masonry panels composed of claybricks, strengthened with different FRP configurations and subjected tocompression diagonal load. In particular, some of the panels have beenstrengthened considering both vertical and horizontal CRFP and GFRPstrips, while other panels have been strengthened arranging FRP stripsalong diagonal directions of the panels. Different levels of the strengthincrease and different failure mechanisms have been observed. In par-ticular, while localized cracks pattern have characterized both the caseof un-strengthened panels and the case of panels strengthened by FRPonly on one side, a more diffuse crack distribution has been observedin the case of panels strengthened on the two opposite sides. In severalcases the debonding of the FRP strips has been also observed.From the experimental studies conducted on FRP strengthened mason-ry panels, considerations useful for improving the Document CNR-DT200/2004 have been deduced. In particular, it has been observedthat for the evaluation of the design shear strength of FRP strengthenedmasonry panels (eqn. 5.16 of the CNR-DT200/2006), the contributionof the masonry component can be evaluated through the eqn. 5.17 (usedin the case of reinforced masonry elements) only if the FRP shearstrengthening is coupled with FRP vertical elements fixed both at thebase and at the top of the panel. For this reason it has been suggestedto include in the CNR-DT200 specific design indications concerningthis aspect. In fact, it is important to underline that, when the FRPshear strengthening system is not coupled with a flexural FRP strength-ening system with efficient anchor elements, the contribution of themasonry material in terms of shear strength is the same of the case ofun-reinforced masonry elements.A further aspect deduced with reference to tests on masonry panelsstrengthened by FRP has concerned the case of fibres arranged alongto the diagonal directions of the panel. From the experimental evi-dences and from global analyses it has been possible to affirm that thecontribution of the FRP can be evaluated only considering the compo-nent parallel to the shear load.A scale model of a typical tuff masonry buildings has been constructedand tested on the seismic simulation shaking table at the structural lab-oratory of CESI, Bergamo, Italy. In particular, the test procedure hasconsisted in three phases: in the first phase dynamic tests have beenconducted on the un-strengthened prototype in order to induce some


methodologies, such as the pushover analysis, to the case of masonrybuildings. Thus, different modelling approaches able to capture thestructural behaviour of masonry panels strengthened with FRP havebeen examined: a macro-micro modelling approach; a macro modellingapproach; a frame model approach.The first model relies on a homogenization approach combined withlimit analysis suitable for the evaluation of the collapse loads and fail-ure mechanisms of FRP reinforced masonry panels. The application ofFRP strips on masonry has been treated adopting a simplified multistep approach. In the first step the un-reinforced masonry, regarded asa periodic heterogeneous material, has been substituted with a homo-geneous macroscopic material using a homogenization technique. Inparticular, an estimation of the homogenized unreinforced masonrystrength domain has been obtained by means of a micro mechanicalmodel based on the lower bound theorem of limit analysis. In the sec-ond step, FRP strengthening has been introduced on the alreadyhomogenized masonry material.The second model is based on the use of both 2D or 3D nonlinearbehaviour finite elements and interface elements. In particular, specialyield criteria coupled with nonlinear constitutive laws characterized bysoftening response have been selected in order to simulate the behav-iour of masonry material both in tension and compression. A specialnonlinear constitutive law has been also considered for the interfaceelements in order to simulate the debonding mechanism of FRP whichcharacterizes in several cases the failure mechanism of masonry ele-ments strengthened with FRP.Finally, the third modelling approach consists of a simple but effective1D frame element able to predict the response of masonry structures,eventually reinforced with FRP materials. In fact, the proposed ele-ments presents some peculiarities both for converting the geometry of amasonry panel in the geometry of the equivalent frame and for account-ing the nonlinear behaviour of masonry and FRP materials. Specialcomponents have been also provided in the developed model in orderto account for the shear failure modes.Several numerical applications have been carried out using the pro-posed modelling approaches and considering experimental casesdeduced from literature. The obtained results, regarding both un-strengthened and FRP-strengthened simple panels and masonryfaçades, have shown the reliability of the proposed models to reproducethe experimental behaviour of masonry elements underlying somepeculiarities due to the presence of the FRP strengthening system. Infact, the micro-macro and the macro modelling approaches have par-ticularly underlined the role of the FRP strengthening system to changethe stress path and, consequently, to induce different damage statescharacterized by a diffuse crack path along the strengthened elements.

This aspect has revealed the important role of the FRP system toincrease the energy dissipation capacity of masonry structures. Theincrease both in terms of strength and ultimate deformation capacity ofstrengthened elements has been confirmed by all the modellingapproaches evidencing the capability of the simple frame model to cap-ture the global response of masonry elements strengthened with FRPand consequently the possibility of using this simple model in a designprocess of the FRP strengthening system for masonry structures.

4.9 Task 8.9: the strengthening of masonry vaulted with FRP systems.The aim of the research task is the development of models and proce-dures for the analysis of vaulted structures reinforced with FRP. In par-ticular the behaviour of arches, vaults and domes subjected to seismicaction is studied in order to investigate the behaviour of these typolo-gies as built and strengthened with FRP materials. The study is performed with simplified analytical procedure and withnumerical methods. Experimental tests are further expected for the val-idation of the models. The research activity has been focused on the experimental tests on thetopic available in literature for a comparison with the analytical andnumerical results.On the basis of the acquired data, a simplified analytical model for theevaluation of the ultimate load of arches and portal frames reinforced atthe intrados and/or extrados has been developed. The main results are related to the definition of a methodology for theevaluation of the ultimate load of one-dimensional vaulted structuresreinforced with FRP and subjected to vertical and horizontal forces.Starting point of the procedure is the assessment that the FRP presencedoes not allow the formation of the collapse mechanisms, which char-acterise the ultimate behaviour of the un-reinforced structure. As a con-sequence the cinematic approach of the limit analysis, usually adoptedfor this kind of structure cannot be applied. The proposed analyticalmodel, starting from the analysis of the ultimate behaviour of the un-reinforced structure, identifies the location of the hinges, up to make thestructure statically determined, and then to be solved with equilibriumcondition. Obviously in this case and contrarily to the limit analysis, themechanical characteristics of the masonry and of the FRP and mainlythe debonding phenomenon of the composite material, significantlyinfluence the structural response. The model has been developed and applied to assigned geometries ofarches and portal frames, subjected to static vertical and horizontalloads, with FRP reinforcement at the intrados. The obtained resultshave been expressed in terms of interaction domains in the plane of thevertical and horizontal forces. Afterwards, the procedures for the evaluation of the ultimate behaviour


In the Laboratory of Structures and Materials of the Department of CivilEngineering of the University of Rome “Tor Vergata”, has been realizeda masonry portal frame, constituted by two columns with rectangularsection (24cm x 49cm) and height of 2m, and an arch with internalradius Ri = 130cm and square section with side equal to 24 cm. The portal frame was constructed with masonry block on mortar beds,and then reinforced with a layer of FRP at the intrados of the arch andon the internal surface of the columns. No anchorage of the compositehas been given at the foundation level.A constant vertical load (v) has been applied on the arch key and twohorizontal increasing loads, with equal intensity (H/2) are given at thearch abutments, which simulate a seismic action. The experimental results on both the arches and a portal frame appearto validate the simplified models and the procedures developed in thisresearch program for the design and check of masonry vaulted systemsreinforced with FRP.

4.10 Task 8.10: the quality control and monitoring of FRP applicationsto existing masonry and RC structures.Several masonry and reinforced concrete structures have been utilizedfor the application of fiber reinforced composite materials, with the aimof carrying out quality control and monitoring tests, in accordance toCNR DT 200/2004. For each structure, special working sheets havebeen developed for a proper characterization of the building from a geo-metrical, mechanical and logistic point of view. On these structuressemi-destructive tests, non-destructive tests and delamination testshave been performed.

Semi-destructive tests

This test consisting in pull-off and shear tearing tests conducted on dif-ferent types of FRP materials, applied on the in-situ structures.Moreover, pull-off tests and shear tearing tests have been conducted inthe laboratory of the Department of Structural Engineering of Universityof Calabria on concrete elements as well, reinforced by carbon fibersheets and laminates. The latter tests results have been utilized for acomparison with a number of experimental tests conducted at theUniversities of Bologna, Naples, Sannio and Milan, in the framework ofTask 8.2 Round-Robin tests, with the aim of obtaining a standard testprocedure for delamination tests and evaluating the scattering betweenexperimental results derived from tests conducted on specimens real-ized by the same worker, but tested in different laboratories. The testshave been conducted using the same device used for outdoor tests. Pull-off tests, used to assess the properties of the strengthened substrate,have been carried out attaching a thick circular 75 mm diameter steel

of masonry arches reinforced with FRP have been extended to othertypologies, such as the masonry portal frames reinforced at the intradosor extrados. In particular has been defined the formulation of a secant linear rela-tionship defined by the maximum load and by the related displacementat the brittle failure, to be adopted for the estimate of the seismic behav-iour of the structure. This model, even if simplified, can be suitable forsimulating the response of vaulted structures subjected to vertical andhorizontal loads, strengthened with FRP sheets. In this case, indeed, ifthe FRP is applied at the intrados, the failure is due, generally, to itsdelamination, while if the composite material is glued at the extrados,the failure is due to compression of shear crisis of the masonry. Anywaythe global behaviour of the structure is often characterised by an almostlinear response up to the maximum load, ad generally with a brittle fail-ure. Numerical FEM procedures in bidimensional field, have been furtherdeveloped. The comparison between the results obtained with the twomethodologies has allowed the validation of the simplified models. Parallel to the analytical-numerical models, an experimental programhas been set up, on vaulted structures reinforced with FRP. Three circular arches, subjected to a key load, have been tested. Thefirst one represents the reference un-reinforced, the other two havebeen strengthened with FRP at the intrados. The tests have been per-formed in displacement control, with the aim of evaluating the post-peak behaviour and the softening branches. The results obtained have allowed the preliminary validation of themodel related to arches reinforced at the intrados. The last phase of the research was devoted to the validation of thedeveloped models, through experimental tests on masonry vaulted ele-ments. In particular, besides the preliminary tests in masonry archesreinforced with FRP at the intrados, a test has been performed on amasonry portal frame, in full scale, reinforced with FRP and subjectedto vertical and horizontal loads. The obtained results allowed the vali-dation of the analytical models and assessed the validity of the assumedsimplified hypotheses.


Fig. 34- Portal frame under construction. Fig. 35- Portal frame during the test.

plate to the FRP and isolating it from the surrounding FRP with a coredrill, taking particular care in avoiding heating of the FRP system whilea 1-2 mm incision of masonry substrate was achieved. The test consistsin pulling off the steel plate by means of an ad hoc device (Figure 36-a), obtaining the ultimate pull-off strength value expressed in kN(Figure 36-c). Whereas, shear tearing tests are used to assess the qual-ity of bond between FRP and masonry substrate. These tests can beconducted only when it is possible to pull a portion of the FRP systemin its plane located close to an edge detached from the masonry sub-strate. The tests have been carried out using the same ad hoc deviceused for pull-off tests. In particular, metallic elements have been set uponto the masonry wall and through the FRP strip, with the aim of con-necting the entire test device. Then, the FRP element has been tight-ened until collapse (Figure 36-b), obtaining the failure tearing force,expressed in kN (Figure 36-d). For what concerns in situ tests, 16 rein-forced concrete structures and 17 brick and stone masonry buildinghave been considered, for a total number of more than 300 tests. TheFRP materials have been applied in the form of strips having thedimensions of 500x200 mm and 50x200 mm in the case of r.c. struc-tures and the dimensions of 500x300 mm e 50x300 mm in the case ofmasonry structures for the execution of pull-off and shear tearing test,respectively. Both carbon, glass and natural fiber composites have beenutilized.According to CNR DT 200/2004, FRP application may be consideredacceptable if at least 80% of the tests return a pull-off stress not lessthan 10% of masonry support compressive strength, or not less than0.9-1.2 MPa in the case of reinforced concrete structures, provided thatfailure occurs in the support itself. For what concerns shear tearingtests, FRP application may be considered acceptable if at least 80% ofthe tests return a peak tearing force not less than 5% of masonry sup-port compressive strength, whereas it has to be not less than 24 kN inthe case of reinforced concrete structures. For what concerns masonry structures, taking into account the com-pressive strength of the support, both pull-off and shear tearing experi-mental results respect the limit values suggested in CNR DT 200/2004.In the case of concrete structures, pull-off results are in accordance toCNR DT 200/2004, whereas in shear tearing test results the limit valueof 24 kN suggested for reinforced concrete structures, has never beenreached. This is probably due to the small dimensions of the FRPstrips. In fact, CNR DT 200/2004 Guidelines don’t provide any instruc-tion about the FRP dimensions that should be utilized to reach theabove mentioned limit value.In the case of pull-off tests, failure has always occurred in the substrate,for each type of FRP material applied, as expected. The semi-destructive tests conducted in the laboratory on concrete ele-

ments reinforced by CFRP sheets and laminates have been realized.From mechanical characterization tests, a value of 25 N/mm2 for con-crete compressive strength was found. The Research Unit of Universityof Calabria tested 3 prisms reinforced by 3 CFRP strips and CFRP lam-inates applied on their surfaces. Pull-off and shear tearing tests havebeen conducted, with the aim of studying the failure mode and debond-ing of FRP from the substrate.The results of pull-off tests conducted on both strips and laminatesrespected the limit values suggested in CNR DT 200/2004, and also inthese cases failure occurred in the substrate, showing the effectivenessof the FRP application. On the other side, shear tests showed a differ-ent behaviour of the composite materials for strips and laminates. Inparticular, in the case of laminates, the collapse occurred suddenly andthe whole composite debonded from concrete, with an ultimate value ofthe shear force higher than 24 kN (limit value suggested in CNR DT200/2004). In the case of CFRP strips, a partial delamination of thecomposite occurred and the limit value was never reached.

Non-destructive tests

Non-destructive tests, consisting in thermographic tests were conduct-ed on both brick masonry elements built in the laboratory and on thesame real masonry structures used for the above mentioned semi-destructive tests, reinforced with different FRP materials. These testsare usually carried out to characterize the uniformity of FRP applica-tion. Both, the active and the passive thermography technique havebeen adopted, in which thermal energy is applied externally onto thetest object or the natural infra-red radiation emitted by the object dueto a sufficient exposure to sun light can be utilized, respectively.


Fig. 36- Pull-off test (a) and shear tearing test (b); Pull-off test result (c) and shear tearing test result (d).

the surface preparation. The profilometer used is the DRSC producedby Miami University, which gives quality and quantity information. Thisinstrument using laser striping, highlights the rough concrete surface bythin slits of red laser light at an angle of 45 degrees, and the surface isobserved at 90 degrees by an high-resolution (tiny) board CCD camera(Figure 38a). The video image of the laser stripes is digitized with aPCMCIA frame-grabber. The projected slit of light appears as a straightline if the surface is flat, and as a progressively more undulating line asthe roughness of the surface increases. Lasers with one to eleven stripeswere used. The roughness degree can be identified by several parameters obtainedby laser profilometry and each one can give specific information onroughness and its particular properties. They can be classified in ampli-tude parameters (Ri of Figure 37) and slope parameters (ia of Figure37a). The first group is sensitive to roughness morphologies, where thesurface is either stepped or slotted and might be described as a discon-tinuous roughness, the second group is insensitive to roughness mor-phologies and is more useful for characterizing continuous roughness.

The first Infra-Red thermographic tests have been conducted in the lab-oratory of the Department of Structural Engineering of University ofCalabria on brick masonry macro-elements, which have been rein-forced by carbon fiber strips placed in different directions onto thespecimen surfaces. From the test a mortar joint was clearly visible dueto the non plane substrate surface.The in situ masonry and reinforced concrete structures have been uti-lized for thermographic tests as well, to verify the quality of bondbetween FRP and the substrate, before and after the conduction of thedescribed semi-destructive tests. For instance, some tests have beenconducted on a reinforced concrete structure on which FRP strips havebeen previously applied for shear tearing tests. From the thermograms,a crack could be noticed corresponding to the strip subjected to thesemi-destructive test. Such a technique is then useful for the detectiondeterioration and damage in the structures.

Influence of roughness surface on the debonding force of FRP

In order to investigate the effect of the concrete surface preparationmethod on the roughness surface and debonding force of FRP, an exper-imental campaign has been carried out. The specimens have been pro-duced with different formworks (staves, panels) and different com-paction types (beating, vibration). Moreover, two different concretestrengths have been used in realizing the specimens in order to evalu-ate also their influence on the surface preparation method efficacy.Thirty 15 cm-length standard cubes have been also poured and used toevaluate the mechanical properties of concrete (according to Italianstandards). Mean compressive strength (Rcm = 15 or 20 MPa) from thecompression tests has been obtained by standard cube at an age of 50days, corresponding to same period of the first profilometer tests. Thenumber of specimens considered in the present experimental campaignand the two different casting processes are shown in Table 1.After curing, four different methods for surface preparation have beenapplied on concrete prisms in order to study the effect of the treatmenton the FRP-concrete bond strength: Grinding – the upper surface of theconcrete block has been grinded with a stone wheel to remove the toplayer of mortar, just until the aggregate was visible; Sand Blasting – theconcrete surface has been sand blasted in order to remove the wholemortar over the aggregates, so obtaining a very rough concrete surface;Brushing – the surface has been brushed with a twisted steel cord bond-ed to a rotating disc; Scabbling – impacting the substrate at variableangle with a metallic tip to create a chipping and powdering action. Thedriving mechanism is compressed air.In order to examine exhaustively the concrete surface, an extensivecampaign of laser profilometer analysis was carried out before and after


Fig. 37- Roughness parameters.

Table 1 Specimens and summary of test sampleN° Specimen Dimension Mean Compressive Formwork Type of

(mm) Strength (N/mm2) (wooden) Compaction20 ? 160 x 400 x 600 15 staves beating20? 20 panels vibration

Figure 37b shows some output parameters provided by the profilome-ter: Rmax – maximum peak and valley, rough measure of the vertical dis-tance between the highest peak and the lowest valley; Re measures thevertical distance between the highest peak and the centerline of theprofile; Rv measures the vertical distance between the lowest valleysand the centerline of the profile; R measures the average of all individ-ually measured peak to valley heights, Rp – roughness profile index,defined as the ratio of the true length in the fracture surface trace to itsprojected length in the fracture plane; iA is the micro-average inclina-tion angle, representing the average of the pixel to pixel angles of thestripe profile.In order to define a unique parameter for describing the surface rough-ness, profilometer parameters were analyzed and correlated. For eachparameter given by the profilometer, average value and covariance have

been calculated for evaluating the quantity and quality factors of rough-ness. The covariance can provide for information on surface homogene-ity; both quantities are interesting especially regarding the efficiency ofconcrete surface preparation methods. In the following, the roughnessis described by coefficient IR = R�ia, where R and ia have beendescribed before. The parameter IR is used to give information on theabsolute value of the roughness and on its specific shape. Profilometeranalysis allows to correlate the casting methods with various degrees ofroughness. In fact, the specimens casted with staves are more roughbecause the disconnection of the staves increase the surface irregulari-ty. The specimens compacted by means of the vibration are more roughdue to different positions of aggregates and the presence of vacuum pro-duced by air bubbles.The roughness of the concrete surface has been investigated before andafter the preparation; its difference is a way to evaluate the efficiency ofeach surface preparation method. In Figure 39a,b are shown the aver-age values and the covariance of the IR parameter for all the differentsurface preparation methods. Figure 39a shows that all the surfacesprior to the treatments have very similar roughness while after them themean value of the IR parameter is particularly high in the case of scab-bing and sand-blasting. On the contrary, all the surface preparationmethods strongly reduce the statistical dispersion of IR parameter(Figure 39b); the scabbing and sand-blasting provide for the higherlevel of homogeneity.

Delamination tests

Delamination tests have been carried out on bricks reinforced with FRPmaterials, with the aim of studying both the collapse load value and thefailure modes that can occur if FRP materials applied during strength-ening interventions collapse. In particular, delamination tests have

been conducted on two half bricks placed inside a properly designedsteel frame and connected between them by means of two carbon, glassor natural FRP strips glued on both sides of the specimens. The framehas been designed in such a way as to avoid any hindrance to the spec-imen collapse. The specimens have been subjected to uniaxial tensiletests under displacement control by means of an electro-mechanic test-ing machine, with a capacity of 100kN, connected to a personal com-puter. Some specimens have been also monitored by means of unidi-rectional strain gauges applied both on the brick surface and on thefibres.From the experimental tests, the specimens failure mode has beenanalysed. Collapse has occurred for delamination in almost all cases.However, some specimens have reached failure due to fibres crack, as inthe case of glass fibers and natural ones due to the different stiffness ofthe FRPs. More than 30 specimens have been tested. In the case of spec-imens reinforced by carbon and glass fibers, the load-displacementcurve relative to failure for delamination obtained from the strain gagesapplied on FRP and brick surfaces was almost linear until failure whichoccurred suddenly and for debonding of the fabric from the reinforcedbricks, with consequent removing of part of the substrate surface. Thepresence of relevant traction stresses is observed at the attachment of theFRP strip with the bricks where the delamination phenomenon begins.For what concerns the tests conducted on bricks reinforced by naturalfibers, the results, obviously, cannot be compared to the ones derivedfrom brick reinforced by CFRP or GFRP; however, the tests have showninteresting properties of natural fibers for not bearing applications suchas on ancient masonry structures where FRP mechanical behavioursensibly affects the global behaviour of the structure.Finally, for each tested specimen, an accurate study of the substrateafter failure has been conducted for the exact definition of the delami-nated surface dimensions; the crack begins at a depth of about 10 mm


Fig. 38- (a) Laser profilometer, (b) geometry of FRP-strengthened prism and (c) experimental set-up.

Fig. 39- Average values (a) and covariance (b) of IR parameter for different surface preparations.

Fig. 40- Failure mode of bricks reinforced with CFRP strips subjected to delamination tests.

been focused on both masonry and reinforced concrete members. Inparticular, advanced FRP materials made of basalt fibers (bondedwith traditional or inorganic cement matrix) have been used for theconfinement of RC and masonry columns. In addition, the mechanicalproperties of several FRP systems at fixed environmental conditionsand/or under cyclic actions, the behaviour of RC members confined orstrengthened in flexure and shear as well as of beam-column andbeam- foundations joints have been deeply analyzed. Design criteriafor the seismic retrofit of masonry members and structures have beenalso provided. The outcomes of experimental activities have, in mostcases, confirmed the reliability of CNR-DT200/2004 provisions, espe-cially for RC members. Regarding to masonry members, test resultsindicate that some coefficients of theoretical expressions provided byCNR-DT200/2004 have to be refined and some other experimentalresults could be necessary to derive theoretical relationships specifi-cally targeted at different masonry substrates and failure modes.


Innovative materials for the vulnerability mitigation of existing struc-tures have been largely analysed in the research activity. Further, newtypologies of FRP systems, also made of several fibers and matricestypes are spreading; on these new typologies further investigations arestrongly necessary in the future. The study of the feasibility of usingcomposite systems made of inorganic matrix strengthened with naturalfibers fabrics is needful. In fact, the idea of using natural fibers is dueto economic and environmental sustainability suggested by the use ofsuch materials. These studies could allow new FRP systems to beinserted in the actual guidelines in order to increase their use in thecivil engineering applications. The indications provided by CNR-DT 200/2004 for the vulnerabilitymitigation of existing structures have been accepted by recent Italianguidelines, but further experimental tests will need, especially to miti-gate the vulnerability of existing masonry structures.

and this distance obviously depends on the experimental equipment,whereas the depth and the form of delaminated substrate appear almostconstant and dependent on the testing modality (Figure 40).From both in-situ and laboratory tests, important information wereobtained regarding composite materials behaviour and experimentalprocedures. In particular, it was noticed that shear tearing tests areextremely affected by instrumental errors that can take place during thetest conduction. In fact, while FRP laminates allow a perfect alignmentof the composite with respect to the hydraulic pull machine, in the caseof FRP strips the applied force is not perfectly aligned with the FRP –concrete interface, and peeling stresses are generated, so causing a sig-nificant reduction of the debonding force.Then, on the basis of the above described experimental results and inaccordance to workshops organized during the Research Activity, somevariations to the CNR DT 200/2004 guidelines, with particular regardto shear tearing limit value and experimental procedure, were proposed.In particular, shear tearing test should be carried out following two dif-ferent procedures, namely direct and indirect procedures. The directprocedure is preferred if the load can be directly applied on the FRPglued onto the specimen surface, especially in the case of laminates. Ifthe FRP system is made of strips to be prepared in-situ, the indirect testprocedure is preferred, in which load is applied by means of steel platesglued onto the FRP surface.

5. DISCUSSION

In this research task a large number of experimental tests has beenperformed in order to validate the design equations provided by CNR-DT200/2004 and to calibrate some coefficients included in theseequations to better fit theoretical predictions and experimentalresults. Experimental tests have been carried out on both real scale or scaledmembers by using traditional FRP systems or innovative typologies ofFRP systems and advanced materials. The experimental tests have


7. MAIN REFERENCES

- Aprile A., Benedetti A., Cosentino N., (2006), “Seismic Reliability ofMasonry Structures Strengthened with FRP Materials”, 100th

Anniversary Earthquake Conference, San Francisco, paper n° 1677.- Aprile A., Benedetti A., Steli E., Mangoni E., (2007), “Seismic RiskMitigation of Masonry Structures by Using FRP Reinforcement”,FRPRCS-8 University of Patras, Patras, Greece, paper n° 424.- Anselmi V., Aprile A., Benedetti A., (2005), “Safety and reliability ofstructures including ductile and brittle elements”, ICOSSAR 2005,Augusti, Schuëller, Ciampoli (eds), pp. 2183-2188, Rotterdam, ISBN90 5966 040 4.- Ascione F., Feo L., Olivito R.S. and Poggi C., “La qualificazione del-l’esecuzione degli interventi di rinforzo strutturale con FRP a marginedelle recenti ”Istruzioni per la progettazione, l’esecuzione ed il control-lo di interventi di consolidamento statico mediante l’utilizzo di com-positi FRP””, Proceedings of National Italian Conference “Ambiente eProcessi Tecnologici – La certificazione di Qualità dei materiali e deiprodotti da costruzione” (in Italian), Naples (2005).- Bastianini F., Olivito R.S., Pascale G. and Prota A. (2005), “Controllodi qualità e monitoraggio dei rinforzi in FRP”, L’Edilizia (in Italian),139, 66-71.- Benedetti A. and Steli E. (2007), “Analytical Solution of the Shear –Displacement Curve for Reinforced Masonry Panels”, The Tenth NorthAmerican Masonry Conference, St. Louis, Missouri, ISBN 1-929081-28-6.- Benedetti A., Camata G., Mangoni E., and Pugi F., (2007), “Out ofPlane Seismic Resistance of Walls: Collapse Mechanisms and RetrofitTechniques”, The Tenth North American Masonry Conference, St. Louis,Missouri, ISBN 1-929081-28-6.- Benedetti A., Mangoni E., Montesi M, Steli E. (2007), “Verifiche diSicurezza ed Interventi di Consolidamento Della Chiesa di S. Martinoin Casola”, INARCOS, 680, pp. 411-423.- Benedetti A., Steli E., (2008), “Analytical models for shear–displace-ment curves of unreinforced and FRP reinforced masonry panels”,Construction and Building Materials, 22, pp. 175-185,doi:10.1016/j.conbuildmat.2006.09.005.- Bianco V. (2008), “Shear strengthening of RC concrete beams bymeans of NSM CFRP strips: experimental evidence and analytical mod-eling”, PhD Thesis, Dept. of Struct. Engrg. And Geotechnincs, SapienzaUniversity of Rome, Italy, submitted on December 2008.- Bianco V., Barros J.A.O., Monti G. (2009a), “Bond Model of NSM FRPstrips in the context of the Shear Strengthening of RC beams”, ASCEJournal of Structural Engineering, in press.- Bianco V., Barros J.A.O., Monti G. (2009b), “Three dimensional

mechanical model for simulating the NSM FRP strips shear strengthcontribution to RC beams”, Engineering Structures, Vol. 31 n. 4, Else -vier.- Bruno D., Greco F., and Lonetti P. (2005), “A 3D delamination mod-elling technique based on plate and interface theories for laminatedstructures”, European Journal of Mechanics A/Solids, 24, 127-149.- CNR DT 200/2004, “Guide for the Design and Construction ofExternally Bonded FRP Systems for Strengthening ExistingStructures”, Italian National Research Council, Rome (2004).- De Lorenzis L., Rizzo A., (2006), “Behaviour and capacity of RCbeams strengthened in shear with NSM FRP reinforcement”, 2nd Int. fibCongress, Napoli-Italy, June 5-8, Paper ID 10-9 in CD.- De Lorenzis L., Galati D. (2006), “Effect of construction details on thebond performance of NSM FRP bars in concrete”, Proceedings fibCongress, Napoli, Giugno 2006.- Dias S.J.E., Bianco V., Barros J.A.O. (2007), “Low strength concreteT cross section RC beams strengthened in shear by NSM technique”,Workshop Materiali ed Approcci Innovativi per il Progetto in ZonaSismica e la Mitigazione della Vulnerabilità delle Strutture, Universityof Salerno, Italy, 12-13 February.- Galati D., De Lorenzis L. (2006), “Experimental study on the localbond behavior of NSM FRP bars to concrete”, Proceedings CICE 2006,Miami, USA, December 2006.- Galati D., De Lorenzis L. (2008), “Effect of construction details on thebond performance of NSM FRP bars in concrete”, Advances inStructural Engineering, Multi-science, in stampa.- Milani G., Rotunno T. , Sacco E. and Tralli A. (2006), “Failure Loadof FRP strengthened masonry walls: experimental results and numeri-cal models”, Structural Durability & Health Monitoring, 2 (1), 29-50.- Olivito R.S. and Zuccarello F.A. (2006), “Indagine sperimentale per ilcontrollo dell’applicazione di materiali FRP a strutture murarie medi-ante prove semi-distruttive e non distruttive”, Proceedings of NationalItalian Conference on Materials and Structures Experimentation (inItalian), Venice.- Rizzo A., De Lorenzis L. (2006), “Analytical Prediction of DebondingFailures in RC Beams Strengthened in Shear with NSM FRPReinforcement”, Proceedings CICE 2006, Miami, USA, December2006.- Rizzo A., De Lorenzis L. (2007), “Modelling of debonding failure forRC beams strengthened in shear with NSM FRP reinforcement”,Proceedings FRPRCS8, Patras, Luglio 2007.- Rizzo A., De Lorenzis L. (2009a), “Behavior and capacity of RCbeams strengthened in shear with NSM FRP reinforcement”,Construction and Building Materials, Elsevier, Vol. 23, No. 4, pp. 1555-1567.


- Savoia M., Ferracuti B. and Mazzotti C. (2003), “Delamination of FRPplate/sheets used for strengthening of r/c elements”, Proceedings ofSecond International Structural Engineering and Construction (ISEC-02), Rome, Italy.

- Rizzo A., De Lorenzis L. (2009b), “Modeling of debonding failure forRC beams strengthened in shear with NSM FRP reinforcement”,Construction and Building Materials, Elsevier, Vol. 23, No. 4, pp. 1568-1577.



bad seismic behaviour. Industrial one storey precast buildings weredefined as inverted pendulum systems to which a very low value of thebehaviour factor was recognised. This was in conflict with the natio-nal codes that did not make any difference between cast-in-situ andprecast frames. This was in conflict also with the experience of pastearthquakes, where precast structures, except for the dry supports,showed a very good ductile behaviour despite their non seismic desi-gn. The lack of experimental data was adduced to justify the heavypenalisation of precast structures in this code.In order to verify the validity of this penalisation the ItalianAssociation of prefabrication industry promoted a campaign of analy-tical investigations that have been developed at Politecnico di Milano.The investigation have been started with a set of cyclic tests on pre-cast columns. Figure 1 shows one of the prototypes of precast columnstested at ELSA Laboratory of the Joint Research Center of theEuropean Commission at Ispra, Italy. Cyclic and pseudodynamic testswere performed for different reinforcement amount and axial actions.An example of force-displacement diagrams obtained from cyclictests is shown in Figure 2.a. The energy dissipated over the half-cycles was compared with the maximum value of dissipated energyassociated to a perfect elastic-plastic cycle, as shown in Figure 2.b.The results of these tests confirmed:- a good ductile behaviour with specific dissipation around 0.4, astypical for cast-in-situ columns;- a more reliable behaviour due to the absence of bar splices and tothe stable position of stirrups during the casting of concrete (precastcolumns are cast in an horizontal position);

Over the last two decades an extensive experimental and theore-tical research activity aimed to investigate the seismic beha-

viour of precast structures has been carried out at European scale.The results of this activity allowed to consolidate a good knowledge ofthe seismic behaviour of precast structures and contributed to theachievement of prefabrication in Europe with outstanding realizationsin terms of both quality and reliability.

The research was developed within six research programmes. Thefirst stage developed between 1992 and 1996 during the drafting ofthe first ENV version of Eurocode 8 (EC8). The initial draft of the spe-cific rules for precast structures gave them the presumption of a very

Experimental Research on Seismic Behavior of Precast Structures

Fig. 1- Prototype of precast column.

Fig. 2- (a) Force-displacement diagrams obtained from cyclic tests. (b) Energy dissipated over the half-cycles compared with the maximum value of dissipated energy associated to a perfect elastic-plastic cycle.

Dim

ensi

onle

ss v

alue

s

Half-cycles(a) (b)


- the fundamental importance of a narrow spacing of the stirrupsagainst the early buckling of longitudinal bars;- displacement ductility ratios between 3.5 and 4.5 consistent with thecode provisions for cast-in-situ frames.

The ENV EC8 was therefore published with precast structures nomore treated as inverted pendulum, but still penalised with a lowerbehaviour factor with respect to cast-in-situ systems.

The seismic behaviour of cast-in-situ and precast structures has beenalso investigated by means of proper numerical models on probabili-stic bases. This investigation was carried out by means of non lineardynamic analyses reproducing the real vibratory behaviour of thestructures under earthquake conditions. The aim was to demonstratethat, under the same seismic action, the monolithic frame shown in

Figure 3.a, with four critical cross-sections dimensioned for a momentm�Fh/2, may dissipate the same amount of energy which the hingedarrangement shown in Figure 3.b dissipates in its two critical cross-sections, dimensioned as they are for a double moment M=Fh�2m. Infact it is the global volume involved in dissipation, and not the num-ber of plastic hinges, that gives the total amount of energy dissipatedby the structure.To demonstrate this assumption, the two prototypes shown in Figure 4were considered, the first cast-in-situ with monolithic connections,the second precast with hinged connections. They have the same ove-rall dimensions, with the size and reinforcement of the columns cho-sen to achieve the same vibration periods and the same design sei-smic capacity in terms of base shear strength. With the combinationof the different heights and cross-sections (Figure 5) a number oftypes of frame were selected for both the frames.

The seismic response of the prototype was investigated in probabili-stic terms for lognormally distributed material strengths and underartificial accelerograms, randomly generated so to comply with thedesign response spectrum. A Monte Carlo simulation based on a large

Fabio Biondini, Giandomenico TonioloDepartment of Structural Engineering, Politecnico di Milano, Milan, Italy

Fig. 3- Energy dissipation in one-storey frames: (a) monolithic and cast-in-situ; (b) hinged and precast.

Fig. 5- Cross-sectional details of the columns. Fig. 6- Statistical distribution of overstrength ratio k: (a) monolithic and cast-in-situ; (b) hinged and precast.

Fig. 4- Prototypes of one-storey frames: (a) monolithic and cast-in-situ; (b) hinged and precast.

(a)

(a)

(b)

(b)


sample of incremental nonlinear dynamic analyses taken up to colla-pse was therefore carried out for each prototype to compute the stati-stical parameters of the overstrength �, ratio of the computed valueover design value of the seismic capacity. Figure 6 shows the distri-bution of overstrength computed for the two prototypes for a set of

1000 accelerograms in one of the cases studied. These results provethat precast structures have the same seismic capacity of the corre-sponding cast-in-situ structures, and confirm the correctness of thevalues given by the code to the behaviour factor of concrete cast-in-situ frames (q=4.5).

Fig- 7- View of the structural prototypes with (a) monolithic and (b) hinged beam-column connections.

Fig. 8- Displacement time-histories for one of the pseudodynamic tests: numerical (thick lines) versus experimental results (thin lines).

(a) (b)


of the seismic capacities of cast-in-situ and precast structures, andat the same time the validation of the analytical model used in theprevious numerical investigation. Figure 7 shows a view of the fullscale prototypes. The comparison of the results obtained for theseprototypes highlights the expected large seismic capacities (aboutag=1g) of this type of structures and confirms the overall equivalen-ce of the seismic behaviour of precast and cast-in-situ structures (see

The third stage of research developed during the revision of EC8 forits conversion to the final EN version. The preceding analyticaldemonstration was effective, but an experimental confirmation wasstill necessary. Therefore, taking advantage of the Ecoleader pro-gramme for the free use of the large European testing facilities, twopseudodynamic tests on full scale prototypes have been performed atELSA Laboratory. The aim was the direct experimental comparison

Fig. 9- View of the structural prototypes with roof elements with axis (a) parallel to the direction of the seismic action (Prototype 1), and (b) orthogonal to the direction of the seismic action (Prototype 2).

Fig. 10- Experimental tests carried out on a connection. (a) Test set up. (b) Failure mechanism.

(a)

(a) (b)

(b)

Growth research projects showed the good seismic performance ofpre cast structures under condition that the connections are properlyover-dimensioned. The last aspect to be still clarified is therefore theactual behaviour of connections under seismic excitation. Based onthese needs, the European research program Safecast has been recen-tly launched to investigate the seismic performance of connections inprecast systems. This project will involve a large campaign of experi-mental static tests carried out on single specimens, such as the con-nection shown in Figure 10, as well as pseudo-dynamic tests on athree-storey full-scale prototype shown in Figure 11. The results ofthe Safecast project are expected to complete the large research pro-gram developed in Europe over the last two decades which providedsignificant advances in the understanding of the seismic behaviour ofprecast systems and in the definition of reliable design criteria for thistype of structures.

ACKNOWLEDGEMENTS

A number of partners participated to the research, coming from theprincipal European countries subjected to seismic hazard. The natio-nal associations of prefabrication industry were involved (ANIPB forPortugal, ANDECE for Spain, ASSOIBETON for Italy, SEVIPS forGreece, TPCA for Turkey). Also the Italian associations of cementindustry AITEC and ready-mix concrete ATECAP participated tosome stages. The “research providers” were: JRC – Joint ResearchCentre of Ispra (of the European Commission), LNEC – LaboratorioNacional de Engenharia Civil of Lisbon, Politecnico di Milano,University of Ljubljiana, NTUA – National Technical University of

Figure 8).

The fourth stage of research was developed within the Growth pro-gramme. Two further prototypes consisting of six columns and a meshof beams and roof elements were designed to investigate the seismicbehaviour of precast structures with roof elements placed side byside. Figure 9 shows a views of these prototypes and of the testingplants. The prototypes differ only for the orientation of the beams androof elements with respect to the seismic action. Common hinged con-nections are used between roof elements, beams and columns. Thecontrol of the pseudodynamic test is based on two degree of freedoms,associated with the top horizontal displacements of the lateral frames,and of the central frame. Also the effects of cladding panels on thestructural response has been investigated.

The measured top displacements of lateral and central columnsduring the pseudodynamic tests resulted practically coincident. Thisresult proves that double connection between beams and roof ele-ments gives a rotational restraint in the roof plane which enables theactivation of an effective diaphragm action, even if the roof elementsare not connected among them. After the pseudodynamic tests bothprototypes have been subjected to a cyclic test under imposed displa-cements up to collapse. With a ultimate displacement du�360 mm anda yielding displacement dy�80 mm, a global displacement ductilityequal 4.5 is deduced, as assumed by the new final version of EC8 forthe behaviour factor of precast frame systems.

The results of the investigations carried out under the Ecoleader and


Fig. 11- Tree-storey full-scale prototype. (a) Transversal and (b) longitudinal section.

(a) (b)

researches coordinated by prof. Giandomenico Toniolo of Politecnicodi Milano. Many other theoretical and/or experimental researcheshave been performed on precast structures by initiative of single com-panies for their specific interests. And this wide activity qualifies theprefabrication industry as one of the most advanced sectors in growthand innovation.

Athens, ITU – Istanbul Technical University, and the private labora-tories LABOR and LUGEA (I). The “users” were some producers ofprecast structures (Magnetti Building – I, Gecofin – I, Civibral – P,Prelosar– E, Proet – GR), and some auxiliary companies (Halfen – D,DLC – I).What described in the present report refers to the series of European


80 RESEARCH - Concrete

lary tension caused by the formation of water menisci developed in cap-illary pores and responsible for the shrinkage of the cement paste(Figure 1).

1.INTRODUCTION

The research activity carried out in Italy on the structural concretesis very busy in both the academic compartments and in the indus-

trial operations. The present report summarizes some of the most impor-tant works in this area including three aspects of this activity:- shrinkage compensating concrete in the absence of wet curing;- properties of concretes with recycled aggregates;- use of bottom ash from municipal solid wastes incinerators.

2. SHRINKAGE COMPENSATING CONCRETE IN THEABSENCE OF WET CURING

Shrinkage-compensating concretes have been extensively used in thelast forty years to minimize cracking caused by drying shrinkage inreinforced concrete structures.The first and most diffused system to produce shrinkage-compensatingconcretes involves the use of expansive cements, according to ACI 223-98, instead of ordinary portland cement. All these special binders arebased on a controlled production of ettringite. Another effective method to produce shrinkage-compensating con-cretes, not covered by ACI 223-98 but commonly used in some coun-tries, like Italy or Japan, lies in the use of a CaO and/or MgO basedexpansive agent. This technology seems to be more advantageous withrespect to that based on the ettringite formation from an economical aswell as from a practical point of view.Recently, the addition of a shrinkage-reducing admixture (SRA) hasbeen found to improve the behavior of CaO based shrinkage-compen-sating concretes especially in the absence of an adequate wet curing [1].Although the actual cause of this synergistic effect has not been com-pletely explained, the use of this technology in construction industryhas been increased, in the last five years, particularly in Italy, with veryinteresting results.In the present report three remarkable examples of special reinforcedconcrete structures are presented in which the use of CaO-SRA basedshrinkage-compensating concretes was successfully carried out in orderto prevent shrinkage related cracks and/or joints excessive opening inthe presence of adverse curing conditions which are normally not suit-able for the use of this technique.SRAs (Shrinkage-Reducing Admixtures), are generally based on propy-lene-glycol ether, neo-pentyl glycol or other similar organic substances,that are able to reduce the drying shrinkage of concrete up to 50% ifused in 1-2% by mass of cement.According to Berke et al. [2] the effectiveness of SRA must be ascribedto the decrease in the surface tension of water .This reduces the capil-

State-of-the-art on the Research onStructural Concrete in Italy

Recently [3], the combined addition of a shrinkage-reducing admixturewith a CaO-based expansive agent has been found to be very success-ful in producing restrained expansion of laboratory specimens protect-ed from water evaporation for just 1 day by using a plastic sheet andthen exposed to air (60% R.H).The influence of the SRA on the length change behaviour of a shrink-age-compensating concrete includes two different aspects: - the � effect in Fig. 2 due to a reduction in shrinkage when the con-crete is exposed to drying, as expected for the presence of a shrinkage-reducing admixture; - the unexpected � effect, which is an increase in the restrained expan-sion when the concrete is protected from drying with respect to thatobtained without SRA, all the other parameters being the same.By using a combination of CaO and SRA, then, it is possible to reducethe amount of expansive agent needed to obtain a fixed restrainedexpansion. This reduces the risk of residual un-reacted lime in the con-crete.Furthermore, the performance in terms of initial restrained expansionand final restrained shrinkage (or residual expansion), of SRA + CaO-based shrinkage-compensating concretes is less dependant on the cur-ing efficiency so that the practical use of this technique is easier andthe results are more reliable.The synergistic effect in Figure 2 has been confirmed by Maltese et al[4] who have found that the use of a CaO-based expansive agent with ashrinkage reducing admixture allows to obtain mortars less sensitive todrying. These authors hypothesize that the synergistic effect of the

• Fig. 1- Water menisci interact with C-S-H fibers determining the shrinkage on cement paste, The New Concrete.

81RESEARCH - Concrete

SRA-CaO combination must be ascribed to the massive formation ofCaO elongated crystals during the first hours of curing.The same authors in [5] propose another mechanism of action: since theSRA is an organic hydrophobic molecule, it could reduce the water sol-ubility of CaO, retarding its reaction and, then, increasing therestrained expansion according to Chatterji [6].Otherwise, Tittarelli et al. [7] have found that SRA doesn’t affect thespeed of CaO reaction with water.

London, U.K.) had proposed the construction of several architecturalconcrete walls (20 meters high and 60 meters long) having a sinuousshape and no contraction joints (Figure 3).A special CaO-SRA based shrinkage-compensating self-compactingconcrete (SCC) was studied in order to assure a marble-like look, asrequired by the designers, even in the presence of a very congestedreinforcement (Figure 4) and, in the same time, to avoid the formationof shrinkage related cracks along the surface.

Mario Collepardi** ENCO, [email protected]

• Fig. 2- Schematic view of the influence of SRA on the length change behavior of a shrinkage-compensating concrete.

• Fig. 3- View of bent and joint-less walls of the MAXXI, Rome, Italy.

• Fig. 4- Example of steel congestion in a typical wall of MAXXI, Rome, Italy.

Although this synergistic effect has been confirmed by several authors,the actual mechanism of action needs further investigations in order tobe completely understood.Notwithstanding this lack of knowledge, the use of this technology, inthe construction industry, has been growing in the last 5 years withmany successful and very interesting results.In the second part of this paper, three remarkable case histories of spe-cial reinforced concrete structures are presented in which the use ofCaO + SRA-based shrinkage-compensating concretes was successfullycarried out in order to prevent shrinkage-related cracks and/or jointsexcessive opening in the presence of adverse curing and thermal con-ditions. The difficulties encountered in using this technique, in each case, willthen highlight describing the countermeasures which have been takento overtake them.

2.1 Case History of shrinkage compensating concrete in the absence ofwet curing: MAXXI of Rome

The Museum of Arts of XXI century (MAXXI) in Rome was the first rel-evant Italian construction in which a SRA + CaO-based shrinkage-compensating concrete has been used (2004-2006). For this very prestigious building, the designers (Zaha Hadid Limited,


In order to demonstrate the effectiveness of this type of concrete in off-set the formation of shrinkage cracks, its performances were comparedto those of an ordinary CaO-based shrinkage-compensating concrete(without SRA) and of a plain SCC mixture without expansive compo-nent and SRA. Table 1 shows the composition of these three SCCs having the same w/c(0.48) and approximately the same cement dosage (350 kg/m3).

Table 1 – Composition of three different SCCMix CaO+SRA CaO Plain

Cement CEM II A/L 42.5R (kg/m3)* 350 348 347

Limestone filler (kg/m3) 150 149 183

Gravel 4-16 mm (kg/m3) 847 884 871

Sand 0-4 mm (kg/m3) 908 916 903

Water (kg/m3) 167 167 166

Acrylic superplasticizer (kg/m3) 6.3 6.2 6.3

CaO-based Expansive Agent 35 35 \

Viscosity modifier (kg/m3) 4.2 4.1 4.3

SRA 4.0 \ \

(*) Blended Portland-limestone cement according to EN 197/1

Figure 5 shows the strength development with time of the three com-pared SCCs (CaO-SRA, only CaO and Plain). The strength of the expan-sive concretes was higher than that of the plain mix. This is probablydue to the consumption of a small part of mixing water caused by thetransformation of CaO into Ca(OH)2 which happens when the concreteis still in the plastic state and to the consequent reduction of the actualw/c.On the other hand, a slight decrease in the compressive strength of theSRA + CaO mix was recorded if compared to that of the CaO mix dueto the presence of SRA as experienced in [8].Although it was specified to protect the concrete surface for at least

three days (to assure a correct hydration of the concrete cover) shrink-age compensating concrete was designed in order to warranty a resid-ual restrained expansion of about 200 mm/m even in case of deficientcuring consisting in just 24 hours of protection by the formwork. Figure 6 shows the length change of the reinforced prismatic specimensmanufactured with the three different SCCs according to ASTM C 878.Specimens were not put under water for 7 days as specified in ASTM C878 test method but were protected with a plastic film for just 24 hours(to simulate the protection offered by the formwork) an then exposed tounsaturated air (60% R.H.) at 20°C.This curing condition was later introduced as “curing method B” in thelast version of the Italian standard UNI 8147 in addition to the “curingmethod A” previously specified, consisting in a total immersion inwater for 7 days as in ASTM C878. Actually, the curing method Bappears to be more realistic and similar to jobsite conditions. Even under these un-favourable conditions of curing, the CaO-SRAshrinkage-compensating concrete performed very well since therestrained expansion after 24 hours of protection with a plastic film wasas high as 560 mm/m and, even after 140 days of exposure to unsatu-rated air, a residual restrained expansion of about 250 mm/m wasrecorded. On the contrary, the conventional CaO-based shrinkage com-pensating concrete showed a lower initial expansion (at lest 320 mm/m)which completely disappeared after a week of exposure to air afterwhich, the concrete started to shrink. Obviously, the plain concrete showed the worst performance reaching arestrained shrinkage of about 550 mm/m after 60 days when somecracks appeared on the specimen surface.Comparing the behaviour of the CaO+SRA-based mix to that of the con-ventional shrinkage-compensating concrete, both the � and � effect ofFigure 6 can be detected.On the basis of the above results, the customer and the contractordecided to adopt the SRA+CaO-based shrinkage-compensating SCC

• Fig. 5- Strength development of three different SCCs. • Fig. 6- Length change with time of the three different SCCs.


age, of the various specimens manufactured. As expected, the specimens manufactured and cured at 20°C per-formed well showing a residual restrained expansion in the range of210÷280 mm/m after 28 days of exposure.The prolonged mixing (at the same temperature of 20°C) caused adecrease of the initial as well as in the residual expansion as reportedin [2].A little higher decrease was recorded in the expansion of the specimensmanufactured and kept at 30°C and cast after 5 minutes of mixing.Anyway the behaviour of these specimens can be considered accept-able.On the contrary, the combination of a high temperature of manufactur-ing and curing and a prolonged mixing cause a strong reduction in theinitial restrained expansion which was completely cancelled after justone week after which the concrete started to shrink. The problem was not eliminated by increasing the amount of expansiveagent up to 45 kg/m3 so that, being impossible to assure a transporta-tion time lower than 60 minutes, the contractor decided to delay thebegin of the main wall construction to the autumn and to stop it duringthe whole next summer.

3. PROPERTIES OF RECYCLED CONCRETES

The problem of recycling industrial wastes is of vital importance for asustainable progress in order to avoid disposals in the environments andpossibly to save resources for the next generations. Such a problem hasalready been faced in using fly ash, silica fume and blast furnace slag,all wastes coming from industries other than cement and concrete. During the last decade a similar problem has been found for wastescoming from the construction industry and from the concrete in partic-ular [9] [10]. These wastes can be recycled as aggregates for concretes with twoadvantages: – first, to save the environment specially in countries, like Nederlandand Belgium, where the available area to build is very limited;– second, to recycle this waste as aggregate specially in areas wherenatural or artificial aggregates are scarce. Therefore, concrete recycling, by using the readily available concrete asan aggregate source for new concrete or pavement subbase layers, isgaining importance because it protects natural resources and eliminatesthe need for disposals. Concrete recycling is a relatively simple process. It involves breaking,removing, and crushing the existing concrete into a material with aspecified size and quality.The quality of concrete with recycled aggregate depends on the quality

for the manufacturing of all the architectural concrete walls of MAXXI.Since it was the first time the contractor used an SCC, it was decided tocarry out several field tests, before starting with the manufacturing ofthe actual walls, in order to optimize all the casting procedures and testthe suitability of formwork. It was, then, a good chance to test on a realscale the effectiveness of the expansive technique.Two field tests were successfully carried out in March and April of 2004with no cracks formation in two long minor walls of the basement. A third test carried out in June in order to verify the behaviour of theexpansive concrete in the presence of high temperature failed sinceafter two weeks, some cracks appeared on the wall surface. The maxi-mum temperature during the casting operation was as high as 35°C andchecking the transport documents of the trucks mixer it was verifiedthat, because of the congested traffic of Rome, the time elapsed betweenthe starting of mixing, in the batching plant, and the casting of concreteinto the forms had been in the range of 60-90 minutes, notwithstandingthe batching plant were located near the jobsite.For this reason the cause of the failure was ascribed to a combinedeffect of the high temperature and of a too prolonged mixing time. Thishypothesis was confirmed by laboratory tests in which some ASTMC878 prismatic specimens were manufactured at 20°C (with raw mate-rials kept at 20°C for 24 hours before the use) whereas other similarspecimens were manufactured at 30°C (with raw materials kept at 30°Cfor 24 hours before the use). In both cases, some specimens were putinto the forms after 5 minutes of mixing whereas the others were kept inthe mixer (in movement) for 60 minutes before casting at the same tem-perature of manufacturing (20 or 30°C). After setting time (about 6hours) the specimens were demoulded and protected with a plastic filmtill 24 hours, at the same temperature of manufacturing (20 or 30°C).Successively, the specimens were exposed to unsaturated air (60%R.H.) at the temperature of manufacturing (20 or 30°C).Figure 7 shows the behavior, in terms of restrained expansion or shrink-

• Fig. 7- Restrained expansion or shrinkage in different manufacturing and curing condition.


of the recycled material used, the most important aspect being the ori-gin of the recycled material such as concrete or demolition, the latterincluding waste from brick walls and other type of rubbles. The crushedand sieved material, must be deprived by contaminating products suchas wood, paper, plastic, and bitumen. This recycled material can beused for pavement subbase layers.The process of recycling demolished concrete is based on four steps:

– selection of wastes;– crushing concrete blocks;– removing of contaminating products;– mixing with virgin aggregates.Reinforcing steel and other embedded items, if any, must be removed,and care must be taken to prevent contamination by other materials, suchas: asphalt, soil and clay balls, chlorides, glass, gypsum board, sealants,paper, plaster, wood, and roofing materials which can be troublesome.The mechanical plants where to recycle concrete structures are not verydifferent from those adopted to treat crushed virgin aggregates. If the material is devoted to concrete production, further crushing andsieving are needed before mixing it with virgin aggregate [11].The crushing characteristics of hardened concrete are similar to thoseof virgin rock and are not significantly affected by the quality of theoriginal concrete. Recycled aggregates can be expected to pass thesame tests required for conventional aggregates. The recycled concretecan be batched, mixed, transported, placed and compacted in the sameway as conventional concrete. Special care is necessary when usingrecycled fine aggregate. Only up to 10% to 20% recycled fine aggregateis beneficial. The aggregate should be tested at several substitutionrates to determine the optimal rate.

3.1 Properties of recycled fresh concretes

The amount of mixing water of the coarse recycled aggregate is about5% more with respect to that of virgin aggregate at given size. Thisvalue becomes as high as 15% when the recycled aggregate containsalso the fine fraction. This effect is due to the rough texture of the aggre-gate and the cement paste surrounding the recycled aggregate.However, the use of superplasticizers and mineral additions can com-pletely overcome this drawback [12].

3.2 Properties of recycled hardened concretes

Due to the higher porosity, related to the lower density, the recycledaggregates are responsible for the lower strength of the concrete withrespect to the concrete manufactured with virgin aggregates. Due to the lower rigidity, recycled aggregates are responsible for the

lower modulus of elasticity of the concrete with respect to the concretewith virgin aggregates. For the same reason, drying shrinkage and creepof concretes with coarse recycled aggregates are much higher (25-50%)with respect to the virgin aggregates. The difference can be still higherif also fine recycled aggregate is used.The permeability of a concrete with recycled aggregate is higher thanthat of the corresponding concrete at a given water-cement ratio. Again,the cement paste of the recycled aggregate is responsible for this draw-back because the cement paste is more porous and permeable of thevirgin stone.The frost-resistance of the concrete with recycled aggregate is stronglyreduced by the amount of fine fraction of the recycled aggregate.Therefore, in concrete exposed to freezing and thawing cycles the frac-tion of recycled aggregate smaller than 4 mm should be removed.The fine fraction of recycled concrete smaller than 4 mm is responsiblefor all the above mentioned limits of the concrete with respect to thecorresponding concrete manufactured with virgin aggregates. However,if the fine fraction is ground very finely (smaller than 0.1mm) it can beused advantageously in manufacturing self-compacting concrete asfiller to improve the cohesiveness of the concrete [13, 14].

4. USE OF BOTTOM ASH FROM MUNICIPAL SOLIDWASTES INCINERATORS

Mineral solids in form of fly ash and bottom ash are produced by burn-ing municipal solid wastes in incinerators (MSWI). Fly ash is negligi-ble and it is so chloride-rich that it cannot be used as mineral additionin cement-based mixtures for reinforced concrete structures. On the other hand, bottom ash is about 25% with respect to MSWI andits chloride content is negligible, so that it could be potentially used asmineral addition for manufacturing concrete mixtures. However, groundbottom ash (GBA) from MSWI does not perform as well as other miner-al additions (silica fume or fly ash produced by coal burning) due to thepresence of aluminium metal particles which react with the lime formedby the hydration of Portland cement and produce significant volume ofhydrogen in form of gas bubbles which strongly increase the porosity ofconcrete and reduce its strength.Due to this drawback, a new process was developed to completely sep-arate the aluminium metal particles through a mechanical removal ofmetals and a special wet grinding of bottom ashes. At the end of theprocess GBA was used as an aqueous slurry to replace Portlandcement. Some researches have actually shown the pozzolanic activity of groundMSWI bottom ashes showing their reactivity with lime or portlandcement clinker [15,16]. Nevertheless, no successful use of MSWI bot-


tom ashes as mineral addition in concrete has been reported, becauseof the side effects of this addition. According to Bertolini et al [17], themain side effect is related to the evolution of hydrogen gas after mixingdue to the presence of metallic aluminium. In the alkaline environmentproduced by the hydration of portland cement (pH around 13), corro-sion of some metals (mainly aluminium) produces a great amount ofgaseous hydrogen. After placing and compaction of concrete, this gas isentrapped in the fresh material, producing a network of bubbles thatleads to significant reduction in the strength and increase in the per-meability of the hardened concrete.The present report summarizes the results of a research [18] aimed atdeveloping suitable treatments to allow the use of MSWI bottom ashesas mineral additions for the production of structural concrete withoutthe evolution of hydrogen gas due to the presence of metallic alumini-um particles.Ground bottom ashes (GBA) from municipal solid waste incinerators(MSWI) were manufactured according to a new technology based on ahigh degree of separation of metals including the heavy ones, the wetgrinding process, and other specific technical solutions to completelyremove the aluminium metallic particles. At the end of the process, afluid slurry was obtained with particle size in the range of 1-5 mm. Bychanging the wet grinding time three GBA were produced with a meanparticle size of 5 mm, 3 mm and 1.7 mm.Compressive strength and durability measurements were carried out inconcretes where Portland cement was replaced by 20% of ground bot-

tom ashes from MSWI in comparison with concretes containing 20% ofcoal fly ash or 10% of silica fume.The performances of GBA with mean sizes of 3 and 5 mm were higherthan that of the coal fly ash particularly at 1-60 days. The finest groundbottom ash (with a mean size of 1.7 mm) performs as well as silica fumein terms of compressive strength, water permeability, chloride diffusionand CO2 penetration. These results appear in particular to be very interesting from a practi-cal point of view since it will be possible to manufacture big amounts ofa pozzolanic material as effective as silica fume (which is not availableand very expensive) in producing high-performance concrete in agree-ment with a sustainable progress for the re-use of waste materialsinstead of a their disappointing disposal.

5. CONCLUSIONS

The research on the progress of structural concrete in Italy is veryactive. Three sections of this activity have been illustrated in the pre-sent report:– shrinkage-compensating concrete for crack-free structures even inthe absence of a wet curing;– recycled concrete with special application of a fine powder materialdevoted to self-compacting concrete;– bottom ash from municipal waste incinerators as pozzolanic materialsfor high-performance concrete.

MET/ACI International Conference on Superplasticizer and Other

Chemical Admixtures, Sorrento, 2006, pp. 103-115.

[9] N. Kashino and Y. Ohama, RILEM International Symposium

Environmental Conscious Materials and Systems for Sustainable

Development, 2004, pp. 179-186.

[10] J.Zh. Xiao, J.B.Li, Ch. Zhang, “On relationship between the

mechanical properties of recycled aggregate concrete: An

overview”, Materials and Structures, 39, pp. 655-664, 2006.

[11] M. Bassan, M. Menegotto, G. Moriconi, “Precast Structural Con -

crete with Recycled Aggregates”, Proceedings, the 2nd fib Con -

gress, Naples, ID 6-19, 2006.

[12] V. Corinaldesi, G. Moriconi, “Role of Chemical and Mineral

Admixtures on Performance and Economics of Recycled-

Aggregate Concrete”, Proceedings of the Seventh CANMET/ACI

International Conference On Fly Ash, Silica Fume, Slag and

Natural Pozzolans in Concrete, Editor V.M. Malhotra, ACI SP

199, Vol. 2, pp. 869-884, 2004.

[13] V. Corinaldesi, G. Orlandi, G. Moriconi, “Self-compacting con-

crete incorporating recycled aggregate”, Innovations and De -

velopments in Concrete materials and Construction, Proceeding

of the International Conference, Editor Ravindra Dhir, Dundee,

UK, pp.9-11, 2002.

[14] V. Corinaldesi, F. Tittarelli, L. Coppola G. Moriconi, “Feasibility

and Performance of Recycled-Aggregate Concrete Containing

Fly Ash for Sustainable Building Development”, Proc. of the

Three-Day CANMET/ACI Int. Symposium On Sustainable

Development of Cement and Concrete, SP 202-11, pp. 161-180,

2001.

[15] K.A. Paine, R. K. Dhir, V.P.A. Doran, “Unprocessed and

processed incinerator bottom ash as a cement bound material” in

R.K.Dhir, T.D.Dyer, K.A. Paine (Eds.) “Use of incinerator ash”,

Proceedings of the International Symposium organized by the

Concrete Technology Unit, University of Dundee (UK) 20-


REFERENCES

[1] M. Collepardi, The New Concrete, pp. 347-362, Second Edition,

Ed. Tintoretto, Villorba, Italy, 2010.

[2] N.S. Berke, L. Li, M.C. Hicks and Bal, “Improving concrete per-

formance with Shrinkage-Reducing Admixtures”, 7th CAN-

MET/ACI International Conference on Superplasticizer and Other

Chemical Admixtures in Concrete, Berlin, Germany, Ed. V.M.

Malhotra, 2003, pp. 37-50.

[3] M. Collepardi, A. Borsoi, S. Collepardi, J.J. Ogoumah Olagot, R.

Troli, “Effects of Shrinkage-Reducing Admixture in Shrinkage

Compensating Concrete Under Non-Wet Curing Conditions”,

Cement and Concrete Composities, 6, 2005, pp. 704-708.

[4] C. Maltese, C. Pistolesi, A. Lolli, A. Bravo, T. Cerulli and D.

Salvioni, “Combined Effect of Expansive and Shrinkage

Reducing Admixtures to Obtain Stable and Durable Mortars”,

Cement and Concrete Research 12, 2005, pp. 2244-2251.

[5] C. Maltese, A. Lolli, C. Pistolesi, A. Bravo, and T. Cerulli,

“Combined effect of expansive and shrinkage reducing admixture

on microstructure of mortars and concretes”, Proceeding of the

International Conference on Durability of Concrete, Editor: V.M.

Malhotra, Montreal, 2006, pp. 781-796).

[6] S. Chatterji, “Mechanism of expansion of concrete due to the pres-

ence of dead burnt CaO and MgO”, Cement and Concrete

Research 25, 1995, pp. 51-56.

[7] F. Tittarelli, S. Monosi, G. Moriconi, R. Troli, “Effects of a shrink-

age compensating Admixture in Shrinkage-Compensating

Concretes”, presented in “9th ACI International Conference on

Superplasticizer and Other Chemical Admixtures”, Seville, 2009.

[8] M. Collepardi, R. Troli, M. Bressan, F. Liberatore, G. Sforza,

“Crack-Free Concrete for Outside Industrial Floors in the Absence

of Wet Curing and Contraction Joints”, Suppl. of 8th CAN-

lepardi, “MSWI ashes as mineral additions in concrete”, Cement

and Concrete Research, 34, 1899-1906, 2004.

[18] M. Collepardi, S. Collepardi, D. Ongaro, A. Quadrio Curzio, M.

Sammartino, “Concrete with bottom ash from municipal solid

wastes incinerators”, paper accepted for publication on “Second

International Conference on Sustainable Construction Materials

and Technologies (SCMT)”, Ancona, Italy, June 2010.

21/3/2000.

[16] A. Macias, E. Fernandez, S. Goñi, A. Guerrero, “Valorizacion de

las cenizas de inceneracion de residuos solidos urbanos en los

materiales de construccion”, Papel de los sectores cementero y

de la construcciòn en la gestiòn y reciclado de residuos, CSIC,

Madrid, 2001.

[17] L. Bertolini, M. Carsana, D. Cassago, A. Quadrio Curzio, M. Col -


88 RESEARCH - L’Aquila Earthquake

construction may undergo severe damage but must avoid collapse inorder to protect the lives of its occupants. The earthquake on April 6 oflast year was of an intensity comparable to that found in the newTechnical Regulations of Construction, and many buildings in L’Aquilahistoric center sustained damage compatible to those norms (the acce-lerograms recorded by instruments present in the area of the “seismiccrater” revealed a spectrum only slightly higher than that in the buil-ding plans).Therefore, one can attest that well-constructed masonry buildings, inother words those built with connections that allow the structure to actas a unified organism (clamping between orthogonal walls, juncturesbetween walls and floors) are able to sustain damage without manife-sting fragility collapse.This article will not deal with monumental structures, but with minormasonry ones, common in the many historic centers in and aroundL’Aquila, where severe damage and collapse was diffuse. Wishing totake into consideration the number of victims as a parameter of vulne-rability, outside of L’Aquila there were 97 victims in masonry buildings,compared to 14 victims in r.c. buildings (it should be said that in thesetowns the percentage of r.c. buildings is much lower than in L’Aquilaand the buildings were also lower).What are the factors that justify the different behavior between theapartment buildings in L’Aquila and the buildings in the surroundinghistoric centers? Effectively, there are two reasons: construction qualityand the presence of incongruous subsequent restoration.In the masonry buildings of the historic center of L’Aquila, most ofwhich were rebuilt after the tragic earthquake in 1703, a series of struc-tural modifications characteristic of the L’Aquila rules of thumb wereeasily recognizable: wooden beams (elements built into the wall thick-ness and connected externally by way of small tie rods), to improve con-nections between walls; the connection of roofs to the tops of the walls,utilizing external wooden posts. These rules were adopted even in thesmaller surrounding historic villages, but often with lesser constructionknow-how and lower quality building materials. Even after the earthquake in 1703, the historic centers of the AternoValley were struck by significant earthquakes, in particular, the one inAvezzano in 1915. In fact, the repairs and seismic reinforcements areeasy to discern (scarp walls, buttresses, and tie rods adjacent tomasonry walls) and in many cases, the partial reconstruction of colla-psed portions. These interventions often functioned well, but in othersthe vulnerability remained, proving once again how difficult it is toperform truly efficient seismic restorations and improvements. As far as recent restorations were concerned, (only in certain cases donefor consolidation purposes) it should be said that while L’Aquilaappears to be better preserved (due to both the large number of protec-

1. INTRODUCTION

This chapter summarizes the main findings of a chapter of the spe-cial issue of Progettazione Sismica devoted to L’Aquila earthquake

the author edited and which is about to appear. In particular, in the fol-lowing the papers of Carocci and Lagomarsino (2010) regardingmasonry buildings, Cosenza et al. (2010) for reinforced concrete buil-dings, Di Ludovico et al. (2010) about school buildings, Casarotti et al.(2010) for hospitals, Menegotto (2010), Faggiano et al. (2010) for indu-strial structures, Dolce et al. (2010) for lifelines, are reported.

2. MASONRY BUILDINGS

Are masonry buildings able to withstand a strong earthquake such asthe one that struck L’Aquila on the 6th of April, 2009? Is it possible torepair damaged buildings, guaranteeing an adequate safety level totheir inhabitants in an area with such a high level of seismic risk?These are the questions asked by researchers and government techni-cians, but especially by those people who lived through this tragic eventand long to see the restoration of historic centers and return to theirhomes but are also afraid.The answer to these questions must be complex and detailed becausethere were so many factors that influenced the seismic behavior ofAquilan constructions. In this earthquake, more than others, the effects of localized seismicamplification played an important role. If one analyzes the macro-sei-smic consequences of many of the historic centers in the Aterno Valley(south-east of Aquila), one will immediately observe that the villagesthat were struck the hardest (Onna and Villa Sant’Angelo, I=9-10) werenear other villages where the damage was limited: Onna (I=9-10) whichis only 1500m as the crow flies from Monticchio (I=6). Near the town ofVilla Sant’Angelo, the distance between the chief town (I=9) and the vil-lage of Tussillo (I=8) is less than 700 m. Even within the town ofL’Aquila, there were zones where the damage was clearly concentrated.Especially in L’Aquila, it was noted that most victims were found inreinforced concrete (r.c.) buildings (135 versus 52 in masonry buil-dings). In the historic center, damage to churches was severe, in somecases with extended collapse, but the increased vulnerability of thesestructures is well-documented by history (in the past, churches wereoften found to have the highest concentration of victims). Some severedamage was rather diffuse in apartment buildings, but did not lead tocollapse except for localized cases in loggias or stairwells; almostalways found in abandoned or poorly maintained buildings.Modern criteria for structural safety, based on tests of differing limit sta-tes (performance-based design), state that during a rare earthquake the

Damages of L’Aquila Earthquake

89RESEARCH - L’Aquila Earthquake

ted buildings as well as perhaps the number of uninhabited buildingsin certain areas), the smaller historic centers were subject to frequentchanges: enlargements, transformations, added floors, restored floorsand roofs. These interventions were often done with building materialsand techniques incompatible with the original structures: alterations inweight displacement, differing rigidity between elements, and dange-rous increases in mass. The final important question is if it is possible to repair severely dama-ged constructions and rebuild collapsed parts of aggregates with tradi-tional construction techniques (i.e. stone), eventually modified on thebasis of the experience of this earthquake. To this end, an examinationof construction damage to masonry is proposed based not only on theidentification of vulnerable areas, but also on resistance (anti-seismicregulations) as a means of truly learning something from the test of anearthquake.

2.1 Typological aspects of masonry buildings in the historic centers ofL’Aquila

The historic centers of the L’Aquila territory are principally made up ofsimply organized masonry buildings. Even the grouping within connec-ted buildings seems to be to simply follow the rules imposed by the oro-graphy of the terrain. Due to the areas extended (including seismic) history, constructionpeculiarities and specific characteristics for the arrangement of aggre-gates for each of these centers have been adopted, and deserve to betaken into consideration, but will not be dealt with in this article.Nevertheless, with reference to aggregates, one can affirm that theurban centers are characterized by smaller dimensioned blocks of buil-dings that develop along the morphology of the terrain where they werebuilt.In the centers located on mountainsides (for example: Fossa, Casentino,Tussillo, Castelnuovo) or those built on plateaus (for example: PoggioPicenze, Sant’Eusanio Forconese, Villa Sant’Angelo), both cases arecharacterized by sloping terrain. One can observe the two types of typi-cal aggregates: parallel blocks and blocks built orthogonally to theincline as well as a wide range of variants in between.Buildings were arranged according to the conformation of the terrainand the existing buildings, and the relation to seismic vulnerability canbe listed schematically: height differences between walls on oppositesides of the street; scaling of contiguous building fronts; number andheight of external exposed walls, and the placement of the building inthe aggregate.In the oldest centers where the building aggregates developed alongside streets, the aggregates are characterized by the continuity of the

street front and the contextual absence of building compactness in thedepth of the fabricated bodies (for example: San Demetrio Colle). Such generalized characteristics are largely identifiable in centersanalyzed herein, even if in our conscience it is not possible to estab lisha correlation between the type of building aggregate and the damagesustained. In any case, the observations relative to the scale of theaggregate are important not only for identifying where the main vulne-rability lies and differentiating it from that of single buildings, but espe-cially for establishing the project phases for restoration and reconstruc-tion.The diffusion of arches placed between facing walls along the streets isa repeated tendency in these villages subject to earthquakes. However,it is not possible to prove that the volume of overpasses (built onto theinhabitable surface area) though widely present in many of these cen-ters (for example: Santo Stefano di Sessanio), is the result of the evolu-tion of the arches, even if it is obvious that both tend to create a rathergood connection between the facing blocks and constrict wall move-ment where they are located. Street names such as Contrada dell’Arco,Chiassetto dell’Arco, Via sotto gli Archi, present in nearly all of the cen-ters we visited, indicate that this construction technique dates wayback. Buttresses, built-on scarp walls and tie rods should be counted amongthe pre-modern anti-seismic defenses that systematically characterizethe building range of these centers. Each technique can be identifiedby materials analysis, production, function and use after the manyearthquakes that have struck this area throughout history. In fact, manyof these elements have been utilized in restoration phases while in othercases, they seem to have become the rule of thumb for the partial or totalreconstruction of buildings.Evidence of the damage done by historic earthquakes is still often visi-ble, such as the often seen out of axis external walls of buildings, andthe presence of stone elements, often hewn after being taken from oldercollapsed buildings and reutilized in post quake reconstruction.. Given the close relationship that links these centers to the territory(agricultural/pastoral economy), the inhabited buildings still show tra-ces of their rural past with ground floors usually for storage or stalls andthe living spaces in the upper floors accessed by means of externalstaircases along a balcony. The structure of these stone stairs is usually surrounded by pillaredmasonry that tends to create volume towards the street front; in caseswhere the building has another floor, the additional volume is used as aloggia while access between the two inhabited upper levels is solved byway of an internal wooden stair.In some of the centers visited, the houses showed more “urban” cha-racteristics in how they were set up on the street front while at times

Gaetano ManfrediDepartment of Structural Engineering, Università degli Studi “Federico II”, Naples, Italy


maintaining the external stone stairway to access the inhabited upperfloor. In any case, a common peculiarity seemed to be the add-on ofadvancing volume on the front of the house for more living space.Along with the mono-celled buildings on the street front, which we havereferred to up till this point, larger and more complex structures existwithout being large enough to be considered real apartment buildings(with the exception of sporadic cases in the smallest villages). Thesewere derived mostly from the fusion of pre-existing smaller buildingswith the addition of an upper floor. The “stall-hayloft” is another very common type of building which com-pletes the overview of the minor architecture in these centers. Its func-tion determines its appearance, and though it is completely distinctfrom the living area, it is often placed contiguously to the house and the-refore inserted into the building mesh. The most important differencefrom a construction and structural point of view is that these construc-tions represent the largest part of the wall structure (where some wallsreach up to 10 m in length) with the presence of dilated wall light whichcorresponds to the systematic use of roof coverings connected to exter-nal walls.From our observations, it was possible to conclude that up until now, thewall thickness of the cell walls were adequate for the dimensions of thebuildings (3 levels above ground and usual light openings); in fact theseprove to be between 50 and 70 cm except at ground level or in rarecases where stone vaults are present where up to 1 m of thickness canbe found. Congruous tapering was also observed in upper floors; scantwall thickness was observed only on rare occasions, always in the pre-sence of more recent interventions.As far as horizontal diaphragms are concerned on the first floor (thedivision of lower floors used for storage or stalls and inhabited upperfloors), there were always barrel vaults with the generatrix perpendicu-lar to the façade. The construction technique calls for a mesh of roughhewn stone elements and the presence of compact buttresses.The upper floors usually have simple wooden flooring. However, brickcovered metal beams were frequently noted, certainly from substitu-tions in the last century (for example after the Avezzano earthquake).Metallic beams, hollow flat blocks covered by flooring were morerecent. In some cases, the beams were connected to bars or welded pla-tes and anchored to the masonry.Another type of horizontal diaphragm is made of thinly layered brickvaults (two thin layers of interwoven bricks) which have proven to bevery vulnerable; these vaults are also relatively recent (from the end ofthe 19th century to the first decades of the 20th century). The roof coverings are made of wood with the constant presence ofoverhangs made of dripstone or wooden elements. In buildings in aggre-gates, the ridge of the roof is parallel to the street front and may have

one pitch towards the street or two slopes with a central ridge (depen-ding on the configuration of the whole aggregate). The wooden beamsare always positioned without overhang; even in the case of buildingsplaced on corners or at the head of a block where the configuration ofthe pitch is a triangular pavilion shape (therefore with the presence oftwo directions of slope). The attention paid to not creating overhangingstructures and nevertheless the wooden elements were almost alwaysconnected to the walls. The attention to the way rooms were construc-ted for reducing seismic vulnerability is also observed in many smalldetails, such as the “light” eaves made from wooden and bamboo struc-tures and the balconies entirely made of wood. Such construction mea-sures were evidently realized in order to limit the lethal effects of dama-ge on the lives of the inhabitants. To conclude this chapter which focuses on the characteristics of thebuildings in minor centers of the L’Aquila area, it is necessary to men-tion two aspects that should be considered in the observation and eva-luation of damage: the techniques adopted in recent restorations andthe conservation state of the buildings. In relation to the first of the two aspects mentioned above, it should beobserved that as long as masonry walls were the dominant techniqueused for building, transformations on existing buildings took into consi-deration this vulnerability and often brought about seismic improve-ments within the limits of the techniques available, eliminating weak-nesses and introducing protective measures.Unfortunately, recent transformations appear mostly linked to damageafter the earthquake. The most common intervention is the substitutionof the original roof covering with a new one which generally followed thesame configuration of the preceding one, but was at times made from anr.c. structure with hollow or infill panels or with metal beams instead ofwooden ones. The result of these changes with regards to seismic vul-nerability was quite often negative.As far as the second aspect is concerned, it seems clear that the con-servation state of the building played an important role in whether ornot it was damaged, but also with reference to the buildings near it inthe aggregate. In fact, in cases where the adjacent buildings were poorlyconstructed due to many decades of abandon, there was less damagethan expected due to the stabilizing contribution of the contiguous cells.In some of the centers visited, the presence of many restoration workzones was observed (for example at Villa Sant’Angelo).As far as we could discern, in most cases the techniques these workerswere utilizing were far from those that should have been adopted for theknowledgeable recovery of historic heritage. On a positive note, also inview of the reflections in the next chapter, we also encountered that therestoration and reuse of abandoned masonry structures had been goingon for some time, and was probably linked to a search for available


floors. At times, the new building eliminated certain walls orthogonal tothe façade yet present in the lower floors when raising the height, thusaltering the passage through transverse walls (and generally rotating theorientation of the floors).Figure 2 shows the case of a building where one of the walls orthogonalto the façade was absent on the top floor, yet traces can be observed onlower floors. The excessive distance between the transversal walls ren-dered overturning nearly inevitable in the case of seism.Such construction defects, observed on many occasions, can have easilybeen revealed in the buildings in the historic centers and the vulnera-bility derived from such configurations found and resolved preventati-vely.

living space close to the city center but also due to a slightly higher tou-rism interest (also foreign).

2.2 Damage observations: vulnerability and resistance

The description of the damage observed will be discussed by topicaccording to the most important building aspects (at times re-examiningthe topics mentioned above) and by referring to the principal damageand collapse mechanisms.Subdivision is necessary in order to shed light on the probable causesthat facilitated or more aptly limited damage. In this way, characteri-stics of vulnerability will be highlighted as well as the strengths whichshould lead to further reflection for future reconstruction.

2.2.1 Structural organization of buildings and position in the aggregate As far as the placement of the building in the aggregate is concerned, ithas been revealed that the configuration of the corner or the head of theblock has proven to be the most disadvantageous as widely noted befo-re. In Figure 1, we can observe three buildings each placed at the headof a block. All of them sustained the collapse of the front wall and agreat portion of the two side walls. Good organization and regularity ofthe vertical load elements of the building represents a point of strengthwith regards to seismic action. The orientation of orthogonal walls to thefaçade, the placement and quantity of openings and the connection withthe horizontal elements determines the greater or lesser flexional trim(vertical and horizontal) of the walls. Irregularity or changes introducedin the configuration of the whole often proves fatal during an earth-quake.One often noted case is that of the buildings created by the fusion ofpre-existing contiguous buildings and the later addition of one or more

Fig. 1- Tempera: collapse of buildings placed at the head of a linear aggregate. Fig. 3- Poggio Picenze: local collapse due to discontinuity.

Fig. 2- L’Aquila: the absence of transverse walls on the upper floor of the building.

The vulnerability derived from the absence of clamping between wallsof buildings built at different times was confirmed (growth phases andevolution of the buildings in an aggregate). Figure 3 shows the collapseof the portion of a wall due to construction discontinuity by the simpleplacement of a wall next to the pre-existing corner.

Beyond typical growth or urban congestion, many violations of buildingcodes were found in the transformations of aggregates in historic cen-ters. In other words, demolition and reconstruction of weight-bearingwalls in different positions in order to obtain small extensions or recon-figurations of the block produced negative alterations of cells. In thesecases, the original clamping was lost and nearly impossible to restoreon the newly placed walls, and thereby eliminated connecting elements.In the case shown in Figure 4, a cut off wooden tie rod is visible rightinside the internal wall, probably during restoration of the building orits apartments. This alteration, which eliminated an anti-seismic con-struction norm put in place by previous craftsmen certainly contributedto the collapse that was verified in that building unit.

the house, and consequently, the use of spaced trusses are necessary(Figure 5-6). From a construction point of view, it is clear that the con-nections between roof and walls were put into place at the time of ori-ginal construction as demonstrated by the position of lintels whichnecessarily implies their placement before mounting the boards and thesubsequent roof covering. Finally, it should be noted that the use of con-necting trusses to the masonry walls by means of wooden lintels wasalso consistently found in churches in the L’Aquila region, and there -fore one can assume that this technique was generally associated withconstruction configurations with great light.Connections by means of wooden lintels were also observed in largerand more important buildings like that in Figure 7 which gives anexample of the positive contribution offered by the roof covering. Thecorner building shows systematic tie rods between floors and the top of


Fig. 4- Villa Sant’Angelo: violations: the external wooden connection to the metallic tie rod was cut during fusion when thewall direction was altered.

Fig. 5- Villa Sant’Angelo: a wooden truss leans on a wooden lintel.

Fig. 6- Casentino: masonry cyma of a hayloft and the truss touching the wooden lintel (note how the internal connection isthanks to a board nailed to the actual lintel).

2.2.2 Roof CoveringsAs already mentioned, wooden roof structures generally do notoverhang, demonstrate well-placed orientation, and are usually well-connected to the masonry walls evidenced by metallic or wooden con-nections visible from the external walls. Such connections are associa-ted to the very common presence of pavilion type roof coverings or thetriangular portion of the pavilion roof, where the presence of ridge raf-ters or slope necessitates the organization of the mesh of the wood to eli-minate thrust. In roofs whose main direction is limited by masonrywalls they are posed upon, it was not rare to find that an effective con-nection to the wall rims contributed to limiting the damage and protru-sion of the external walls. The habit of connecting, often by way of wooden lintels, roofs tomasonry walls seems to be linked to the dilated dimensions of the cells;in fact this modality is constantly found in uninhabited spaces such asstalls-haylofts. In the latter, the floor area is usually larger than that of

the external walls without involving the architrave structure of the ope-nings on the highest floor (Figure 9). In both cases, damage may havebeen caused by movement relative to the placement of wooden elementsand the effect of the addition of wall openings which weakened wallstrength.Localized damage in the tympanum walls (which is one of the mostcommonly found types of limited damage observed in the L’Aquila cen-ters) may be more or less extended depending on the surrounding con-ditions, where the tympanum wall was placed in the aggregate (differingheight to the contiguous buildings), or in relation to the quality of themasonry pattern at the top of the wall.Extensive damage involving the top of the external walls may be attri-buted to where the wooden mesh elements (primary and secondary)were not interconnected which led to greater displacement at the toppart of the wall during the quake.Localized collapse to a portion of the masonry cyma illustrated inFigure 10 is presumably due to thrust in the rafters of the slope of thepavilion roof. Localized cracks are also visible near the corners, per -haps caused by horizontal displacement transferred to the walls by theangular rafters of the pitch. One can also observe the collapse of theobstruction of an arched opening.It was dramatic how many cases of damage were unmistakably causedby the substitution of traditional wooden rafters with structures whichalthough appeared to imitate the original had vast differences in weightand rigidity.Both in the cases of heavy stringcourses as in those where the weightis due to infill roof panels, the effect produced by the seism was fatal.The external walls below were seriously damaged and the roof cove-

the wall put into place when the masonry walls were elevated. One canalso observe the presence of wooden tie rods which are connected to thetrusses which create the pavilion roof structure (the presence of a tie rodmay indicate the absence of an orthogonal wall to the façade). In anycase, it is interesting to note how the building has cracked in the floorwith walls (second mode mechanism) while initial cracking is entirelyabsent on external walls.


Fig. 7- Poggio Picenze: a small corner building of a block built on a slope.

Fig. 9. Villa Sant’Angelo: diffused collapse of the masonry cyma without involving the window openings.

Fig. 8- Poggio Picenze: local collapse of tympanum walls due to holes punched in the rafters.

Below, the Figures demonstrate how damage can be attributed to a lackor modification of the roof structure. Damage caused by the poor qua-lity or advanced state of decay of the wooden elements of the roof wasvery common. This was due to the lack of connection to the masonry(along with the scarce conditions of placement), the lack of connectionbetween main and secondary elements of the roof covering, or the badconditions of the wooden boards (which if in good condition guaranteea very limited bracing effect). In these cases, local collapse was recorded in the tympanum of the cellwalls (Figure 8), in other words extended collapse to the fascia above

An analogous effect seems to have resulted from the presence of an r.c.floor posed in order to consolidate and increase the rigidity of the origi-nal floor in metallic beams with thin layered brick vaults (Figure 14).In this case, the wall collapsed from the base to the summit with theexclusion of the r.c. stringcourse and the roof structure which remainedin position.The head of the building had a rounded corner and the collapse of the


ring structure remained on top of what remained of the building(Figure 11-12).

2.2.3 Horizontal diaphragms and vaultsIn the case of floors, it was generally observed that modifications playeda negative role in seismic behavior; recent substitution interventions oforiginal horizontal structures were quite diffused, in particular in thelargest centers, and they seem to have led to the type of damage discus-sed below. The use of floors with metallic beams and brick (small brick arches orthin hollow flat blocks) was common for nearly all of the last centuryand may improve the overall behavior in some cases when well-made.However, such interventions require localized clamping to the masonry(of less impact than that necessary for the insertion of r.c. stringcourses)and in a certain increase in the vertical load on the masonry. Theseloads inevitably influence one of the masonry parameters favoringdamage to the transverse connections in the wall and phenomena oflocalized instability.In Figure 13, observe how the portion of the wall collapsed from theground floor up to the height of the lintel of the window on the top floor.The collapsed portion was limited to the vertical alignment of twocolumns of puncturing and the nesting seems to have been favored bythe presence of an r.c. floor positioned over the original thin brick vaultand by the presence of the addition of a staircase leading to the attic.

Fig. 13- L’Aquila: a second floor and a staircase in r.c. were placed on top of the original one.

Fig. 10- Vallecupa: localized collapse of a portion of the masonry cyma.

Fig. 11- Tempera: the heavy substituted roof caused the crumbling of the walls of the floor beneath it.

Fig. 12- Villa Sant’Angelo: the r.c. roof with infill panels induced the collapse of the masonry walls beneath it.

mon horizontal structure for most of the buildings. It is interesting toobserve that in the range of current seismic damage, the vaults reactedwell. In fact, even in the cases involved in extensive collapse, the vaultsof the ground floors appeared intact (Figure 16). This is certainly due tothe correct dimensions of the masonry piers as well as the good place-ment of the vault within the complete configuration of the building andthe whole aggregate.

walls may have been determined by the rigidity of the floor panelswhich formed a diagonal point which tended to expel at the corner.The case documented in Figure 15, merits some attention due to thefact of its common reccurrence within the complete range of damage.The deformation localized in correspondence to the size of the horizon-tal diaphragms, absent at the top of the external wall, may indicateexcessive floor weight and/or the lack of any connection between thehorizontal diaphragms such as a barrel vault or flooring built on beamsparallel to the façade.


Fig. 14- Paganica: a thick cement layer for rigidity was posed on the horizontal diaphragm.

Fig. 15- Poggio Picenze: evident deformation of the masonry walls of the facade with the worst damage at the level of thehorizontal diaphragm.

Fig. 16- L’Aquila: the vaults remained intact notwithstanding the collapse of the facade and the upper horizontal diaphragm.

As previously mentioned, the use of vaulted structures for horizontaldiaphragms is rather common in luxury buildings as well in minor onesin the Aquilan centers.In the latter, the vaults are generally in stone with thickness varyingbetween 20 and 30 cm, while in the city of L’Aquila, there were predo-minantly brick head vaults (12 cm). Barrel vaults create the most com-

Fig. 17- Paganica: collapse of two thinly laid brick vaults (horizontal diaphragm on the first floor and a false ceiling on thesecond floor).

Both in the role of horizontal diaphragms as well as in false ceilings forinhabited attic spaces, the brick vaults were not always spared fromdamage (Figure 17). This fact is easily proven both in smaller con-structions as well as in important and monumental buildings (the dama-ge is visible to these types of structures in the upper floors of many buil-dings in the historic center of L’Aquila). To the contrary of stone vaults,the reduced thickness of the brick vaults rendered them more sensitiveto even the smallest movements of the imposing walls. Obviously, for anaccurate evaluation of vulnerability of this type, the many cases where

these vaults collapsed due to the fall of the roof covering structures orsingle elements placed above them should be excluded.

2.2.4 Out of plane and in plane response in masonry wallsThe topic of masonry quality in the Aquilan territory is the center ofmuch debate, and not only technical, due to its relevance for interpre-ting damage but also and most importantly in the choices that must bemade for reconstruction. It is the author’s opinion that judgment on thequality of masonry cannot be expressed in a univocal manner as of yet. However, one can certainly attest that most of the building patrimonywas built based on earthquake knowledge, in particular for the syste-matic use of clamping and tie rods between walls (besides the previou-sly mentioned use in the connections of non-thrusting roof coverings).The organization of construction systems and the realization of masonrytechniques in many cases calls for the presence of tie rods put in placewhile the wall is raised. The construction of such tie rods is easily visi-ble today by the mass of partial collapse dating back to the reconstruc-tion after the earthquake in 1703, when the technique now used for overtwo centuries was first put into use. The clamping consists in the pla-cement of a wooden element built inside the wall and connected at theextremities by an iron plate and nails (Figure 18) and then anchoredexternally to the corner by way of a small tie rod (Figure 19). It is veryefficient as long as the wooden element is not placed under too muchtension causing weakness nears the nails.Obviously, the necessity of balancing costs and the availability of mate-rials called for the elaboration of variations which were not always effi-cient (Figure 20).More often than not, the progressive deterioration of these elements wasin correspondence to connections between the wood and the iron plate,especially in abandoned or poorly maintained buildings.Generally speaking, the limited presence of out of plane displacement

within the vast array of damage found can be attributed to the systema-tic use of tie rods in building walls. Nevertheless, it was possible to encounter certain types of out of planedisplacement after this earthquake. Figures 21 and 22 show crackswhich caused detachment from the façade. Both were caused by poorclamping at the corners and involved of a portion of the orthogonalwalls. In both cases, the portion of the isolated façade was limited to theheight of the horizontal diaphragm.Figure 23 illustrates a case of the start of overturning of a wall portionrelative to the two floors laid over the ground floor; here the out of planedisplacement seems to have been caused by a lack of clamping with themasonry adjacent to the contiguous building and a lack of efficient con-nections at the level of the second floor.Also damage ascribed to localized displacement of portions of themasonry cyma were common, both in the form of fractures (Figure 24)


Fig. 18- San Demetrio: metallic tie rod with a nail connecting the rod to the wooden beam.

Fig. 19- L’Aquila: an elegant 17th century tie rod connects to a wooden beam placed within the masonry thickness (note theadjacent strengthening tie rod, inserted after construction on the internal side of the wall).

Fig. 20- Villa Sant’Angelo: a poor example of wooden beams placed in the middle of the wall thickness without externalclamping.

as well as the collapse of limited portions due to cracking near openings(Figure 25).In plane damage to walls is present in all of its characteristic variants.In fact the form and placement of the shearing varied according to theefficiency of the connections between the elements it was built with andthe quality of the masonry pattern.Figures 26 and 27 illustrate two cases; in the first, the inclining fractu-res cross the external wall, traveling indifferently to the horizontaldiaphragms; while in the second, the fracturing is concentrated in themasonry piers between the openings of the first floor.These situations are found wherever there was a systematic presence ofclamping or for more recent buildings the presence of stringcourses. Regarding the validity of anti-seismic regulations, which had beenadopted in most of these buildings, one can only say that in cases wherethe workmen had the technical capacity to follow the rules of good con-


Fig. 21- Castelnuovo: detachment of a wall caused by defective clamping to the lateral wall.

Fig. 22- Paganica: overturning which involved a portion of the orthogonal wall.

Fig. 24- San Demetrio: localized fractures above the lintel of an opening.

Fig. 25- Paganica: collapse of a portion of a wall over the lintel.

Fig. 23- Tempera: overturning of the top portion of an external wall.

the necessary clamping to the pre-existing masonry; a defect which infact greatly limited its seismic resistance.As previously mentioned, the traditional connecting systems betweenconverging masonry walls are the most effective: both those put intoplace at the time of construction (clamping at the corners, anchoredwooden beams) as well as later strengthening interventions or repara-tions (tie rods placed in walls). The limited presence of collapse due tooverturning of the façade (compared to damage caused by major inplane wall displacement) leads one to believe that these systematic con-necting measures played an essential role in conservation (Figure 29).In this sense, the diffuse presence of shear fractures in masonry wallsshould not be interpreted as a demonstration of poor masonry quality,

struction and utilized these norms, such measures functioned to limit ormodify damage mechanisms within the limits of their efficacy. Instead,Figure 28 shows the frequent case of a buttress added to a wall without


Fig. 26- Villa Sant’Angelo: shearing which crosses the entire external wall.

Fig. 27- Sant’Eusanio Forconese: shear cracks limited to the masonry piers. Fig. 29- L’Aquila: a building where an overturning mechanism initiated but was effectively contrasted by the connections atthe summit thereby causing in plane resistance behavior of the walls.

Fig. 28- Castelnuovo: masonry buttress built without clamping to the pre-existing wall.

but as a positive consequence of the anti-seismic measures which pro-duced a box-like behavior, impeding out of plane mechanisms. In awell-built stone wall, the formation of even deep shearing followingsuch a violent earthquake is inevitable and allows the activation ofsignificant dissipative capacities, limiting the risk of collapse.Over all, much of the damage observed can certainly be attributed toinsufficient quality of the masonry pattern, whether in ruinous collapseor the loss of only the external façade. In general, the root of the pro-blem is a lack of building quality: bad equipment for the façade, poorquality cement (and its excessive proportion to stone elements), absen-ce of transverse connections between walls. Such cases are commonlycaused minor damage to adjacent buildings.Figure 30 illustrates the case of a building where the entire upper floorcollapsed. One can observe how the seismic activity disarranged thewalls to such a point that it crumbled to the ground in rubble notwith-standing the light roof covering (and therefore not able to induce theshearing particularly prevalent in upper floors). On the contrary, it


Fig. 30- Castelnuovo: total collapse of the top floor due to defective masonry quality.

Fig. 31- Villa Sant’Angelo: collapse caused by the lack of transverse clamping in the masonry mesh.

Fig. 32- Casentino: detachment of hewn stone elements around an opening.

Fig. 33- L’Aquila: localized collapse around the borders of the openings.

Fig. 34- Vallecupa: a chimney exposed by the seism.

should be noted that the external wall of the building to the left was leftwhole. In Figure 31, the very common case of the collapse of only the externalwall is shown. In these cases, the modality of damage declared itscause: the lack of transverse compactness of the masonry panels. Smallor badly placed stone elements, lack of regularity in the pattern andclamping between the facing sections rendered the walls vulnerable tohorizontal displacement and set off autonomous behavior in both wallsthat then very often led to collapse of this kind.The collapse of a single wall occurred more frequently where r.c. string-

courses were found at the top of a building, in particular when the enti-re roof covering was heavy and rigid. The detachment of the wall at thelevel of the stringcourse is linked to the local increase of compressionwhich originates in out of plane flexion from the condition and blockingvertical displacement (caused by the flexional rigidity of the roof cove-ring).

2.2.5 Localized damage Always on the subject of masonry walls, localized damage was found tostructural and nonstructural elements, justifiable in a situation wherethe force of the seism is extremely violent. In fact, even in cases where

the masonry buildings behaved well, it was natural that there was somesmall specific vulnerability.A frequent case observed was the constructive vulnerability of hewnstone above doorways and windows. The hewn stone elements used toborder openings were rarely clamped to the masonry walls within thewall thickness which inevitably led to their detachment (Figures 32 and33). This glaring omission to the attention given to seismic vulnerabi-lity in local anti-seismic construction deserves to be examined moreclosely. The problem was probably overlooked considering that suchelements do not determine the stability of the building as a whole.Further damage was found in consequence to local irregularities ofvarious types, such as continuity solutions which weaken masonry wallslocally, or the presence of wide parameter walls with no structural func-tion for the building.In the case shown in Figure 34, one can observe the collapse of the clo-sure of a chimney along the external wall. Figure 35 shows the collapseof a portion of the top of a wall characterized by scarce thickness, sinceit did not bear the load of the roof and was conceived as a simple lightclosure.

3. REINFORCED CONCRETE BUILDINGS

3.1 L’Aquila reinforced concrete building stock

ISTAT 2001 data, representing the official source for information about


Fig. 36- 2001 census ISTAT data for L’Aquila town: (a) building typology, (b) ageof construction, (c) storey number

Fig. 35- San Pio alle Camere: localized collapse of the top portion due to scarce thickness.

building stock in Italy, indicate in L’Aquila city a 24% of reinforcedconcrete structures, 68% of masonry structures and an 8% of structu-res whose typology is not identified (Figure 36a). Data that identify theage of construction of the buildings (Figure 36b) indicate that 55% ofthe whole patrimony was realized after 1945. Considering low inciden-ce of RC structures on the total it is possible to infer that after 1945 newmasonry structures were still realized and that RC structures numberincreased gradually over years.From distribution of storeys number (Figure 36c) it can be observed thatonly 5% of the buildings have more than three storeys. Assuming thatall buildings with more than three storeys are RC structures, it is stillpossible to infer that the great majority of L’Aquila RC buildings haveno more than three stories.Assuming an interstorey height that can vary between 3.0 and 3.5meters, period approximate formulation given by the Italian Code,referred to RC frame structures, gives for three storeys buildings a fun-damental period equal to 0.4 seconds.In the following section, when comparing different Italian Code spectralshapes, it will be possible to focus on period value ranges minor orequal to 0.4 seconds, thus leading to a comparison between constantacceleration parts of the spectra.

3.2 Structural and non structural damages

In this section main structural and nonstructural damages to RC struc-tures after L’Aquila earthquake are presented. Figuregraphic documen-tation was produced in the following few days after the 6th April 2009mainshock. Generally speaking, damages to structural elements are notso frequent and they seldom involve the whole structural system, on theother hand damages to nonstructural elements such as internal andexternal infill panels involved the main part of RC structures inL’Aquila.

3.2.1 Columns and wallsMain structural damages involving RC columns can be easily interpre-ted as failures caused by mechanisms which capacity design principlescan avoid or at least limit.It is worth to observe that during an earthquake columns put up with

a b c

high flexural and shear demand. Maximum flexural demand joined withaxial force produced by gravity loads and seismic loads are located atthe end of the element; in these zones (critical regions) rotational duc-tility demand can concentrate. Therefore, it is necessary to provide anadequate rotational capacity and to avoid buckling phenomena on com-pressed longitudinal reinforcements.Actual Italian seismic code provides prescriptions aimed at increasingsection rotational capacity: upper limit on longitudinal reinforcementpercentage, fixed the flexural resistance of the section, leads to a higherultimate section curvature; a proper spacing between hoops and cross-tie presence give, by a more efficient confinement action on concrete,an additional increase in section deformation capacity. Additionally, aproper hoop spacing avoids buckling phenomena in longitudinalreinforcement or at least fixes an acceptable upper bound limit forwhich this phenomena occur.On the other hand, prescriptions and structural detailing presentedabove are typical of modern design concepts that in Italy appeared forthe first time in 1997 with explicative document to the 1996 code butwere adequately ruled only in 2003 with OPCM 3274, and finally offi-cially adopted in 2008 by DM 14/01/2008 and its subsequent explica-tive document.So, according to codes that were in force before 1997, it is possible tofind RC columns with longitudinal reinforcement percentage thatexceeds 4% limit, section dimensions not conforming to actual limita-tions, hoops with a not sufficiently thick space (15-20 cm) and closedwith 90° hooks.Figure 37(a) presents a corner column of a RC building in L’Aquilahistorical centre, probably realized between 1950 and 1960, wheredamages occurred at the bottom end section of the element. Presence ofsmooth bars and small diameter of the hoops (6 mm) closed with 90°hooks can be observed, but most significantly the absence of any tran-sversal reinforcement in the first 30.40 cm of the elements immediatelyadjacent to the beam-column joint region.


Fig. 37- Column with smooth bars and poor transversal reinforcement (a); damage to a column due to axial - bendinginteraction (b).

a

Fig. 38- Shear failures of rectangular (a) and circular (b) columns.

Figure 37(b) reports a circular column belonging to a building in theresidential zone of Pettino, realized during ’80s, which shows a typicaldamage due to axial force and bending moment; concrete cover wasspalled due to high compression strain and longitudinal bar buckling.In this case too hoop spacing is not thick enough but probably designedin perfect accordance with code prescriptions at the age of construction.In analogy, shear demand can produce brittle failures with an outstan-ding dissipative capacity reduction of the column. Minimizing shearresistance mechanism to transversal reinforcement spacing only can benot exhaustive. In modern design rules shear design has to involvecapacity design principles such as fixing a proper hierarchy betweenbrittle and ductile failure mechanisms, that is shear and bending beha-vior of the element.In order to prevent brittle failures during post-elastic behavior, sheardemand has to be evaluated based on maximum flexural demand of theelement. When this criterion is applied to column, it consists in a rota-tional equilibrium, obtaining shear demand by the ratio between thesum of the bending moments at the end sections and the total height ofthe column. It is possible to prevent a brittle failure occurrenceapplying an amplifying coefficient to the shear demand. When sheardemand is evaluated, possible local interaction with adjacent infills isconsidered; in fact when masonry panels do not completely fill the RCbay, evaluation of shear demand is based on the height of the columnsubtracting infill panel height.Adopting proper capacity models it is possible to take into account con-crete degrading resistance mechanisms due to ductility cyclic demand.These prescriptions are provided by the Italian code since 2003, befo-re OPCM came into force shear design of columns used to be madeassuming shear demand from linear analyses; this procedure can easilylead to shear capacity underdesigned respect to flexural capacity; so nocontrol of the failure mechanism can be applied and a priori the failu-re mechanism can be either brittle or ductile.All above considerations can be confirmed by column damages repor-

b a b

ted in Figure 38.Considering the rectangular column in Figure 38(a), whose section isprobably (30x100) cm2, belonging to a building realized in 80’s, a shearfailure involving top end section is evident. Transversal reinforcementhas a hoop spacing approximately equal to 15-20 cm, that is completelyunderdesigned respect to column section dimension and consequentlyrespect to the section inertia, leading to a premature shear failure of theelement. The brittle failure mechanism is to be noted, underlined by thecrushing of the concrete within the reinforcements. Third and fourthhoops from the top end of the element that are completely opened.Figure 38(b) shows shear failure of a circular column characterized bya 30 cm diameter; in this case too it is possible to detect the hoop spa-cing not thick enough, that leads to diagonal cracking typical of shearmechanisms and to instability of longitudinal bars in the column.In order to highlight the non secondary role played by column – infillinteraction in determining shear failures in the elements, Figure 39 pre-sents some damages to columns. In particular in Figure 39(a) it is pos-sible to recognize the brittle failure in the column due to the local inte-raction with the concrete infill that partially covers the bay frame get-ting to 1/3 of the total height of the column. Partial infilling effectivelyinteracting with the column reduces the effective height of the element,producing a higher shear demand that exceeds column shear capacity.This kind of phenomenon involves all of the columns that interact withthe concrete partial infilling.Figure 39(b) shows an example of buildings with a partly below groundlevel. This basement level is characterized by walls, often realized withconcrete, aimed at a retaining function of the adjacent embankment;concrete wall height is limited respect to column height to allow the rea-lization of windows. This structural solution leads to a strict reductionof column effective height with a consequent increase in shear demand.Moreover shear span decreasing of the element can modify shear spanratio up to a modification to a squat column behavior. This situation isnot of secondary importance, because shear resistant mechanism of a

squat column is different respect to the typical behavior of a slenderelement. This is why, if local interaction between the column and theconcrete wall is not taken into account, a premature brittle failure dueto concrete excessive compression can more often occur.Even columns belonging to staircases can show brittle failures. Mostcommon staircase typologies, generally, are characterized by disconti-nuity elements in the regular RC frame scheme composed by beamsand columns. In fact, on a side staircase is composed by inclined axiselements (beam or slab), on the other side squat columns are created bythe intersection of inclined axis element with the column.Staircase elements considerably contribute to lateral stiffness of thewhole structural system due to axial stiffness of inclined axis elementsand to higher lateral stiffness of squat column. This contribution is sim-ply estimable via linear analyses. On the other hand, staircase elementsare characterized by higher shear demand that can lead to brittle failu-re mechanisms.Figure 40 shows a staircase composed by inclined axis beams, in par-ticular the squat column in the corner, due to intersection with thebeam, is characterized by a typical shear failure, due to an unfavorableshear demand capacity ratio. Poor transversal reinforcement in terms ofspacing and hoop diameter is to be noted.Shear failures characterized reinforced concrete wall performances too.


Fig. 39- Shear failures of column adjacent to partial infilling panels (a), shear failures of squat column adjacent to basementlevel concrete walls (b).

a b

Fig. 41- Failure in reinforced concrete walls.

Fig. 40- Shear failure in squat columns of the staircase.


Fig. 42- Joint failure with evident longitudinal bar buckling (a), diagonal cracking failure in concrete joint panel (b).

a b

As an example, Figure 41 reports damages detected in two reinforcedconcrete walls, respectively characterized by two different shape fac-tors; damages consist of a spread diagonal cracking. Low longitudinaland transversal reinforcement percentage can be observed, especially ifcompared with minimum values prescribed by actual design code.

3.2.2 Beam - column jointsBeam-column joints can completely modify structural complex beha-vior and their failure should be necessarily avoided in a proper seismicdesign, considering that these element are characterized by brittle fai-lure mechanisms. In these zones, geometrically very small, demandfrom beams and columns is concentrated and concrete panel with lon-gitudinal bar is subjected to high gradients of shear and flexuraldemand. Beam-column joints influence structural behavior in terms ofboth ductility when concrete cracking and bar sliding occur and resi-stance when brittle failures occur.Failure mechanisms of joints are principally governed by shear andbond mechanisms. In fact force distribution which allows shear andmoment transfer produces diagonal cracking and consequently joint fai-lure due to diagonal compression in the concrete can occur, producinga reduction in strength and stiffness in the connection. Cyclic degradingof bond mechanism, on one hand, produces a reduction in bending resi-stance and in the ductility of the elements meeting in the joint; on theother hand an increase in lateral deformation of the level is detected.Therefore, capacity design rules essentially aimed at avoiding brittlefailure mechanisms point to shear failure prevention in joints by meansof design rules and proper reinforcement detailing. In fact if the jointcollapses, a strict limitation results in resistance and deformation capa-city of the adjacent structural elements.Generally speaking, joint design is defined by the condition of a diago-nal stress induced by the elements meeting in the joint not exceedingallowable concrete compressive stress. Furthermore, in order to keepstructural continuity when concrete cracking occurs, a proper transver-sal reinforcement along the whole element should be provided.Transversal reinforcement allows to transfer stresses even if concretecracking phase has been overtaken, by means of a strut and tie mecha-nism that can develop if longitudinal, transversal reinforcement andconcrete struts contribute to a truss formation. By these design pre-scription in the joint a ductile mechanism in beam elements can deve-lops, avoiding a brittle failure in the joint.Joint design rules appeared in the Italian design prescriptions only in2003, in fact in 1997 explicative document to 1996 code a transversalreinforcement in joints at least equal to the hoop spacing in the adja-cent columns was simply necessary.It can be gathered that all structures realized before 1996 are charac-

terized by beam-column joints without transversal reinforcement. Inthis kind of situation, when cracking occurs no truss mechanism candevelop, leading to a strict reduction in strength capacity of the joint.Considerations presented above are confirmed by damages observed inRC structures after 6th April 2009 earthquake. Figure 42(a) shows anexternal beam-column joint, characterized by an extensive cracking inconcrete belonging to the joint panel. The absence of transversalreinforcement in the joint, probably because of the vertical componentof the seismic action, led to local buckling of the longitudinal bars thatconsequently brings to concrete cover spalling. It is worth to observethat the absence of a proper transversal reinforcement in the joint con-ducted to a loss of anchorage in beam longitudinal reinforcement.Figure 42(b) shows a typical diagonal cracking failure in concrete panelbelonging to an external joint. Crack begins at the intersection betweenjoint and upper column and ends at the intersection between joint andlower column producing the loss of monolithic connection. Hoopsabsence, in this situation too, leads to a buckling in the external longi-tudinal bar and it involves lower column not provided of transversalreinforcement in the first 30-40 cm.Other remarkable aspect in RC damage observation after L’Aquilaearthquake is a peculiar loss of connection at joint-lower column inter-face, probably due to a not prepared cast surface that can lead to a shearfriction failure. Generally, there are no code prescriptions, nor Italian orinternational, providing a shear friction verification at joint-columninterface, because some execution details, such us preparation of thecast surface and a proper diffusion of longitudinal reinforcement alongthe perimeter of the column section ensure that this failure conditiondoes not limit or influence design procedure. In fact, main resistantmechanisms in post-cracking condition are referred to (i) concrete toconcrete interface shear, (ii) dowel action in column longitudinalreinforcement and (iii) clamping action produced by a local yielding inlongitudinal bar that contributes to transfer main part of the shearstrength.

Shear friction mechanism is evidently influenced by axial force amountand roughness (friction coefficient) of the casting surface; a not properpreparation and control of the casting surface can reduce shear frictionresistance.Clamping action – a friction mechanism too, integrating above contri-butions – is proportional to longitudinal reinforcement amount.Dowel action, not negligible in post cracking phase, is strictly connec-ted to longitudinal reinforcement percentage but mainly, due to theirperipheral position, to effectiveness of transversal reinforcement in thezone adjacent to the sliding surface. Hoops, in fact, work as a transla-tional restraint to longitudinal bars involved in the mechanism.Figure 43(a) reports a failure mechanism due to the loss of continuity atthe intersection between joint and lower column. A poor treatment ofthe casting surface can be easily detected and it can lead to a lower con-crete to concrete friction coefficient in correspondence with this weaker


Fig. 44- Infill panel failures: diagonal cracking (a), (b) and corner crushing (c)mechanisms.

Fig. 43- Failure mechanisms at joint – column interface surfaces.

surface. Reinforcement probably composed by only four longitudinalbars and the absence of hoops in the zone strictly reduce dowel action.In fact, this limitation is substantially due to a low longitudinal reinfor-cement percentage not properly spread on the section perimeter, joinedwith the spalling of the concrete cover that is the only transversalrestraint to horizontal bar sliding. A strong reduction in the axial forceessentially due to a non ordinary strong vertical component of L’Aquilaevent has further reduced shear friction capacity. Figure 43(b) shows,as well as Figure 43(a), a clear separation in concrete at joint-columninterface. Absence of hoops in the joint and axial force reduction due tovertical seismic action strictly reduce, respectively, dowel actionmechanism and friction mechanism at joint-column interface.

3.2.3 InfillsIn the previous section it was emphasized how damage limitation limitstate prescriptions and verifications are essentially aimed at avoiding orreducing infill damage and most remarkably that this kind of prescrip-tion was firstly introduced in the Italian code only in 1996 and betterdetailed and completed in 2003 with OPCM 3274.Therefore, it is reasonable to assume that the main part of RC buildingsin L’Aquila was realized without any deformability control and verifica-tion. On the other hand it should be emphasized that, even if a designprocedure according to 1996 or better according to 2008 code had beenemployed, involving damage limitation verification, the strong PGAcharacterizing L’Aquila event would have equally produced a spreaddiffusion of damages to nonstructural elements such as external infills.As a general rule, infill failure mechanisms can be classified in: (i) hori-

a

b

c


Fig. 45- External infill panel failures and discrete (a) or lined up (b) connections between layers.

a b

Fig. 46- External infill panel failures without connection between layers (a) and with ineffective connection (b).

a b

Fig. 47- Infill failure mechanisms differing form opening position and percentage.

a b

c

zontal sliding in the central zone of the infill panel, (ii) diagonalcracking due to tensile stress in the central zone of the infill panel, (iii)corner crushing in the direct contact application zone.Figure 44 reports two building facades in which infill panels are cha-racterized by a diagonal cracking mechanism. In the first case, seeFigure 44(a), it is worth to note how cracking diffusion involves infillingadjacent to window openings and damage is concentrated at the firstlevels of the building; in the second case, Figure 44(b), diagonalcracking is more emphasized by plaster layer because the external layerof the infill is composed by solid clay bricks. Figure 44(c) shows a typi-cal corner crushing mechanism. Out of plane failure of the infillingexternal layer gives the possibility to detect corner crushing mechanismof the internal layer; other evidence is the crack, that visibly involvesthe plaster but probably is deeper, localized at the top of the columnadjacent to the infill panel as a consequence of column-infill local inte-raction.The great majority of external infill panels are composed by doublelayer infill panel, internal layers are generally realized with clay bricks;connections between the two layers are realized by the interposition ofbrick elements discretely, Figure 45(a), or lined up, Figure 45(b). Notreliable efficacy of this system should be stressed.Furthermore, in most of the observed cases, internal infilling layers arerestrained at the four corners of the RC frame while external ones areconstrained only by the upper and lower beam by means of a little pawl.This executive solution leads to a reduction in the interaction mechani smbetween RC frame and external infill panel in both plane and out ofplane seismic forces. In fact, the low efficacy of the restraint applied tothe external panel, coupled with ineffective connections or completeabsence of connections between the two layers, produces a damagerestricted to the external infill panel, which shows an out of plane fai-lure due to seismic action in both directions as it can be detected inFigure 46.Windows or door openings represent a discontinuity in the infill panel,modifying its performance and capacity by a reduction in terms of stiff-ness contribution and by a modification in the failure mechanism.Figure 47 shows damages detected in L’Aquila RC buildings characte-rized by a different opening position in the panel or a different openingpercentage respect to the total area of the bay.Both local and global interaction effect between infill and RC structureare not negligible. As it was previously emphasized, local interactionbetween infill panel and adjacent column can bring to (i) a reduction inthe effective length of the column, an increase in shear demand and aconsequent brittle failure of the column when the panel partially fillsthe frame bay; (ii) to a concentration of shear demand at the end of thecolumn and to a consequent brittle failure when diagonal compression

is applied by the panel to the RC element.As a global phenomenon, infill-structure interaction increases globalstiffness of the complex system and consequently spectral accelerationdemand, besides it can represent a source of irregularity in plan or inelevation when a non uniform distribution is present.

3.2.4 Criteria for regularity in plan and in elevationRegularity criteria in plan or in elevation were introduced in the Italiancode in 1997 by the explicative document to the 1996 code as a quali-tative definition. Only in 2003 by means of OPCM 3274 introduction a

quantitative definition of regularity in plan and in elevation was furni-shed.A compact shape in plan that implies geometrical limitation in theappendixes, a uniform distribution of resistant systems in plan, slabsthat can be considered as rigid respect to vertical resistant elements canensure a regular structure in plan.Conversely, a verification in mass and stiffness distribution, verifyingthat storey lateral strength is not characterized by unexpected reduc-tion, leads to a regular structure in elevation.In spite of a quantitative definition of regularity criteria, not necessarilythese criteria are enough to avoid other irregularity sources not consi-dered in the design procedure of the RC structure. It is the case of infillsthat by means of interaction with the structure can strictly modify stiff-ness and strength distribution in plan and in elevation.It is necessary to avoid non uniform distribution of infills or converselyit is necessary to explicitly take into account infill contribution in theanalysis procedure. For example, Italian seismic code provides someprescriptions aimed at considering infill irregularity contribution inplan by an increase of accidental eccentricity value or in elevation byan increase of the shear demand at the storey characterized by an irre-gular distribution of the infills.

Some peculiar cases of structural failure after L’Aquila event, mainlycaused by irregularities in plan or in elevation, are reported in Figure48. Figure 48 shows a view of the structures before the earthquake, sowithout any damages, on the left, and after the event where collapse isevident on the right.The first structure, Figure 48(a) and 48(b), was placed in the centre ofL’Aquila city (Porta Napoli street); it was characterized by elevationirregularity due to the non continuity of the seismic resistant schemeover height, additionally, second level was characterized by an evidentdiscontinuity in terms of infill distribution, in fact on the left wing of thebuilding there is a sort of porch. Observing building collapse, a dama-ge concentration at the second level is to be noted, that consequentlyproduced complete failure of the upper levels.The two other structures proposed in Figure 48 are both placed in theresidential zone of Pettino (Dante Alighieri street), close to L’Aquila,and present the same shape in plan similar to a T, so they both can beconsidered as irregular in plan. In addition to plan peculiarities, a dif-ferent distribution of infills at the first level respect to the others, due tothe presence of garages entrances, can be observed. Both buildings pla-ced in this street showed a soft-storey mechanism at the first level thatcan be explained by infills irregularities in elevation and presumably bya local interaction between infills and adjacent columns, leading to abrittle failure of some columns at the first level.

4. SCHOOL BUILDINGS

One of the main objectives of the Civil Protection in the immediatepost-earthquake of L’Aquila of April 6th, 2009 has been scholastic buil-dings’ damages assessment as well as the fast repair of the ones withnon structural damages only. These activities have been developed bya joined work of Function 1 of Emergency Management and Quarter ofDepartment of Civil Protection (DPC), Consortium ReLUIS, SeismicRisk Competence Centre of DPC and Public-Works Office of LazioSardegna and Abruzzo.The structural safety assessment of L’Aquila scholastic buildings star-ted on April 8th, 2009. The in-site inspections have been coordinatedby ReLUIS under the supervision of Function 1 of DPC at ReissRomuli; the inspections have been related to both L’Aquila and its pro-vinces scholastic buildings.The activity involved 62 scholastic buildings of L’Aquila: 53 under theadministration of L’Aquila municipality (6300 students on the total ofabout 7000) and 9 of province (4000 students on the total of about5000). A total of 156 structures have been investigated. The results ofstructural safety assessment is summarized in Figure 49.The in-site inspections on scholastic buildings in the L’Aquila provin-


Fig. 48- Soft storey mechanisms examples in L’Aquila: Porta Napoli street (a), (b), Dante Alighieri street (Pettino) (c), (d), (e)and (f) before and after collapse occurrence.

a b

c d

e f

and about 21% masonry structures (see Figure 51)In terms of damages the results of in-situ inspections showed that about31% of RC framed structures were assessed as A (i.e. no significantdamages), about 43% as B (i.e. no significant damages on structuralmembers) and about 26% as E (i.e. significant damages on both struc-tural and non-structural members). The structures with a lower level ofdamage (i.e. A and B) have been mainly built between ’60s and ’90syears while the structures recorded as E were mainly built between ’20sand ’70s years. The RC framed structures mainly showed damages onnon-structural members (i.e. partitions and ceiling); the elementaryschool of Paganica is a typical example of such kind of structures (seeFigure 52-53). An example of RC framed structure with significantdamages on structural members is the school “Celestino V”, see Figure54-55. The RC shear wall-type structures were assessed only as A(27%) or B (73%); no significant damages on structural members werefound at all. The masonry structures, mainly built before ’60s years,were assessed as: 30% A, 24% B, and 46% E. Figure 56-57 show somesignificant damages on masonry members of the elementary school“S.Elia”.

ces were performed in 64 different municipalities on 224 buildings fora total of 309 structures; the results of such activity is summarized inFigure 50.The scholastic buildings of L’Aquila are mainly reinforced concrete(RC) or masonry structures; in particular, about 66% are RC structures(56% RC framed structures and 10% RC shear wall-type structures),


a b c

Fig. 49- Structural safety assessment on L’Aquila scholastic buildings.

Fig. 50. Structural safety assessment on scholastic buildings of L’Aquila provinces.

Fig. 52- Damages on partitions at elementary school of Paganica.

Fig. 51- Structural safety assessment results on scholastic buildings of L’Aquila: a) A; b) B; c) E.

In the immediate post-earthquake several teams with members fromReLUIS, Seismic Risk Competence Centre of DPC and Public-WorksOffice of Lazio Sardegna and Abruzzo were involved in a further stageof in-situ inspections on schools in order to investigate the repair pos-sibility before the new scholastic year official opening (foreseen inSeptember). In some cases several destructive and non-destructive testswere performed in order to investigate on the materials mechanical pro-perties. As a result of this second round of in-situ inspections, a costestimate to fully repair these schools was also performed. The repairinterventions were planned on schools assessed as A and B; the repairstrategy and the interventions design were provided by engineers ofmunicipality or province under the supervision of ReLUIS and Public-Works Office. The Public-Works Office also managed the bids for worksexecution (see Figure 58).


Fig. 53- Cracks and plaster spalling on partitions of elementary school of Paganica.

Fig. 54- Column crack at elementary school “Celestino V”.

Fig. 56- Roof collapse of elementary school “S. Elia”.

Fig. 57- Diagonal cracks on masonry panels of elementary school “S. Elia”.

Fig. 55- Damages on ceiling at elementary school “Celestino V”.


Fig. 58- Table of works at elementary school of Torrione. Fig. 61. Intervention on partition to avoid the overturning.

Fig. 59- Strengthening of joints on elementary school of Paganica. Fig. 62- Investigation on masonry corner of elementary school “Villa Grande” of Tornimparte.

Fig. 60- Strengthening of joints on elementary school of Torrione.

According to Ordinances 3789 and 3790 and commentaries the worksinvolved not only the repair of non-structural members but also localstrengthening interventions on structural members (i.e. strengthening offront and corner joints of RC structures (Figure 59-60), insertion onmasonry members of chains and braces) and non structural members(interventions on partitions in order to avoid their overturning, to con-nect their internal and external faces, and application of zinc coatedsteel grids on partitions), Figure 61. The execution of repair andstrengthening works has been developed together with the execution ofmaterials non-destructive tests (rebound and sonic tests, tests on con-crete cores and steels specimens, (Figure 62) as well as tests by usingflat jacks on masonry structures). Finally, load tests were performed onslabs and flights. The analysis of several cases of study allowed to optimize a series ofrepair or local strengthening intervention on existing RC or masonry


structures; a detailed description of standard repair interventions torestore damaged buildings due to seismic actions has been reported ina document developed in collaboration between DPC and ReLUIS,“Guidelines for Repair and Local Strengthening of Structural and NonStructural Members”.The synergy between DPC, ReLUIS, Public-Works Office as well asmunicipality and provinces allowed scholastic buildings assessed as Ato be opened with urgency in order to regularly perform the exams at theend of the scholastic year. Further, the whole stock of scholastic buil-dings assessed as B has been opened in the period between September21st and October 5th.

5. THE CASE OF SAN SALVATORE HOSPITAL OF COPPITO

The recent earthquake of 6th April 2009 significantly hit the city ofL’Aquila and its surroundings both for the serious number of casualtiesand for the damage suffered by residential and important structures.Among the latter, one of the most important is surely the San SalvatoreHospital of Coppito, the crucial point of the hospital system in the areaof L’Aquila, which was completely evacuated during the emergency duethe damage at various floors of the buildings.A thorough analysis of the hospital facilities has shown besides theessentially non-structural damage also some aspects of the constructionof the hospital complex, which might be crucial for the seismic respon-se to future shocks and therefore open more critical damage scenarios.Other smaller hospitals located in neighbouring urban areas showedminor damage, allowing partial absorption of emergencies, thus redu-cing the enormous overload on the field hospital set up close to the SanSalvatore Hospital. Recent valuations, in fact, have shown that morethan 1500 injured received assistance during the days after the emer-gency, which confirms the tremendous impact on the hospital system.The hospital structure has been designed in 1966 and took about 30years to build entirely. Such an information is particularly important ifrelated to developments in domestic seismic regulation. In fact, the firstItalian seismic law, considered a forerunner of the most modern ones,was n. 64 of 2/2/1974, issued after the 1974 Ancona earthquake.Previously, references to seismic law came from Royal Decrees (e.g.Regio Decreto Legge n. 2105 of 22/11/1937), while during the Re pu -blic era there was Act n. 1684 of 1962 that followed the Campaniaearth quake of 21/08/1962, and was later completed by Act. n.1224 of5/11/1964, and by the Act n. 6090, 1969 issued after the Belice earth-quake. However, such references were oriented towards defining heights,thicknesses, executive methods and quality of materials rather than cal-culation methods and design criteria.After the 6th April earthquake, only 3 buildings out of 15 in the San

Salvatore complex suffered considerable structural damage. This waslimited to small areas and primarily was due to evident issues that willbe described in detail later in this work. There was slight and relativelylimited non-structural damage and significant non-structural damage inonly a few buildings. In these latter inner partitions significantly helpedthe lateral resistance by dissipating the earthquake’s energy and suffe-ring critical damages (Figure 63).The basement and semi-basement part of the complex, mostly made upof reinforced concrete walls, showed a stiff box response without signi-ficant damage. Finally, no evident damage was observed in the founda-tion structures.From the point of view of the human safety, the most widespread andrelevant non-structural damage was that of exterior façade bricks, cove-ring the entire surface of all the buildings. Such a coating, not linked tothe interior walls, in many cases was partially or totally detached. Nosignificant damage was observed on the equipment and internal mecha-nical devices inside.

5.1 Description of the structures of the San Salvatore Hospital

In terms of typology, the San Salvatore Hospital complex consists of aseries of reinforced concrete frame structures, with interior and exteriormasonry walls, built from the mid ‘70s on, and put into service in thesecond half of the ’90s. Some of the buildings of the complex are nothospital property.The buildings differ in typology, materials and heterogeneous construc-tion details depending on the different age of construction. A coveredwalkway connects the various blocks on four floors, two above groundand two underground.There are several building typologies: L-shaped of 2 or 3 storeys, towerblocks of 3 or 4 floors, in-line buildings of 2 or 3 storeys and someground-level buildings. Most of them have one or two basement floors.The approximate date of construction and the main function of thevarious blocks, identified on the basis of the numbering given in Figure64, was provided by the Technical Department of the hospital and isshown in Table 1.

Fig. 63- Evidence of the reinforced concrete frame structure through the damaged partitions: Building 9 (a); Building 10 (b).

a b


Fig. 64- Site plan with different blocks (the groups of buildings identify theclassification of usability as described in section 4). (Gruppo/Group; Edificio/Building)

5.2 Usability surveys and structural response

The buildings of the hospital complex have been repeatedly checkedbecause of a succession of significant after-shocks related to the so-cal-led seismic swarm, and were grouped into categories depending on theassessed damage.All the buildings have an exposed brick wall covering, which is notenough or not at all connected with the infill panels and the reinforcedconcrete frame structure. These are therefore critical for human safetysince they can be dangerous and might fall onto the walkways (Figure65). Another type of frequent and potentially dangerous non-structuraldamage is the complete detachment of the coating of many of theground floor tiled walls.

In all the buildings of the complex there are also several structural jointsof sizes and characteristics not appropriate to shock induced move-ments. The resulting local pounding between adjacent bodies causedlocalised damage, somewhere particularly evident. Figure 66, for exam-ple, shows the damage to a structural joint ending on the top of a column,which have caused an abnormal concentration of pressure on the joint.Figure 67 shows the cracking continuing in the ceiling from the dama-ged joint in the wall, found in a connecting walkway (Building 2).Some local damage have been caused because of improper construction

Table 1 - Buildings of the San Salvatore Hospital complex, Coppito,L’AquilaDenomination Function Construction ageBuilding 1 Thermal Power station 1977/78

and refectoryBuilding 2 Analysis laboratories 1976/77Building 3 Diagnostics and radiotherapy 1976/77Building 9 Emergency room 1978Building 10 Pharmacy and operatory rooms 1978/79Building L1 Direction 1983/84Building L2 Obstetrics and gynaecology 1983/84Building 6 Wards 1987Building L3 Oncology 1979/80Building L4 Infectious Diseases 1979/80Building L5 Neurology 1983/84Delta 6 Wards 1987Delta 7 Medical Delta 1985Delta 8 Surgical Delta 1980

Fig. 65- Damage to the coating of Building 2 (a), Building 9 (b) and Building L3 (c).

Fig. 66- Improper structural joint ending on the top of a column.

a b c


detailing, at least according to current seismic design criteria, such asthe not always appropriate confinement of the structural elements, theinsufficient concrete cover (Figure 68), the presence of columns madesquat by the infill masonry walls.

Group 1: usable buildingsRegarding the basement, 6 metres underground, where the thermalpower station is located, Building 1 shows moderate non-structuraldamage, particularly in the offices at the lowest floor. The portion hou-sing the power plant shows a moderate damage in some beams, mainlyas a result of the relative movement of the Gerber half joint, and it istherefore usable. The rest of the building, however, shows greater damage in the upperstoreys, even if it does not constitute a danger to the floor below, and itis therefore to be considered unusable.

Group 2: buildings that can be made usable with short term countermeasuresSome buildings with slight non-structural damage can be reopened

shortly, and therefore can be usable by means of simple short termcountermeasures, among which: removing falling cladding, plaster anddetached coatings, making safe damaged false ceilings, repairing lightdamage to claddings and partitions, checking all the fixings hangingfrom cladding and false ceilings, in order to evaluate the possible deta-chment risk, creating localized barriers to protect walkways, and remo-ving unsafe portions of the outer coating. Buildings L1, L2 and L5belong to this category, which show widespread light non-structuraldamage, particularly at the ground level. These three buildings will bethe first to be reopened at the end of May 2009, less than two monthsafter the main shock.

Group 3: partially usable buildingsEach of the Buildings 2 and 3 is composed of two different blocks sepa-rated by a structural joint, which show different levels of damage. Inparticular Buildings 2A and 3A, which face Building Delta 7 (Figure69), are not usable, whereas light non-structural damage is common tothe rest of Buildings 2B and 3B (Figure 69).Unlike group 1, in this group of buildings the unusable part featureshigh structural risk, whereas the less damaged part requires modestintervention before to be used again.

Fig. 67- Damaged structural joint on wall, extending in ceiling.

Fig. 68- Examples of improper transversal reinforcement and insufficient concrete cover in Building 2A and in the externalconnecting walkways.

Fig. 69- Buildings 2A/3A and 2B/3B (Edificio/Building).

Fig. 70- Damage of the stairs of Building 2 (a) and Building Delta 8 (b).

a b

Buildings 2B and 3B can be made usable with short term countermea-sures, like those described in the previous paragraph. At the last flightof the inner stairs between portions 2A and 2B, mild localized structu-ral damage was detected, probably due to concentrated rotation of the


Fig. 71- Structural damage (plastic hinges and shear failures) at the top of co -lumns (ground floor of Building 2A) caused by insufficient transverse rein -forcement.

Fig. 72- X cracks on the external coating of Building 2A.

Fig. 73-. Important damage to partitions in Building 9.


knee beams (Figure 70). In this case limited restoration interventioncan be provided, such as repairing cracks with resin or anti-shrinkagemortar, and later if necessary a seismic retrofit can be designed usingfiber-reinforced materials.Blocks 2A and 3A are unusable because of the considerable non-struc-tural damage and the structural risk caused by the failure of somecolumns of the covered walkway over the driveway to the EmergencyRoom, facing the Building Delta 7. This is the first area of significantstructural damage. As shown in Figure 71, at the top of the columns itis possible to recognize clear plastic hinges, shear failures, spalling ofthe concrete cover with the consequent instability of the longitudinalreinforcement. All these effects are due to the scarcity or absence oftransversal reinforcement, i.e. of confinement. This has led to a nearstate of collapse, requiring urgent propping intervention. In the samearea there is widespread damage to infill panels, with the formation ofX-cracks (Figure 72) typical of shear failure. Although such failuresmay affect only infill panels, potential shear failure to the structural ele-ments under the damaged coating cannot be excluded, and thereforethey must be checked.

Group 4: buildings requiring more extensive non-structural measures A number of buildings can be made usable only after significant localdemolition and reconstruction of the most damaged partitions and remo-val and restoration of all unsafe or detached parts. The restoration of con-ditions of temporary usability in this case requires more time, and the-refore cannot be included among short term countermeasures.Buildings 9, L3 and L4, in this group, have suffered moderate to severenon-structural damage to many of the ground floor partitions (Figure 73)and a widespread light non-structural damage at the upper storeys. On the more damaged part, the usability restoration requires demolishingand rebuilding some infill walls, and propping a stair in Building 9.

Group 5: unusable buildingsThe last group of buildings is classified as unusable, due to the signifi-cant structural damage or because they are close to dangerous buil-

dings, or because they feature a high percentage of severely damagedpartitions compared to the volume of the building. Building 10 has severe structural damage at the ground floor, becauseof the shear failure of all the columns on the side in front of the church.In this part the columns have been made squat by the infill panels inter-rupted by the belt windows along the north-east side (Figure 74). It isclear that for most of these columns the bearing capacity is seriouslycompromised. The building also has widespread moderate to severenon-structural damage, particularly at the ground level. To ensure this building will not collapse, short-term countermeasuresare arranged, by means of a series of cement block infill panels betweenthe columns of the external porch and by propping the beam supportedby the damaged columns.In Buildings Delta 7 and Delta 8 there is widespread structural dama-ge more important at the lower levels. The damage to partitions is non-structural but widespread and of moderate to severe intensity. The onlylight structural damage is found in one of the stairs. The two buildings are considered unusable, due to the significant exten-sion of damage, together with the structural irregularity and the proximityof buildings with structural problems (Building Delta 8 is next to Building10 and Building Delta 7 is adjacent to Buildings 2A and 3A).

5.3 Usability of walkways and connections in basements

The basements of all the inspected buildings, where accessible, have no

Fig. 74- Structural damage caused by the presence of belt windows, which makethe columns squat in Building 10.

Fig. 75- Connecting walkways: hollow clay block collapse of the intrados of the basement at 3 metres below ground.


significant damage. There are rare and limited examples of hollow clayblock collapse in the ceiling of basements at 6 and 3 metres belowground (Figure 75).At the level of 6 metres below the ground there are unused spaces,accessible only by technical personnel, without significant damage,except for localized seepage and settlement/shrinkage cracks.At the level of 3 metres below the ground, in the underground passageconnecting the buildings, the damage is predominantly non-structuraland not very extensive. It mainly consists of the damage to existingjoints and the formation of ‘natural’ joints after the event, with limitedplaster and coatings detachments, and some hollow clay blocks col -lapse. In such basement levels the damage is more significant towards Build -ing 1, where some localized modest damage is present. At the first floor level (3 metres above ground) there is damage to exi-sting joints, formation of cracks, primarily to the completion of incom-plete joints, and the subsequent detachment, sometimes only partial, ofplaster, false ceilings and coatings. At the ground level the most critical situation is found since the con-nection walkways between the various buildings are not protected,sometimes there is just a covered porch. The risk of falling of loose partsof partially detached coating or new portions of coating falling off, evenafter minor shocks, makes the need to remove partially detached oralready collapsed coating urgent, to protect all the paths adjacent to thebuildings (when there is no covered porch) from falling objects and toclose off the riskiest paths. At this level there is some structural dama-ge: a column behind Building 9 stressed by a structural joint, andanother column on the corner of Building 2 with a plastic hinge at thetop.

6. INDUSTRIAL STRUCTURES

6.1 Building-like industrial structures

Industrial buildings were built for many years as an assemblage of pre-cast reinforced concrete elements. The April 2009 L’Aquila earthquakehas struck, for the first time in Italy, industrial structures on a largescale. In fact, the Irpinia 1980 earthquake hit an area with few indu-strial sites; and similarly happened in the Umbria and Molise earth-quakes, which moreover were felt within a limited area. On the otherhand, the Friuli 1976 earthquake damaged industrial structures, butthey were designed with no regard with respect to the seismic action; ifany design rule was used, this however belonged to inadequate seismiccodes. L’Aquila and its surroundings are instead undergoing a ratherstrong industrial development. Precast reinforced concrete buildings,

with a column-beam structure, are the usual type in the industrial areasin Pile, Bazzano, Monticchio and Ocre (AQ). Buildings have generallyone storey; two storeys are seldom observed, moreover only on a morelimited area than the first storey. Beam supports on the columns may beof the saddle or bracket type. Deck beams are usually transversally ali-gned (with respect to the longer side of the structure plan) and they areI-shaped, with variable depth; longitudinal alignment and inversed T-shapes are less commonly observed. In the latter case, the shape is usedto support the tiles. The roof is generally built with tiles �-shaped; lessoften U-shapes are observed. Skylights are sometimes present. �-tilesare also used for the intermediate deck, when there is one. External par-titions are either made with bricks, or with precast reinforced concreteshells, with no stiffeners.Structural nodes are those typical of the Italian constructions: beakerfooting for the foundation to column nodes, simple support for the beamto column nodes (with neoprene bearings and steel pins). Tiles aregenerally directly resting on the beams, with neither horizontal restraintnor neoprene bearings. Pin connections are seldom present. Tiles maybe connected each other with the upper reinforced concrete layer, orsimply linked via steel restrainers (partly poured within the tiles con-crete, partly welded with the next tile restrainer). Partition panels areeither supported by the eaves beam or by the column, via links of manytypes. They are also sometimes supported by the deck tiles. A typicalshell-eaves beam connection is via a steel plate partially put within theconcrete shell during pouring. A bolt connects the plate and a steelangle, which is restrained to the eaves beam edge. The shell to columnconnection is often built with a steel plate within the column and abayonet with bushing linked to the shell, via a long bolt. This techno-logy is used also to connect the partitions to the deck tiles; the steelangles and bolts connection is less frequent. It is worth to note thatstructural shell buildings are much less common than beam-columnbuildings. A few precast industrial buildings were under completion onApril 6th, 2009, when the earthquake struck; so it was possible to verifythe seismic behaviour of these structures under variable degrees ofcompletion.

6.2 Damage and seismic performance analysis

The response of the structural elements of the industrial buildings to theApril the 6th 2009 earthquake was generally in accordance with theirdesign level: no column collapsed, even though, in many cases a plastichinge was observed, due to the high intensity of the seismic action(Figure 76). In some cases such plastic hinge was not observed at thecolumn base, i.e. at the column-foundation joint, but even one meterabove, where the longitudinal reinforcement decreases. Furthermore no

plastic hinge was observed in beams or tiles due to the increment of thevertical action. However, the damage of the precast industrial buildingsshould be well analysed; indeed, it was characterised by collapse ofparts of the buildings, which, if the mainshock had happened during theworking time instead of at 3 a.m., it would have caused victims.The static scheme of such structures is characterised by large deforma-bility; consequently, the most of the observed damages of structural ele-ments (made by reinforced concrete) depend on the relative displace-ments between the elements. Indeed, many cases of pounding betweenelements of the same structure were observed. Furthermore, poundingbetween adjacent buildings was frequent, in the case of both precastand cast in situ structures, due to the insufficiency of separation joints.In Figure 77 the pounding between the tiles and the beam of an indu-strial precast building placed at Bazzano is shown.Confirming the numerical studies performed in the last years, the con-nections represented the weak parts in terms of seismic performance ofboth old and new precast buildings. Some buildings have shown dama-ges at the beam-column connection: the only observed case of precastbeams collapse was due to the damage of such connection and to thefollowing support loss. Indeed, as shown by numerical analyses, thesplitting of the joint bar cover happened where the thickness was mini-mum. In other frames of the same structure, it is also possible to obser-

ve the collapse of the beam due to support loss at the side without jointbar, caused by too large displacements, and the pounding between thebeam and the column top fork (Figure 78).The phenomenon of the joint bar cover splitting can be also noted at theintermediate level of some two-storey precast buildings, where, asalready written, the beam-column joint is on corbel. The same pheno-menon has also characterised the collapse of some tiles. In this case,indeed, even where the joint had been fastened by a steel bar, the littlethickness of the bar cover of the beam, also characterised by the lack ofstirrups, collapsed, causing the tile support loss (Figure 79a).Obviously, such support loss easily happened where the tile-beam con-nection was not fixed and/or there was no connection between tiles; par-ticularly unlucky situations were characterised by buildings in phase ofassembly, where the floor slab, joining the tiles, was not made yet.


Fig. 76- Plastic hinge in columns of industrial buildings: (a) FIAT garage at Pile; (b) building used for bovine-breeding at Fossa. Fig. 78- Collapse of beams due to loss support of a building at Fossa used for bovine-breeding.

Fig. 79- (a) Tiles collapse due to cover splitting and support loss of a FIAT garage at Pile (b).

Fig. 77- Effects of pounding between the cover tiles and the beam.

a b

a b

Collapse of perimeter panels due to the breaking of the angle stirrup orto the bolt head going out from the profile happened in a building usedas material and machine deposit at Bazzano.However, the most important and spread damages of precast industrialbuildings caused by the April 6th earthquake are those concerning theelements on the perimeter; indeed, the large damage of such elements,even though the structural typologies are different, associates precastbuildings to the in situ cast ones. The top connection of the verticalpanels to the side of the gutter beam, made by a profile drowned in thepanel, bolt with nut and angle stirrup, in some cases gave way due tothe angle stirrup breaking and/or to the bolt head going out from theprofile (Figure 79b). This last phenomenon also caused the collapse ofpanels connected to columns by a profile drowned in the column andbayonet with bushing joined to the panel by bolt; some of these last con-nections, instead, collapsed due to the bayonet breaking at bushing

which have induced strong deformations of the shells of the sili. Suchtype of damages are a clear effect of the earthquake vertical component,whose importance they highlighted (Figure 82). The Vibac sili have a metallic structure. Generally for their conceptionsili have a very low structural weight, normally significantly lower thanthe weight of the contained material. Such a characteristic implies avery slender structure. It is evident that such structures are sensitive toboth local and global buckling phenomena. In fact the most commonfailure mode is the instability of the wall panels due to the effects of theaxial force in compression. Such actions are due to the friction betweenthe silage material and the walls. The horizontal radial pressure, actingon the cylinder surface from the silage material, has a stabilizing effectagainst the buckling of the silos’ walls, giving rise to a tension stressfield of membrane type. The distribution and intensity of the internalforces in every constituting part of the silos, the cylinder and the hop-per, are strongly influenced by the material extraction behaviour, whichin turn depends on the shape of the silo. The Vibac sili have an elongated shape typically used for the storage ofplastic material. Therefore the predominant extraction mode is of the socalled “mass” type, having the characteristic that the first materialcoming out is the one inserted as first in the silos, all the material massis in movement at the leakage. In case of sili with a stocky shape, theextraction behaviour of “funnel” type prevails, it has the characteristicthat a central tube forms in the material mass, which is sucked by thehopper. Such a “tube” is fed by the silage material all along the height,the part of material external to the tube rests during the leakage. In par-ticular, in elongated sili, when completely full, along the height of thecylinder, from the higher ring bands the radial pressure grows towards

position. Some other failures where due to the going out of the wholeprofile from the panel where it was drowned. A better seismic responsewas shown by panels joined to the structure by angle stirrups and bolts.In the case of perimeter elements made by bricks, the seismic actiondetermined their out of plane deformation, in many cases up to theexpulsion of bricks and the consequent partial or total collapse of theperimeter element. Finally, among the carrying out mistakes, it isnoteworthy, for precast structures, the local failure of the beam support.In Figure 80, a near collapse condition is shown, caused by the largecolumn cover due to the fire protection provisions; indeed, due to suchprovisions, a volume of concrete without reinforcement works as beamsupport.

6.3 Non-building-like structures: the case study of the Sili Vibac at Baz -zano

The sili of the Vibac multinational (a chemical company which produ-ces plastic films), located at Bazzano, close to Onna (Figure 81) repre-sent an exceptional case of damage to steel constructions. They alsorepresent an emblematic case of damage induced by the earthquake ofApril 6th. The sili are used for the storage of polypropylene pearls, andthey were full when earthquake struck. Some sili collapsed, some otherremained standing even though strongly deformed, both locally at somerings and diffusely (Figure 82).A more close visual inspection indicated that the collapses occurred foroverturning due to the crushing of the base rings and the hopper.Moreover, along the sili height, deformations induced by buckling phe-nomena of the wall panels are apparent. In some cases an effect ofpounding on the adjacent precast reinforced concrete constructionstook place, the latter have achieved the partial failure of the infills,


Fig. 80. Local failure of the beam support.

Fig. 82- Pictures of the sili after the seism (Figure by G. Verderame, ISPRA, F.M. Mazzolani).

Fig. 81. (a) Localization of the plant VIBAC in the Bazzano municipality (AQ). (b) Sili before the earthquake.

a b

the base up to assume a constant value, then at the section variation, itreduces from the ring where the hopper is installed, and high stresslevels arise. Obviously in case of empty or partially full silos the beha-viour is different, the stabilizing effect of the radial pressure is lost inthe empty part with a consequent abrupt variation of the critical stress.Pressure variations inside the silos depend also by the leakage of thematerial through the hopper, which causes a backwash effect and thusdepressions. In order to control and regulate such an effect, sili are pro-vided with pressure valves. Given that it is plausible that on one sidethe effect of the seismic vertical component provoked a sharp andimportant increment of the actions in compression in the sili walls, cau-sing buckling, the contemporary seismic action in all the componentsaccentuated the effect of possible asymmetrical distributions of pressu-re, due either to structural eccentricities, or to the silos filling method,or to the anisotropy of the silage material, causing a reduction of the sta-bilizing effect of the radial pressures themselves. Furthermore, buck-ling could also occur due to constructional imperfections at the jointsbetween the coating ring bands of the silos, where joints in any caserepresent a discontinuity in the flow of longitudinal stress in compres-sion, with high concentrated stress. The above mentioned considera-tions fully justify the collapse behaviour observed during the L’Aquilaearthquake.

7. EMERGENCY MANAGEMENT FOR LIFELINES ANDRAPID RESPONSE AFTER L’AQUILA EARTHQUAKE

7.1 Road Network

ANAS S.p.A. is the agency that manages in the Abruzzo Region, as wellas in the rest of the national territory, the state road network. The resi-dual functionality and safety investigation of the road network were thefirst priorities identified by ANAS for the management of the first phaseof the emergency. Physical and human resources were deployed toachieve the following goals: 1) rapid survey of the road network to ensu-re, at the largest possible extent, the regional mobility; 2) activation ofemergency contracting procedures (“somma urgenza” agreements) toimmediately begin, where possible, activities for the restoration of nor-mal mobility conditions; 3) damage survey of the road-network compo-nents; 4) short term planning for the repair of damaged components. At the same time, physical and human resources were deployed in sup-port of the Civil Defence for a first partial debris removal and for theexcavations works necessary for the installation of relief campsites. It isworth mentioning that, further to the local resources, additional oneswere used to manage the emergency. These resources were availablefrom few ANAS’ Regional compartments differently located on the

national territory, with an average daily commitment of 80 men and 70vehicles.Rockfalls (Figure 83a) and landslides triggered by the earthquake andaggravated by the heavy rain that hit the area in the days following theevent, were identified as the most problematic situations affecting thenetwork mobility. However, the rock falls and landslides occurredmainly in mountainous areas around L’Aquila, while the main roadnetwork in the city was not affected by the aforementioned phenomena.In the urban area, mobility limitations were caused by debris followingdamaged and/or unsafe residential and monumental buildings adjacentto the roads. Immediate activities for the restoration of normal mobility conditionsincluded: 1) removal of rocks and soil from the roads; 2) rock slope con-solidations; 3) enhancement of soil slope stability. These activities wereconducted employing, where possible, internal resources or activating,alternatively, emergency contracting procedures with external organisa-tions. Securing of unsafe buildings adjacent to roads was carried out byfiremen. Temporary traffic management measures were extensively implementedin order to minimize road closures; these measures included traffic flowrestrictions; alternating one-way; lane and velocity restrictions (Figure83b).The only significant damages occurred to the road network componentswere the structural failure of the viaduct “Corfinio” on the nationalroadway SS5 and the collapse of a bridge on the main road SP36“Forconese”. No further significant damages were reported to the com-ponents of the road networks including the numerous tunnels present inthe Region that performed well. The urgent need for a standardized and structured survey form to reportdamages and disruptions in the road networks was highlighted whileperforming safety investigation and damage survey operations. A rapidsurvey form and an ad-hoc procedure were therefore identified and for-malised while the survey work was in progress. The timely information on the mobility conditions was a key componentof the effective emergency management. The Civil Defence issued dailya report summarising road closures, mobility restrictions and repair


Fig. 83- Impact of the earthquake on the road network: (a) SS80 “Gran Sasso d’Italia” road affected by rock falls, but featuringrock-proof tunnels. (b) Distribution of traffic management solutions (updated to 01/05/09) for the 61 road tracts affectedby the earthquake (red = road closed; dark green = passable with limitations; yellow = alternating one way; light green =lane and velocity restictions).

a b

morning of April 6, a significant and sudden change in the water flowfor a main pipeline in Paganica. The immediate closure of the relativeshutters for that pipe was operated directly from the GSA headquarters,before the technician team reached the affected site. The cause of therupture was identified in the fault crossing the Paganica pipe. Becauseof that, the steel joint of the pipeline (diameter = 600cm; pressure 25-30atm) slip-off, causing a violent escape of water (Figure 84a). A connection portion at the joint, however, was still grasped for a lengthof 6cm. In order to quickly respond to the emergency, the repair waslimited to the welding of the pipes at the joint. Exception made for the aforementioned joint slip-off, no significantdamage was observed to the main distribution and storage system.Following the repair of the damaged joint it was, therefore, possible torestart the provision of potable water for all municipalities administeredby the G.S.A. SpA since the evening of April 6. As a lot of ruptures wereexpected in the minor water distribution system, in order to preventflooding and deterioration in the buildings already damaged, the deci-sion was made, not to restore the water distribution in L’Aquila histori-cal centre and in the most affected villages. For these areas, the resto-ration of the water provision was gradually operated starting from theless affected zones and/or the zones with a strong need for reactivation;priority was given to the strategic services, secondly to the commercialand industrial activities, including the hotels to be reopened for the G8meeting, and finally to the residential buildings classified safe, after thespecific AeDES survey. The partial restoration of the water distributionwas possible because of secondary networks and of a shutter systemthat allowed the exclusion of areas where the water supply was noturgently needed. A few days after the earthquake (19 April), due to afurther slip of the fault, the welded joint of Paganica pipe broke, requi-ring a further repair intervention. The priorities identified in the second phase of the emergency manage-ment were, on one hand, the provision of the water service to the reliefcampsites and, on the other hand, the management of all the activitiesfor restoring the water provision in L’Aquila City. To carry out the worksfor the water network connection in the relief campsites, the technicalstaff of the company (fully operative since the third day after the earth-quake) was supported by the “Genio Civile” staff. On the other hand,the works for repairing damages and restoring the functionality of thewater service in L’Aquila were operated, where possible, by the G.S.A.SpA technicians, or activating emergency outsourcing procedures forthe most demanding operations. Relationships with external organiza-tions have been unfortunately, nowadays, interrupted because of thefinancial difficulties that the company is undertaking due to the lack ofincome.Most commonly observed damages in the minor distribution system

works carried out in the road network. Using a Geographic InformationSystem, GIS, the technical compartment of the Direction of Commandand Control, Di.Coma.C represented this information in a cartographicformat. Road closures and other temporary traffic management measu-res were overlaid to aerial Figuregraphs, technical regional maps, etc.providing maps that had a fundamental role in supporting many emer-gency management operations. As for the public information, emergency bulletins were regularlyissued to update in real-time the end-users about the mobility situationin the Abruzzo Region. Communications and timely news were, as well,posted on the ANAS website. Once the firth phase of the emergency was managed, efforts and resour-ces were concentrated, on one hand, to handle the modified traffic con-ditions in L’Aquila city due to the closure of the main road that ranthrough the city and, on the other hand, to respond to the new mobilityrequirements created by the relief camps, and by the construction of theprovisional accommodation: Temporary Housing Modules M. A. P, andC.A.S.E project.

7.2 Water distribution network

Gran Sasso Acqua G.S.A. SpA is the water provider for L’Aquila cityand for 37 municipalities in the earthquake area. The organisationoffers an integrated water service including potable water supply, sewe-rage and wastewater treatment. The G.S.A. has 3 major supply systems (Chiarino, Gran Sasso, WaterOria) in addition to some secondary ones. The water supplied is tran-sported by a network consisting of approximately 900km of large dia-meter pipes and is stored in a huge number of tanks (about 200) thatrequire continuous functional and hygienic monitoring and maintenan-ce. The water is distributed from the tanks to approximately 100000customers through a 1100 km distribution network made of quite oldcast iron and steel pipes. The pressure inside the main pipeline networkis quite high, reaching 30-50 atm., as well as in the distributionnetworks where it can reach 6-8 atm. Thanks to a remote control service and guided valves connected, throughcables or wireless connection, to the main reservoirs and supplysystems, it is possible to check the water flow inside the pipelinenetwork and to manage partial or total opening/closing operations direc-tly from the Gran Sasso Acqua headquarters. In particular, electroma-gnetic sensors, measuring input low pressure, and electromagnetic gau-ges (or “Clamp on”), measuring output differential pressures, are instal-led in the tanks. The remote control service allows furthermore theassessment of the water level in the tanks. The equipment connected to the remote control system revealed, on the


were the slippage/breakage of the joints and the breaking of cast ironpipes (Figure 84b). It is important to emphasize, however, that in largepart of the “red zones” (damaged zones with prohibited access) thewater network is still closed. Because of that, it has not yet been possi-ble to completely estimate the extent and the spread of the damage suf-fered by the network1.Finally, it is worth mentioning that the drinking water purity and qua-lity has been officially tested and certified daily since the early daysafter the seismic event. Because the G.S.A. official testing laboratorywas severely damaged after the earthquake, this service was guaranteedvia mobile laboratories of a neighboring water organization, C.A.M..The third phase of the emergency management focused on the con-struction of the water distribution network and connections for the sitesidentified for the construction of the provisional accommodation:Tempo rary Housing Modules M. A. P, and C.A.S.E project. Both thedesign and the new construction of the reservoirs and of the distributionnetwork for these areas were committed to external organizations andcontractors. The costs for both the design and the construction of thenew reservoirs and networks for the temporary accommodation werecovered by the Civil Defence. The G.S.A. SpA will continue to be incharge of the management of the water provision for the temporaryaccommodation areas.

7.3 Wastewater treatment plant

The technical visits at the wastewater treatment plants serving L’Aquila(AQ), in the resorts of Ponte Rosarolo, Pile and Arischia, and at thatlocated in the City of Corfinio (AQ ) have shown that examined systemshave similar technical characteristics, as they have the same practicalfunctions. Each plant was equipped both with the structures necessaryfor the treatment of wastewater (primary clarifier tank, aeration tank,digestion tank, settling tank, thickener, sludge dewatering band press

and chlorinator system) and with those for management purposes(build ings used as offices, rooms for technical equipment and laborato-ries).The facility in Ponte Rosarolo is located near the historical center ofL’Aquila (42°20'18.10''N - 13°23'39.09''E). Structures were built the’60s-’70s. The reinforced concrete digestion tank suffered partial colla-pse of a longitudinal wall (Figure 85a), several vertical cracks on a tran-sversal wall and the separation of orthogonal walls at the edges (Figure85b). The partial collapse of the wall also involved the steel pipe adduc-ting wastewater that was connected to it. In buildings used as offices,local technological and laboratory equipment (RC framed structure)were also found cracks of both internal partitions and external walls.However, there were no evidences of damage to structural elements: thecracks detected on non-structural elements did not represent signifi-cant damages and did not prevent the use of building. The inspectedfacilities were therefore useable at the time of inspection, except thedigestion tank that was useless. Due to this damage the tank has lostwater and the plant were partially closed by reducing the disposal capa-city of about 60%. The remaining functionality was still sufficient toface the demand, which was significantly reduced due to the large num-ber of evacuated people (approximately 30,000), housed outside thecity.


1 The water consumption was reduced by 30% as a result of water shut off into the ‘red zones’.Mobile water tankers were used to serve the relief camps in the first days after the quake.

Fig. 84- Impact of the earthquake on the water distribution network: (a) Joint slip-off in a main water network pipeline inPaganica. (b) Repair on a cast iron pipe in a Paganica at the moment when some of the evacuated people were returninghome.

Fig. 85- Ponte Rosarolo Plant. Digestion Tank: (a) partial collapse of a longitudinal wall and of the pipe connected to it. (b)Detail of the detachment of the orthogonal walls at the edges. (c) Displacement of the pump in the control room.

a

a b c

b

The structures of facility in Pile (42°21'3.25''N - 13°22'13.41''E),which is situated between the town and the industrial area of L’Aquilabeing the second plant serving the city, were realized in two differentperiods (’80s and 2000) with RC walls and slabs. Structural damageswere not detected, only some damages to the partitions of local officesoccurred. With regard to the older settling tanks, characterized by a cir-cular cross section, a deterioration of the curbing RC beam was detec-ted due to significant corrosion of the steel reinforcements.The inspected structures, therefore, were viable and fully functionaldespite the damages (of non-seismic origin), due to degradation of mate-rials descending from a insufficient maintenance of the settling tanks.However, in the control room, a tube connected to the pump (not ancho-red) was damaged due to a displacement of 15 cm, Figure 85c. Finally

it has shown some structural damages;- The Corfinio plant was not damaged because distant from the epicen-ter (approximately 50 km): the maximum acceleration recorded by theaccelerometers of Sulmona station (Sul) located near the plant, is in -deed equal to 0.34 m/s2, approximately one-twentieth of the maximumrecorded at AQV Station of L’Aquila.

7.4 Gas distribution network

Enel Rete Gas S.p.A. is the gas provider for L’Aquila city and for other5 municipalities in the earthquake affected area, namely Lucoli,Tornimparte, Ocre, Rocca di Cambio, Rocca di Mezzo.The gas is distributed via a 621 km pipeline network, 234 Km of thatwith gas flowing at average pressure (2.5-3 bar) and the remaining 387Km with gas flowing at low pressure (0.025-0.035 bar). The medium pressure network is connected to the high pressure nationalone (namely SNAM network) through 3 reduction cabins while, about 300reduction groups allow for the transformation of the gas transport pressu-re (2.5-3 bar) into the gas distribution pressure (0.025-0.035 bar). The gas network is mainly made of steel pipes, with an average internaldiameter of �intenal =125cm (external diameter �external =139.7cm) andthe joints are mainly welded. The first priority identified for the management of the gas network, inthe first phase of the emergency immediately after the earthquake, wasthe timely securing of the network in order to avoid explosions, gasleaks and fires and to allow the emergency vehicles and the USARteams to act in the safest possible way. To ensure this priority, the entire network managed by Enel Rete GasS.p.A. in the affected area was shut off via the closure of the 3 reduc-tion cabins. Thanks to this decision, and to the rupture of a pipelinenear Onna (Figure 86a), it was possible to timely and significantly redu-ce the gas pressure and to avoid the occurrence of secondary effects.The subsequent closure of the 300 reduction groups ensured the fullsecuring of the network in less than two hours after the earthquake. Inthe days following the event, the gas valves external to each residentialbuilding were as well closed. The pipeline damaged in Onna was repla-ced with a new one that was too rigidly connected to a reinforced-con-crete support. It is worth highlighting that, as a result of the earthquake,the Enel Rete Gas headquarters in L’Aquila resulted unusable. Becauseof that the chief executive and the staff had to manage the emergencywithout the support of their data, software and maps. Luckily, the natio-nal society Enel Rete Gas has, at a national level, an integrated infor-mation system, including a data base and a geographical informationsystem GIS. Making reference to the closest Enel Rete Gas headquar-ters in Teramo and Pescara, it was possible to reprint the maps and all

it should be noted that this plant has been out of energy for three daysafter the earthquake, so it worked through its own backup generator.The plant located in Arischia (42°24'49.02''N - 13° 20'25.48''E) pre-sents reinforced concrete structures with the exception of the circulartanks for leaching, consisting of circular walls of artificial masonryblocks connected with a RC curb at the top of the tank, and a gravityretaining wall. The structures date back to the ’70s with the exceptionof RC curb which was more recently constructed. Cracks on the wallsof a distribution trap and damages to the retaining stone wall, which ledto the partial obstruction of the hydraulic groove drain at the base of thetank, were observed. With regard to the circular tanks, one of the tworotating distributors was put out of service for damage to its support; thecracks found on some blocks of the structure were dated before theearthquake. Therefore, the inspected facilities were functional, althoughthe restoration of the full functionality of the hydraulic facility requiredsome minor rehabilitation and repair of the tank distributor. In any case,the age of the plant suggests a constant monitoring even after the reme-dial action.The treatment facility in Corfinio (AQ) situated not far from the centerof the same town (42°7'25.74''N - 13°50'31.78''E) is a RC constructionbuilt in the ’90s. The central part of the longitudinal walls of the aera-tion tank, separated from lateral walls, shows a rotation very probablyoccurred in large part before the seismic event, as witnessed by thecomparison of the positions of monitoring slides before and after theearthquake; such slides were applied two years before the event: thedisplacements due to the earthquake did not compromise the hydraulicseal of the joint, nor the functionality of the structure.A comprehensive analysis of the observed damages was carried out inrelation to the position of each facility with respect to the epicenter ofthe earthquake of April 6th, 2009 (UTC 01.32 hours) and to the recordsprovided by the National Network accelerometric (RAN) available. Itcan be observed that: - Ponte Rosarolo facility is located near the epicenter and close to theAQK accelerometric station, which recorded ground accelerationsequal to 3.7 m/s2 equal to about 50% of the maximum value recordedfor the same seismic event (station AGV - 6.6 m/s2); after the earth-quake, the plant has shown damages to the tanks with rectangular wallslarger than those found in circular tanks of the Pile plant, despite thegeographical proximity. The structural behavior of the circular tankswas essentially better than that of the rectangular ones, mainly becau-se of the lack of structural details ensuring effective connection bet -ween the orthogonal walls;- Arischia plant lies about 5 km from the L’Aquila accelerometric sta-tions AQV, AQG and AQA, which recorded maximum ground accele-ration values; even if distant from the epicenter (approximately 10 km),


the documentation necessary to operate. The second phase of the emergency response was focused on the acti-vation of the physical and human resources in support to the CivilDefence. The timely provision of gas to the strategic structures was thefirst priority identified and was operated via mobile reduction cabinsand gas wagons. H24 shift were organized for the local technical andadministrative teams, as well as for the teams coming from other areasof the national territory including the Enel Rete Gas national head-quarters in Milan. In the first month after the earthquake, the dailycommitment of physical and human resources resulted on averageapproximately equal to 70 men and 35 vehicles, including equippedtrucks, gas wagons and gas-leak detectors. On the same time, activities for the reactivation of the gas provision werestarted. The reactivation of the shut gas network required to operate gra-dually restoring, first of all, the gas flow into the medium pressurenetwork, secondly the gas flow in the low pressure network, up to eachexternal valve pertinent to each residential building previously closed. Reactivation of the service was managed according to the following foursteps: 1) seal verification; 2) nitrogen check; 3) repair of damaged pipesand/or valves; 4) reopening. In the seal verification phase, the detectionof broken pipes and/or the possible joint slip-off was made, acting in thefirst instance, from node to node, and further segmenting the networkwhen necessary. The material and equipment needed for the repair was immediatelyavailable from the integrated logistics system which Enel Rete Gasuses; actually, the material normally in storage in the Battipaglia inter-harbour to perform ordinary repairs and maintenance works, was sim-ply diverted to L’Aquila. The adopted strategy ensured the remediationand testing of more than 90% of the gas network in three month timeafter the earthquake. The diagram in Figure 86b shows how, threemonths after the quake, it was possible to restart the gas distribution forall the end-users with a safe home, exception made for L’Aquila city.It is worth mentioning that the reconnection of the individual user sup-plies required, on one hand, the definition of the priorities to be fol-lowed and, on the other hand, the definition of the testing procedures tobe carried out to certify the safety of the gas systems that were subjec-

ted to the action of the earthquake. As for the priorities, those identifiedby the Civil Defence were followed; namely, the service was providedfirst of all to the strategic buildings, secondly to the manufacturing andindustrial plants, and finally to the residential buildings identified assafe after the AeDES ispection. As for the testing procedures, in accor-dance with the procedures used by Enel Rete Gas for routine checks,an ad hoc protocol was defined in collaboration with the Civil Defenceand the Firefighter Department. It was decided to reconnect each sin-gle user following the fulfillment of four conditions: 1) safe dwelling(classified as A following the AeDES survey); 2) leak-tightnesschecking; 3) operative test of the equipment; 4) smoke test. It is worthmentioning that the Civil Defence fully covered the cost of the wholeprocedure to reconnect the individual users to the gas service and thata dedicated phone line (Line Amica Abruzzo) was specifically set up tofacilitate and support the end-users in this operation. As a final note it is worth remembering that no damages were detectedto the gas storage facilities.

7.5 Electric power distribution network and telecommunications

It was reported that two substations serving the greater L’Aquila haddamaged connections between a rigid bus and insulator, Figure 87a.That was due to shifting of the un-anchored transformers during theearthquake. Also due to sloshing of the cooling oil within the transfor-mer, cooling oil pressure increased, and actuated the safety shut off fea-ture to avoid costly damage. One of the transformers moved about 14cm. In the distribution system, 30 posts were damaged causing severedlinks that resulted in service disruption. More than 180 pedestal typeconnection boxes were dislocated and severed cable connections at thetermination lugs that resulted in localized power failure (Figure 87b).The Electric Power Control Center at L’Aquila sustained severe dama-ge, both building and equipment, and it had to be moved to a temporarybuilding in the yard of the building premise. It took three days to com-plete the move, while the essential part of the system was functional by9 AM the day after the earthquake (Figure 86). Transformers in substa-tions were not anchored. We noted that steel angles were welded on the


Fig. 86- Impact of the earthquake on the gas distribution network: (a) Onna (AQ),damaged pipeline. (b) End-user gas connections activeted on June 8, 2009(Green = end-users that can be potentially reconnected; Bleu = end-userreconnected with respect to total that can be potentially reconnected).

a b

our investigation. Since tenants were not allowed back to their housesor apartments, most landlines were not used. Hence the demand on thiscircuit became much lighter.

7.6 Temporary housing

The Italian government organizations and NGOs (Non-GovernmentOrganization) were to be commended on a great effort providing the vic-tims with relief services and care. The military and fire brigade set upservice camps to provided needed services to the victims. Some of therelief campsites provided the victims with Internet services in additionto daily necessities such as medication, food, and water. In general thevictims were very satisfied with the relief service. Many residents wereafraid to get back to their houses even when their houses (marked asclass A or B) were not condemned, due to their fear of future earth-quakes and the potential for damage to their homes. Temporary housingis scheduled to be completed by September 2009 (before winter arrives)for the victims, Figure 89. These houses will be on a base isolationsystem to protect residents from future earthquakes. There were morethan 30,000 victims settling in more than 160 campsites.

tracks that the transformers were supported to stop sliding, Figure 88a.This was done after the earthquake. However the steel angles seemedto be under sized. In the control house of substations, the batteries werenot anchored or tied to the racks, Figure 88b. There was no batteriesdamage reported at these substations. Some locations were withoutpower for three days, e.g. wastewater treatment plant.Telecommunication service performed reasonably well. It went off airfor a couple of hours right after the earthquake. Cellular phones seemedto be the main means of telecommunication in this small community.Although there was no reported damage to the physical equipment andequipment building, we saw a number of temporary cellular sitesdeployed within the earthquake impacted areas. The increase of cellsites might have reduced the circuit overload that commonly occursafter an earthquake. Both Fire Fighters and Police used their own radiosystem as the primary communication tool. Cellular phones were alsoused to compliment the radio system. With a good backup power gene-ration plant, their communication was not interrupted. The Fire depart-ment had three repeater stations, which were not damaged. A numberof landlines were damaged or severed, as repairs were evident during


Fig. 87- (a) Damage to rigid connection of a transformer. (b) Typical damage to pedestal box.

a b

Fig. 88- (a) Steel anchors installed after the earthquake to avoid sliding of transformers. (b) Unanchored batteries’racks insubstation.

Fig. 89- One of the relief campite in L’Aquila set up by the Civil Defence.

a b

REFERENCES

C.F. Carocci, S. Lagomarsino, Masonry Buildings in the historic centers of theL’Aquila area, Progettazione Sismica, 2010. IUSS Press, Pavia.E. Cosenza, G. Manfredi, G.M. Verderame, Reinforced concrete build ings,Progettazione Sismica, 2010. IUSS Press, Pavia.M. Menegotto, Observations on precast concrete structures of industrial buil-dings and warehouses.M. Di Ludovico, G. Di Pasquale, M. Dolce, G. Manfredi, C. Moroni, A. Pro ta,Behavior of scholastic buildings after L’Aquila earthquake, Proget tazione

Sismica, 2010. IUSS Press, Pavia.C. Casarotti, A. Pavese, S. Peloso, Seismic Response of the San SalvatoreHospital of Coppito (L’Aquila) during the 6th April 2009 earthquake,Progettazione Sismica, 2010. IUSS Press, Pavia.B. Faggiano, I. Iervolino, G. Magliulo, G. Manfredi, I. Vanzi, Post-event analy-sis of industrial structures behavior during L’Aquila earthquake, ProgettazioneSismica, 2010. IUSS Press, Pavia.M. Dolce, S. Giovinazzi, I. Iervolino, E. Nigro, A. Tang, Emergency Ma na -gement for lifelines and rapid response after L’Aquila earthquake, ProgettazioneSismica, 2010. IUSS Press, Pavia.


should actually be considered permanent, since they had a lifetime lon-ger than 10 years (ignoring the fact that apparently between 10 and 50years works can neither be called provisional, nor permanent).If then the provisional does not exist from a durational point of view, itwould be useful to wonder whether it makes sense that it would existlooking at energy consumption, sustainable environment or pollution. Itwould also be useful to wonder whether buildings could be constructedwith environmental characteristics and safety level similar to that requi-red for permanent ones on a temporary basis and with cost per unitsimilar to provisional ones. If this should be the case, it would be logi-cal to propose to build provisional houses with characteristics of thepermanent ones. These ideas and others were discussed in the days directly following theAquila earthquake with Guido Bertolaso for the political, administrati-ve and economical aspects, with Mauro Dolce, Edoardo Cosenza andGaetano Manfredi for the technical and scientific aspects.A first complete conceptual proposal, with 3D-rendering and prelimi-nary calculations was submitted on April 16th, together with severalcomments. It was hypothesized to deliver the buildings for 3,000 inha-bitants within 5 months, guaranteeing seismic safety by means of an

THE IDEA

What is the time difference that distinguishes a temporary or pro-visionally home from a permanent or final? It is not easy to

respond to this question, if you consider the seemingly enduring eter-nity of what in Italy is built with the objective to last for months, or fora maximum of few years.With reference to Italy, it is enough to consider what happened after theearthquakes of Belice and Irpinia (or even in Friuli), there is thereforeno need to further elaborate the concept. On the other hand, we could refer to the technical code of 2008 [1], inwhich the nominal lifetime of a structure is defined as the number ofyears in which the structure – normally maintained – can be used for thepurpose it was built for, it is indicated in a table and it needs to be spe-cified in the design documents.It is interesting to note that the code only indicates a maximum for pro-visional works (10 years) and two minima for ordinary and importantworks (50 and 100 years respectively). If one sticks to these data, itshould be concluded that all the provisional works that were construc-ted in the aftermath of the earthquakes that took place after WWII

Reconstruction between temporary anddefinitive: the CASE project

Fig. 1- One of the first sketches of the project illustrating the logic of the buildings constructed on isolated plates.

Fig. 2- A plan sketch made by architects Ragazzi e Hoffer as to illustrate the logic of the infrastructure in a courtopen to pedestrians.

Fig. 3- One of the many 3D renderings used to illustrate design hypotheses.

1

3

2


isolation system at the level of an urban block, and proposing elevatestandards of living, technology and environment protection. The pursuitof these objectives, apparently impossible, was based on the construc-tion of large isolated plates and the subsequent assembling of pre-fabri-cated three-storey living units. The need for the project to be as muchas possible independent from local soil conditions and from the un -known construction technology (many different ones would have beennecessary to meet the deadlines) became immediately clear. To this endit was stated the need of urgently identifying the possible building tech-nologies compatible with the timing programme and the technical con-straints, of selecting technical and commercial partners and of explo-

ring the production capacity of the market. The time programme wasdefining in four weeks the date to open the construction sites, i.e. tostart construction by mid-May, to deliver houses to 3,000 inhabitants bySeptember. The economic analyses indicated an estimated cost of 120 million euro,VAT excluded, for 3,000 inhabitants, with a 20% uncertainty rate andwithout considering furniture, purchase of the terrain and photovoltaicinstallations. In preliminary calculations it was assumed to use friction pendulumdevices [2-8], with a radius of curvature of 4m, a vibration period of 4s,a displacement capacity of about 300 mm, a friction coefficient between

Gian Michele Calvi1 and Vincenzo Spaziante21 Eucentre Foundation - Centro Europeo di Formazione e Ricerca in Ingegneria Sismica, Pavia.

www.eucentre.it2 Department of Protezione Civile, Rome. www.protezionecivile.it

Fig. 4- Images used in the preliminary phase to illustrate possible technologies for the assemblage of the buildings.

3 and 5% and an equivalent viscous damping between 20 and 25%.The alternative of using rubber isolators was also taken into considera-tion, but appeared in this specific case to be less competitive, conside-ring the relatively low axial forces and the large horizontal displacementdemands.In the days immediately following, several aspects that would have per-manently defined the project were discussed and clarified: - The reduction of each one of the isolated plates to about 20 by 60 m,suitable to sustain three-storey buildings with each floor surface ofabout 600 m2, with a capacity of about 80 inhabitants in 25 to 30 apart-ments. Plates of this dimension should allow an adequate flexibility inrelation to the plane altitude conditions of the areas to use (at that


moment unknown) and the construction technologies, also unknown; - The definition of 150 as the approximate number of plates to constructand therefore of about 12,000 inhabitants to settle in; - The division of the intervention in numerous small villages, consistingof 4 to 20 plates and hence a number of inhabitants between 300 and1600;- The definition of a serial timesheet in which a group of 30 platesshould be finished about 15 days after the previous group, whichimplied a forecast of delivery of the apartments in 5 tranches for 2,400inhabitants a time, with deadlines spread out between 30 Septemberand 30 November;- The decision to manage the entire project directly, without interven-tion of a general contractor, setting up a non-profit technical structurethat responds directly to the Civil Protection Department (DPC). It wasthought that this way it would be possible to save substantial economi-cal resources, mainly on general additional costs and to have a moreaccurate control on deadlines and quality of the project.

THE ORGANISATION

The definition of an operational, management and outsourcing structu-

re, of personnel roles, activities and their interaction, time programmeand milestones required several days of intensive work and was com-pleted and formalised by May 8. The way the project is managed is veryinnovative with respect to the schemes that are normally adopted, andnot only in Italy. In fact a single–purpose consortium was created(named ForCASE), formed by Eucentre (a non-profit foundation, centreof competence for seismic risk of the department of civil protection,founded by four public institutions and with a nature of ‘public com-pany’ in Europe) and two construction companies, (ICOP and Damiani).The two companies agreed to operate in this context as non-profit enti-ties and not to participate in any other reconstruction activity inAbruzzo. Their role would have been that of a technical office, and the-refore to facilitate the consortium to act on behalf of the CPD as a gene-ral contractor, with the capacity to manage directly supplies purchasing,to coordinate activities on the construction site, to arrange and verify allaccounting matters.Obviously, the consortium had as well the main task of carrying out alldesigning and construction management activities, under the responsi-bility and coordination of the authors of this article. Coherently withwhat has been briefly described, the operational organogram demon-strates five main areas of actions: two being related to design activities,

Fig. 5- The personnel and work organization plan set up in the preliminary phase.


pletion of the project;- Finalisation and publication of call for bids;- Stipulation of contracts;- External checking and control; - Relations with institutions and obtainment of permissions;- Identification of the intervention construction sites, expropriations oflands and related activities.

INFRASTRUCTURES AND ARCHITECTURAL DESIGN

The architectural project of a building unit, as briefly presented, favou-red the development of different types of apartments, as a function offamily compositions, which resulted in 109 different shapes after theselection of 16 contracting bids, as discussed later.Regarding the choices on infrastructure, it needs to be highlighted thata first guiding concept was that of placing the settlements in the neigh-bourhood of existing villages that had suffered severe damages becau-se of the earthquake, to be able to relocate the people within their ownterritory, to preserve the close ties that people have with land and neigh -bours. This general principle was confronted with technical difficulties deri-ving from non-ideal geomorphic, hydrological and geotechnical cir-cumstances of the areas, to finalise the best possible selection of theareas of intervention. Once the settlements had been defined and sized as a function of quan-titative needs and land capacity, considering the dimensional andmorphological characteristics of the location, the problem of existinginfrastructures (roads, pipelines, sewing system, etc.) and of theirimprovement and integration had to be faced. Finally, the population indices could be defined, starting from figuresbetween 100 and 150 inhabitants per hectare, for location in more ruralor more densely populated areas. Such figures imply a rather sparse set-tlement typology, marked by large green areas.A final infrastructure index had been identified by assigning 30% of theland surface to services and facilities, such as leisure, sport, shoppingcentres or education and religious structures.Based on these premises the final urban design of the areas was com-pleted, obviously combining the building units previously described(essentially consisting of three inhabited floors above a covered park -ing), also considering exposure to sunlight, valley and mountain views,steepness of terrain.Driveways and walking paths were kept separated to the maximum pos-sible extent, generally locating vehicles roads on the outer skirts of eacharea, with access limited to parking lots and ground floors of the buil-dings, also used as parking. The walking paths were designed elimina-

two to management and accountancy activities and one to project coor-dination. In order to obtain maximum efficiency, in terms of time and costs, andto ensure quality control, three different operational modalities for con-tracting and execution of the work were identified:- For the activities of preparation of the construction site and infra-structure works, it was decided to mainly use local contraction compa-nies;- For the foundation and isolation systems, it was opted to act directlyas a general contractor acquiring materials and supplies, such as con-crete, welded wire meshes, steel columns, isolation devices, formworkpositioning, concrete casting, etc.;- For the construction of the housing structures it was decided to launcha call for bids that included final design and global construction,allowing the use of any building technologies compatible with the needsand available time, and selecting the proposals with the highest qualityand the lowest cost. The economic quantification of the costs for the management of all acti-vities was estimated based on the pure cost of the staff assigned to thistemporary job (in months), on a monthly cost, in general between 3,000and 12,000 euro (these are costs for the consortium, not net salaries),and on a sum to cover cost of accommodation and travel, that could inany case never exceed 3,000 euro per person per month.As all the activities would be executed within the non-profit frameworkthat characterizes the Eucentre Foundation and the ForCASE consor-tium, the estimations were considered as a maximum not to be excee-ded, while the real costs would be subjected to accountability checks. The Department of Civil Protection would directly execute, in coopera-tion with the consortium, all the activities related to:- Definition of Civil protection ordinances, possibly needed for the com-

Fig. 6- A simplified version of the extremely detailed and complex time schedule that allowed daily overviews on eachaspect of the project and the construction.


Fig. 7- An example of the plans that weredesigned for the bids of housing con -struction, with an underground parking bet -ween the two plates.


Fig. 8- Location of the construction sites, all within the municipality of L’Aquila.

Fig. 9- Examples of the urban plans for some site.

Bazzano

Preturo Sassa

Coppito 2

Fig. 9- Examples of the urban plans for some sites.

in which H is the height of the building and Cs is 0.05 for wall structu-res, 0.075 for reinforced concrete frames, 0.085 for steel frames. It ishowever well known that equations of this sort tend to underestimate thereal vibration period resulted from a secant stiffness to yield, that for theexamined buildings could arrive at values between 0.8 and 0.9 s [9, 10].Based on these considerations, the design period of vibration of the iso-lation system was selected in the range of 4 s. It was also preliminarily observed that even an extreme temperaturevariation of ±30 °C, leads to variations in length of about 8,5 mm oneach side of the axe of symmetry, that would not induce excessive hori-zontal loads into the columns.

Seismic actionSeismic action and in particular spectral demands in acceleration anddisplacement are discussed in detail elsewhere in this volume [11].Here it is however important to note that the fundamental parameter tobe assessed for a proper design of the isolation system is the maximumdisplacement demand at a period of about 4 s. The spectra derived fromthe registrations of April 6th show generally displacement demand ofless than 120 mm, with one exception, the AQK registration, in whichspectra values are close to 250 mm. The code spectra for events withreturn periods of 1000 years, to be used for the design of the isolationsystem, have values of about 300 mm for soil type B and 400 mm forsoil type E. These values can be significantly reduced in presence ofenergy dissipation, as a function of an appropriate equivalent damping,according to the � factor:

where � is the equivalent viscous damping value, that could be in theorder of 10-15% for rubber bearings and of 20% for friction sliders. Thevalues obtained from the reduction coefficients are between 0.6 and0.7, with consequent estimations of displacement demands of about 250mm for soil type E. For the non-linear analyses the code spectrum for vertical actions hasalso been considered, while for the building phases it was defined as a‘construction event’ consistent with what indicated in addendum A ofEurocode 8, part 2 [13]. Such an event appeared to be consistent withregistrations corresponding to a magnitude of 4.0, and was thus consi-dered reasonable. While the demand in terms of acceleration was signi-ficant (in the order of ag = 0.10 on stiff soil), the displacement demandwas negligible. Eight sets of spectrum compatible accelerograms have been used fornon-linear analyses, derived from registrations made in L’Aquila (3records), and during the events of Imperial Valley in 1979, Loma Prieta

� = � 105+�


ting all architectural barriers, connecting green areas and inhabitedlevels to road systems and parking lots with external elevators whenneeded.The final character of the new settlements tried to combine in an opti-mal way people needs, environmental and landscape requirements, useof existing infrastructures and construction of new ones, in an integra-ted vision. Later, another problem had to be faced, i.e. how to combine each one ofthe specific building units (at this stage 150, in 20 different locations)to each one of the 16 different typologies proposed by the companieswho won the call for bids. Choices had to be made in relation to con-struction technology and material, external aspect, number of buildingsawarded to each company, construction plan and schedule proposed byeach company.Finally, considering the high environmental value of the landscape, thedesign and realisation of the green areas was again the subject of apublic, international call for bids, where again cost, time and quality ofthe proposal were considered to select the winning bids.

STRUCTURAL DESIGN

Preliminary considerationsThe structural design of the buildings constitutes the fundamental ele-ment that allowed the development of the entire project and is extre-mely simple in its basic logic: two reinforced concrete plates, separatedby columns and isolators, the lower one being in contact with soil andthe upper one with the building. The plates were designed withoutknowing the local soil properties, nor the weight and plan distributionand structure of the buildings. Therefore for both aspects conservativeassumptions were used, to be verified later. In a few cases, some poten-tially selected construction location had to be discarded because thesoil properties appeared to be unsuitable.It should be noted that the two plates are characterized by similar flexu-ral actions induced by gravity, if it is assumed a uniform distribution ofthe building load and of the soil reaction. Preliminary evaluations, basedon a column span of 6 m in both directions (convenient for park ing arran-gements), lead to a required thickness of both plates of 500 mm. The weight of each building, with three floors of about 600 m2 each, wasestimated in a maximum of 21 MN, with a consequent total maximumweight of slab, dead loads and building details of between 30 and 40MN (or an average weight per column of about 1 MN). The first vibration period of the building can be estimated betweenTs = 0.25 and Ts = 0.45 s, using the equation:

Ts � CsH0,75

soil properties.Obviously other combinations may be possible, also related to thevarious displacement demands for isolators placed in different positions(because of the eccentricity of the loaded mass, even only accidental,the demand at the perimeter is larger than that closer to the slab cen-tral area). It was therefore allowed to bidders to propose different solu-tions, provided that they were respectful of design performances andinput. The result of the call for bids, in which FPS systems were pre-ferred, should not be considered as a general demonstration of superio-rity with respect to elastomeric devices, but rather as a consequence ofthe specific conditions of this project, characterised by relatively largehorizontal displacement demands, low vertical forces on the devicesand relatively low horizontal stiffness (as discussed, vibration periods ofthe order of 4 seconds were assumed). This was the reason why elasto-meric isolators had to be coupled with pot bearings: the use of rubberbearings alone would have resulted in stiffness values incompatiblewith the requirements of the project. In the case of the FPS devices, the force corresponding to a displacedposition is defined by the following equation:

In which Mg is the axial action (M is the mass and g the acceleration ofgravity), R = 4 m the radius of the spherical surface, � = 3% is the fric-tion coefficient and d the displacement of the isolator. The least favourable conditions for the verification of displacementcapacity of the isolation system versus the corresponding demand arelikely to be those of a rigid and heavy superstructure, i.e. those of alarge participating mass and deformations concentrated in the isolationsystem. With a configuration of this sort, the system global characteri-stics (40 pieces) resulted to be as follows.

F = Mg� + dMgR( )

in 1989, Northridge in 1994, Kobe in 1995 and Taiwan in 1999 (one foreach of these events).

Isolation systemThe design and the verification of the isolation system was carried outconsidering the possibility of adopting two different configurations, cha-racterised by different devices, one based on the use of 12 elastomericisolators, together with 28 multi-directional sliding pot-bearings andthe other on the use of 40 isolators sliding on spherical surfaces, uni-versally known as friction pendulum [FPS, 2].Both choices are compatible with the project requirements, in differentways. Actually, the smaller dissipation capacity of the system with ela-stomeric isolators (estimated to be equivalent to 12% damping) withrespect to the one with FPS isolators (estimated damping 20%,) requi-res a larger displacement capacity; in the order of 300 to 360 mm forthe elastomeric isolators, versus 260 mm for the FPS, depending on the


Fig. 10- Acceleration and displacement spectra of an event with a 1000 year return-period in L’Aquila, according to the Italian code [1], soil category B and E, damping 5%.

Fig. 11- Comparison of several spectra recorded on April 6th on soil type B, code spectra for an event with a 1000 yearreturn-period according code [1] and results of a recent research project (DPC-INGV-S5 [2]).

Effective stiffness, secant to the design displacement:

Keff = 14615 kN/m

Corresponding period of vibration of the isolation system (note that ingeneral heavier structures are also stiffer, therefore characterised bylower vibration periods):

Corresponding equivalent damping:

�FPS = = 0.201=20.1%2��M�g��Keff�d

T = 2� =3.29s � MKeff

Verification of slabs and columnsThe foundation and isolation plates have been subjected to numerousfinite element analyses, that allowed to calculate the maximum bendingand shear demand levels for several load combinations, to design thereinforcement, generally made by welded wire meshes to favour a fastpositioning, and to verify the resulting action combinations with appro-priate strength domains. Local verifications for loads concentration at the column ends were alsoperformed on both slabs, considering the consequences of the substitu-tion of a bearing as well. This operation was needed in hundreds of caseduring construction, when the isolators were not yet available at thetime of casting the upper slabs. The columns have been designed and verified considering either thecase of reinforced concrete and of steel, again to allow the use of various


Fig. 12- Force – displacement response of a system of 40 isolators and heavy superstructure. Fig. 13- Force – displacement response of the system considering axial force variation due to vertical acceleration and globalinteraction response [2, 4].

Fig. 14- Examples of displacement histories for an elastrometic isolator (left) and for a FPS isolator (right), subjected to events with a 1000 year return period derived from 3 registrations in L’Aquila, compared with capacity circles of 360 mm (left) and260 mm (right).

tions on the characteristics of the buildings needed to be defined, inorder to keep them within the parameters assumed for the analyses andtherefore avoiding unexpected responses and jeopardizing the verifica-tion of plates, foundations and isolation system. A summary of theseprescriptions follows:1. The load resulting from the building structures shall not induce inany element of the slab – foundation system local actions larger thanthose resulting from a uniformly distributed live load equal to 50 kN/m2

(i.e. excluding the slab self weight).2. The load distribution on the plates shall exclude concentrationspotentially resulting in local collapses.3. The maximum vertical action on a single bearing should be less than2800 kN, either for the seismic load combination and for the gravitycombination at the ultimate limit state, including the weight of theplate.4. Bearings shall not be subjected to tensile forces in any load case. 5. The main period of vibration of the building (considered fixed at thebase) shall not exceed 0.5 seconds. 6. The eccentricity between centre of mass of the building and centre ofmass of the plate shall be less than 5% of the total length of the plate(57 m) in the longitudinal direction, and less than 10% of the length ofthe plate (21 m) in the transversal direction.7. The maximum seismic mass of the building alone (i.e. without consi-dering the weight and loads of the slab), calculated including self weight, dead load and the fraction of live load to be considered for seis -

technologies and thus reducing the operation time. For the same reasonsteel columns were in general preferred, even if more expensive, usingconcrete only when no steel elements were ready to be mounted.

Prescription for building designAs already mentioned, the final design of the home buildings was left tothe bidders, to allow the use of any building technology. However, thespecifications to which the projects would have anyway to conform nee-ded to be defined as to assure an appropriate safety level to the globalstructural system. The seismic demand was defined in terms of design acceleration of thebuilding masses, calculated with reference to the maximum value of theratio between base shear and weight of the building, obtained in theworst loading conditions, corresponding to those of a stiff building (T =0.19 sec) with the lowest mass (1500 t).For analyses performed with accelerograms compatible with the designspectrum at a collapse limit state (SLC, return period � 1000 years), thebase shear always resulted less than 0.11 times the weight of the buil-ding. It was therefore prescribed to assume a design acceleration equalto 0.1 g to verify the buildings at a life safety limit state (SLV, returnperiod � 500 years). For the same conditions, the average inter storey drift resulted on therange of 0.1%, a value that certainly allows a full use of the buildingseven after a high intensity event. Together with these extremely simple design data, a series of prescrip-


Fig. 15- Examples of reinforcement distribution in a section of the isolated plate.


Fig. 16- Example of bending stresses in the foundation plate, for gravityloads (1st row, moments around the two axes of symmetry in kNm/m)and for seismic loads (2nd row: maximum values, and 3rd row: minimumvalues).

Fig. 17- Example of local reinforcement of the foundation plate at column bases.

Fig. 18- Example of bending moment – axial action strength domain, for a section of plate (atcolumns centres).

Fig. 19- Example of bending action on the isolated plate during bearing substitution at differentlocations.

17

18

19

for bids for excavations, supply of concrete (initially about 200,000 m3,with peaks in delivery of more than 5,000 m3 per day, self compactingand aerated), supply of welded wire meshes (initially about 260,000 kN,in general with diameter 14 mm at 100 mm), supply of steel columns(initially 180,000 ton, diameter 800 mm), supply of isolators (initially6,000 pieces, including assistance to positioning) and supply of castingforms (initially for about 336,000 m2) and on-site assistance for reinfor-cement positioning and pouring of concrete. All quantities were latersignificantly increased, since the number of buildings passed from 150to 184. The prices per unit obtained through bidding have been the fol-lowing:

mic verification shall be less than 2100 t. 8. The buildings shall be designed in accordance with the technicalcode of 14/01/2008. It is accepted to represent the horizontal load equi-valent to the seismic action by means of a static force vector, to beapplied to the building floors, according to equations given in the code,assuming a design acceleration Sd (T1) of 0.1 g.

CONSTRUCTION OF THE PLATES

As previously discussed, for the production of the plates, the ForCASEconsortium has directly taken the role of general contractor, with calls


Fig. 20- Reinforcement in a foundation plate with concrete columns and steel columns with isolators on a casted plate.

Contractors for the production of the plates with initial price and offers


Fig. 21- Rendering and floor plans of some buildings, proposed by the bidders.

the existing norms (that already represent a high standard). The maxi-mum time allowed for completing each building from the availabilityof the upper plate was fixed at 80 days, a proposed reduction was alsoconsidered in the evaluation, together with a reduction of the proposedprice.Following the presentation of the 58 proposals and an accurate review,16 contractors were selected, with a total average amount per lot ofabout 10,500,000 euro, which means an offered price reduction ofabout 5%.On a total of 150 buildings, timber structures were proposed for 75(50%), concrete structures for 45 (30%) and steel structures for 30(20%).

INFRASTRUCTURES, FURNITURE, ELEVATORS, MECHA-NICAL AND ELECTRICAL INSTALLATIONS, GREEN AREAS

To complete the project it was necessary to prepare and launch other5 groups of bids, in order to satisfy various needs:- The upgrading and integration of the external infrastructure(networks of any type) with the difficult problem of interacting with the

• Self-compacting concrete 82,55 €/m3

• Welded wire mesh 0,49 €/kg• Steel columns 2,09 €/kg• Isolators 1,427 €/piece• Forms and on-site assistance 91,7 €/m2

CONSTRUCTION OF THE BUILDINGS

A public call for bids was launched for the construction of the buil-dings; including final design. The 150 buildings to be built were grou-ped in 30 lots, each one of 5 buildings, allowing a bidder to present aproposal for a maximum of 10 lots. Depending on the final ranking of the offers, it might have been possi-ble to have from a minimum of 3 contractors (in case the first 3 wouldeach propose 10 lots) to a maximum number of 30 contractors (in caseeach one would have proposed 1 lot). The basic price for any lot of 5buildings (about 160 covered parking spots, 3,000 m2 of outside pave-ment and 9,000 m2 internal living area) was fixed at 11 million euro.The evaluation of the proposals was essentially based on the proposedimprovement of the minimum performance characteristics foreseen by


Contractors, structure material, number of buildings offered and price per building

Structure

Walter

e

di

construction sites. Twenty bids were released, one for each area ofintervention, inviting companies located in Abruzzo and preferably inthe province of L’Aquila. Five companies randomly sorted out wereinvited to bid for each site. - The furniture and supplies necessary to immediately use the apart-ments. In this case a public competition was set up, on four lots ofabout 1,000 apartments each. The foreseen time to assemble the fur-niture on site was 6 days from the moment an apartment would be fini-shed. 18 companies presented an offer, with the following four resul-ting winners: Deltongo Industrie spa, Mobilificio Florida srl, RTIEuropea spa – P.M. International Furnishings srl – Martex spa andEstel Office spa. The average price reduction offered was about 34%,which corresponds to an average cost for the interior furnishing of anapartment of 9,500 euro. It has to be underlined that the specifics ofthe bid requested the highest possible standards also for the electricaland mechanical equipment included in the offer, such as dish washer,washing machine, tv set, etc. - 309 elevators to connect the various floors of the buildings and 129elevators to connect the buildings to the parking ground floor. Thisneed derived from the specific choice of completely eliminating allpotential architectonical barriers case, in excess of what compulsoryfor legal requirement. A call for bids was released for three lots of 146elevators each; 12 companies participated in the competition.Marrocco elevators srl, ATI S.A.S. srl – Grivan Group srl, Schindlerspa resulted winners with an average reduction of price of about 16%.- The opportunity of producing electric energy on site, collocating pho-tovoltaic panels on the roofs. An estimate of about 45,000 m2 of roofsurface was considered adequately exposed to sunshine and conse-quently another bid was released for the design, construction, mana-

gement and maintenance of a photovoltaic system, capable of produ-cing about 4,500 kW. The call for bid assumed that there should havebeen no cost to the administration, and was based on technical meritand on one fundamental economical parameter, i.e. a yearly fee to bepaid, as a percentage of the public incentives provided to favour theuse of alternative, renewable energy sources. The winner and therefo-re contractor was Enerpoint spa, Ener Point Energy Srl and Troiani &Ciarocchi Srl., who offered to refund 9,01% of the incentives.- Finally, two last calls for bids were launched to complete the greenareas, simply grouping the eastern and western construction sites. Theoffers should obviously include land preparation, grass, bushes andtrees, walking and cycling paths, but also irrigation and drainagesystems, external furniture, sport and leisure fields. 19 companies par-ticipated in the bid and contractors were selected on cost and on eva-luation of landscape beauty, environmental sustainability and mainte-nance and management characteristics of the offers. The selected con-tractors were 3A Progetti S.p.a, which lowered the estimated price of39% and the Sestante Consortium, that managed to offer a 35,16%reduction.

WORK MANAGEMENT, QUALITY CONTROL, SAFETY MEA-SURES

The extremely limited time available for the completion of the projectrequired an extreme level of control and programming, with a conti-nuous flow of information between engineering work management andconstruction companies and daily reports and checks. A coordination and management system was therefore set up, focusingon the definition of priorities and of the main activities consequently


Fig. 22- Same complete buildings.


Fig. 23- Work in progress and completedworks.

required and on the identification of potentially critical elements andwork phases that could threaten the fulfilment of the programmed timeschedule. As previously discussed, the work activities and their management andcontrol were organised in five sectors, corresponding to production ofthe plates, buildings (including interior design and furniture), mecha-nical and electrical installations, infrastructures roads and green areas.For each technical sector a technical coordination structure was defi-ned with a responsible for programming, coordination with the con-struction companies and management of works. The general timeline ofthe works was accordingly subdivided into the same five sectors. Dailyupdates on the work progress and comparison with the time planningguaranteed that each sector was closely monitored in terms of work pro-gress, as well providing all technical personnel with an overview of thegeneral picture, fundamental to manage the coordination between dif-ferent sectors. The graphical visualization of the daily progress of theworks resulted to be particularly useful for a rapid interpretation of thecomplexity of the data that needed to be managed. An idea of the large quantities of materials and labour that needed to bemanaged, can be obtained by considering the example of the foundationand isolation system, that on a daily basis needed an average concrete


Fig. 24- Example of a global daily overview form.

Fig. 25- Example of a production summary overview daily form (25 October, general overview on the left, production and delivery of plates on the right).


Fig. 26- Example of a daily report on the general development of the projectworks (19 October).

Fig. 27- Example of a daily report on the general developments of the works ina specific area (19 October).

supply of the order of 5,000 m3, welded meshes and reinforcing bars forabout 10,000 kN, about 200 steel columns (with a diameter of 800 mm),to be provided in general in 20 different construction sites. The efficiency of the team that was set up to program and coordinatethe activities, allowed such a proper and precise forecasting of thework progress that all construction sites proceeded always on time andactually all works were completed ahead of time, despite of the diffi-culties inherent in the number of workers (more than 8,000 in somephases) and in the complex interaction between different work activi-ties. An example of the time schedule programmed for the constructionof the two slabs systems, with foundation and isolation, can be sum-marised in Fig. 23. Finally, the great efficiency of the team in charge of controlling allaspects of safety in the work process should be noted. The extremelydetailed and continuous checks allowed the completion of hundreds ofmillions of euro worth of work in just a few months without any nota-ble accident and with the appreciation of all external controlling insti-tutions, from those aimed to assure workers health to the unions.

THE COSTS

The total cost of the project is split in the table below, considering the

different category of work and giving the average cost per building, perapartment and per square metre of living space.The total cost of 655 million euro refers to a total of 164 buildings,while 150 were foreseen at the starting of the project and 184 wereactually built at the end. Applying a criterion of linear proportion it canbe inferred that the original 150 buildings would have cost 599 millioneuro, which is in line with the 700 million that were initially estima-ted, since the sums indicated do not include the cost for land expro-priation and V.A.T. The cost of the double slab foundation system iscompensated to significant extent by the value of the covered parkingspots, each one of them have the size of a large garage box (6 by 3metre). The number boxes exceed that of the apartments. It is thus rea-sonable to assume that the real cost of the foundation system is actual-ly a fraction of that indicated in about 30,000 € per apartment. If afraction of 30% would be assigned to the foundation itself, the cost persquare metre of the living space should result to be less than 1,400 €,including foundations: a reasonable value, especially considering thecompressed time of construction, that was made possible by a conti-nuous work over the 24 hours, with three turns of 8 hours each, day andnight and considering as well the very high quality of the buildings foraspects related to energy consumption, environment and detailing qua-lity.


Parametric costs of the whole intervention, based on 164,29 equivalent buildings, v.a.t. not included

will be delivered within February 2010, with the possibility of notableanticipation if the meteorological circumstances should be favourable. The delivery of the houses started Tuesday 29 September with about500 apartments and will continue at a pace of approximately 300 apart-ments a week. The property of the buildings will be eventually assignedto the city of L’Aquila who will be responsible for the management andmaintenance based on pre-defined procedures, specified in detail in theproject documents. Political and economical choices, with relation tothe progress in the repair, strengthening and reconstruction of the buil-dings damaged in the historical centre and in the city outskirt, will drivethe decision on rent costs and use of the new villages. The users of the houses are being carefully selected jointly by the city ofL’Aquila and the DPC, taking into account the preferences expressed bythe homeless people, parameters connected to the family situation (num-ber of components, age, economical capacity, etc.) and the localisation ofthe original place of living. A prerequisite to be considered is that afamily previous home should have been classified in the category ‘noneasily to be repaired’ (type E and F in the classification of damage).

It is interesting to note that the seismic isolation cost only about 1,5%of the total, or rather just a bit over 2% when only the building cost isconsidered.The modest cost of general and technical activities has to be noted,made possible by the way the project was managed, extensivelydiscussed in the previous sections. The real share of the technicalcosts of the ForCASE consortium (design, management, security, etc)has been around 8 million euro, i.e. not much more than 1% of the totalcost. The costs of the furniture includes everything, from TV sets tobed sheets.

THE FUTURE

At the moment this article is being completed (September) the last 20buildings are being constructed, with a significantly lower cost becau-se they are built in already inhabited areas. The decision of the CivilProtection Department to build additional houses was motivated byupgraded in the population census. It is foreseen that these buildings


Forecast of the number of apartments and beds available in function of the foreseen completionperiod (forecast of September 22nd, including the twenty buildings that had just been added, ofwhich forecasts are cautious)

The houses will anyway become part of the city’s heritage and in thefuture it will be therefore possible to reuse them to host vulnerable cate-gories of population (such as the elderly people) or to host students, aneed particularly relevant in L’Aquila, where a significant fraction of

the 25,000 students come from other regions. In the near future the availability of student housing at a controlledprice could become a relevant peculiarity of the university, modifyingits attraction capacity in a positive way.


Those who have contributed to the success of the C.A.S.E. project:

Department of Civil ProtectionGiacomo Aiello (legal responsible),Emilia Aloise, Giovanna Andreozzi, EnricoArdito, Vincenzo Ardito, Arianna Bertelli, Mariano Bonvegna, Angelo Borrelli(administration responsible), Fabrizio Bramerini, Cristina Capriotti, Maria TeresaCartolari, Mario Cera, Claudia Ciccone, Pietro Colicchio, Alessandra Conti, MarcoConti, Maria Laura Conti, Chiara D’angelo, Antonella De Felice, Giulio De Marco,Giovanni Di Achille, Giovanni Di Mambro, Mauro Dolce (technical responsible),Riccardo Fabiani, Maria Cristina Ferroni, Salvatore Fiengo, Claudia Fiore,Mariasilvia Gianneramo, Beatrice Guerra, Gerarda Iannarone, Federica La Chioma,Luisa Marinaro, Lucia Palermo, Francesca Paneforte, Ada Paolucci, RobertoPesolillo, Giancarlo Piccione, Patrizia Picuti, Immacolata Postiglione, GiuseppinaSementilli, Vincenzo Spaziante (general coordinator), Tiziana Tarduini, VergilioTidei, Fabiola Toni, Angelo Vici.

CASE ConsortiumFabio Aldrovandi, Francesco Ambrosi, Francesco Amici, Maurizio Ardingo (respon-sible for construction safety measures), Luciano Baglione, Giovanni Bastianini, PaoloBat tegazzore, Giuliano Bellini, Maria Teresa Dolores Bertelegni, Federica Bianchi,Saverio Bisoni, Gaia Boggioni, Filippo Bonali, Barbara Borzi, Maria BenedettaBossi, Matteo Bottari, Vittorio Bozzetto, Roberto Brandimarte, Piero Burba,Maurizio Calde rari, Andrea Caligari, Maura Castellani, Gian Michele Calvi (projectleader, designer and construction director), Salvatore Caroli, Christian Caroli, PaoloCaroli, Francesco Ceribelli, Antonio Coccia, Andrea Colcuc, Oliviero Comand,Massimiliano Cordeschi, Filippo Dacarro, Michele D’Adamo, Alberto Damiani(responsible for building construction), Pietro Damiani, Edi Danielis, Simonetta DiNicola, Maurizio De Santis, Pasquale Di Marcantonio, Dante Di Marco, StefanoD’Ottavio, Ettore Fagà, Mario Fanutti, Carlo Florio, Pierluigi Fontana, FabrizioFrau, Renato Fuchs (organisation coordinator), Nicola Gallina, Marco Gasperi,Fabio Germagnoli, Federico Gianoli, Daniele Gimnetti, Sergio Giordano, StefanoGrasso, Carlo Lai, Massimo Lar dera (responsible for infrastructure), Ignazio Locci,Giuseppe Lombardi, Mauro Ma ga netti, Giovanni Magenes, Claudio Maggi, CarloMagni, Fabrizio Magni, Michele Magnotti, Gabriele Mantini, Antonio Marcotullio,Paola Marotta, Sara Martini, Ema nuele Mea go, Paola Migliazza, Enrico Misale,Marta Molinari, Federico Monutti, Matteo Mo ratti (site responsible for structures),Vincenzo Pane, Vincenzo Paolillo, Alessandro Pa pale, Carmine Pascale, PierluigiPascale, Moreno Pavan, Fausto Pedet ta Peccia, Gian franco Peressutti, EdoardoPeronace, Michele Pescina, Paolo Petrucco (site responsible for plates and infra-structure), Piero Petrucco, Nereo Pettenà, Dario Pietra, Ro berto Pitolini, FedericaPolidoro, Alessandro Pollini, Stefano Pozzi, Salvatore Provenzano, Bruno Quadrio,Nadia Rizzardi, Enzo Rizzi, Fabio Roiatti, Cristiana Ruggeri (responsible for mecha-nical and electrical installations), Gaetano Ruggeri, Mario Rusconi, DanieleSambrizzi, Valentina Scenna, Matteo Schena, Michele Schiabel, Paolo Scien za,Fabiola Sciore, Roberto Scotti, Domenico Sgrò, Martino Si gnorile, Danilo MarcoSiviero, Luigi Spadaro, Davide Tagliaferri, Piergiuseppe Tam burri, Alessandro To -sello, Stefan Trenkwalder, Roberto Turino (site responsible for buildings), Diego Ur -

bani, Marco Vecchietti, Paolo Verri, Stefano Vitalini (site responsible for plates), Ro -berta Viviani

Checker team, administrative aspectsGiovanna Andreozzi, Maria Laura Conti, Alessandra Conti, Michele D’adamo, Gio -vanni Di Mambro, Salvatore Fiengo, Giorgio Grossi, Emilia Aloise, Mariano Bon -venga, Carlo Bordini, Cristina Capriotti, Maria Teresa Cartolari, Carluccio Code -ghini, Fabio Compagnoni, Dario Compagnoni, Massimo Criscuolo, Antonella DeFelice, Giu seppe Fasiol, Maria Cristina Ferroni, Arturo Furlan, Achille Gentile,Alessandro Greco, Gerarda Iannarone, Giuseppe Ianniello, Giovanni Infante, EttoreIorio, Paolo Mar chesi, Luca Pagani, Lucia Palermo, Roberto Pesolillo, SalvatoreProvenzano, Ro sario Romano, Gianni Strazzullo, Fabiola Toni, Daniela Ursino,Michele Villani

Checker team, structuresEdoardo Cosenza, Gaetano Manfredi, Claudio Moroni, Paolo Pinto (President),Paolo Zanon (assistenti: Massimo Acanfora, Claudio D’Ambra, Antimo Fiorillo)

Companies and organisationsExcavations: CO.GE.FER. s.p.a.; Midal s.r.l.; P.R.S. Produzione e Servizi s.r.l.Concrete: Colabeton s.p.a.; Società Meridionale Inerti SMI s.r.l. Steel reinforcement:La Veneta Reti s.p.a. Steel columns: A.T.I. Edimo Metallo s.p.a. /Taddei s.p.a.;Cordioli & C. s.p.a.; Formwork and assistance: Consorzio Edile C.M. Gruppo Bison;Sacaim s.p.a.; Zoppoli & Pulcher s.p.a. Isolators: Alga s.p.a.; FIP Industriale s.p.a.Buildings: A.T.I. Consorzio Stabile CONSTA s.c.p.a./Sicap s.p.a.; A.T.I. Donatis.p.a./Tirrena Lavori s.r.l./Dema Costruzioni s.r.l./Q5 s.r.l.; A.T.I. Eschilo Unos.r.l./COGEIM s.p.a./Alfa Costruzioni 2008 s.r.l.; A.T.I. Ille prefabbricatis.p.a./Belwood s.r.l.; A.T.I. Impresa Costruzioni Giuseppe Maltauro s.p.a./Taddeis.p.a.; A.T.I. Iter Gestione e Appalti s.p.a./Sled s.p.a./Vitale Costruzioni s.p.a.;A.T.I. COGE Costruzioni Generali s.p.a. /Consorzio Esi; Consorzio Etruria s.c.a.r.l.;Consorzio Stabile Arcale; Cosbau s.p.a.; D’Agostino Angelo Antonio CostruzioniGenerali s.r.l.; Impresa di Costruzioni Ing. Raffaello Pellegrini s.r.l.; Meraviglias.p.a.; Orceana Costruzioni s.p.a.; R.T.I. Ing. Armido Frezza s.r.l./Walter FrezzaCostruzioni s.r.l./ Archilegno s.r.l.; Wood Beton s.p.a. Furniture: Del TongoIndustrie s.p.a.; Estel Office s.p.a.; Mobilificio Florida s.r.l.; R.T.I. Europeo s.p.a./PM.International Furnishing s.r.l. Infrastructure: CO.M.AB. Appalti Pubblici ePrivati s.n.c.; Codimar s.r.l..; Codisab s.r.l.; Conglo merati Bituminosi s.r.l.;Facciolini s.r.l.; G.C.G. s.r.l.; I Platani s.r.l.; Impresa Edile Di Cola Michele; Ing.Armido Frezza s.r.l.; Molisana Inerti Conglomerati s.r.l.; Produ zione e Servizi s.r.l.;Ridolfi Idio e Figli s.r.l.; San Giovanni Inerti di Pietro Mascitti s.r.l.; ValentiniCostruzioni s.a.s.; Elevators: Marrocco elevators s.r.l., ATI S.A.S. s.r.l./Grivan Groups.r.l., Schindler s.p.a.; Photovoltaic panels: R.T.I. Ener Point s.p.a./Ener PointEnergy s.r.l./Troiani & Ciarrocchi s.r.l.; Green areas: R.T.I. 3a Progetti/Gsas.r.l./O.Ci.Ma. s.r.l./Bellomia-Sebastianini-Euroengineering s.r.l., ConsorzioSestante. Demolition: CODISAB SRL, A.S.M. s.p.a.; Connections to external pipe -line networks: ENEL Rete Gas, ENEL Energia, GranSasso Acqua.1


REFERENCES

1. NTC (2008) - Norme Tecniche per le Costruzioni, D.M. 14/01/2008, Gazzetta Uffi -ciale 04/02/2008, Italia.2. Zayas V., Low S. (1990) - A Simple Pendulum Technique for Achieving SeismicIsolation, Earthquake Spectra, Vol. 6, No. 2.3. Almazan J.L., De la Llera J.C. (2002) - Analytical model of structures with fric-tional pendulum isolators, Earthquake engineering and structural dynamics, Vol.31, 305-332.4. Calvi G.M., Ceresa P., Casarotti C., Bolognini D., Auricchio F. (2004) - Effects ofaxial force variation on the seismic response of bridges isolated with friction pen-dulum systems, Journal of Earthquake Engineering, Vol. 8, SI1, 187-224.5. Christopoulos C., Filiatrault A. (2006) - Principles of Passive SupplementalDamping and Seismic Isolation, IUSS Press, Pavia.6. Priestley M.J.N., Calvi G.M. (2002) - Strategies for repair and seismic upgradingof Bolu Viaduct 1, Turkey, Journal of Earthquake Engineering, Vol. 6, SI1, 157-184.7. Tsai C.S. (1997) - Finite element formulations for friction Pendulum seismic iso-

lation bearings, International Jour. for Num. Methods in Engineering, Vol. 40,29-49.8. Wang Y., Chung L.L., Liao W.H. (1998) - Seismic response analysis of bridgesisolated with friction pendulum bearings, Earthquake engineering and structuraldynamics, 27, 1069-1093.9. Priestley M.J.N., Calvi G.M., Kowalsky M.J. (2007) - Displacement based designof structures, IUSS Press, Pavia.10. Crowley, H. and Pinho R. (2004) - Period-height relationship ofr existingEuropean reinforced concrete buildings, Journal of Earthquake Engineering, Vol. 8(SP1), 93-120.11. Crowley H., Stucchi M., Meletti C., Calvi G.M., Pacor F. (2009) - Uno sguardoagli spettri delle NTC08 in relazione al terremoto de L’Aquila, capitolo 1.7 in que-sto volume.12. AA.VV. (2007) - Definizione dell’input sismico sulla base degli spostamenti,progetto S5 INGVDPC, http://progettos5.stru.polimi.it.13. Comité Européen de Normalisation, Eurocode 8 part 2 (2006) - prEN1998-2,CEN, Brussels.

Here and in the next pages, some pictures of the completed buildings areas.







CONSTRUCTION

154 CONSTRUCTION - Civil engineering Works

Italian High-Speed NetworkA special focus on concrete structures

T he new Italian high-speed Network, due to the needs of increasing thecapacity of the actual railway operating lines, nearly doubling the

actual number of trains running daily, and decreasing travel time, has involv -ed in the last fifteen years, large engineering resources, construction skillsand strict planning, managing organization, railway engineers supervision,apart from the necessary economical huge investments.The main characteristics of the new lines and the strategic choices from theinfrastructural and structural points of view are presented in the next chapter. The new lines are designed according to the national code for the designand construction of railways bridges [1], and to the most advanced techno -logical standards in order to achieve the safest conditions of service, speedand interoperability with operating railways and with the European high-speed lines of railway transportation of passengers and freight.More than 90 % of bridges and viaducts of the new high-speed lines arerealised with simply supported spans of prestressed concrete (PC) decks,so attention will be paid on their typical structural solutions. Simply supported composite steel and concrete spans, few continuous brid-ges and some special structures (arch bridges and the cable-stayed bridgeover Po River [2]) compose the remaining part.Main concepts of durability of concrete structures required in the design ofrailway bridges are then focused.Construction processes are then described, taking examples from the late-st viaducts with PC decks.Finally, main concepts in the design of continuous PC bridges are outlined.

Strategic choices

Traffic analysis carried out on existing Italian railway network at the end ofthe 1980 remarked the following needs: quadrupling the main passengertransport routes, upgrading and increasing freight transport, reducing timeof travel for passengers trains, integrating Italian network with Europeannetwork.So, the first infrastructural choice was the realization of a new mixed pas-senger/freight high-speed network with a close integration with existinglines and with interchange centres (interports, ports, airports).The close integration with the existing conventional network will produce anincrease of freight transport capacity on the “historical” lines, clearing theexisting network from the long distance passengers traffic, and an increaseof freight traffic using new lines during specific time bands (usually at night).The choice of a mixed traffic meant low ruling gradients (less than 12 ‰)and heavy design loads (SW0/SW2) adopted in the new Italian standardfor railway bridges [1]. For this reason, the standard was rewritten in 1995,then revised in 1997. The old standards had been written 50 years beforeand were related only to conventional lines. The Italian standard for railwaybridges [1] has introduced LM71 (passengers traffic as showed in leafletsUIC 702 and 776-1), SW/0 and SW/2 (heavy traffic) models of loadsaccording to ENV 1991-3: Actions on structures, Part 3: Traffic loads on brid-ges (Ed. 1996). These new standards are in perfect agreement with the European TechnicalStandards for Interoperability of the trains in the European High-speedNetwork. One of the main prescriptions asks for a structural design respectful to allprescriptions for seismic areas (at least III category – the minimum consi-dered in the 1996 Italian seismic code), even in no-seismic areas, apartfrom Sardinia. A proper standard was written and recently revised for thedesign of railway bridges to be built in seismic area [4]. It deserves to bepointed out that many no-seismic areas became recently seismic, in thelatest proposals of codes, giving interesting confirmation to the conservati-ve railways code approach. According to general seismic design principles, the adoption of special rulesand technical details is requested to guarantee a minimum ductility of thestructure, and it has direct consequences on the care for the details ofreinforcement design of piers and foundations. Two main characteristics of the applied national code [1] are the conceptsof train-track-structure dynamic interaction and train-rail-structure staticinteraction.The dynamic interaction analysis is evaluated to check the safety of thetrain and the comfort of the passengers, with an analysis of the followingparameters: all decks for high-speed railway must respect the limit value of2.5 as maximum dynamic amplification of static deflection (“impact factor”j real=j dyn /d stat) and the value of the vertical acceleration at deck mid-span, induced by real trains running at different speed (from 10 km/h upto 1.2 maximum speed of the line), must be lower than 3.5 m/s2.For standard simply supported beam or continuous bridges with totallength shorter than 130 m, a simplified analysis can be adopted accordingFig. 1- Italian New High-speed Network: lines under design, construction and the operating line Roma-Firenze, built in the

70’s-80’s for a design speed of 250 km/h [3].

155CONSTRUCTION - Civil engineering Works

to Annex A of [1], with a preliminary check of flexural - frequencies of vibra-tion modes: the simply supported prestressed concrete decks alwaysrespect the limit values specified in [1], being first flexural mode frequencybetween 4 and 8 Hz. For non-conventional structures, as arch bridges, cable-stayed bridges, etc.,a “Runnability” analysis is required. The analysis has to consider all dynamiccharacteristics of the system: railway structures, suspension system of thevehicles, rail fasten system etc., track and wheel irregularities. The static interaction analysis studies the effects on rail and bridge struc-ture due to variation of thermal conditions in the structures, to the longitu-dinal forces associated to braking and traction, and to the longitudinaldisplacements due to vertical loads. For simply supported prestressed con-crete decks, which respect a maximum length of 65 m and small variationsof the longitudinal stiffness of piers and foundations, a simplified methodcan be adopted as indicated in Annex B of [1]. In any other case, it is necessary to analyse advanced Finite Element modelsto evaluate these effects. The analysis has to check rail stress limits withmaximum value of compressive stress 60 MPa and maximum value of ten-sile stress 70 MPa, the relative displacement between deck bridge and therail, and the forces acting on the bearings. Also reliability under service conditions is required: comfort limit state hasto be verified for a maximum midspan deflection with the load of oneLM71 load model, increased with dynamic factor. This deflection must notexceed l/2400 for design length of the span l<30m, l/2800 for30m<l<60m and l/3000 for l >60m. Maximum deformability of structu-res under train load is checked to keep the contact rail-wheel safe and sta-ble: deck torsion, rotation at supports and horizontal deflection have to beevaluated. The limits of deformation of the structures are similar to thosepointed out in the same Eurocode, and are widely respected by commonsimply supported spans. Special attention has been put in the concepts of durability of structures forrailway bridges, introduced in [1]. As the subject deserves wide illustrationand details, a full paragraph has been devoted to the scope.As high-speed network is designed for the use of long welded rail, the struc-tural system for the viaducts must avoid rail expansion devices. In all high-speed network, only along the Milano-Bologna line, for the crossing over PoRiver, composed by two continuous bridges and the cable stayed bridge [2],two joints in the rails have been necessary to keep the expansion lengthwithin allowable limits. Other rules and prescriptions for design and construction are taken intoconsideration in [1] and are illustrated in the following paragraphs. All theseare finalised to have low costs of maintenance of the infrastructure, to mini-mise the irregularities of the track and to reach a high performance levelin the field of the durability and reliability of the system.

Simply supported prestressed concrete bridges

Simply supported spans of prestressed concrete deck realised more than90% of the new lines. It is undoubtedly a traditional choice of Italian Rail -way Company (Ferrovie dello Stato – FS) for ordinary viaducts: to better fit

with long welded rail, to avoid rail expansion devices, to ease maintenanceoperations and minimize maintenance costs. Besides, this solution is usual-ly preferred in those cases when bridges have to be designed in areas withcompressible soils or in river channels.Figures and statistics of this paper are based on more than 600 km longhigh-speed lines. Attention will be paid to double-track decks, with a distan-ce between tracks of 5.0 m, designed for a train-speed of 300 km/h, andfor both heavy and passengers traffic load models [1], with rails on pre-stressed concrete sleepers on ballast. They count nearly 2300 spans of sim-ply supported prestressed concrete bridges, and they are composed ofnearly 6400 precast beams or monolithic decks.In order to reach this frame of prestressed concrete decks, analysis will con-cern the structural characteristics such as use of pre-casting, tensioningsystems, bearings, expansion joints, all durability issues as multi-layer pro-tection systems, monitoring, methods of construction and costs.Deck The pre-stressed concrete elements are realised with both pre-ten-sioning and post-tensioning systems. The post-tensioning systems are alwaysdesigned with bonded internal cables, even if Italian standard for railwaybridges [1], generally speaking and under severe controls, allows also exter-nal post-tensioning. According to [1], post-tensioning cables composed bybars should be preferred for viaducts along railways with electric traction ofdirect current, and both solutions with cables composed by strands or barscan be used in structures for railways with electric traction of alternatingcurrent as the high-speed network.In [1], special attention for durability and limiting or avoiding cracking ofconcrete is introduced: undoubtedly most limiting verifications deal with limi-tation of maximum compressive stresses and, especially, strong limitation oftensile stresses during construction and final conditions. In particular, no lon-gitudinal tensile stress in PC structures is admitted, with maximum designloads and both Allowable Stress or Limit States methods of verification. Besides, cracking of concrete must be verified towards no decompressionlimit state for verification under track equipment, where inspection is notpossible. The experience of existing railway lines with concrete structures with pos-sible beginning of corrosion of the reinforcement and spalling of the con-crete, which leads to easier access to the pre-stressing tendons for aggres-sive agents, has been translated into design prescriptions. The required con-crete cover to reinforcement, tendons and pre-tensioned strands has beenincreased, compared to Italian standard for design of structures. Minimumconcrete cover thickness is required to be 3 cm for PC decks, increased to3.5 cm under track equipment, at least one external diameter of duct incase of post-tensioning, and 3 strand diameters in case of pre-tensioning.Mix design of concrete for PC deck has to respect a 0.45 water to concreteratio, a S4÷S5 concrete consistency class of at least 45 MPa characteristiccubic strength. A quality assurance system and testing before and duringevery casting operation reveals the quality of mix design, which is recogni-sed as an important factor for life and durability of PC structures.Bearings and expansion joints Under simply supported railway brid-ges, only one kind of bearing is generally present: spherical bearings withpolished stainless steel and PTFE plate.


Italian High-Speed Network

With this kind of bearing, rotations can occur till ±0.0167 rad in all direc-tions, in order to place the bearing without inserting packings. In order toavoid parasite forces arising with only one train on a double line bridgedeck, a new kind of fixed bearings has been studied: it has a special devicewhich controls horizontal stiffness.The cover of expansion joints are realised with dielectrical elastomericcushion joints, composed by neoprene reinforced with vulcanized steel pla-tes. They allow fast bearings changing with a maximum differential lifting of50 mm between decks, operated by hydraulic jacks between deck and piercap (all simply supported or continuous decks have to pass this design veri-fication) without any operation under rails. Actually, the use of mechanicaldevices instead of resins avoids disease to daily train in case this operationbecomes necessary.On pier caps and abutments of every bridge span, reinforced concrete orsteel devices (“stroke end device”) are required in order to avoid deck slip-ping out of pier cap and falling, because of accidental breaking of fixed bea-rings e.g. in case of devastating earthquakes. There are pillows of reinforcedneoprene where decks may hurt against these provisions and their main-tenance or changing operations has to be assured by proper design. Italian standard for railway bridge bearings and cover of expansion jointsrequires these devices underpass preliminary homologation tests led by F.S.technicians, through prototypes testing, in order to assure quality of everysingle component and of the final assembled products.Piers and foundation Piers have usually circular or rectangular, full orempty, cross-sections, while foundations are usually realised with plinths withlarge diameter reinforced concrete piles.In case of piers in riverbed, even if empty structural sections are adopted,low class of concrete is always poured inside till the river maximum level, inorder to avoid unexpected water inside.As previously mentioned, in order to increase structural safety, all bridges aredesigned considering at least low seismic condition: it focuses the designers'attention especially on reinforcement details, very important for piers andpiles. Good number of stirrups and loops for longitudinal bars and concre-te confinement, use of hooks for good stirrup behaviour, limitation of maxi-mum compression stress in pier concrete, no junction or superposition oflongitudinal bars in the length of 3 m from foundation, etc. are consequen-ces of above-mentioned prescriptions.The minimum reinforcement areas for both piles and piers is fixed to the0.6 % area of concrete section, and spirals are admitted as stirrups inreinforced concrete piles only if welded to longitudinal bars in every inter-section.

Typical cross sections

The most common cross-sections of prestressed concrete decks are showedin Fig. 2, 6 and 11; in the following, a brief description of main features ispresented for each typical cross-section. Type “a” is a box girders deck, spanning till 34.5 m, generally composed bytwo precast box girders, prestressed with longitudinal steel strands and con-nected with small second step casting in the slab and with transversal

beams with post-tensioning cables. In Roma-Napoli line it was also realisedwith two V-beams and cast in situ slab. Transversal beams are usually pre-stressed with straight cables of strands or bars.The number of transversal beams is prescribed in [1]: for a deck with twoor more girders, at least two prestressed concrete transversal beams haveto be designed out from supports and more in case of decks longer than25 m. Strands getting out from the heads of the box girder are cut, isolated andprotected with the use of dielectric resin. Decks' deformability is largely veri-fied for comfort limit state: maximum deflection at midspan for Type “a” isless than l/5600.Type “b” is composed by four precast V-beams and cast in situ slab: actualmaximum length is 33.6 m. Beams are steam cured, pre-tensioned with longitudinal steel strands and

Type aWeight of one precast box: 455 ton (33.1 m)19% of total length of viaducts915 ton one deck weight (34.5 m)343.196 ton is total weight

Type bWeight of single precast V beam: 88 ton40.5% of total length of viaducts650 ton one deck weight (25 m)698.476 ton is total weight

Fig. 2- Two box girders deck (a) and four precast V beams and cast in situ slab (b).

Fig. 3- Prestressed concrete V beam (type b) in stocking area, Torino-Milano line.


transversally connected with cables in transversal concrete beams; it is themost common deck: it has been chosen for 40.5 % of total length of sim-ply supported prestressed concrete deck. The maximum deflection at midspan is largely verified: for type “b” span-ning 25 m (22.3 % of length of all prestressed concrete decks) it is lessthan l/6000.Because of the prescription of complete absence of tensile stress duringconstruction and life of the bridge, and of the large amount of pre-tensio-ning strands in V-beams, the technique of strand passivation for few metresalong beam ends, over supports, has been introduced for a portion ofstrands, in order to reduce even minimal cracking on heads of the beam.From an aesthetic point of view, shortest spans of box girders (both V-beamor cellular deck) can be put at a disadvantage, because in [1] a free hei-ght of at least 1.6-1.8 m inside box girder is prescribed to be left to easeinspection, leading to relevant height of the deck even for short span brid-ge. Anyway, these spans can be agreeably inserted in case of viaducts withshort piers.Type “c” is a single box girder deck, realised in two different ways and

sversal beams or second step castings of concrete in head anchorages oftendons in the required transversal beams, which always become visiblewith time.Anyway, in some case of grillage deck, when perspective had to be impro-ved, concrete noise barriers have been usefully adopted, covering secondstep casting or empty spaces on pier-cap between decks for inspection. Type “e” is composed by four precast I beams and cast in situ slab.Beams are longitudinal post-tensioned with cables composed by strandswith straight and parabolic profiles.Some are tensioned in precasting plant, then, after completing the bottomslab and tensioning of transversal cables, the second part of longitudinalcables is tensioned over the piers and slab is casted. Longest span of type“e” is also the longest span for simply supported prestressed concretedecks: 46.2 m. Type “e” has the advantage to manage precasting and launching of onebeam at time instead of full deck, so requiring simpler technology, but, atthe same time, the operation of assembling formworks and casting con-nection of lower slab and transversal beams, and the huge transversal post-tensioning (no. 47 4-strands cables for 46.2 m long deck) may put it to adisadvantage. Type “f” is the original Modena viaduct: the first case of lower way U deckfor high-speed lines; the double track is realized by two independent decks,piers and common foundation. It has two single track decks spanning 31.5m, and a total width of 18.4 m; each deck is pre-stressed with 20 longitu-dinal post-tensioning tendons of 12 strands. 566 km of corrugated plastic

Type cWeight of single precast deck: 567 ton (25 m)Weight of cast in situ box deck: 1043 ton(43.2 m) 11% of total length of viaducts173.856 ton is total weight

Type dWeight of single precast deck: 970 ton7.2% of total length of viaducts145.500 ton is total weight

Fig. 6- Box girder with single cell (c) and Box girder with two cells (d).

Fig. 4- Box girder (type a) on carriers towards launching operations, Milano-Bologna line.

Fig. 5- The first precast 25m long single box girder (Torino-Milano), during launching operations.

lengths: 25 m long precast box girder with longitudinal pre-tensioned steelstrands on Torino-Milano line (3.78 km long Santhià and 1 km long Carisioviaducts) and cast in situ post-tensioned deck spanning 43.2 m (2.8 kmlong Padulicella viaduct) on Roma-Napoli line. Type “d” is adopted in 5.1 km long Piacenza viaduct: 150 precast spanswith two cells and curved transversal profiles. It is a single monolithic boxgirder of 970 ton, with a maximum length of 33.1 m.Piacenza viaduct has been provided with 119 km of corrugated plasticducts and it represents the first application of electrically isolated disposalsfor the anchorages of 12 and 19 strands for longitudinal post-tensioningcables. Compared to the grillage decks, the monolithic decks have theaesthetic advantage of clean prospects and even deck sides, without tran-



ducts is used. In Milano-Bologna line this structural solution for prestressedconcrete deck is used for 10.7 km long double track viaducts. It is also usedin Junctions, for another 3.33 km of single-track, for a total number of 767precast spans. In particular, Modena viaduct is the longest viaduct of allhigh-speed lines, with its length of 7.1 km. The structural solution of lower way deck has been usefully adopted tominimize structural height plus noise barrier, because the noise barriersbecame a part of the structure. Modena system of viaducts is one of thefew cases, where aesthetics and environment impact have so strongly ledthe process of design and the solutions for structural and constructionneeds, in order to have quite original deck, unique in its kind.Durability

Through periodical inspection of old railway lines and particularly of theRoma-Firenze railway line, an analysis of the most common defects on brid-ge deck have been made in order to work out new guidelines of design andconstruction of the high-speed lines. Inadequate access for inspection,leaking of waterproofing system of the expansion joints between decks,insufficient cover, not efficient bearings towards maintenance operations aslifting of decks in order to change bearings, or reduced space to inserthydraulic jacks on pier-caps, etc. have been found as most frequent defects.These circumstances led to the introduction of new or more detailed pre-scriptions in the Italian standard for railway bridges [1], in order to gua-rantee better behaviour of structures with time, care for durability of con-crete structures, tensioning cables, system of waterproofing, devices for gooddrainage and anything else whose purpose is to ensure the overall long-term integrity of bridge structure. But, first of all, great care is put to inspec-tion ways to check the conditions of the structures and medium-life ele-ments. Access for inspection All bridges are designed assuring access forinspection, testing, maintenance and possible replacement of medium lifeelements. It must be always possible to walk over bridge decks because awidth of min. 50 cm on both sides is left for maintenance people. Inside cel-lular deck or closed box girders, as previously mentioned, a minimum hei-ght of 1.6-1.8 m must be always guaranteed; fixed stairs from deck to piercap must be provided every 3 spans or 100 m and from piers to theground every 500 m, for viaducts longer than 1000 m. Over pier cap itmust be always possible to pass from one deck to the following and stairsor landings are fixed to ease the movements of maintenance people. Finally, it must be also possible to inspect bearings and stroke end devicesor to operate in case of replacement of bearings or neoprene pillows, so afree height of 40 cm is left between lower side deck and top of pier-cap. Drainage It is a key issue about durability; deck slabs are provided withprovisions for good drainage and great attention is put to design, testing andlayout of all devices. Drainage of expansion joints is assured by a flashingtray of elastomeric material stuck with resins to slabs’ ends, in order toavoid leaking over piers and to drain water out of deck sides. Over bridgedeck, in order to protect from atmospheric agents, a thick layer of water-proofing is extended, also beneath the footways. In case of sensible pre-stressing system, pre-stressing strands or post-tensioning cable system justbeneath deck slab or critical drainage system (types “a”, “c”, “d” and “f”),a sprayed polyurethane waterproofing of 3 to 5 mm is extended. Checksare been carried out on site for adhesion and thickness by F.S. technicians.This surface treatment has proven to be very long life cycle performant. In post-tensioned concrete structures, deck anchorages are to be avoidedon deck slab; anyway, for all anchorages, design has to avoid leakage to getaccess to anchorages, providing protection against leaking expansion joints,as water drips.Post-tensioning tendonsThe grouting of the sheaths of prestressed con-crete bridges is always done with vacuum technique with a depression of0.2 bar during injection. It is standard for prestressed concrete deck withpost-tensioning tendons because grouting has been recognised as a key

Fig. 7- Piacenza viaduct precast deck, 33.1 m, in stocking area.Fig. 8- Perspective of curved profiles of Piacenza viaduct.

Fig. 12- The first one of 750 Modena precast decks in stocking area, Milano-Bologna line.Fig. 13- Beam head of Modena precast deck in stocking area, Milano-Bologna line.

Fig. 9- Four precast I beams in stocking area, from Roma-Napoli high-speed line.Fig. 10- Four precast I beams launched over pier caps, from Milano-Bologna high-speed line.

Type eWeight of single precast I beam:270 ton (46.2 m)Weight of four I beams deck:1.400 ton (46.2 m)7.6% of total length of viaducts137.274 ton is total weight

Type fWeight of single precastdeck: 689 ton15% of total length ofviaducts446.472 ton is totalweight

Fig. 11- Four precast I beams and cast in situ slab (e) and Lower way U deck (f).


operation to ensure durability of prestressing. Usually, it is difficult to ensu-re complete filling of the ducts: pathology teaches us that the prestressingtendons are more vulnerable in the case of post-tensioning than in the caseof pre-tensioning. Vacuum injection should avoid having air bubbles near high points of ten-don profile and in case of simply supported structures high points coincidealways with anchorage locations. Besides, all end caps are filled with groutand surrounded by concrete held in place by reinforcement, with no-shrinking concrete and the same compressive strength as deck concrete.Latest tendencies about durability of anchorages are High DensityPolyethylene ducts with plastic end cap left in concrete and the use of elec-trical isolated tendons as a further protection of the tendon and mean formonitoring: in Milano-Bologna and Torino-Milano railway lines, a large scaleapplication of plastic ducts and electrical isolated tendons has been under-taken. Italferr has taken the Technical Report from fib (fib 2000) and Swissguidelines on electric isolated tendons (2001) as standards, asking formechanical and chemical tests, field measurements to be undergone; aSystem Approval testing has to be conducted on site on every first applica-tion of a prestressing system.Deck equipment Every bridge deck is furnished of railings on both sides,anchorages of noise barriers for their future assembly, electric tractionpoles, stairs to pier cap and from piers to the ground. Great attention hasbeen put to deck equipment: every steel finishing is installed with linkageselectrically isolated from deck reinforcement and connected to a dissipati-ve end in the earth for safety reasons. Stainless steel is preferred for themost sensible connections.Electrical isolation of structures Every bridge deck is electrically iso-lated through isolated bearings and expansion joints, from piers and theother decks, besides, in order to prevent and protect bridge reinforcementagainst strain currents, few disposal for every deck are disposed in anaccessible area in order to measure strain current and isolation grade aftertraffic activation. This is probably a minor problem on high-speed line decks, but felt deeplyin every normal bridge deck and in the Junctions of high-speed line.Whenever problems of potential differences should arise, structures will beelectrically connected to earth or a cathodic protection should be eventual-ly adopted.In case of post-tensioned structures, all anchorages are electrically connec-ted (when no electrical isolated tendons is adopted) and the terminal isdrawn out of the structure in order to provide eventually in the future thesame provisions as for deck reinforcement, otherwise, in case of pre-tensio-ned decks, the head faces of the beam are protected with synthetic dielec-trical resins.Monitoring and Maintenance In order to improve deck’s behaviourknowledge and control it with time under the influence of external agents(environmental actions, traffic loads, seismic events or exceptional hydro-geological events), a complex monitoring system integrated with the high-speed line has been designed. At least one section (deck, pier, foundation, piles) per viaduct and, in case oflong viaducts, one section every 1000 m is instrumented: it means a large

number of strain-gages, inclinometers, thermocouples, instrumented bea-rings, load cells, foundation settlement meters, piezometers etc. Seldomaccelerometers are provided in order to evaluate dynamic response of thestructures also in case of seismic actions.Maintenance program of high-speed lines is essentially based on mainte-nance actions followings inspection visits: in the code 44/c of Italian railwaysabout lines maintenance [5], frequencies, ways of inspection and followingcheck schedules are prescribed. These check schedules have the doubleaim to check the safety of structures towards train traffic, and to keepmemory of time evolution of the behaviour of structures. According to [5]every year a program of action has to be adopted to eliminate anomaliesencountered in structures or to face critical situations.

Construction process

As all bridge designers know, construction process may deeply influencedesign choices, also in case of pre-cast prestressed concrete beams, whichcompose the majority of our new bridges and viaducts: the constructionscheduling, the technologies of lifting, transporting and lowering over thepiers are investigated.To improve the overall quality of the infrastructure design and production,a quality control system is implemented during both the design phase andthe construction phase. In the following, main construction features of eachstructural solution are described in the text and representative constructionphases are showed in the pictures.Precast beams and cast in situ slabWhen talking about constructionfeatures we must divide our decks in few major families: one of these is theprecast V or I beams with cast in situ slab. The short spans of four V-beamsdeck of type “b” (Fig. 14-15) and the shortest I-beams of deck of type "e",of an average load of 100 tons each, are the only ones which can be castedand pre-tensioned in pre-casting plant, moved on ordinary roads and led onsite, where each beam is lifted to its final position. Then predalles or formworks are assembled, reinforcement is laid and theconcrete slab is casted over the piers. As it is not necessary to pass overcompleted decks, ordinary roads can be used and there’s no obligedsequence of spans’ layout, this method of construction has the importantproperty of flexibility; besides, relatively simple technology is necessary forits realization. Cast in situ box girder In the only case of single box girder Padulicellaviaduct (Fig. 16), the deck is casted and pretensioned over the piers on self-launching formworks and special casting equipment is used.To accelerate the production, the reinforcement cage had to be pre-assem-bled, transported and lowered in the formwork before casting.Two box girdersTwo precast box girders, each weighting about 450 tons,are precast and prestressed with pre-tensioned strands, then lifted over theviaduct where they are transported by two small carriers on tyres, towardslaunching operations (Fig.17).Deck is completed over piers, with second step casting in central slab andin transversal beams, then transversal strands cables are tensioned andgrouted. The case of two precast box girders is half way between the pre-



The use of pre-assembled reinforcement cages, independent casting linesand several formworks, different phases of tensioning operations and theuse of storage areas for cables’ injection can speed up the process.On the other hand, if for any reason one deck has to be stopped in thestocking area, the process may stop for days. So this method doesn’t havethe flexibility pointed out before for beams and slab decks.

Fig. 17. Launching operations of a box girder of S. Rocco al Porto 2 viaduct, on Milano-Bologna line.

Fig. 18- Modena precast deck lifted and transported on tyres from stocking area, Milano-Bologna line.

cast beams and the full-span pre-casting.Actually launching operations similar to those of full span pre-casting arenecessary, but second step casting and transversal prestessing in transver-sal beams are needed. As the full span precast decks of the following chapter, these box girdersare realized in a plant near the viaduct: all these plants are dismantled atthe end of the works. Full span precasting In those cases when much more spans had to bebuilt in short time, full span pre-casting has been preferred (Figs. 18-21).According to this process of construction, decks are totally pre-casted; nopost-tensioned transversal beam is needed to complete the deck over piers.Afterwards, one by one they are moved towards launching operations.A «carrier» and a «support beam» always compose the launching system.The carrier slowly moves on tyres or steel wheels, lifts a stored beam, andtransports it along the viaduct, moving on the placed girders. The device for transport forms an integral part of the device for the laun-ching of the girders: the carrier drives then into a second steel girder calledsupport beam, suspends the beam over its final position. The support beam is drawn back and the beam is lowered. With four orsix bearings, hydraulic jacks or load cells are used in order to check weightload distribution, and then the beam is lowered on its final bearings.

Fig. 14- Precast V beam lifted up from stocking area.

Fig. 15- Precast V beam towards final position over the piers.

Fig. 16- Reinforcement cage transported in the formwork over the piers, Padulicella viaduct, Roma-Napoli.


Fig. 20- Modena precast lower way U deck during launching operations, on Milano-Bologna line.

Fig. 21- Santhià precast single box deck during launching operations, Torino Milano line.

Fig. 23- Span length distribution (e) among the a-f typologies (figures based on high-speed lines: Roma-Napoli; Bologna-Firenze; Milano-Bologna; Torino-Milano).

Fig. 22- Concrete, Reinforcement and Strands loads of different typologies of deck versus span length, (figures based on high-speed lines: Roma-Napoli; Bologna-Firenze; Milano-Bologna; Torino-Milano).

Fig. 19- Piacenza precast transported on steel wheels from stocking area towards launching, Milano-Bologna line.

Peak cycle is variable: Modena viaduct has a casting and launching speedof two precast girders per day. Others, as Piacenza viaduct, have the desi-gn of the spans and of the casting yard centred on a target peak cycle oftwo double-track decks per week.

Comparison between a-f typologies

Last data about PC simply supported spans are in Figg. 22 and 23 wherethe most important figures about double-track decks of the new high-speedlines are presented. There’s good uniformity between different typologies dealing with deck load

and reinforcement and prestressing steel amounts for each span, apartfrom few exceptions.Talking about deck load, first exception to be mentioned is Modena deck:as a single way deck it results heavy solution for a double line, but, at thesame time, the simple “U” profile, easy to manage from a design point ofview, doesn't cause similar examples of exception for the load of reinforce-ment and prestressing steel. Decks composed by beams can never minimi-ze the use of steel because of their transversal connections, while single boxgirders seem, even if based on few examples, to behave more efficiently.Anyway, many other factors are to be considered in the choice of the beststructural and technological solution for a new railway bridge deck: theymay depend on construction method, workmanship factors, number ofspans to be built and required time scheduling of construction, as previou-sly mentioned.Another factor for evaluating the solution for a bridge deck is, of course, itsaesthetic impact: even if mentioned at the end of the analysis, in some caseit has strictly led the design choices.Generally speaking, the span length over deck height ratio can be conside-red one of the simplest measure to evaluate the grade of slenderness: incase of simply supported PC decks, it ranges between 9 and 12; even forshort spans, the prescription of minimum free height inside cellular deckscauses small ratio. Anyway considering that the average piers’ height is not more than 7÷8 m,usually long spans are not used with very short piers because of aestheticreasons too. Monolithic decks are to be preferred to precast beams decks because tran-sversal beams are always impacting on the sides' prospects.Finally, a good example of agreeable bridge aesthetics in the field of simply



supported spans was obtained changing the structural solution: the Savenaarch bridge on Bologna-Firenze high-speed line, designed to overpassSavena River. It is composed by reinforced concrete arch, steel hangers con-nected by spherical hinges to the 2930 ton prestressed concrete deck, withboth transversal and longitudinal post-tensioning cables. It has the lowestheight under rails (1.60 m) for a span length of 62.5 m.

Continuous prestressed concrete bridges

In few cases, for riverbed or embankment crossings, highway or railwayover-flyings, multi-span continuous PC bridges have been designed. The mostcommon choice for a single long span in a sequence of shorter simply sup-ported spans is the composite steel and concrete deck, from about 40 mto more than 70 m, but they result often strongly impacting with longer PCviaducts perspective, consisting generally of a single exception, with differentstructural height, colour and side profile.In other case, the need to harmonize approach viaduct spans length to abigger structure as arch bridges or cable-stayed bridge has led to multi-spanPC viaduct with spans of 60÷70 m.According to ref. [1], all previously mentioned design and durability pre-scriptions have to be applied to continuous bridge: access for inspection ofevery pier cap is more stringent and difficult to obtain, leading often to com-plex systems of stairs and landings around central supports. Besides, everydeck has to be verified for lifting in case of bearings' replacement and itmay result structurally demanding, while, from a technological point of view,it can require a specific design to give disposals for 20% additional pre-stressing in each span longer than 40 m.PC continuous beams are often casted in phases, it is quite rare to have asingle casting operation for two or three spans and in [1] precast segmen-tal construction is forbidden. For cast in situ segmental PC bridge, a mini-mum reinforcement area through every joint of 3.0% area of concrete crosssection of deck is prescribed and minimum compressive stress of 1.0 MPa(rare load combination, also during construction stages) is expected fromdesign. In the following, because of lack of space, only two examples of continuous

PC bridges are mentioned. The first example is composed by the river embankments approach via-ducts to the cable stayed bridge over Po River [2]: five spans on the left sidefor a total length of 260 m (Fig. 24) and three spans on the right side fora total length of 130 m. The decks are three cells box girders, built by balan-ced cantilevers from central piers: a couple of segments at one time iscasted in situ and post tensioning cables are tensioned.Then the remaining gaps of 1.0 m long are concreted and post-tensioningcables are laid in the spans to join together, two by two, the cantilevers. Forthe Italian State Railways, this is the second example of this kind of struc-ture and method of construction (the first one being built for the Rome-Firenze line).The second example refers to the Modena continuous bridges to overpasstwo rivers, a highway and a railway Junction in the Modena System ofViaducts, for a total number of nine single-track continuous beams.All nine bridges have span lengths of 40-56-40 m, the same outer cross-section of the lower way U decks of Modena simply supported spans (typef), with higher webs on central piers. They are built in three segments castedin situ e prestressed with 40 mm bars, coupled at the end of each segment.Two methods of construction were experimented for the same structuralsolution, because of the different environmental conditions: for two of thesebeams, the segments were casted on a formwork, scaffolding from theground with two temporary bars for each joint, in order to help the partialstructure during construction conditions. The other seven obtained the sameeffect with a formwork hanging from steel box beams (over Panaro River)or truss beams (over Secchia River and on Brennero highway, see Fig. 25)and the end support of the beam applies nearly the same reaction of thetemporary bars.

Conclusions

This paper has described the strategic choices and the most relevantstructural and infrastructural features for the design and construction ofthe new Italian high-speed railway network. Most topics concerning sim-ply supported prestressed concrete decks for new high-speed lines have

Fig. 24- First balanced cantilever of Left Embankment viaduct, approaching cable-stayed bridge over Po River. Fig. 25- Continuous PC beam of Modena viaduct over Brennero highway, Milano-Bo logna line.


been analysed, while some information on continuous PC beam havebeen introduced at the end. Deck with precast beams and cast in situ slab is the most common choi-ce due to its flexibility (type “b” is actually the most used structure for brid-ge decks) but full span pre-casting is the most important future trend.

The mentioned principles of design, the so-called multi-layer protectionsystem, the method of construction, which deeply influences design choi-ces, and the tendency to experiment models and tests before every firstrealization, have been recognised as strategic factors for good results indesign, construction and management of railway infrastructures.

References

[1] Italian standard for railway bridges: Istruzione F.S. n. I/SC/PS-OM/2298 del 2.6.1995 “Sovraccarichi per il calcolo dei ponti fer-roviari - Istruzioni per la progettazione, l’esecuzione e il collaudo”,Final review 1997.

[2] Petrangeli, M. P., Traini, G., Evangelista, L., Della Vedova, M. The cable-stayed bridge over Po River: design and construction. Proc. of the2nd Inter na tional fib Congress, 5/8 June 2006, Napoli.

[3] Figures, deadlines and picture in §1.Introduction are drawn fromwww.tav.it, revision April 2004.

[4] Italian standard for railway bridges in seismic areas: IstruzioneFS 44/b “Istruzioni tecniche per manufatti sotto binario da vostrui-re in zona sismica”, Final Review 14/11/1996.

[5] Italian standard for railway line maintenance: Istruzione FS 44/c“Visite di controllo ai ponti, alle gallerie ed alle altre opere d’arte delcorpo stradale: frequenza modalità e relative verbalizzazioni”, FinalReview 16/02/1994.



The new railway linking Bologna to Milan is part of the High Speed Lines Italian network. It cros-ses the Po near Piacenza in a section where the river is usually about 350 m wide, up to 1 kmbetween the main embankments. The bridge will be 1200 m long, 400 m to cross the ordinaryriverbed, an obliquity of 22° resulting between the tracks and the river. Two approach viaducts,respectively 6 and 4 km long, complete this work, the most important of the whole line.Four main spans of 96 m were proposed in the preliminary design to satisfy the navigation requi-rement, and two solutions, both with prestressed concrete decks, were selected after a first study,but the competent Authority for the Environment insisted in eliminating the central pier, a 192 mmain span resulting for that. It is one of the longest prestressed concrete railway span in the world operated at a speed up to300 km/h. Three types of structure are present in the crossing in addition to the standard 14 km long approa-ch viaducts placed outside the upper banks: the cable stayed bridge, 12 simply supported deckson the right bank and two continuous p.c. box girders necessary to overpass the main embank-ments.The decks are subdivided in such a way that two joints in the rails are necessary to keep the expan-sion length within the allowable limits. This is the only exception along the jointless HS RailwayItalian Network.The relevant part of the crossing has a 192 m central span and two 104 m long side spans. The deck is a p.c. continuous box girder with the fixed point at one tower, sliding bearings at thesecond tower and at the transition piers. Expansion lengths of 296 and 104 m derived from thisarrangement of the bearings, joints in the rails so being required.The height of the cross section isconstant and equal to 4,5 m (L/42,7) along the central span; it varies and decreases to 3,70 m inthe side spans, in order to fit with the other decks. The towers are 60 m high from the footing, 51 m from the deck.The top of the towers, wherethe stays are anchored, is a steel-concrete composite structure. The stays are made of 55 to 91 zinc-coated, singularly greased and sheathed 0,6” super strands.The total amount of steel for the stays is 410 tons, corresponding to about 66 Kg per square meterof deck.The foundation of each tower has the footing (shaped to reduce the drag force) supported by28 piles, 2m diameter and 65 m long.Derailment of railway vehicles. Two accidental design situations have been considered withrespect to the stays:- collapse of two consecutive stays along one side due to the derailment of a vehicle: the bridgemust remain in service with one design train over the track nearest to the injured side and onepassenger train ( 40 KN/m ) over the other, the thermal effects being excluded;- the consecutive stays collapsed are three: only the effects due to one design train is taken intoaccount. Dynamic analysis.Three different trains (ETR 500, TGV, ICE) have been considered for the dyna-mic analysis considering the dynamic behaviour of the vehicle as well as the irregularities of thetrack.Combined response of structure and track. The effects resulting from variable actions havebeen taken into account according to prEN 1991-2 and for two limit stiffness of the foundations. Seismic analysis have been carried out in the elastic range according to the Italian RailwaySpecifications and EC 8. Because of the low seismicity, seismic actions did not influence the designof the bridge but in a few sections in the upper part of the towers, while they were relevant for thebearings and the joints. All the decks have been built by cantilever method with cast in situ segments, but the 13 simplysupported spans.

ProjectHS railway cable-stayed bridge over the Po river

LocationBologna Milano High Speed Line near Piacenza

ClientT.A.V. S.p.A. Concessionaire for the design and construction of the Italian HS system for RFI

DesignProf. Ing. Mario Paolo Petrangeli

Structural engineerProf. Ing. Mario Paolo Petrangeli

ArchitectsProf. Ing. Mario Paolo Petrangeli

Management ContractorCEPAV UNO- Syndicate formed by ENI per l’AltaVelocità

General ContractorA.S.G. Scarl – Syndicate formed by Aquater,Snamprogetti, Grandi Lavori Fincosit

HS railway cable-stayed bridge over the Po river

Winner A.I.C.A.P. Award 2009 for Structural Concrete Works - Category “Civil Engineering”


A number of physical tests have been executed to assess the theoreti-cal assumption, the most outstanding being: (i) test on a half scalemodel reproducing a segment of the deck with a stay anchorage, car-ried on in the yard; (ii) fatigue test on a full scale model of the steel boxembedded in the upper part of the tower to anchor the stays and (iii)fatigue tests on three stays (composed by 55, 73 and 91 0,6” strands)complete of anchorage. Both fatigue tests were carried on in the ECJoint Research Centre of ISPRA.Due to the importance of the bridge, a large number of sensors have

been permanently placed on it. The monitored quantities are: loads onthe piles, stress and temperature in the most representative sections ofthe deck and the towers as well, forces transmitted by a number of staysand bearings, geometrical data like the angular rotation of towers andthe deflection of the decks and, finally, the scour near the piers in theriverbed. Both sonar and magnetic devices have been installed to detectscour. All the data will be collected inside the cable-stayed deck and fromthere automatically transmitted to a remote office located in Bolognathat will manage the monitoring system of the bridges of the line.

• 1- General view of the whole bridge.

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2 3

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• 2- Completion of the towers and access viaducts. 3-4 Construction of the deckby segments. 5- Pouring the crown segment. 6- Tower elevations.

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“Piacenza” viaduct

In the Milan-Bologna stretch of the high-speed/high load railway line the Piacenza Viaduct runsa total length of 5103 m. The structure breaks down into two stretches, Piacenza1 andPiacenza2, their lengths 2522 m and 2581 m respectively.The first stretch features a two-light caisson deck and a steel-concrete mixed-structure span 50m long. The second stretch, like the first, also comprises a two-light caisson deck plus two mixed-struc-ture spans of lengths 38 and 40 meters.The prestressed-concrete spans consist of monolithic precast two-light caissons with post-ten-sioned prestressing cables. The caissons weigh 975 tons each and are 33 meters long. The section depths/thicknesses are 45 cm for the webs, 37 cm for the deck slab and 30 cmfor the bottom slab with increased depth at 60 cm from the ends.The prestressing cables are 24 post-tensioned cables, electrically insulated by the use of insula-ted head ends and HDPE sheaths. Of them, fifteen 19-strand cables are placed in the threewebs, with parabolic trajectories, while the other nine cables, 12-strand, are placed in the bot-tom slab on a straight-line trajectory. The strands used are class fptk �860 MPa, having a ratedsection of 139 mm2.Maximum pier height is 12 meters.The foundations, deep, are built on piles.The caissons were worked up in the plant, in the following phases: creation and placement ofthe slack reinforcings and of the forms, a concrete pour carried out with the aid of truck mixersand of two truck-mixer pumps positioned nearby the head ends, threading and tensioning of thecables and, finally, hauling the caisson to the launch area. This last phase was carried out by aspecial launch car weighing 900 tons.

ProjectPiacenza viaduct

LocationPiacenza, Italy

ClientFerrovie dello Stato S.p.A.

ConcessionaireT.A.V. S.p.A.

Higher supervisionItalferr S.p.A.

DesignDott. Ing. Villa

General ContractorCepav uno


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c

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• 1- Preparing the launch caisson. 2- Hauling the caisson. 3- Launching the cais-son. 4- Mounting the side wings. 5- Viaduct cross section. 6- Construction phases: a)pour of the bottom slab; b) pour of the septums; c) pour of the top slab; d) foldingback the internal forms for their removal. 7- Cross section through the two-light cais-son. 8- Longitudinal section through the two-light caisson, showing prestressing ca-bles.



The Modena Viaducts system belongs to the high-speed/high-load Milan-Bologna line and has atotal length of 25,182 meters, of which 23,958 m built with precast segments and 1224 mwith in situ-poured continuous beams. The Table gives the names and the principal characteri-stics of the viaducts belonging to the “Modena” System.All these viaducts have in common a type of deck known as the Omega, featuring a U beambearing the track on its flange for each track. This particular cross-section conformation has theadvantage of mitigating noise emissions by partially enclosing the train within the structurethrough the bearing walls, which also act as noise barriers.What is special about these viaducts is the particular precasting process used, which enabledinstallation, during full production (between 2003 and 2004) of more than one segment perday with a peak of 52 segments launched in July 2004.Besides the 767 plant-precast (isostatic) segments, the structures call for an additional nine sta-tically-indeterminate continuous-beam stretches of length 136 m, with spans of 40-56-40 m(statically indeterminate), for a total of 27 spans. In the statically-indeterminate stretches too thecross section is still the Omega type, with projections in the zones where the bending momentis negative. The statically-indeterminate members are poured in forms hanging from overlyingribs.The typical viaduct span in the isostatic stretches is a precast prestressed-concrete memberhaving an open cross profile. The precasting, for a single-track, is 9 m wide and 3.5 m deep, wei-ghing a total 690 tons of which 33 tons are steel (slack reinforcing and prestressing). The out-side surface of the deck features longitudinal channels and four r.c. support elements solidly joi-ned to it, the internal surface is smooth and encloses the trackway. The decks were all built usingan Rck 45 MPa concrete.The (circular section) piers are 3.5 m in diameter with heights from 5.45 m to 12.20 m. Theirsections display three typologies: hollow with a 50 cm wall-depth, hollow but filled, and full-sec-tion. In the statically-indeterminate stretches the central piers are 4.50 m in diameter with fullcircular sections. The piers are built of Rck 35 MPa concrete.The foundations were built in two ways: on 1500 mm diameter piles of lengths between 35 and50 m, and on diaphragm septums, 120 cm thick and 36 m long.The footings have dimensionsof 9.5 m x 16.7 m x 2.2 m and were built of Rck 30 concrete.

ProjectModena viaducts

LocationModena, Italy

ClientFerrovie dello Stato S.p.A. (State Railway System Inc.)

ConcessionaireT.A.V. S.p.A.

Higher SupervisionItalferr S.p.A.

DesignProf. G. Macchi (Final design), Ing. Sangalli(Construction design)

General ContractorCepav uno

“Modena” system viaducts


Viaduct name

Brennero - Odd track

Brennero - Even track

Modena Interconnection - Odd track

Total lenght (m)

Isostatic beams, no.Statically

indeterminatebeams

Modena Interconnection - Even track

Modena - Odd track

Modena - Even track

Panaro - Even track

Secchia

TOTAL

Panaro - Odd track

• 1- Longitudinal section through deck. 2- Cross section through isostatic portion.3-4 Sections through statically-indeterminate portions showing enlargements nearbythe areas of negative moment. 5- Precast segment for the isostatic span. 6- Laun-ch of the isostatic segment. 7- Statically-indeterminate portion.

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The Savena viaduct is the crossing of the river of the same name by the Florence-Bologna High-Speed Railway Line on a single double-track deck. This structure’s main characteristic lies in thedeck’s structural depth (below the ballast), which is just 80 cm for a 63.5 m span. To achievethis slenderness design opted for a through-arch in which the deck is the structural system’s ten-don. The deck plate is thus so shaped as to create in its central zone the basin containing theballast, below which the structural depth is 80 cm. Laterally the depth was suitably increasedso as to correctly house the prestressing-cable heads.The deck is suspended laterally and two alignments per side of hangers, connected to two r.c.arches, inclined toward the interior of the bridge and connected in their turn by three r.c. cross -pieces. In order to meet the design performance requisites, consisting in total prestressing in ser-vice, at the deck edges in the main directions two-way prestressing had to be applied to theplate. Its effects were evaluated using a finite-elements mathematical model consisting of shellelements. The hangers are connected to the deck through pre-tensioned prestressing bars andanchored in the soffit. The bridge’s construction was carried out in the following phases:• the three-phase deck pour on the embankment behind the abutment and the deck’s pro-gressive crosswise prestressing on the ground; then the creation of the longitudinal prestressing;• construction of two provisional piers in the river;• a forestarling 22 m long is mounted;• thrust of the deck up to the next abutment;• the mounting of centerings and forms for the in situ pour of the arches;• mounting and adjustment of the hangers, with the deck being detached from the provisionalpiers;• demolition of the provisional piers and their foundations.

ProjectSavena Viaduct

LocationBologna – High-Speed Florence-Bologna Railway Line

ClientT.A.V. S.p.A.

DesignStudio Sintecna – Prof. Ing. Giuseppe Mancini

ContractorIMPREGILO S.p.A.

“Savena” Viaduct

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• 1-2- Deck plan, and cross section through viaduct.

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STRUCTION- Civil engineering W

orks

Italian High-Speed N

etwork

•3- Construction of the provisional riverbed piers. 4-5 D

etail of the arch-plate joinand the reinforcing. 6- Installation of the arch reinforcings. 7- Forestarling for theconstruction of the deck. 8- Anchorage of the hanger on the deck. 9- Bridge sideview.

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For the construction of the Rome-Naples High-Speed line through the municipality of Caivano,the railway had to be inserted with the least possible environmental impact on the territory. Thisled the customer, RFI-TAV, to work out “structures” that would be an addition to the territory.The variation, which runs some six kilometers between stations 200 km and 206 km of theRome-Naples High-Speed Line substantially comprises three principal structures:- the conical-capital viaduct;- the multiple-arch viaduct;- the two cut-and-cover tunnels Caivano 1 and 2.

The conical-capital viaductThe structure is 203 m long and serves to overpass three interfering lines having centerlinesstrongly skew to the railway line: the Turano-Biferno aqueduct, the Secondigliano sewer line andthe SS87-bis highway. The positioning of the vertical elements follows directrices parallel to thecenterlines of the interferences, the deck is created by an orthoanisotropic plate, woven cros-swise to these directrices so as to minimize the bearing structure’s span. Deck characteristicsand constraint scheme: two-span frame with an orthoanisotropic plate fixed-jointed to thecolumns.Soil conditions, water table: soils of volcanic origin featuring a first 8-10-meter stratum of fairlywell compacted pyroclastites, an underlying stratum six meters thick of tuff having lithoid cha-racteristics and a successive stratum of more deeply densified pyroclastites. The water table liesat an average four meters below site level.Degree of seismicity: the area is classified as a second-category seismic zone, so that its degreeof seismicity, S, is 9.Foundations typology: deep piles 1200 mm in diameter.Construction procedures: the pier foundations and standing structure were in situ poured. Forthe deck, resort was had to precasting long-span predalles; the precast beams were prestressed,and an in situ completion pour was made for the slab.

The multiple-arch viaductThe viaduct comprises a sequence of 74 parabolic-profile arches having a 33.00 m chord, arran-ged in two parallel files spaced 8.40 m apart. The arch height varies between 0.80 m at thecrown, including the depth of the slab, to 4.00 m at the springer. Its depth is instead a constant1.00 m. The upper slab, 13.60 m wide, has a depth varying between 0.60 m at midspan to0.50 m at the ends. Overall viaduct length is 2400 m.Deck characteristics – constraint scheme: reinforced-concrete arch.Soil characteristics, water table level: soils of volcanic origin, the first stratum of pyroclastites ofmedium density varies in depth, being greater at the start of the viaduct and lessening towardsits end. Right below it is a tuff bench six meters deep and then the pyroclastites are taken upagain, more deeply densified. The water table lies at an average four meters below site level.Degree of seismicity: the area is classified as a seismic zone of second category, so that thedegree of seismicity, S, is 9.Foundation typology: deep piles an average of 35 m long and diameter 1500 mm.Construction procedure: to build the standing structure a form was used created especially forthe purpose. It enables building in a first phase both arches and, after their curing, the slab in asingle pour.

The cut-and-cover tunnels Caivano 1 and 2In order to underpass the two interfering roadways, motorway A1 and the Nola-Villa Liternosuperhighway, two gigantic monoliths were designed, weighing 12,000 and 8,000 tons respec-

THE CONICAL-CAPITAL VIADUCT

ProjectRome-Naples High-Speed Line

LocationCaivano, Naples, Italy

ClientR.F.I. S.p.A. (Italian Railway System Inc.) - T.A.V. S.p.A.(High-Speed Train Inc.)

DesignProf. Remo Calzona CE

ContractorSocietà Italiana per Condotte d’Acqua S.p.A.

Year of completionMarch 2006

THE MULTIPLE-ARCH VIADUCT



ClientR.F.I. S.p.A.(Italian Railway System Inc.) - T.A.V. S.p.A.(High-speed Train Inc.)



Year of completionMarch 2006

THE CUT-AND-COVER TUNNELS CAIVANO 1AND 2



ClientR.F.I. S.p.A.(Italian Railway System Inc.) - T.A.V. S.p.A.(High-speed Train Inc.)



Year of completion2007

“Caivano” variation structures


tively, which were translated underneaththe two aforesaid interferences. Complet -ing the job are the cut-and-cover rain tun-nels, which at the mouths prevent theentrance of rainwaters.Soil characteristics and water table: soils ofvolcanic origin. The first stratum of me -dium-dense pyroclastites varies bet ween 6-7 meters of depth; the next 5 m down seesa tuff stratum of good mechanical charac-teristics, and deeper down yet are againdense pyroclastites. The water table lies atfour meters below site level.Degree of seismicity: the area is classifiedas a second-degree seismic zone, so thatits seismicity, S, is 9.Construction procedures: the monoliths,115 m and 70 m long, were poured in situin special forms and then translated usinga thrust system that enabled the transla-tion of the greater of the two caissons injust six days.



The design and construction of the tunnels were carried out under the A.DE.CO-RS (Analysis ofControlled Strains in the Rocks and Soils) design methodology.This approach was used to make up the base-contract project and then the construction de sign.This tunnel design and construction methodology avoids the limitations of traditional approa-ches, by supplying the possibility of industrializing the tunnel advances even when traditionaldriv ing methods are applied, in complex geomechanical contexts and with covers running froma few meters to six hundred. The tunnels drive through sundry Apennine geological formations,consisting mainly of marls, sandstones, scaly clays and limestones. In driving the running tunnelsthe following methodologies were adopted:• “Traditional” driving by the use of traditional machines (jumbos, excavators, jackhammers,power shovels), except for the Ginori service tunnel, the only stretch in which wholly mechani-zed driving was carried out (by a shielded TBM). Overall some 17 million cubic meters of mate-rial were excavated.• Full-section driving after consolidation of the core-face (where necessary), especially under dif-ficult stress-strain conditions.• Limitation of soil decompression by means of pre-confinement operations on the core-face orconfinement of the cavity (sub-horizontal jet-grouting, fibreglass plastic structural elements in thecore or around the cavity, shotcrete and steel ribs, radial bolts, etc.). The operations were defineddepending on the strain behaviour of the advancing core-face, evaluated during design phase,described according to the following three behaviour categories: A= stable front, B= short-termstable front, C= unstable front. Corresponding to each category is a tunnel section type for thedriving and advance. The percentage distribution of the tunnel section types applied is as follows:type-A sections (35%), type-B sections (53%), type-C sections (12%).• Final concrete and reinforced-concrete lining. The final linings (roof, sidewalls and inverted-arch)were in situ-cast concrete of class C25/30 in depths of 60-100 cm. On the whole seven millioncubic meters of concrete were applied.Duration of the construction phase: 13.5 years (from July 1996 to December 2009) with advan-ce rate of one to five meters per day and up to 2000 meters per month on thirty faces at thesame time.

ProjectTunnels in the Florence-Bologna stretch of High-Speed Line

LocationFlorence-Bologna stretch

ClientR.F.I. S.p.A. (Italian Railway System Inc.)

GaranteeT.A.V. S.p.A. (High Speed Train Inc.)

Higher SurveillanceItalferr S.p.A.

DesignMAIRE Engineering Rocksoil S.p.A.

General ContractorFIAT S.p.A.


Main featuresTotal length of the High-Speed Florence-Bolognarailway stretch: 78.5 kmUnderground works:- 9 double-track driven tunnels having multi-centered

sections with a driven area of 140 m2 and a netarea of 82 m2, their lengths running from 693 m to18.2 km;

- 13 access windows for a total length of 9.3 km; - one service tunnel 10.65 km long; - 2 single-track interconnection tunnels (total length

2.2 km); - 3 cut-and-cover double-track tunnels 2.8 km long; - one large underground chamber for switching

space, one underground interconnection chamber.

Tunnels in the Florence-Bologna stretch ofHigh-Speed Line

1


• 1- Pianoro Nord area: Savena bridge and the portal of the Pianoro Tunnel. 2-Northern portal of the Monte Bibele tunnel. 3- Firenzuola chamber. 4- From thenorthern Morticine portal to the southern Borgo Rinzelli portal. 5- Southern portalof the Sadurano tunnel. 6- Southern portal of the Raticosa tunnel.

2

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In all cities the construction of a new railway station is a special event, if only in that unlike an airport or other man-made transportation facilities, itconstitutes a “port within the heart” of the city itself. In Italy station construction has marked the stages in the modernization of the country; the stations of the main Italian cities were the starting pointof the “journey” and as such were often constructed as authentic monuments.

The new stations that are emerging due to the advent of high-speed rail transportation are being located in the cities according to different criteriathan in the past, often being designed to reunite parts of the city separated by the railway lines and to contribute to their redevelopment and enhan-cement. Some of the world’s leading contemporary architects have been commissioned to design stations that are full of life, light and sound -in aword, stations with a definite soul, and one that they are able to project: no longer an anonymous “non-place”, a desolate point of transit and metaphorfor lone travel, the new stations are “the place”, and, similar to how a town square encourages aggregation, dialogue and leisure, they embody thevery concept of urban community life. Railway stations will be home to bookstores, cafes and shops and become venues for music, art and cultureso that they come to be looked on by the public as places in which social relations are played out.

On the horizon are the new HS stations of Naples Afragola designed by Zaha Hadid, Rome Tiburtina designed by ABDR-Paolo Desideri; FlorenceBelfiore designed by Norman Foster; Bologna designed by Ricardo Bofill; Reggio Emilia designed by Santiago Calatrava and Turin Porta Susa designedby Arep Group. Meanwhile, plans and works are in progress for the redevelopment of the historic mainline stations to adapt them to new manage-ment and operational requirements and enhance the cultural heritage they represent.

New stations for Italian High-SpeedNetwork


1. BOLOGNA HS STATION

ProjectBologna High-Speed train station

LocationBologna


Design and structural engineerITALFERR S.p.A.

Management Contractor Italferr S.p.A.

ArchitectsRicardo Bofill

ContractorAstaldi S.p.A.

The Bologna HS Station extends under a sec-tion of the track array of the existing centralstation into a large underground chamberapproximately 640 m long, 42 m wide and 23m deep. The new underground station will besplit into three levels. The HS tracks will be laidon the bottom level, passenger services andcommercial activities will be located on themiddle level, while the upper level will be reserved for vehicular traffic to thestation and as parking space. The middle level will also be used as an exhi-bition area for the numerous archaeological finds (in cluding a road, a fur-nace and burial chambers) found during excavation. The HS Station wasdesigned by the Ca talan architect Ricardo Bofill who adopted innovativesolutions of high architectural value. The supporting structures of the ex -cavation, for example, are designed as “facing arches” that transfer earththrust to “spurs” placed at 12 m intervals: this results in large bulkhead sec-tions free of structural obstacles both horizontally (12 m of clear space) andvertically (24 m of clear space). This innovative solution exploits the facingarches as architectural elements to project daylight onto the tracks creatinga vertical illumination effect similar to that found in Gothic architecture andalso allows the insertion of vertical communicating elements between levels.An international design competition for the completion of the central stationand the redevelopment of the urban area in which it is located has beenrecently awarded to the Japanese architect Isozaki.

Main featuresPlan dimensions: 640m x 42mOverall height: 23m (underground)Pile perimetral bulkheads: 20,000 m2

Bulkheads, rider arches and spurs = 37,000 m2

Foundation concrete: 57,000 m3

Foundation reinforcing steel: 10,000 tConcrete in standing structure: 100,000 m3

Reinforcing steel in standing structure: 15,000 tStructural steel: 31,000 tFloor-structure area: 80,000 m2

• 1- The new HS Bologna Station – the map. 2- Structural sketch of the new Bo-logna HS Station (drawing by P. Bellotti). 3- The transversal section. 4- The new HSBologna Station work site.

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2. FLORENCE HS STATION

ProjectFlorence High-Speed railway station

LocationFormer Macelli/Belfiore area of Florence, Italy


DesignATI Foster and Arup

Structural engineerDefinitive design: ArupFinal design: Nodavia

ArchitectsFoster & Partners

Management ContractorItalferr S.p.A.

General ContractorNodavia, Ati Coopsette-Ergon engineer

The new Florence HS station will be built in anineteenth-century district of the city, that ofthe former Macelli (slaughterhouses) area, andis part of a wider-ranging project for the recu-peration of abandoned industrial sites in va -rious parts of the city. The new station will not replace but form aninterchange node with the existing Santa Ma -ria Novella station. The two stations will belinked by a new tramway that will extend toPeretola airport and by shuttle trains betweenthe Circondaria regional train station andSanta Maria Novella station. Like the great stations of the past, the new HSstation will feature large, mainly glass-coveredvolumes. While construction technology andtrains have radically changed since the nine-teenth century, large spatial dimensions andquality remain indispensable requirements formodern stations. The project was developed bythe architect Norman Foster. The station will bebuilt on two underground levels in a vastchamber 454 m long and 52 m wide, and therails of the high-speed line will be located at adepth of 25 m from ground level. From a structural point of view the station ischaracterised by 1.6 m thick vertical supportpanel bulkheads reinforced by perpendicularspine walls and strutted in such a way as toallow ample clearance for the passage of light.

Main featuresPlan dimensions: 450m x 52mOverall height: 26.5mBulkheads: 40,000 m2


Reinforcing steel in foundation: 22,800 tStanding-structure concrete: 75,000 m3

Standing-structure reinforcing steel:23,700 tStructural steel (tons): 8,000 tFloor-structure area: 35,000 m2

• 1- Overall view. 2- Detail the steel-glass canopy. 3-4- Internal view. 5- Cross sec-tion.

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3. ROMA TIBURTINA HS STATION

ProjectThe new Rome Tiburtina High-Speed train station – 1st functioning job segment

LocationRome, Italy


DesignABDR Architetti Associati – Paolo Desideri arch.(Chief of job coordination)

Structural EngineerEzio Maria Gruttadauria CE

ArchitectsStudio Tecnico ABDR architetti associati

Contract Management Italferr S.p.A.

ContractorATI Coopsette / MECoop

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• 1-2- Overall view. 3- Detail.

Main featuresOverall height: 36 m of which 26 m aboveground. Bulkheads: 29,150 m2

Piles: 62,190 mFoundation concrete: 42,450 m3

Reinforcing steel in foundation: 14,425 tStanding structure concrete: 24,675 m3

Standing structure reinforcing steel: 7,415 tStructural steel: 11,750 t of which 1,475 t for theroof space-lattice structureFloor-structure area: 54,000 m2



4. TURIN PORTA SUSA HS STATION

The underground stationThe station was built in two phases undertwo integrated projects, developed in differentperiods. The first project is tied to the con-struction of the Turin through line for puttingunderground the tracks running through thecity and for quadrupling them; this to be donewith the construction of the Porta Susa sta-tion underground. This station, essentially anunderground chamber conceived as an un -der ground station, is in an ad vanced stage ofconstruction and is partially operating since2009.In what follows the principal characteristics ofboth projects are set forth.

ProjectThe Turin Porta Susa High-Speed train station

LocationTurin Porta Susa, Italy


DesignITALFERR S.p.A. (Definitive), Ing. Campa (Final)

Structural engineerITALFERR S.p.A. (Definitive), Ing. Campa (Final)

ArchitectsITALFERR S.p.A. (Definitive), Ing. Campa (Final)

Management ContractorITALFERR S.p.A.

ContractorA.T.I. – Astaldi S.p.A. (Mandatary Group Leader) –Vianini Lavori S.p.A. (assignor) - Impresa diCostruzioni Rosso geom. Francesco & Figli S.p.A.(assignor) – Di Vincenzo Dino & C. S.p.A. (assignor).– Turner & Townsend Group Limited(assignor)

Main featuresPlan dimensions: 465 m x 45 mOverall height: 12 mBulkheads: 2,500 m2


Reinforcing steel in foundation: 2,4 ktConcrete in standing structure: 23,500 tReinforcing steel in standing structure: 3,4 kt

• 1- General view. 2- Transversal section of the first part of the station. 3-4-5-6-7-8- The work site.

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The new station building

The second project for completion foresee astation building integrated into the new cityplan. The station also acts as an exchangenode with the underground MRT system.

ProjectThe Turin Porta Susa High-Speed railway station

LocationTurin Porta Susa, Italy


Design/Structural engineer/ArchitectsJoint venture of designers: AREP J.M. Duthilleul, E. Tricaud (Group leader), arch. Silvio D’Ascia,prof. arch. Agostino Magnaghi

Contract Management R.F.I. S.p.A.

ContractorJoint venture of companies: Guerrino PivatoS.p.A. (Group leader), BIT S.p.A.

Main featuresPlan dimensions: 385 m x 33,60 m Overall height:• portion below street level: 10.19 m• portion above street level: varying from 3 to12 m

Floors above ground: 1Floors below ground: 3Gross pavement area: 39,800 m2

Underground parking area: 8,700 m2

• 9- The new station. 10- The structure completed. 11- Detail. 12- Overall view.

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5. REGGIO EMILIA HS STATION

ProjectReggio Emilia High-Speed railway station

LocationReggio Emilia, Italy

ClientThe Emilia-Romagna Region and T.A.V. S.p.A.

DesignSantiago Calatrava

Structural EngineerImpresa Cimolai S.p.A. is making use of the following designers: SETECO (steel) – SGAI(concrete)

ArchitectsImpresa Cimolai S.p.A. – RPA (architecture) withthe supervision of architect Calatrava for theartistic aspect


ContractorImpresa Cimolai S.p.A.

Main featuresPlan area: 483 m x 50 mOverall height: 20 m average, since the roof has asine-wave profile and h varies. Foundation concrete: 16,734 m3

Foundation reinforcing steel: 1,700 tStanding structure concrete: 2,250 m3

Standing structure reinforcing steel: 430 tStructural steel: 9,140 t (portals, arches andbeams) Floor structure area: 7,728 m2 (loading platforms)and 200 m2 (ground-floor business stalls)


6. AFRAGOLA HS STATION

ProjectNaples – Afragola High-speed railway station

LocationAfragola district - Naples, Italy


DesignDefinitive design - Zaha Hadid arch.Final Design: SAIR – GEIE (EuropeanArchitecture, Town-planning and EngineeringGroup)

Structural EngineerDefinitive design - Zaha Hadid arch.Final design - SAIR – GEIE (EuropeanArchitecture, Town-planning and EngineeringGroup)

ArchitectsZaha Hadid arch.


ContractorATI DEC S.p.A.

Main featuresPlan dimensions (project area, including parking):450 m x 350 m Overall height: 26.45 mBulkheads: not present Foundation concrete: 38,000 m3

Foundation reinforcing steel: 4,000 tStanding structure concrete: 45,000 m3

Standing structure reinforcing steel: 5,200 tStructural steel: 4,200 tFloor structure area: 23,000 m2

After the floods of October 2000 that caused enormous damage, the province of Turinappropriated funds for setting the riverbeds to rights and for reconstructing damagedinfrastructures, among which the bridge over Sangone creek. The impossibility of changingthe existing highway route, owing to the numerous constructions lined along it, compelleddesign to conserve the old bridge’s position in both line and grade, and this was a heavyconstraint on the design of the new one. In making up the design, the alternatives evalua-ted had to eliminate the intermediate supports (their presence would mean the construc-tion of deep-lying foundation structures) and increase bridge length by twenty metres, whilemeeting the hydraulics clearances required and providing a wide range of choices of con-struction methods and an architectural value suited to the site’s environmental value. At theend the choice fell on a cable-stayed bridge eighty meters long. The deck cross section, atotal 15.10 m wide, is the classical one for small-span stayed bridges: there are two maingirders, of rectangular section 2.0 m x 1.20 m, in situ-poured and lightly prestressed, onwhose centerlines are anchored the stays, with their heads placed in special places. The twolongitudinal beams bear the secondary crosswise weave, created with adherent-strand pre-stressed precastings placed at 2.50 m spacings. The pylon, a total 36 m high from its sprin-ger, consists of two blade uprights inclined by ten degrees to the vertical and connected byan upper prestressed crosspiece, placed immediately below the stay anchorages. The pylonbears eight pairs of stays that sustain the eighty-meter span and is itself anchored to theground by another four pairs of stays that are anchored on the mooring block. The pylonbase is a caisson foundation structure, such as to directly transfer the stresses below thecreek bed directly onto the bedrock.


ProjectThe Colletta cable-stayed bridge

LocationSangone creek, Giaveno (Turin), Italy

ClientThe District of Turin

DesignA.T.I. [GEODATA SpA, Mario Petrangeli & AssociatiSrl (Prof. ing. Mario Paolo Petrangeli, ing. L. Fieno, ing. L. Pinchiaroglio)]

General contractorA.T.I. (SISEA, COGEIS, FIP Industriale)

Stays (supply and work) TESIT SpA

PhotographsMario Petrangeli & Associati Srl

“Colletta” cable-stayed bridge


• 1- Plan, in which is seen too a bridge built for provisional traffic flows. 2- The brid-ge, front view.

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• 3- Pour of the deck slab. 4- The completed pylon structure, awaiting removal ofthe scaffoldings. 5- The stays system in its end configuration: in view are the twomooring blocks anchoring the stays, since the single-span bridge scheme is dis-symmetric. 6- Cross sections through the deck: typical (left) and nearby the pylon(right).

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The structure, enabled the elimination of several kilometres of tortuous road that made con-nectionswith the city of Macomer, in the heart of Sardinia, inconvenient. This was effected by a via-duct that crosses a valley 180 m broad. The nature of the place, imposed on design choicesaimed at maximizing the oeuvre’s harmony with its context. The structure comprises a single-light box deck of variable section set on two slender piers, each composed of two septumsset close together. The deck breaks down into three spans of lengths 45, 90 and 45 metres. The deck depth runsbetween 2.30 and 4.50 metres, while the depth of the box soffit slabvaries between 0.22 and 1.00 m. The thickness of the side walls is 0.50 m and the extradosslab is 0.22 m deep. The crosswise width is 13.00 m at the upper slab and 6.50 m at thelower. The piers each comprise two septums having a section of dimensions 1.00 m x 7.50m and heights of 21.70 m and 25 m respectively. The two septums are connected together at the base by crosswise septums 0.50 m thick.Deck construction proceeded by symmetric balanced advance cantilevered from the pierswith pours of successive segments; these were then prestressed in pairs by the application ofpost-tensioned cables. The deck was built of 43 segments. The two approach spans compri-se ten segments each, and the centre span comprises two facing halves, each of which breaksdown into ten segments; there are also the crown segment and the two pier-head segments.Each segment is 4.20 m long except for the centre span crown segment, which is 1.00 mlong. The deck is prestressed with post-tensioned cables. The pierhead segments were pouredon centerings, to create the base platform for the launch cars. From these segments con-struction went ahead symmetrically with two isostatic tees being built, joined by the crownsegment.


ProjectRoad viaduct

LocationRio S’Adde – SS 129, Macomer, Sardegna, Italy

ClientANAS spa

Agency letting contractorProvince of Nuoro

General designIng. Gian Paolo Gamberini

Structural designIng. Pietro Paolo Mossone

Chief of supervision of construction Ing. Gian Paolo Gamberini

General contractorCONSCOOP, Forlì – Coop. Edile “Edile di Orgosolos.c.r.l.” (awarded contractor)

Rio S’Adde viaduct


• 1- Deck cross section at the pier. 2- Cross section through deck at midspan. 3-The procedure for constructing the deck cantilevered from the pier. 4- The septumscomposing the pier structure.

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• 5- Bridge grade profile.


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The bridge over Vajont creek, which stands five kilometres in an air line from the dam, wasliterally swept away by the wave, running uphill, generated by the slide of 260 million cubicmeters of material that detached from Mount Toc on October 9th 1963. Some three yearsago the Erto and Casso city administrations decided to start up reconstruction work of thebridge and to improve the existing city road on the left bank of the creek. The solution appro-ved was a prestressed-concrete box-girder bridge built cantileverwise by successive segmentsstarting from the right bank of the creek. In its definitive configuration the bridge is a total87.25 m long and 7.50 m wide. The net span of the bridge between its bearings is 74.50 m. The rest of the deck correspondsto a stretch built on the right bank, that acts as counterweight, and anchorage for the staycables. The route in plan is in tangent, and its grade profile lies at 731 m above sea level. Theroadway section has a 4.50 m wide carriageway and a bicycle and foot path 1.50 m wide.The deck consists of a prestressed-concrete single-light box whose depth varies from 6.94 mat the bearings on the right bank, to 3.25 m in midspan. The width of the box is a constant3.00 m. The upper slab is 7.00 m wide. The depth of the extrados slab is a constant 0.25m in the center, between the (vertical) webs. It descends to 0.20 m in the cantilevered parts.The soffit slab depth varies from 0.25 m in span to 0.70 m at the starting section of the firstsegment. The webs are 0.30 m thick. The statics scheme of the structural system as a wholeis thus a fixed-jointed or simply-supported beam. The deck comprises a total of fourteen insitu-poured segments. The abutment segment, 14.00 m long and weighing 930 tons, was pou-red in a specially built form bearing on the ground. The last segment, 2.75 m long and wei-ghing 55 tons, was poured directly on a form mounted on the left-bank abutment.


ProjectRoad bridge

LocationSan Martino, Erto (Pordenone), Italy

ClientErto and Casso Municipality

Structural designIng. Paolo Giovenale, ing. S. Rossi, Roma

Geothecnical works designProf. F. Colleselli, Padova

Chief of supervision of construction Ing. P. Sommavilla, Belluno

Safety coordinatorIng. A. Tenani, Belluno

Impresa costruttriceMonti S.p.A., Auronzo di Cadore (BL)

Bridge over Vajont creek

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• 1- Longitudinal section through the bridge. 2-3 Phases in the cantilevered con-struction of the deck by successive segments. 4- The last segment connecting withthe tunnel.

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• 5- The bridge in plan. 6-7 Testing and load test ing.

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The “Cesare Cantù” cable-stayed bridge is a roadway bridge, crossing the Adda river betweenthe municipalities of Olginate and Calolziocorte (about 60 km north-east of Milan). The bridge,opened to traffic on February 13th 2009, is the seventh crossing of Adda river in the Provinceof Lecco and represents an important commercial link between the opposite banks of theriver in a very intensely industrialized area; the conceptual and executive design of the newbridge were completed in 1999 and 2003, respectively.The bridge consists of two H-shaped concrete towers, double-plane cables and a pre-stressedconcrete girder deck; the bridge girder is formed by a central span of 110.0 m and two late-ral spans of 55.0 m, for a total length of 220.0 m.The deck, made in C40/50 pre-stressed concrete, was cast in place and post-tensioned. Thedeck is 11.50 m wide and consists of 2 two-cell box girders, 1.50 m high, connected by acentral slab and by a series of 24 transverse cross-beams, providing the lower anchorage ofthe stay cables. The equally spaced cross-beams are 8.00 m apart and their width exceedsof 1.25 m per side the width of the deck. The deck was designed to have a depth of 1.5 m,corresponding to about 1/150 of the total length, so that a good transparency of the bridgegirder was attained from aesthetic stand point.The deck is suspended from 48 stay cables, arranged in two planes in a semi-fan, held at thetop of the two main reinforced concrete portal towers in special welded steel tower-headassemblies. The cast-in-place concrete towers (C32/40) are about 38.0 m high and consist oftwo concrete piers, a lower concrete wall connecting the piers and supporting the deck, uppersteel devices providing the anchorage for the stay cables and a transverse steel truss con-necting the upper part of the piers.The bridge stays use 19 and 31 16 mm diameter strands of steel cable; each strand com-prises seven 5 mm wires and is protected by a high-density polyethylene sheath.The construction of the deck was carried out by using an unusual technique. First, a series oflarge diameter steel pipes were placed on the bed of Adda river in order to allow the waterflow corresponding to the maximum expected flood level. Subsequently a stabilized embank-ment was constructed over the steel pipes and the river bed, so that the embankment wasused to continuously support the formworks for the casting of the concrete deck. It should benoticed that the above construction technique, usually adopted for very small spans, turnedout to be cost effective; in addition, the stability of the embankment and the effectiveness ofthe water pipes were successfully checked by the severe flood occurred in Italy during Spring2008.It is further noticed that, according with the seismic classification introduced in 2003, the sei-smic hazard at the bridge site was described by the ground design acceleration ag=0.05g andthe importance factor was assumed as gI=1.3. The seismic design of the bridge was carriedout in order to ensure: (a) elastic behaviour in the vertical and transverse directions; (b) non-linear dissipative behaviour (by using base isolation hysteretic devices) in the longitudinaldirection.It is worth mentioning that the reception tests of bridge included not only the procedures thatare mandatory in the Italian Code (i.e. extensive characterization of the materials and seve-re load tests) but, according to the international practice, also dynamic tests (involving deck,towers and stay cables) during the construction phases and before the bridge opening.


ProjectCable-stayed bridge, crossing the Adda river

LocationBetween Calolziocorte (LC) and Olginate (LC), Italy

ClientProvince of Lecco

DesignEng. Angelo Valsecchi (Department of Transportation,Province of Lecco)

Structural engineeringProf. Carmelo Gentile, PhD, Eng. Roberto Gentile

Management ContractorDepartment of Transportation, Province of Lecco

General contractorVitali S.p.A., Cisano Bergamasco (BG)


“Cesare Cantù” cable-stayed bridge


• 1- The cable-stayed bridge on the opening day (February 13rd, 2009). 2- Eleva-tion, plan and deck cross-sections of the bridge (dimensions in m).

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• 3-4-5- Construction phases of the bridge. 6- Aerial view of the “Cesare Cantù”cable-stayed bridge. 7- View of the bridge central span during the reception load te-sts. 8- Details of the transverse cross-beams providing the lower anchorage of thestay cables.

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The bridge replaces a temporary Bailey bridge connecting the islands of La Maddalena andCaprera. The design bridge length is 52 meters, with three spans symmetrically arranged (twolateral spans 13.5 m long and a central one of 25 meters). The rise of the arch is about 5m. The bridge is composed by double cantilever truss elements, conceived to conceptuallyallow the rotation of each side on its foundation to open the central span. The variable thick-ness of the arches that form each cantilever element is such that the vertical dead loads reac-tion is centred in each foundation. The steel truss elements connecting the upper and lowerarches have circular hollow sections (D=150 mm). The lower central span arch and the upperdeck merge at the centre in a flexural hinge transmitting shear forces. Axial forces and ben-ding moments reactions are provided at each one of the central arch foundations, by threebearings with tensile reaction capacity. Tension and compression axial force restraints are pro-vided at each abutment. The use of high performance concrete allowed to reduce the struc-tural thickness and to use post-tensioning tendons in the upper deck, to equilibrate part ofthe tensile forces developed in the upper chord of the spatial trusses. The whole bridge hasbeen protected with white polymeric resin, to provide durability and at the same time toimprove the aesthetics of the bridge.All design choices are essentially governed by environmental constraints:1. the geometry of the fixed points derived from the historical heritage, as well as the idea ofan arch bridge and of a conceptually rotating structure;2. the central span height was the best compromise between deck slope and boat clearancespace (5 m (h) by 8 m (w));3. the arch shape and the reduced element thickness minimizes the environmental impact; 4. the horizontal reactions at the abutments and foundation had to be limited because of thesoil characteristics;5. an improved structural durability was dictated by the marine environment;6. the on-site work duration had to be kept at a minimum to minimize the interference withthe local traffic;7. two longitudinal concrete ribs (20 x 50(h) cm) have the double purpose of protecting thepedestrian traffic without inserting heavy guard rail elements, and of contributing to the upperchord structural capacity;8. the external light parapet complete the sailing boat reminiscence of the whole structure,with only white resin and stainless steel used as finishing materials.


ProjectBridge between La Maddalena e Caprera Islands

LocationLa Maddalena Island (Italy)

ClientItalian Government

DesignProf. Ing. Gian Michele Calvi

Structural detailingProf. Ing. Gian Michele Calvi, Lombardi Reico srl (Ing.Giorgio Pedrazzi, Ing. Carlo Beltrami), Ing. MatteoMoratti

ArchitectProf. Ing. Gian Michele Calvi

Contract ManagementIng. Valter Frascaroli, Ing. Matteo Moratti

General contractorA.T.I. Dott. Mario Ticca S.r.l. – Sassari, S.C.I.R. S.p.A. –Cagliari, Novaco S.r.l. – Sassari


Bridge between La Maddalena and CapreraIslands

La Maddalena

Caprera


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• 1- Historical bridge in the same location (destroyed). 2- Details of the parapetsof the pedestrian walkway.


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• 3- Construction of the piles cap. 4- Detail of the active-anchorage zone of thelongitudinal tendons. 5- Temporary support and scaffolding of the central arch. 6-Joint at midspan: fixed ends of the tendons on both sides.

The Via Arenaccia is one of the Naples street system’s main arteries, a role it has maintainedas the centuries passed.In 1999 the City of Naples decided to replace the old bridge with a new structure that wouldre-establish, in its entirety, the city’s old connection – as borne witness to by 18th century maps– with its northern districts.The project is highly complex and has many critical features, arising from the bridge’s posi-tion in relation to other existing buildings, in an old urban area that is today densely popula-ted. A few of its interactions with existing infrastructures may be mentioned: two 1000 mmpipes under pressure, two sewer trunks six meters wide, 24,000 telephone cables, many elec-trical cables of medium and high voltage, and gas lines.The very slender deck (its depth at midspan is 50 cm) has the shape of a strongly depres-sed skew vault. The net span between the bridge abutments, measured on the skew, is 26.90m, with a width of 19,85 m. The slenderness coefficient, L2/f, is 1450, a very high value for aconcrete arch, putting it at the top of the list of constructions of this type internationally.The heavy thrust exerted by the arch deck is taken by the very sturdy caisson-form abut-ments, which constitute two control rooms needed to create the hydraulic bypass.Special care was taken with durability, by choosing the best material that could be used forthe purpose: self-compacting concrete, and by seeing to its physical and mechanical protec-tion through the use of a cladding, long-lasting and easily maintained. The concrete protection was thus converted into a design and architectural point of depar-ture to best fit the infrastructure into its urban context. The choice was to wrap all surfaces:piers, abutments, vault, shelf girders, etc. with glass-mosaic tesseras forming a design, its pur-pose also to give the structure uniformity and continuity.Bridges built in urban areas too often feature elements lacking order, disconnected andvarious: beams, crosspieces, bolts, welds, aprons, sails, little shelves, shelf girders, railings, guard -rails, meshes. These elements often form a disorderly universe that allows no possibility ofcleanness of form in creating an insertion into the urban fabric. The infrastructure thus beco-mes, too often, synonymous with poor architectural quality. Concrete, with its vocation forthe continuous structure, lends itself to eliminating the diversity of elements: piers, main andsecondary girders, crosspieces, shearbracings etc. The mosaic thus offered the possibility ofcreating a uniform whole. The decision to clad the bridge with mosaic tesseras gave the new infrastructure complex avery obvious character, making it readable from the road crossing it even at long distances.Design had, in this case, set up, after consulting the painter Mariangela Levita, an abstractdesign of great size in tones of black, grey and golden-yellow on a white background, the ideato transform the passageway into a true promenade amid art and architecture.The designon the down-slope abutment bends then towards the new stairway both to mark the thick-ness of the abutments (six meters deep) and to suggest the continuation of the walk towardsthe city’s upper level.The protection of the concrete surfaces, with a view to the durability and sustainability of apublic work, formed the point of departure for an artistic operation and for the best fit of theinfrastructure into its urban context, where the bridge becomes something that does not arro-gantly state its presence but that rather seeks to fit its various parts in, discreetly. The mosaicis like a mantle in Christo’s installations, which enwraps all parts of the bridge to give it confor-mity and continuity. The whole (mosaic, railings and light poles) as white as light.


ProjectDon Bosco bridge at Arenaccia

LocationNaples, Italy

ClientCity of Naples – RUP Giuseppe Pulli arch.

Design and yard managementProf. Antonio De Luca CE

AssistantsGiuseppe Mautone CE, Alfredo Sasso CE, geom. Corrado Esposito

Architectural consulting Prof. Fernanda De Maio arch., Gianluca Marangi arch.

Artistic consultingMariangela Levita

Structures consultingProf. Attilio De Martino CE

Geotechics and foundationsProf. Carlo Viggiani CE

General proof testing Prof. Roberto Ramasco CE

General contractor Joint venture of 3 parties: Fico Costruzioni srl,Amato Trivellazioni srl, Fico Giuseppe

Concrete supplierIMECAL s.r.l.

Mosaic supplier TREND GROUP S.p.A.


Project co-financed by the European Union FERS -European Fund for Regional DevelopmentP.O.R. Campania 2000-2006 Measure no. 5.1

“Don Bosco” bridge at Arenaccia.Architecture, white as light


• 1- The project design. 2- Cladding of the upslope abutment. 3- View of the down-slope abutment.

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• 4- View of the bridge with the stairway connecting Via Arenaccia and Via Don Bo-sco. 5-6 Bridge longitudinal section and plan.

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The viaduct is part of the upgrading operation on the regional highway Laghi di Avigliana, the form -er State Road 589. This secondary extraurban artery (class C in the CNR standards) is being con-verted into a principal extraurban highway (class B), in the stretch in the City of Pinerolo from sta-tion 27+800 km to station 30+250 km, by improving its exchange systems with the main com-munication ways (principally State Road no. 23). At the intersection bet ween SS 589 and SS 23 theconstruction of a traffic circle was envisaged, having three lanes and an internal radius of 91.00 m,to distribute the traffic into the various directions. Called for too is the construction of an overpass,SS 23 flying over the new traffic circle, thus freeing the bypass traffic from all the other flows. At thecrossing with Corso Torino there was a grade intersection. Since the object was to free the statehighway's route of the urban fabric, the construction of a high-speed road in cut was proposed, soas to underpass the crossing. This will be placed above site level and will be a two-lane traffic circlehaving an internal radius of 31.00 m. It will comprise two simply-supported viaducts in curve. Bothviaducts display a section involving a ribbed prestressed-concrete plate of maximum depth 1.30 m.Its outside profile is rounded to make it fit better into the landscape. This solution, which today recei-ves everybody's thorough approval for its safety and durability, offers the following advantages:• it has a significant structural mass that, even with reduced spans, makes it insensitive to the dyna-mic effects induced by the transit of vehicles;• it can harmoniously take on any form in plan;• it has a high guarantee of durability, because of its massive structure devoid of thin elements, becau-se of the reduced extent of its exposed surface as compared with more traditional solutions (beamsplus slabs), and because of the adoption of HDPE sheaths for the prestressing cables, wholly impe-netrable by the chlorides of antifreeze salts, which will certainly be used for a considerable period ofthe year;• it provides the possibility of forming the section at design’s pleasure, and thus of facilitating its fitinto the surrounding environment. Proposed for the case in question is a crosswise profile havingsizeable curved fairings and a central zone of lower width, which gives the impression of high slen-demess to an observer standing on the ground around it, even if his point of observation is but ashort distance from the structures;• there is the possibility of pursuing the structural shape with the finish works (parapets, barriers,protections for the underlying crossings), so shaped as to harmonize the structure-finishings complex. The SS 23 viaduct on the traffic circle grafting it onto SS 589 has six spans. The two side spans atthe end are 28 m long, and the four central spans are 33 m long. The structure is continuous-beamwith intermediate bearings. Considering its height (not far above ground) and the reduced interfe-rence with the road system during construction phase, its construction was envisaged as in “standardshoring”, that is with formwork and strutting bearing on the ground, through a slab to distribute theforces. The construction methods call for a complex centering involving six successive phases, eachof which involving the tensioning of the prestressing cables envisaged for the construction phases.Finally, when the pours have been completed, a further prestressing will be introduced into the finalstatics scheme through the use of a second series of cables, having two different types of trajecto-ries. The piers have a structural geometry with a profile faired into the deck’s and a very limited lon-gitudinal dimension (one meter) so as to not form a barrier effect for the observer glancing’at thepanorama from below. On the outside of the viaduct a service sidewalk is envisaged with a parapetwhose function is also that of noise barrier. It is composed of metal uprights and coloured polycar-bonate panelling having a curved profile so as to fair into the deck section. In order to facilitateinspection and maintenance of the works inspection spaces are provided for the prestressing headends. There is also the possibility of replacing the deck bearings without limiting traffic.


ProjectRoad viaduct

LocationPinerolo – traffic circle grafting SS23 to SS589

ClientAgenzia Olimpica Torino 2006

DesignStudio SINTECNA – Prof. Giuseppe Mancini CE

ContractorTorino Scavi Manzone S.p.A. – General construction


Viaduct for State Road (SS)23


• 1- Cross section through viaduct at pier centreline. 2- Cross sections through deckin span and at abutment centerline: reinforcings.

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• 3- View of the structure. 4- General plan. 5- Form for the pour of the deck. 6- De-tail, showing slack reinforcing and prestressing cables. 7- Centering to support theformwork.

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The viaduct displays a continuous-beam scheme having a single-light caisson section of variabledepth, built cantilevered from the piers with span lengths of 120 m. Only in the end stretchwhere it is expected that landslips might take place was a statically-determinate span built, itslength 45 m. On the pier shared with the continuous deck it utilized special adjustable keepersthat permit the pier to freely move without generating stress states in the decks. Design optedfor a mixed steel-concrete solution with steel webs and r.c. slabs that make possible segmentsof considerable length, although only relatively light pieces need be handled (the webs), the wholebeing completed with in-situ-poured top and bottom slabs. The hollow piers appear as shafts,running from 15 m to 87 m tall, of square cross section whose dimension varies with shaft hei-ght. The minimum width lies at around 25 m from the taller pier tops, so that the piers tallerthan 30 m display a trumpet profile that contributes to fining the structure, while at the sametime ensuring for the taller shafts an optimum sizing against instabilities. The section of leastdimensions displays plan dimensions of 6x6 m. For the pier and abutment foundations circularcaissons filled with concrete were used, their diameters varying from 9 m to 14 m. This solutionwas dictated by the nature of the soils (fractured and scaly oxylites), by the considerable forcesunloaded by the viaduct and by the need to counter landslips in the viaduct's end area. The via-duct in question is the first structure in Italy in which the caisson deck, built of wed segmentscantileverwise from the pier, is of mixed structure, with the use of both outside prestressing andbonded-cable prestressing. This choice enabled construction of the 9-meter-long segments owingto the fact that the elements to be handled consisted of the steel webs only (and were thus ofrelatively low weight), a building-construction crane anchored to the pier being used. Once thewebs were positioned they were connected together with a system of provisional guys and strutsthat made it possible to obtain the section form desired, the distortions necessary in the stret-ch in tangent being imposed as well to create the screw-like behaviour of the section tied to thechanging crosswise slopes. Once the webs were blocked in definitive position the lower slab waspoured, using precast predalles as throwaway forms. In the final phase the upper slab was pou-red, using a metal form sliding within the caisson, precast pilaster strips being used to anchorthe upper prestressing cables. Once the tees were connected the outside continuity cables werethreaded, being deflected by means of two full diaphragms placed at the quarter-spans, andfinally they were tensioned to one-half design value. Thus the construction rises could be redu-ced. Only when the entire deck had been built and solidized, was the tensioning of all the out-side cables completed, starting from the Parma side and ending on the La Spezia side.


ProjectA15 Roccaprebalza road viaduct (North carriageway)

LocationCisa Motorway A15: upgrading of the motorwayroute at the Vigne viaduct

CustomerAutocamionale della Cisa S.p.A.

DesignStudio SINTECNA - Prof. Giuseppe Mancini CE

ContractorL.A.S. S.c.a.r.1.


“Roccaprebalza” viaduct


• 1- Cross section through deck in span (a) and at pier centerline (b).

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• 2-3-4. Cantilevered construction from the piers of the mixed steel-concrete struc-ture deck. 5- Detail of the connector reinforcings for the precast pilaster strips.

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In June 2005 the masonry bridge spanning the Potenza river near the town of Macerata (92metres long with five 18 m spans) was closed to the vehicular and pedestrian traffic becauseof evidence of structural damage, apparently due to significant foundation settlement. The importance of the bridge (average traffic 24,000 vehicle/day) required a rapid intervention,that should have also addressed the problem of seismic safety (with an expected peak groundacceleration of about 0.25 g for a 10% probability of exceedence in 50 years).The architectural and historical relevance did not allow signigicant changes in the aesthetics ofthe bridge. The proposed solution was to built a new concrete bridge inside the old masonry one,only emerging with the new deck, wider than the original one to adequate the geometry to thepresent prescription for vehicular and pedestrian traffic. The final design and the work program were developed in ten days. Works were completed in40 days. The conceptual design was very simple:1. two piles (D=1,2 m, length 30 m) were drilled into the three existing piers and at the abut-ments;2. all material above the arch structures was removed; 3. five transverse cross beams were casted above each pile couple;4. thirteen high damping rubber bearings were installed, 2 on each abutment and 3 on each pier;5. a deck composed of precast extruded beams was placed on temporary supports;6. top slab and six diaphragm beams were casted in situ;7. the temporary supports were removed;8. the structure was completed with a waterproof membrane, protection, finishing, asphalt, para-pets, joints, deck drainage, lighting systems, and traffic signals.


ProjectSandro Pertini road bridge

LocationMacerata, Italy

ClientCounty of Macerata

DesignProf. Gian Michele Calvi CE, Matteo Moratti CE

Structural detailingProf. Gian Michele Calvi CE, Matteo Moratti CE

ArchitectsProf. Gian Michele Calvi CE, Matteo Moratti CE

Contract managementStudio Calvi s.r.l., Pavia, Italy

General contractorA.T.I. Rosi Giancarlo Costruzioni Srl, G.S. CostruzioniGenerali Srl, Cagnini Costruzioni Srl, Costruzioni EdiliSirolesi Srl, Dell’Orso Perforazioni Srl

“Sandro Pertini” bridge upgrade


• 1- Longitudinal section. 2- Deck plan.

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1 - Drilling

2 - Water bed

3 - Original concrete foundation

4 - Siffening foundation with micropiles for masonry piers

5 - Sands, silty sands and sandy silts6 - Gravel, gravel with sands7 - Argillaceous silts8 - Sands with gravel9 - Siltstone clays with thin sandy layers

LEGEND


• 3- Emptying of the bridge, with the heads of the piles already driven. 4- The emp-tied bridge. 5- Longitudinal section through the project site’s geology. 6- Longitudi-nal section. 7- Cross section. 8- Laying the aseismic isolations. 9- Laying the preca-st beams.

The bridge has a single span of about 66 m passing over an important road and a river, iscurved in plan and in elevation, the deck has a variable thickness of about 300 mm and issustained by a single pylon, hinged at the base. The main structure has been mounted in fourdays, without any temporary support, using six identical prefabricated deck sections, each onebeing 12 m long, and a 35 m monolithic steel pylon to which the deck sections are ancho-red, by means of four cables each. The total construction cost has been approximately500,000 €. The bridge is technologically highly innovative, light, beautifully inserted in theenvironment and very cost effective. The design choices were essentially guided by the envi-ronmental constraints:• no intermediate support was really possible, and only on the west side topography and build -ing locations permitted a relatively easy construction of foundations; on the same side, anunderground parking under construction provided some appropriate anchoring mass;• the beauty of the valley and the presence of an ancient stone bridge required a light struc-ture, with minimum interference with the surroundings;• the construction time on site needed to be reduced to a minimum, to mitigate as much aspossible traffic interruption on the main road.It was decided to design an asymmetric steel pylon, rotated both vertically (about 7 degrees)and horizontally, in order to optimize the force distribution. The pylon is made of a monolithic 35 m steel pipe (812 mm external diameter) with asecond external co-axial pipe (850-1100 mm variable diameter) welded to the internal oneby six radial steel wings). The pylon is hinged at the base and its position is essentially governed by the actual loading,with a variable inclination. Eleven tendons (52 mm maximum diameter) restrain the pylon atthe ground. The deck is formed by five precast high performance concrete elements supported on 10couples of thinner cables. Each segment has the same length (12 m) and the same radiusof curvature both in plan (about 300 m) and in elevation (about 1200 m). The in–plane curved shape of the deck is effectively reacting to horizontal loads by archingaction, whilst vertically the deck is free to rotate around a horizontal axis on the west sideand is connected to the east abutments with a double-hinged 6 m long truss that allows ver-tical free movements and rotations of the deck. Pylon foundation, abutments and anchor mass for the fixed cables were constructed on site,taking advantage of the contemporary construction of an underground parking lot. The pylonwas transported overnight in a single piece and mounted with two cranes; within the subse-quent three days it was possible to position and anchor the five deck sections, prefabricatedelsewhere. During construction, a temporary connection between the deck sections was pro-vided by steel self centring couplers, later on included in concrete injections that made thedeck fully continuous. The results of the complex nonlinear time-history simulations carriedout during the design phase were later confirmed by in-situ dynamic testing, with induced ver-tical displacements of ± 180 mm.


ProjectCable-stayed footbridge over the Frodolfo river

LocationBormio, (Italy)

ClientBormio Municipality

Structural detailingProf. Gian Michele Calvi CE, Dario Compagnoni CE,Matteo Moratti CE

ArchitectProf. Gian Michele Calvi CE

Contract ManagementStudio Calvi s.r.l., Pavia, Italy

General contractor G.A.L. costruzioni, Bormio, Italy

Cable-stayed footbridge over the Frodolforiver



• 1- Post-tensioned concrete plinth at pylon base. 2- Temporary self centring cou-plers (clockwise: plan view; section B-B; detail of the steel pin; section A-A RC tran-sversal beam cast on site). 3- Main section of the deck (hatched zone indicates RC,dimensions in mm). 4- Main geometry of the footbridge frontal view from South. 5-Horizontal hinged bearings at the West abutment. 6- Deformed shapes under liveloads of the f. e. model (from left to right: South-East view; East view; plan view).

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The new stretch of the SS16 (State Road) bypass between Pescara and Montesilvano featu-res, besides an 1800 m long tunnel, a new bridge over the Mazzocco creek, described here.Both the contracting agency and the local authorities considered the arch bridge to be a sinequa non for the project underway, even though the creek could have been crossed with a tra-ditional prestressed-concrete-beam bridge, built by successive segments.Another item that could not be ignored was that the projected works had to be compatiblewith the future doubling envisaged, which would upgrade this section to the characteristics ofalready-existing SS16 alternate, thus obviating future demolitions.The construction procedures were so worked out as to not need centerings, which would havehad to stay in place for a long time since they would be in the creek bed and there was noguarantee for the whole period of construction that the centering’s ground supports wouldnot settle differentially, owing to the presence on a creek bank, revealed by soil studies, ofwaste material originating from the driving of an old tunnel.Mazzocco creek was crossed on a four-span bridge a total of 140 m long. The 70 m archspan founds on circular-section caissons 15m in diameter, of maximum depth 25 m. Pier 1 and the abutments found instead on �1200 piles up to 30 m in length, which had toassure, besides the required bearing capacity, a minimum strain so as not to induce drops inthe arch thrust owing to interaction with the not especially good foundation soil.The piers, of lengths up to 14 m, are two-column frames, since deck width varies between17.2 and 26.23 m. The decks were built with 4 precast concrete U-beams 1.6 m deep, pre-stressed with bonded strands, their length between 24 and 24.7 m. The first span was builtinstead with 18 double T beams 1 m deep.The 4 V beams, between piers 2 and 3, bear on the crosspiece connecting the five membersconstituting the arch. Each member, its double T section varying continuously from the baseto the minimum section at the crown which is 1.20 m deep, comprises three precastings join-ted in place. The central element of the three is prestressed with two post-tensioned cablesof ten 0.6” super-strands, anchored on the two ends of the precasting.The connection of the members among themselves and to the foundation footing is effectedby prestressing bars.The precastings were built on site to limit handling. For their launching and mounting two pro-visional struts 15 m tall were set up, each built with five Innocenti steel-pipe towers connec-ted together and shearbraced, bearing on two provisional footings founded on � 800 piles.In the first phase all 5+5 low elements were launched, bearing by means of supports on thecaisson footings and on struts. After launching by means of a 300 ton crane of the central elements, the stitches were effec-ted and the provisional struts unloaded, this being carried out progressively by means of thelowering of a series of large screws and sand-filled boxes placed below the provisional bea-rings of the precastings, the structure’s strain behaviour being constantly monitored to checkon the correspondence of the shifts with theoretical. Thus was completed the launching of the deck beams and the subsequent pour of the slabs.


ProjectSS16 Adriatica - Bypass of the settlements ofMontesilvano and Marina di Città Sant’Angelo – jobsegment 1 – Road bridge over Mazzocco creek

LocationPescara, Abruzzo, Italy

ClientANAS S.p.A.

DesignProf. M. P. Petrangeli CE, E. Cipolloni CE

Management ContractorANAS S.p.A.

General contractorA&I Della Morte (Naples) - Impresa Martella,Pescara, Italy


Bridge over Mazzocco creek


• 1- Bridge plan.

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• 2- Forms for the arch central precasting being struck. 3- Lower arch precastingbeing mounted. 4- Handling the precastings. 5- Detail of the provisional support forthe precasting on the footing. 6- Longitudinal section. 7- Central precasting launchphase. 8- Detail of the precastings’ upper connections.

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The bridge over the Sacco river is one of a series of operations called for by the new highwayssystem setup, itself made necessary by the construction of the new high-speed railway linebetween Rome and Naples.The solution adopted is an arch bridge built using high-performance curved precast-concreteelements. They are positioned on provisional supports and locked in place by completionpours. To balance the thrust, at the abutments the arch ends are connected by diagonal strutsto the deck, which thus takes on the role of chain and is under tension. The moment arisingfrom the eccentricity of this force relative to the arch thrust is balanced by the action of acti-ve tendons placed at the deck ends, so that the forces transmitted by the structure to the soilare practically vertical.To eliminate continuous forms, which significantly affect the structure’s overall cost, in the con-struction of the arches precast curved elements were used, of high-performance concrete, pla-ced in position by the use of provisional supports between the spans and later solidized by insitu pours.In order to cut construction time, a solution was adopted for the deck construction involvingprecast beams completed by an in situ pour. To balance the thrust of the arches, which could be incompatible with the soil’s mechanicaland strain characteristics, their ends at the abutments are connected by diagonal struts tothe deck. It thus takes on the statics role of a chain and is thrown into tension.The moment given rise to through the eccentricity of this force relative to the arch thrust isbalanced by a counter couple due to the action of active tendons anchored to the deck endsand by the corresponding reactions of the piles below the abutments, so that the forces trans -ferred to ground are prevalently vertical.The structure has a total length of 132 meters and comprises two arches, 56 m in span witha 5.6 m camber, which sustain the deck connected to them at the crown and at the abut-ments. The structure’s “permeability” reduces its environmental impact and interference with the flowof water, increasing the flowrate and lowering the crosswise hydraulic thrust. Each arch is created by the assembly of 20 x 5 precast arch segments of high-performanceconcrete, having a 70 cm x 50 cm section. They are set side by side to create a structure 10m wide, then solidized together by the pour of a concrete slab of 25 cm minimum depth.Since the bridge centerline is skew to the river’s flowline, the precast arch segments are moun-ted having a mutual longitudinal slide so as to follow this geometry. In each arch’s centralzone the deck slab is directly connected to the arches by ribs of variable depth held up bythe arches. At the abutments and the pier the support consists of precast inverted-T-sectionbeams .The total width of the deck is 13.10 m. It is sized for a road having two 4.75-meter lanesand two sidewalks raised up and protected by guardrails.The abutment and pier foundations are built on large-diameter piles (150 cm), their lengthvarying from 16 m at one of the abutments to 33 m at the central pier.


ProjectRoad bridge over the Sacco river

LocationMunicipality of Sgurgola, Rome, Italy

ClientConsorzio Pegaso S.C.AR.L.Roma I.T.S. S.p.A.State Railway System – High Speed lines

DesignPROGEEST S.r.l., Prof. Arch. HC E. Siviero CE, Prof.R. Di Marco CE

Structural engineeringEnzo Siviero, Roberto Di Marco

ArchitectsEnzo Siviero

Management ContractorConsorzio IRICAVUNO

General contractorConsorzio IRICAVUNO Consorzio PEGASO S.C.AR.L.

Bridge over the Sacco river


• 1- Bridge side view.

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The design solution and the arrangement in plan ensure simplicity and lightness to the whole,in perfect harmony with the surrounding landscape and fitting with the environment. In plan,the highway bridge has a deck in the form of a parallelogram, whose major side is 64 m long,while the width is 12 m. As regards durability and resistance to weathering the choice of materials assures the struc-ture the capacity to last if properly maintained. The structure’s symbolic significance too maybe considered quite as durable. Costs are in proportion to function, to structure safety, to theadvantages that the new structure produces, to the construction phases and schedules neces-sary to finish it, and to the positive judgement on the view of the whole at the end.The bridge has a basic characteristic: its centerline is skew to the river’s flowline.This design choice gives the structure a special characteristic, creating a very interesting fore-shortening in perspective and a play of light and shadow.Its centerline forms a 36° angle with the river’s flowline, this being perceptible from somepoints of view but absolutely invisible from others. The result is, at times, a forced perspecti-ve that tends to lengthen the image laterally.In the same way, the deck and arch dimensions and the imposingness of the whole structu-re are legible only when compared with the human dimension, bearing witness to a carefulsearch for proportion. Despite the strong dimensions the eye is not especially impressed bythe beam depths, whether before or after the successive pours.Its juxtaposition with the existing bridge is natural: after having undergone restoration theolder structure will be used as a foot bridge, while continuing to be a memory of the past ofthis part of the territory.Structurally speaking the bridge comprises the following principal reinforced-concrete ele-ments:- a central parallelogram-plan arch 1.00 m deep, thrown across a 39-meter chord; - a horizontal parallelogram deck 1.00 m deep and a width perpendicular to the sides of 7.00 m;- along the lateral edges two wings are cantilevered, each 2.50 m wide, thus bringing totaldeck width to 12.00 m;- two diagonal lateral plates, 1.00 m thick, connect the central-arch springers with the endedges of the deck plate; - two support walls in the riverbed, created by continuous bulkheads 1.00 m thick, having aplan length of 12.5 m, thrust down to a depth of 22.00 m below the arch springer;- two support walls at the banks, built from continuous bulkheads 1.00m thick, having alength in plan of 12.50 m, thrust down to a depth of 18.00 m below the springer of the hori-zontal upper deck.


ProjectRoad bridge over the Santa Caterina

LocationSant’Urbano, Padua, Italy

ClientMunicipality of Sant’Urbano, Province of Padua

DesignPROGEEST S.r.l., Prof. HC E. Siviero arch. CE

Structural engineeringEnzo Siviero, Luigi Rebonato

ArchitectsEnzo Siviero

Management ContractorMunicipality of Sant’Urbano

General contractorImpresa Locatelli, Impresa Thiene

Bridge over the Santa Caterina channel

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• 1- Longitudinal section: slab reinforcings. 2- Cross section through deck.

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• 3- Longitudinal section: foundations. 4- Plan of the highway route in which thebridge is fit. 5- Longitudinal section: reinforcings of the arches and of the diagonalslabs.

As part of the general project for upgrading the roadway system to which S.P. 92 belongs, theTreviso Provincial Administration indicated its intention to widen the structures so as to ena-ble them to take the carriageway common to type V CNR roads (extraurban C2 according tothe new operating standards for highway construction, Ministerial Decree of 11 May 2001).This involved upgrading the carriageway to a total width of 9.50 m and inserting servicesidewalks on both sides. During this phase the construction of just the deck holding the vehi-cle way is envisaged, while the sidewalks will have to await new financing. The considerable use of precasting technology for the slab enabled shortening the construc-tion-yard phase, greatly reducing the inconveniences caused to traffic. The system designedenvisaged the construction of six typological slabs combined in four types of segments to defi-ne the bridge plan for a total length of 420 m, broken down into 20 m spans. The deck comprises three principal full-web beams 1000 mm deep, spaced 3500 mm apart,connected to the 26 cm deep r.c. slab by Nelson-type rungs. The slab is constituted of pre-cast plates having a 2 m module, broken down into two parts to cover the entire carriagewaywidth. The plate depth coincides in the central portion with the slab’s finished depth, while insome areas (lateral relative to the plates) and in the zone where it is supported on the steelbeams, the slab is lowered to permit an in situ pour of concrete, in order to bring about sta-tic continuity both longitudinally and crosswise, as well as with the steel beams below. The pla-tes are so positioned as to supply the crosswise slope of the roadway plane, making the depthof the asphalt pavement constant over the entire roadway surface, and consistent with thebehaviour of the centerline in plan. Envisaged are trestlework diaphragms (cross pieces) distri-buted at 4.00 m spacings and connected by diagonal shearbraces placed at the level of thelower flat arch in such fashion as to form a structure of considerable torsional stiffness, ableto distribute the eccentric loads almost uniformly over the three principal deck beams. Thetotal deck width is 10.70 m, of which 9.50 m are carriageway, while 60 cm for each side ofthe cross section are used to hold the guardrail and its cladding. On each pier and on theabutments elastofip-type bearings are called for, made up of a coupling of neoprene, steeland confined teflon, so as to create unidirectional bearings for the central beams, and multi-directional ones for the border beams, able to permit the shifts due to temperature changesor to creep. For dynamic forces of the instantaneous type (earthquake, braking) it is providedthat the bearing for the central beam be fixed longitudinally as well, so as to distribute overeach pier the dynamic forces falling to it. Called for anyway is the insertion of a fixed bearinginto one of the two central piers. Called for on the abutments is the insertion of expansionjoints able to take the shifts due to the summation of slow and dynamic forces. It has beenascertained, on the basis of preliminary calculations and after checking the original staticsrelationships, that the hollow box structure overlying the beams’ current support plane is notessential to the strength of the pier-pulvino complex and will thus be demolished (three sides).Therefore, the construction can be prospected, above the pulvino itself, of a continuous steel-concrete deck outfitted with expansion joints only at the abutments. The structure will appearslenderer than the preceding one (H=1.26 m versus H=1.58 m). User comfort will be increa-sed and maintenance costs reduced owing to the radically reduced number of joints and bea-rings. Under the conditions indicated above, the carriageway can be widened to 9.50 m. Inrelation to traffic conditions (current and future) the placement of lateral containment bar-riers is envisaged, sized for an H2 impact-severity index. On the outside of the curbing theplacement of a metal cladding 2.80 m high is called for, its function to clad the deck as wellas to provide partial protection of the roadway surface.


ProjectRoad bridge over the Cimadolmo branch

LocationCimadolmo, Treviso, Italy

ClientProvince of Treviso

DesignPROGEEST S.r.l., Prof. Arch. HC E. Siviero CE

Structural engineeringEnzo Siviero, Luigi Rebonato, Federico Zago

ArchitectsAlessandro Stocco

Management ContractorProvince of Treviso – Lucio Bottan arch.

General contractorF.lli PACCAGNAN S.p.A., Ponzano Veneto (Treviso)


Bridge over the Cimadolmo branch


• 1- Axonometric view of the plate constituting the floor slab. 2- Axonometric ex-ploded view of the floor slab. 3- The paired piers, placed with a 20 m interaxial di-stance. 4-5 Laying the precast slabs.

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The project presented concerns a flyover which construction was completed in 2007, locatedin Verona - Isola della Scala, Italy. The total length of the structure is approximately of 400mwith 13 spans. According to the designers’ knowledge, this is at the moment the longest IAB ever built. Theconstruction of the bridge, initiated in 2001 as a simply supported flyover, was interruptedafter 2 years because of economical problems. At the time of interruption, all pre-stressed con-crete girders had been nevertheless purchased. At the beginning of 2006, works restarted with a new proposal, aiming to improve the qua-lity of the structure and change the static scheme from simply supported to integral abut-ment without changing the built parts, namely, the rigid abutments and the piers, in the pur-pose of not to increase the cost of the final structure. During refurbishment, in order to achieve an IAB eliminating all bearings and expansion joints,continuity was attained at the pier caps with the casting of concrete diaphragms between thebeams of adjacent spans, in order to achieve negative moment resistance. Hogging momentresistance was also determined with a similar technique at the abutments for the end bays.Connection between adjacent beams was carried out casting the concrete of the diaphragmsalso inside the V-shaped girders for a length of 2 meters. The bridge was opened to traffic in 2007; no mentionable damages have been noticed untilnow, except for some cracks in the approach slabs.


ProjectRoad bridge

LocationIsola della Scala, Verona, Italy

ClientANAS S.p.A. (Italian Road Administration)

Original designANAS S.p.A.

Redraft design and structural engineeringProf. Enzo Siviero CE, Prof. Bruno Briseghella CE,Prof. Tobia Zordan CE

Management ContractorANAS S.p.A.

General contractorNuova Bitumi srl, Trento


“Isola della Scala” bridge


• 1- Typical cross section. 2- Longitudinal section at the pier. 3-4 Detail of the deckreinforcings. 5- Typical crosspiece during construction. 6- The bridge in an advancedphase of construction. 7- Viaduct scheme.

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ProjectThe first most urgent operations to make safe thestructures involved in the earthquake of April 6th

2009

LocationMotorway A24 – Various viaducts between 99+000km and 116+500 kmMotorway A25 - Popoli viaduct

ClientStrada dei Parchi S.p.A.

Firm Responsible for the proceedingStrada dei Parchi S.p.A. – Marco Carlo Rocchi CE

DesignSPEA Ingegneria Europea – Fulvio di Taddeo CE

Supervision of constructionStrada dei Parchi S.p.A. – Luca Bartoccini CE

General contractorTOTO S.p.A.


Main featuresThe project calls for the provisional restoration tofunction of the structures, with a view to re-openingthe following viaducts to traffic:A24 - Fornaca viaduct,A24 - Genzano viaduct,A24 - Raio viaduct,A24 - Aterno viaduct,A24 - S.S. 17 viaduct,A24 - Fosso Vetoio viaductA24 - Pettino viaductA24 - S. Sisto viaductA24 - S.Giacomo viaductA24 - viaduct on the L’Aquila Est interchangeA24 - Le Campane viaductA24 - Palude viaductA24 - Viadotto Vigne BasseA24 - Costa del Molino viaductA25 - Popoli viaduct

The “Strada dei Parchi”

On April 6th 2009 at 3.32 AM the L’Aquila area was struck by a strong earthquake (Richter magni-tude (Ml) of 5.8, Moment magnitude (Mw) of 6.3, 8th-9th degree on the Mercalli scale). The earth-quake sequence continued developing with a great many aftershocks, more than 20,000 of themrecorded as of June 10th 2009. 31 of them had a M1 lying between 3.5 and 5 and three had amagnitude exceeding 5 (April 6th M1=5.8, April 7th M1=5.3, April 9th Ml=5.1)The plan distribution of the aftershocks brings out very well the area concerned by the earthquakesequence, extending for more than 30 km in the NW-SE direction, parallel to the axis of theApennine chain. The earthquakes of the sequence took place for the most part in the upper crust, a depth of 10-12 km. Only the event Ml=5.3 of April 7th to the SE of L’Aquila was as deep as 15 km. The datagathered to date (seismicity, GPS, SAR, geology) agree in identifying the structure responsible for themain shock as a fault having direct movement that extends some 15 km in the NW-SE directionwith a SW dip. Its extension on the surface is located in correspondence with the Paganica fault.The damage in the epicenter zone was due, not only to the size of the earthquake (and therefo-re to its magnitude), but also to the break’s direction of propagation and to the soil geology.In particular, the major damage is observed in the direction along which the faulting propagates(effect of the source directivity) and is amplified in the areas where “soft” sediments (such as allu-vial deposits, earth fill, etc.) lie on the surface.In the case of the L’Aquila earthquake, the break associated with the April 6th event was propa-gated from below upwards (and therefore towards the city of L’Aquila) and from northwest tosoutheast, towards the Aterno valley.Motorways A24 and A25 have a total length of 281.4 km, and feature 174 viaducts of variouslength and typology. Their total length sums to 58.3 km (21% of highway length). They were builtbetween the end of the sixties and the first half of the eighties. Added to these are 77 overcros-sings so that there is a high incidence of crossing structures, owing to the territory’s orography.The seismic event involved an extensive area crossed by motorways A24 and A25, causing muchdamage to the infrastructures.After the shock that struck at 3.32AM of April 6th 2009, Strada dei Parchi SpA activated its ownengineering structures and those of the principal companies in the sector to carry out checks onthe infrastructures managed.The most serious damage observed was that done to the bearings of forty spans on nine viaducts.The consequent discontinuities at the expansion joints exhibited 10-20 cm steps. The damage undergone by the roadway infrastructures varied depending on their zone. Threemotorway stretches (two on A24 and one on A25) can in fact be identified exhibiting very obviousproblems: • A24 –from the Tornimparte interchange to that of L’Aquila West, featuring moderate damage toviaducts and settlements of embankments at the viaduct abutments;• A24 –from the L’Aquila West interchange to that of Assergi, featuring both important damage tothe viaducts (particularly to the S.Sisto viaduct), and breakage of the motorway embankment andits settlements at the abutments;• A25 –from the exit for Pratola Peligna to the exit for Bussi, with motorway embankment settle-ments at the abutments of some viaducts and serious damage to the Popoli viaduct.The activities of monitoring the structures started up by Strada dei Parchi SpA and the very firstrepairs carried out right from the start of the event kept the motorway open for rescue vehicles.On the other hand, the activities necessary to immediately restore the infrastructures enabled reo-pening to traffic the stretch lying between the Tornimparte exit and that of L’Aquila West at 8.00PMof April 6th, without any limitations on speed. On this stretch Civil Defence anyway kept in force thelimitation for vehicles heavier than 7.5 tons in order to handle the rescue vehicle flows.


• 1- Map of L’Aquila area, showing active faults. 2- Raio viaduct: yielding of theabutment body.

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The motorway stretch between the L’AquilaWest station and that of L’Aquila East wasinstead reopened to traffic on Friday April10th at 5.50 PM on the S.Sisto viaduct. Thisinvolved using one two-lane carriageway withtraffic in both directions, but without weightlimits and with the sole constraint that topspeed be limited to 60 kmph. For the leftcarriageway the situation was instead morese rious and more restoration time was need -ed.On A25, in the stretch lying between the sta-tion of Pratola Peligna and the Bussi station,the critical element was the Popoli viaduct,with separate carriageways. In particular theleft carriageway (Pescara to Rome) wasmade transitable on an emergency basisonly for light vehicles right from the firstminutes after the earthquake. Thanks to therestoration operations carried out, on Thurs -day afternoon of April 9th all the weight con-straints that conditioned transitability couldbe removed. Traffic then ran in both sensesof flow on a single carriageway up throughJune 5th 2009, when the detour was remo-ved and circulation returned to normal.In all these operations, contracted to TotoSpA through the Extreme Urgency procedu-re, a daily average number of eighty personswere working with the numerous equipmentitems available (by-bridges, cranes, hoistingequipment, trucks, cutters, finishing machi-nes, rollers, etc.) as well as seventy personsbetween engineers and workmen of Stradadei Parchi SpA.To be highlighted in particular is the un -common spirit of self-sacrifice and emotionalinvolvement exhibited by the executives,engineers and workmen involved in the job,who worked under conditions truly at thelimit. Only thanks to this spirit of sacrificewas it possible to assure full-capacity drivingalong the motorway on April 10th.


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EPICENTRO DEL SISMASEISM’S EPICENTRE


• 3- Areas involved in the earthquake of April 6th 2009. 4- San Sisto viaduct: breaka-ge of the roller and detachment of rack from the rack-roller supports. 5- San Sistoviaduct: step resulting from breakage of the supports. 6- Vigne Basse viaduct: yiel-ding of the embankment on the abutment body. 7- Popoli viaduct: step on theplatform owing to the expulsion of the roller from its support. 8- Le Campane via-duct: complete expulsion of the roller from the roller-rack support. 9- Vigne Basseviaduct: disarrangement of the rack-and-roller support. 10- Popoli viaduct: expulsionof the roller of the rack-and-roller support at the abutment. 11- Popoli viaduct: ex-pulsion of the rollers of the rack-and-roller supports.

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The job involved the completion of motorway A24 in the Villa Vomano-Teramo stretch, by the recon-struction of the east carriageway for a length of 5.7 km. The principal structures included in theproject are:• Vomano viaduct: a structure of 770 m long, on 44 m spans whose first two spans on theRome side run on a continuous beam having a mixed steel-concrete structure. The remaining partruns on simply-supported prestressed-concrete decks having spans from 26.70 m to 35.40 m, anda continuous-slab deck. The viaduct displays over its entire length a left curve in plan. The roadwayplatform of the new east carriageway is a total 13.00 m wide, with two curbings of 65 cm and115 cm, and a carriageway 11.20 m wide.• S. Antonio viaduct: a structure having a total length of 2500 m with simply-supported beamsand 33.80 m and 35.30 m spans, with a continuous slab in 500-meter-long segments, which sub-divides the viaduct into five stretches. The roadway platform is identical to that described above forthe Vomano viaduct.• Carestia Tunnel: the structure has a total length of 824 m with a route in plan and a gradethat follows the existing adjacent west-carriageway tunnel of the same name. The soils involved inthe driving are Miocene marls, mantled by layers of detritus.The stretch in driven tunnel is 731 m long, while at the tunnel mouths there are two stretches incut-and-cover tunnel 49 m long on the L’Aquila side and 33 m long on the Teramo side. The di -stance between the two tunnel centerlines is 40 meters.The average section is 173 square meters with a driving width of 15.85 m and a height of 13.40m, and a tunnel soffit profile having a radius of 6.77 m. The tunnel roof reaches peak heights of60 meters above the soffit in the central stretch, while it stays rather moderate (10.20 m) at thetwo mouths for a total stretch of 300 m. At spacings of 300 m are two emergency foot by-pas-ses. The tunnel was outfitted with modern safety systems.

ProjectA24 – Completion of the motorwayRoma-L’Aquila-Teramo

LocationMotorway A24 - Villa Vomano-Teramo stretch

ClientStrada dei Parchi S.p.A.

Firm Responsible for the proceedingStrada dei Parchi S.p.A. – Marco Carlo Rocchi CE

DesignTOTO S.p.A. – Vincenzo Consalvo CE

Supervision of constructionStrada dei Parchi S.p.A. – Ernesto Maffei CE

General contractorTOTO S.p.A.


A24 – Completion of the motorway Roma-L’Aquila-Teramo

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• 1- A24: Panoramic view of the Vomano viaduct. 2- A24: Panoramic view of the S.Antonio viaduct. 3- A24: Mouth of the new barrel of the Carestia tunnel.

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A24 Towards L’Aquila1. Vomano viaduct2. S. Antonio viaduct3. Carestia tunnel4. Underpass transitable by car5. Subway6. Interchange underpass7. Embankment8. Abutment wall (BVV)9. Beams field.10. Begin job segment.

A24 Towards Teramo

The operation is necessary to doubling the roadway in the motorway’s last stretch of remai-ning single-lane-per-carriageway; it runs from Villa Vomana to Teramo. The job as designed, itsoverall amount being 110 million euros, consists of building all the structures needed for dou-bling the stretch.Among the structures, the S. Antonio viaduct is the most important economically, but also engi-neeringwise and construction-wise. It comprises 72 spans 35 m in length with decks com-prising two precast beams and a continuous slab poured in place. The essential aspects inaddition to those just mentioned concern the study of this viaduct’s foundations.In the stretch in question, the infrastructure in fact runs through many landslip-prone areas.This fact led the designers of the existing viaduct 20 years ago to use caisson foundations forall the piers. In our case, careful study of the site geomorphology led to a differentiation ofthe foundation type. In fact, for the valley bottom areas the choice fell on more economicalfoundations on piles (�1500).The Decks The decks are built of precast U beams 2.0 m deep, prestressed with bonded cables (80T15).The beams are connected in place by the pour of the upper concrete slab 25 cm deep.Continuity under horizontal forces is provided by the slab, which at the beam heads, owing toexpedients aimed at reducing its depth, enables transfer of the longitudinal stresses in its ownplane and hence on the one hand the use of the technique of precasting the beams on asupported scheme (with obvious economic and construction-time advantages) and on theother the possibility of seismically isolating the viaducts (Precast beams: 45 MPa; In situpours: 30 MPa; Concrete incidence: 7.5 m3/ml; Steel: slack for r.c. fy = 430 MPa; incidence:200 kg/m3; Bounded-cable prestressing: fptk = 1860 MPa; incidence: 14.5 kg/m2; Cross pre-stressing bars: fptk = 1230 MPa; incidence: 1,8 kg/m2).Seismic isolation It was mentioned that the site’s strong seismicity (ag=0.25) and the need to limit the forcesat the foundation led to widespread use of seismic isolation of the deck from the substruc-tures. In fact, the site’s orography and the constraints on the design grade profile (the viaductlies at the mouth of an existing tunnel) demanded piers of quite variable height, thoughalways less than 18.01 m, and hence an irregular viaduct, �low ductility and �high found -ation stresses. Thus, seismic isolation both crosswise and longitudinally was obligatory.The viaduct was broken down into five sub viaducts having a varying number of spans (anaverage of 15), whose constraints scheme envisages, longitudinally, connection with elastomerisolations at the five central piers and moveable constraints on all the other piers and abut-ments. In the crosswise direction fixed constraints on the abutments or joint piers and ela-stomer isolations of varying stiffness on all other piers. By acting on the isolations’ stiffnesses(32000kN/m ÷ 80000kN/m), the forces on the foundation could be regularized and diffu-sed practically uniformly over all piers, whatever be their height. The result was the adoptionof joint devices and most especially of slides for the moveable bearings, of significant dimen-sions: (�l � 260 mm; �t � 150 mm).


ProjectMotorway A24 – Rome-L’Aquila Teramo –Completion of the doubling between Villa Vomanoand Teramo – the new S. Antonio viaduct

LocationVilla Vomano, Teramo, Abruzzo, Italy

ClientStrada Dei Parchi S.p.a Holder of the Motorway A24 Rome-L’Aquila Teramoconcession for ANAS

DesignProf. M. P. Petrangeli CE

Management ContractorStrada Dei Parchi S.p.a.

General contractorTOTO Costruzioni S.p.A., Chieti, Italy


“S. Antonio” viaduct


• 1- General plan, from pier 0 to pier 45.

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• 2- Excavation for the foundation caissons (Ø = 12 m; h = 20 m). 3- Construc-tion of a prestressed beam. 4- The beam precasting field was located in the Westconstruction yard. In view is the tensioning head for the adherent-strand prestres-sing cables. 5- Longitudinal profile: new East carriageway works. 6- Cross section th-rough prestressed-concrete beam. 7- Deck construction phase, with the launch ofthe beams by steel launch car bearing on the piers. 8- The new viaduct: in view isthe S. Antonio ditch, repositioned and diverted between the two viaducts. 9- Pre castbeam plan.

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The viaducts were built as part of the construction of a stretch of 15 kilometers of the East-West motorway in Algeria. Both were built of prestressed prefabricated reinforced-concretesegments having a caisson cross section. The structure at station 50.1 kilometers comprised a continuous deck of two separate car-riageways, each of five spans having end spans 45.50 m long and intermediate spans 70 or80 m long. Both carriageways are in curve, its plan radius 500 m. The structure at station 49.2 kilometers also comprises a continuous deck of two separatecarriageways. It has eight or nine spans, each having end spans of 35 m or 45 m and inter-mediate spans of 50, 60, 70 or 80 meters. Both carriageways are in curve, its plan radius500 m. The 15.57 m wide individual carriageway was built having a single-light caisson section, itsdepth varying from 2.80 m (in midspan) to 4.00 m (at the piertop). Both viaducts were sei-smically isolated using elastoplastic dissipators both longitudinally, at one abutment, and cros-swise (at the piers). The degree of seismicity corresponds to a peak acceleration at the groundof 0.35 g. The foundations are on 1200 mm diameter piles. The construction system calledfor precast segments using the “short line” system, the segment earlier cast being used asform wall for the next one. Launching was effected by a launch car, the segments were moun-ted with provisional prestressing bars; the tee prestressing cables being subsequently tensio-ned.


ProjectRoad viaduct

LocationAlgeria, El Affroun-Hoceinia job segment

CustomerA.N.A. National Motorway Agency

DesignStudio SINTECNA – Prof. Giuseppe Mancini CE

ContractorCooperativa Muratori & Cementisti CMC, Ravenna


Viaduct for the Algeria East-West Motorway


• 1-. Cross section through double-deck viaduct. 2- Structural steel for the spansegment.

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• 3-4 Handling and installation of the deck precast segments. 5- Structural steelfor the pierhead segment. 6- Steel reinforcing of the segment. 7- Dissipator for thestructure’s seismic isolation.

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In Rome’s Flaminio district, nearby the capital’s new Auditorium, has risen the MAXXI, thenational museum for the arts of the 21st (XXI) century, conceived by architect Zaha Hadid.Its construction began in 2003, in a construction yard of experiment and innovation. The newstructure, which houses museums and cultural activities as well as workshops and expositionspaces, brought about a great transformation of the entire block. And this even if the designsolution adopted took its point of departure from a reading of the context, configuring a fabricand a volume in continuity with the strictly horizontal lines of the surroundings. Entrance tothe MAXXI is gained in the heart of the block. From the building-high hall access is had to two museums – Maxxi Art and Maxxi architec-ture – and to the reception services, the cafeteria, the book shop and the spaces for tempo-rary expositions. Outside, a foot route insinuated below the overhanging volumes follows thebuilding plan, restoring an urban connection interrupted by the earlier military installationoccupying the lot. The architectural and structural elements connoting the oeuvre basicallynumber two: the walls that delimit the exposition galleries and that determine the interlacingof the volumes; and the transparent roof that naturally lights the rooms. The concrete wall is the element organizing space, while the roof system is the highly inno-vatory technological and systems element. In fact, integrated into the roof are the skylight-frame elements, the devices for controlling natural lighting, the artificial lighting fixtures, andmechanisms for limiting heat from solar irradiation. The roof system comprises a dual glazingand is protected on the outside by a metalgrille sunshade that, besides screening light, actsas walkways for maintenance purposes. Cement concrete (self-compacting concrete, SCC) isthe MAXXI’s true protagonist. In fact of r.c. are the walls characterizing its form and struc-ture, as too are the horizontal surfaces and the roof blades, entirely clad with fibre-reinforcedconcrete (GRC). Concrete also forms a large part of the finishings, such as surfaces in view,floorings and furnishings.

258 CONSTRUCTION - Buildings

ProjectMAXXI – National Museum of Arts of the XXI Century

LocationRome, Italy

CustomerMinistry for Cultural Assets and Activities – Departmentfor architecture and contemporary art

CreationMinistry of Infrastructure and Transport – Public WorksSuperintendence of Regione Lazio: Angelo Balducci -R.U.P./Proceedings’ person in charge: Roberto Linetti –Construction yard management: Roberto Tartaro -Architectural aspects’ operating director: Mario Avagnina

Architectural designDesigners: Zaha Hadid – Patrik SchumacherProject leader: Gianluca Racana (Zaha Hadid Limited)

Structural consultantsAnthony Hunt Ass., OK Design Group

Final construction designStructural design: Studio S.P.C. S.r.l. Giorgio Croci –Aymen HerzallaGeotechnical consulting: V. M. SantoroSpecialist consulting: A. Viskovic, S. Di Cintio, M. FranciniAssistants for the structural design: F. Croci, S. Di Carlo,I. De Rossi, A. Bozzetti, C. Russo.Steel staircase and “monocoque” floor design: StudioE.D. In. s.r.l. - Fabio Brancaleoni, Marcello ColasantiSCC mix design consultant: Mario CollepardiStructural design validator: Antonio Maffey

Construction Syndicate: MAXXI 2006Group leader: ITALIANA COSTRUZIONI S.p.A.(Group Navarra)Assignor: S.A.C. Società Appalti Costruzioni S.p.A.(Group Cerasi)Prime Contractor: Marco OdoardiConstruction site engineering manager: Roberto RossiConstruction yard chief: Gianni ScennaAssistant construction yard chief: Luigi CarducciEngineering office: Daniele Centurioni, Silvia LaPergola, Fabio CeciDesign coordination, execution of systems: ClaudioPassiniAccounting Office: Roberto Cascino, Enrico Bottacchiari

Photographs Iwan Baan

MAXXI – Center for the contemporaryarts

Winner A.I.C.A.P. Award 2009 for Structural Concrete Works - Category “Buildings”

259CONSTRUCTION - Buildings


• 1- Ground-floor plan. 2- Section through building.

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The affairs of the Palavela (“sail hall”) go back to 1958. In that year, as part of the Italia 61exposition (organized in the Piedmontese capital to celebrate the centennial of Italy’s unity), acompetition was promoted for the design of a building to house the Fashion and Costumesevent. The winning solution was a smooth reinforced-concrete box vault, bearing on three pointsand built using procedures like those used in the construction of the Centre des nouvellesindustries et technologies, built between 1956 and 1958 in the Défense quarter. The roofhas a hexagonal plan inscribed in a circle 150 m in diameter, and consists of a self-bearingreinforced – and prestressed-concrete shell. The height at the crown of the arches is 29.00 m,while the composite vault roofs an area of 14,625 square metres; the volume enclosed is332,000 cubic metres. The vault structure consists of two slabs, each 60 mm deep, developingover the entire roof and connected together by continuous longitudinal and crosswise ribs; its1.30 m gross depth includes a transitable interspace 1.18 m high. The continuous monitoringsthe structure was subjected to over the years gave results so encouraging as to include the buil-ding in the list of works potentially useable for the competition outfittings of the 2006 WinterOlympics. In fact, its restructuring aimed at the creation of plant for artistic ice-skating and forshort-track. The new building, wrapped by its original roof, comprises two bodies set closetogether, with a reticular steel roof that, although at different elevations, connects them together.The southeast-southwest body is assigned to spectators of sectors 1 and 2 (7196 seats), andthe northeast-northwest body is assigned to the “Olympic family”, to the athletes and to themedia (1062 seats) for a total of 8258 seats. Structurally, the two building bodies are suppor-ted by parallel septums, on which the floor structures are made to bear as are the precast sea-ting tiers, and by the perimetral walls. The bearing septums were built of reinforced concretewith a fair-face finish.


ProjectThe renewal of “Palavela”

LocationTurin, Italy

ClientAgenzia Torino 2006

Final designing. Arnaldo De Bernardi, arch. Gae Aulenti, ing. G. Siniscalco, arch. C. Roluti, arch. S. Basso, arch. M. Filippi, ing. G.C. Gramoni, arch. F. Quadri, ing. W. Peisino, ing. G. Forte, ing. E. Rosati

Construction designing. Valerio Actis Grosso (Project Manager), ing.Giovanni Vallino Costassa (Structural engineer)

Yard management prof. ing. Giorgio Siniscalco (SI.ME.TE Snc)

General contractorA.T.I. (Maire Engineering SpA, Impresa CostruzioniRosso Geom. Francesco & Figli SpA, KeltermicaCordero)

Self Compacting Concrete (SCC)Unical SpA - Gruppo BuzziUnicem

FormworksDoka Italia SpA

“Olympic Palavela”


• 1-2- Images from the files (1958-1961) of phases in the construction of the Pa-lavela (“Sail palace”). 3- The sail roof was given a preliminary restoration: one of noparticular significance since careful monitoring over the years had brought out thatthe building’s structural integrity and its surfaces had remained very nearly perfect.

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• 4- The elements of the new building’s structure were created with fair-face con-crete, the structure having to perform an architectural and aesthetic task as well.Thus design opted for self-compacting mixes, capable of assuring both high me-chanical performance and aesthetic beauty, besides meeting the specific pour re-quisites tied to so particular a construction job. 5- Design envisaged for the Ice Pa-lace the construction of a galvanized steel reticular roof, given the necessary sound-absorbent and sound-insulation panelling. 6- The sports facility’s plan: the groundprojection of the hexagonal sail roof completely encloses the Ice Palace structure.7- Phases in the construction, by in situ pour, of the tank that holds the ice-skatingrink. 8- Section through the building. 9- Detail of the facade of the new Ice Palaceduring the final finish phases: the mix design and the special care taken in carryingout the pours made it possible to create quality concrete surfaces, which at the sa-me time became the sign characterizing the entire oeuvre.

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The new building housing Milan’s Luigi Bocconi Business University, officially opened on October31st last, completes the historic campus – conceived by Giuseppe Pagano a little more thanseventy years ago and enlarged to a design by Giovanni Muzio and Ignazio Gardella – andbecomes the whole university complex’s new “entrance gate”. The construction stands at theintersection of Viale Bligny and Via Roentgen on a 60 m x 150 m lot. Its functional programenvisages that on the 68,000 square metres of walking surface a quarter of the university’sneeds be housed, among which the offices, the departments, the classrooms, the expositionzone and the great hall seating a thousand. Characterizing the oeuvre too is the complex structural design worked out: the sophisticatedfoundations, which reach down more than 14 m below site level, the traditional continuous con-crete raft 2-3 metres deep, the prestressed-concrete floor structures on the basement levels,and the enormous wall beams, with thicknesses of 400 mm, heights up to 30 m and 24 mspacings. To build the enlargement of Milan’s Bocconi, concrete was used to pour almost thewhole of the bearing structures, which remain in view on the interiors and in some parts onthe outside too. In particular, for the wall beams, the bearing septums, the roof beams and thegreat hall’s roof structure, self-compacting concrete (SCC) was used, a material of long-lastingworkability that ensures greater compactness of the pours and therefore better quality and amore uniform fair face in terms of aspect and colour, as well as an improvement in the oeu-vre’s mechanical strength and durability. Many reasons led to this construction and technological choice: first of all, the design’s archi-tectural and structural complexity, which meant the need for thickly-laced reinforcing units thatwould not have permitted the introduction of vibration equipment, the heavy stresses in thehardest-working areas, the high ambient temperatures and the jobsite’s location right in down-town Milan.


Project“Bocconi 2000” of Bocconi University (enlargement)

LocationMilan, Italy

ClientUniversità commerciale Luigi Bocconi, Milano

Client representativeGeom. Nicolò Di Blasi

ArchitectGrafton Studio, Dublino - arch. Shelley McNamara,arch. Yvonne Farrel

Design team Gerard Carty, Philippe O'Sullivan, Emmett Scanlon

Project ArchitectSimona Castelli

Co-workersLennart Breternitz, Matthew Beattie, PhilipComerford, Miriam Dunn, Andreas Degn, Ann Henry,David Leech, John Barry Lowe, Eavan Meagher, OrlaMurphy, Aoibheann Ni Mhearainn, Kieran O'Brien,Sterrin O’Shea, Eoghan O’Shea, Michael Pike, AnnaRyan, Maurizio Scalera, Ansgar Staudt, GavinWheatley

Structural design and supervision of constructionStudio Pereira, Milano - ing. Emilio Pereira, ing Vincenzo Collina, ing. Massimo Sandrelli, ing. SilvioValloni

On-site building supervisionProgetto CMR - Marco Ferrario, Danila Aimone,Maurizio Cantoni, Claudio Pin

Utility systems designAmman Progetti

Lighting consultantMetis - Claudio Valent, Marinella Patetta

Interior designAvenue Architects - arch. Dante Bonuccelli

Acoustic and electrical consultantARP Service, Paolo Molina

Fire-fighting consultanting. Silvestre Mistretta

General Contractor GDM Costruzioni S.p.a.

PhotographsUNICAL, Studio Pereira, A. Faresin, Redazione iiC

New Bocconi University


• 1- Floor plan. 2- Construction of the basement-levels’ r.c. bearing septums.

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• 3- First-basement-floor plan. 4-5 Box elements composing the above-ground bea-ring structure. 6- Cross section through building. It shows the common undergroundbase and the breakdown of the standing portion into distinct bodies. 7- Installinga floor structure’s slack reinforcings and its prestressing.

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The job is the first part of the Milanofiori 2000 development project. Located nearby the Milan-Genoa motorway, it is also served by the subway’s line 2.It consists of a large plate housing two floors above ground of parking and a raised plaza at theelevation of the subway’s loading platform, i.e. at 6 m above site level. Three nine-storey office buildings stand up from this foot plaza; on it space is found for businessactivities, dining and refreshment, and sundry services.The built area is 70,000 square meters, plus another 19,500 for site-level parking.The structures feature an inclined lay of the plans of the standing buildings, which must be wed withthe orthogonal grid of the underlying parking structures. They are built of solid concrete plates.The buildings’ typical grid (they feature irregular projections on their sides and large free-plan ope-nings too) is 9 m by 6 m, arranged parallelogram-wise with a 73° minor angle.The parking grid is square, 8.1 m on a side.The columns, their dimensions kept limited, are typically in reinforced concrete, with a steel-con-crete composite section for the buildings’ lower floors.Right from the start of the study of possible structural alternatives to serve the buildings’ archi-tecture with economy and speed of execution of the design, the solid-r.c.-slabs solution, cast overindustrialized formwork, was seen to be most advantageous. Analysis of the geotechnical data and the value of the acting loads yielded the selection of an indi-rect foundations on r.c. piles drilled using the continuous helix technology.Three types of piles are distinguished:a. 80 cm diameter piles, 21.5m long, with Pnom = 2500kN;b. 60 cm diameter piles 21.5m long, with Pnom = 1400kN;c. 60cm diameter piles 11.5m long, with, Pnom = 875kN.The parking area columns are of r.c., of typical circular section 50 cm in diameter. Columns ofdifferent dimensions and types are envisaged in the standing buildingsThe fire resistance requirement is R90.The columns in the standing buildings are typically circular in form 50 to 70 cm in diameter, witha composite steel-concrete type, which includes a steel H-section. The outer concrete of the columns is not only requested by fire protection, but is considered asworking with the steel column to provide the necessary bearing capacity.The deck of the first floor above ground (second parking level) and the deck at plaza elevationconsist of solid r.c. plates, one 28 cm deep having REI90 fire characteristics and the other 400mm deep with REI180 fire characteristics. Their construction is envisaged by in situ casting overindustrialized modular forms. The plate fields’ plan dimensions vary depending on architecturallayout. The typical plate dimensions are 8.10 m x 8.10 m.A dual grid of two-way reinforcing bars is necessary, with densely reinforced zones on the extra-dos at the columns and on the soffit in span at points where the fields are most highly stressed.Prefabricated bundles of bars were used and unrolled on site in order to achieve the correct spa-cings, speed up the construction time and reduce the placing price. A specific shear reinforcingagainst punching was included at the columns. The offices floor deck comprises solid r.c. plates300 mm deep with REI90 fire characteristics. Its construction is envisaged by in situ casting overindustrialized modular forms. The plan dimensions of the plate fields vary depending on archi-tectural layout and measure 9.00 m by 8.50 m or 6.00 m by 8.50 m.The need to create terraces for the top offices floor and to maintain the facade module constantas per the inter storey type meant that the mechanical deck and the roof had to be suspendedat the terraces. In fact, since the façade columns had to be set back without increasing the deckdepth for the top offices floor, a part of the mechanical deck and its roof had to be hung fromtendons.


ProjectMilanofiori 2000 – Corporate Center (Segments Band C1)


ClientBrioschi Sviluppo Immobiliare spa

DesignErick van Egeraat, Rotterdam

Structural engineer and supervision of construction Redesco srl, ing. Mauro E. Giuliani, ing. Gianluca Vesa,Milan

General contractorUnionbau srl

Coordination of final design, general supervision of constructionIntertecno spa, Milan

Milanofiori 2000 – Corporate Center


• 1- Cross section. 2- Longitudinal section.

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• 3-4- Construction of the first levels: the pour of the deck plates. 5- Use of pre-fabricated bundles of rebars. 6- East façade. 7- North façade.

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The new industrial installation called for the construction of various buildings on an area of400,000 square meters, for a roofed area of 175,000 square meters. It stands within a gra-vel quarry and runs parallel to the Vicenza-Treviso railway line for a kilometer.The most extensive building complex forms a heterogeneous body of plan 700 m x 120 m,its height varying between 12 m for buildings A1 and A2 and 30 m for the remaining area.Buildings A1 and A2 were built with reinforced- and prestressed-concrete precastings, whilebuildings A3, A3/1 and A5 have steel bearing structures. The production buildings, A4 andA4/1, were built of post-tensioned in situ-cast concrete. Building A4 (production division) has a rectangular plan of 121 m x 164 m. On the railwayside and on the quarry side, adjacent to the building, stand four 26-meter-tall towers, in whichare created the stairwells. The building has two floors, each of net height 8.5 m, and a totalheight of 30 meters. The structural grid is 15 m x 15 m with rectangular-section columns ofsection sufficient to ensure, in at least one direction, a net distance between columns of notless than 14 m. To build the floor structures various structural solutions were evaluated. That best able to meetspecifications and most advantageous appeared to be the in situ-cast, with a grating-type floorstructure having post-tensioned ribs.Construction needs, together with the need to create suitable expansion joints, identified the“ideal” module as a 30 m x 30 m floor structure constituted of four meshes 15 m on a sideeach. Thus was identified the dual typology of a “bearing” floor structure, statically indetermi-nable, on nine columns, and of a “borne” floor structure, on just three columns in line andborne on saddles envisaged at the sides of the adjacent floor structure. Each module is a two-way ribbed floor structure, built of in situ-cast concrete with partial prestressing. The post-ten-sioning was effected with both bonded cables and unbounded cables.The definitive solution calls for the construction of: 12 “bearing” modules of 1100 squaremeters, 12 “borne” modules of 820 square meters, and 12 “hybrid” modules (bearing on oneside and borne on the other) of a thousand square meters each. The floor structure is lighte-ned with re-useable aluminum forms so arranged as to create a two-dimensional ribbed struc-ture.The ribs are spaced 1.50 m apart and are 28 cm wide. The upper cap is 120 mm deep andis surface-treated with quartz powder. A further surface treatment having a modified sodiumsilicate base forms a protective barrier against aggressive agents and limits shrinkage effects.The arrangement of the prestressing cables is such as to create “principal bands” 6 m wide,aligned to the columns and prestressed with grouted-sheath cables of six and eight 0.6”strands. On the remaining ribs are present two unbonded single-strand cables. Further grou-ted cables 6 meters in length are called for in the areas of greatest negative moment andshear.


ProjectBuilding for industrial plant

LocationPaese (Treviso), Italy

ClientAcqua Minerale San Benedetto S.p.A., Scorzè (Venice)

DesignGiuseppe Zago CE

Structural engineeringStudio di ingegneria RS – Stefano Secchi CE, Padua

General contractorSetten Genesio S.p.A.

“Acqua minerale San Benedetto” plant


• 1- The industrial complex’s development in plan and in elevation.

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• 2- Aluminium lightener forms. 3- Pour phases for a bearing module. 4- Installa-tion of the curtain walls. 5- Bearing-module plan. 6- Ribbed floor-structure soffit. 7-Support system with saddles. 8- Support system with shear-connectors. 9- Supportsystem.

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Building 1 is tied to building 2 through a connector body, building 2 in its turn being tied tobuilding 3 through a suspended bridge. The multifloor bridge (30.15 m in span, 16.80 mwide) over the city street is held up by structural-steel work, consisting of two trestlework gir-ders placed on the facade to form a resisting structure as high as the entire (three-story) faca-de. Regarding the structural system’s capacity to stand up to horizontal forces, this is delega-ted to the r.c. structures of the first- and top-level floor structures, which in their plane act asinfinitely stiff beams (plates). The facade glazings transfer, at these levels, the wind loading tothe diagonals of the main trestlework beam, which have elliptical sections (composed of twosemi-elliptical pipes) with greater inertia crosswise to the loads. Except for the stair blocks,the structure is built of precastings. The building floor structures are precast of adherent-strand prestressed concrete, the columns are plant precast (Rck=50 MPa) in forms preparedespecially for the SKY project, with different r.c. sections for transport, assembly and removalfrom the forms in the plant. Their weight was kept down to forty tons. The column was thencast in two pieces, solidly joined together during assembly phase. The central column, its sec-tion varying with height from the base, 0.90 m x 0.90 m, is a single piece twenty meters high.It was jointed with an element whose section varied with height, the element being 20-25 mhigh. The structures were mounted using precise sequences, which permitted construction ofthe structures in shorter times than is usual for traditional r.c. structures. The constructionmodule is based on a typical grid of 8.40 m x 8.40 m, which is adapted to the various situa-tions, the spans reaching 18.00 m (building 1) and 16.80 m (building 2).Building 1 – technological: rectangular-plan, of dimensions 180.5 m x 28.2 m (36,000 m2

of floor structure). Composed of: basement floor structure, ground floor and seven floor struc-tures above ground. Total height: 39.20 m. Building 2 – offices: trapezoidal-plan, with dimensions 103 m x 26 m (22,000 m2 of floorstructure. Composed of: basement floor structure, ground floor and nine floor structures aboveground. Total height: 47.30 m. Building 1B: connector building between buildings 1 and 2(4400 m2 of floor structure). Composed of: basement floor structure, ground floor and sixfloor structures above ground. Total height: 34.45 m. Building 3: trapezoidal-plan, dimensions12 m x 25 m (25,700 m2 of floor structure). Composed of: basement floor structure, groundfloor and eight floor structures above ground. Total height: 44.75 m. The structures (stairwellsand elevator shafts) acting as shearbracing for the buildings under horizontal forces were builtin situ (Rck=37 MPa) and tied in a second phase to the precast structure by various pour-restart and continuity systems. The soils demanded a raft foundation on single-fluid or two-fluid jet-grouting columns, whether interpenetrating or tangent to one another, having a maxi-mum diameter of 1.90 m for the greater vertical loadings and maximum height of 11.15 mbeneath the stair wells. Owing to the high level of the water table, the raft was waterproofedusing the so-called “white tank” waterproofing system. Building 1 required specific structuralanalyses as well as special design choices since it had to possess considerable stiffness (maxi-mum allowable antenna rotation: 0.01° or 36”) under horizontal forces (reference windspeed: 110 km/h), in order to ensure alignment of the trasmitting signals to the satellitesthrough the roof antennas. The displacement and rotation fields were derived by analyzingtwo different structural models. The first is of a generalized type and was used to evaluatethe field of floor structure horizontal displacements at the various floors. From this the hori-zontal rotations of the decks (rotations with axes normal to the building floor structures) couldbe deduced. The second concerns the building’s top floor structure, where the trasmittingapparatus is installed. From this latter model the rotations at the antenna bases (rotationswith axes in the plane of the floor structures) were derived due to the wind pressures on thetras mitting antennas.


ProjectNew Sky Italia headquarters – Top Management andTelevision Production

LocationMilan Rogoredo, Italy

ClientMilano Santa Giulia S.p.A., with the technical direction of Silvio Bernabè CE

DesignByron Harford & Associates – East Sydney

Structural engineerIn situ-built structures design and supervision of construction: MSC Associati S.r.l. (Milan) – DaniloCampagna CE, Andrea Sangalli CE; Precast/prefabricated structures design: GammaEngineering (Lecco) – Gianluigi Fregosi CE, Riccardo Castagna CE

ArchitectsByron Harford & Associates – East Sydney

Management ContractorColombo Costruzioni S.p.A – Lecco

General ContractorColombo Costruzioni S.p.A – Lecco


New Sky Italia headquarters


• 1- Section through buildings. 2- Plan of the complex.

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• 3- Construction of the stairwell core structures. 4-5 Phases in construction of thebuildings. 6- Section through building 2. 7- Foundations plan. 8- Typical-floor plan.

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The Light pavilion stands under Lake Como’s evocative light, in a ten-thousand-square-meterpark between the city’s historic downtown and Cernobbio. The project’s origin was quite sim-ple: the program in fact needed a space as open to the outside as possible, without the natu-ral light’s disturbing the staging of the events. To meet this need the construction’s volume was intersected by several plane surfaces, arran-ged arbitrarily in space in such fashion as to create a series of fragments, suggesting a pro-cess that could continue on to infinity. The intersections thus determined have become the linear voids through which a limited andquite definite dialogue develops, between rule and free will, between structure and light. The hall volume unveils a continual weave of cuts crossing it vertically and horizontally, takingon the landscape’s blue and green tints. The idea behind these cuts came out of the sugge-stiveness evoked by a former Fascist-headquarters building standing not far off, where a cry-stal slash crosses the upper part of the atrium roof, putting the mountains behind the city inrelation with the cathedral nave. Analogously, in the pavilion the irregular network of cuts takes as its aim to reveal the powerand the all-pervasiveness of the corrosive action of the lines that have over hundreds of mil-lennia modelled the present-day forms of the Lombard lakes landscape.


Light Pavilion

ProjectExhibition pavilion

LocationGrand Hotel, Como, Italy

ClientMeta S.p.A. (Paolo De Santis)

Design and yard management Attilio Terragni

Design Team Chiara Assanelli, Luca Mangione, Maja Leonelli

Structural design Amis Milano – Antonio Migliacci CE, GiovanniFranchi CE

Mechanical and electrical equipmentAmman Progetti Milano

Acoustics Paolo Molina CE

Sun-protections Abba Srl, Treviso (Person in charge: LucaFranceschin)

Safety engineering CDR, Carlo Ruckstul CE

General contractorMondelli Battista. Construction yard supervision:Aldo Mondelli CE

Facade PERMASTELISA Construction yard supervision:Alfredo Piccoli


• 1- Pavilion plan.

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• 2- Sections through pavilion. 3-4 Phases in building the structure.

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The Italian Space Agency’s new headquarters stands in the Tor Vergata university district, Eastof Rome’s great ring road, on a sloping lot adjacent to the engineering faculty.The complex comprises ten buildings: the semicircular main building (A), where most of theoffices are concentrated; the rectangular auditorium and atrium buildings (B’ and B”), whicheccentrically intersect the previous one, on the centerline of the faculty of engineering – theyrepresent the monumental aspect and act as connection with the outside; a series of minorbuildings, for offices or services, such as the library (C), the nursery-school and offices building(D), the bank and fitness-center building (E), the doorkeeper’s office (F), the cafeteria-kitchen(G), the laboratories-offices (I), and the underground parking garages (L).Structural design had as reference OPCM 3274, and thus took account of the city of Rome’slisting in seismic zone 3. This is one of the first applications of the new seismic design crite-ria for a complex of considerable size. The design was sifted by the Higher Public WorksCouncil, as a building having a value exceeding 25 million euros.The main building A and buildings C-D are buildings having an r.c. core and a mixed steel-concrete structure, with hanging steel columns and floor structures made up of predalles anda 9 cm deep slab. The other buildings are substantially wholly of r.c., in some cases with floorstructures made of prestressed elements (honeycomb floor structures or omega tiles).Retaining walls of considerable height are present.Vertical seismic joints separate building A into three bodies. In each body there are two r.c.cores, while the remaining structure is mixed steel-concrete. The first two bodies have eachsix floors, one or two of which are below ground depending on the lay of the land. The thirdbody stands five floors above ground.Buildings B” (lobby) and B’ (auditorium) stand along a line that crosses main building A, withwhich they share the joint. They are almost totally built of r.c., with walls of great extent inplan. Both are lower than building A. The zone of intersection, the focus of the complex, where four structurally - independent build -ings converge, is as high as building A and is without floor structures from the second floor tothe roof. In correspondence with the facade of the concave side, on the higher floors, theoverhead passageways run between the two bodies composing the main building, bearing onmore-or-less radial septums. The overhead ways, the first two floor structures and the conca-ve facade have vertical joints in the middle of the atrium.To roof this space an innovatory solution was found. It consists of a steel grid with concreteslab resting through seismic isolators on the two r.c. cores of the two independent bodies ofthe main building which delimit the atrium.The upper part of the convex facade hangs from the isolated roof. At mid height a horizon-tal joint separate the facade from its lower part which sticks out from steel structure of theroof of building B. Vertical joints separate the facade from the cores of building A on whichthe roof of the atrium stands.The latter solution, involving a structure bearing on seismic isolators, was particularly apt. Infact, the stresses induced by independent seismic motion of the cores are moderate, the struc-ture is simple, the support scheme is particularly suited to the mixed steel-concrete structu-re typology and its construction proceeded rapidly.The connection of the mixed-steel concrete floors to the two bracing cores in each of thebodies which compose building A was effected with special metal inserts to which the slab’srestart reinforcing were welded. These special pieces were connected to anchorage platesburied in the cores by welding in place.The vertical reinforcement of the structural walls stick up from the foundations for one floorand a half without interruption, so as to exclude overlapping in the critical zones of the brac -ing walls where yielding may happen.


ProjectHeadquarters

LocationTor Vergata, Rome, Italy

ClientAgenzia Spaziale Italiana

Structural engineerProf. Camillo Nuti CEwith STIN Section chief: Danilo Pierucci CE, Rome

Architects5+1AA Alfonso Femia, Gianluca Peluffowith Annalaura Spalla arch.

Management ContractorInfrastructures MinistryInterregional Office for Public Works for Latium,Abruzzo and Sardinia Superintendent: Giovanni Guglielmi CESupervision of construction: Mario Avagnina arch.

General contractorSAC Società Appalti Costruzioni S.p.A. – RomeProject manager: Bruno Cavallaro CE


Agenzia Spaziale Italiana new headquarters


• 1- General plan of the complex. 2- Main building (A) and Auditorium (B’). (pho-to: Giuseppe Maritati). 3- Main building (A) West facade, concave side.

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Building A – “Crescent” officesBuilding B –Atrium-auditorium systemBuilding C – LibraryBuilding D –Nursery, officesBuilding E –Bank, infirmary, gymnasiumBuilding F –Doorkeeper’s officeBuilding G –Cafeteria, kitchenBuilding H – BarBuilding I – Former workshop officesBuilding L –Underground parking


• 4- Foundations plan. 5- Main building (A): restart steels. 6- Building atrium (B”). 7-Main building (A): body 3, mounting the mixed structure after construction of the r.c.core. 8- Main building (A) and Auditorium (B’). 9- Main building (A) – Concave side.10- Cafeteria building structure (G). 11- Cafeteria building (G).

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The ‘Altra Sede’ Tower, located in the center of the city of Milan, has recently been completed. On January22, 2010 an official ceremony marked the end of the construction phase and unveiled the new buildingto the citizens of Milan, who were also called to cast their vote and decide the new name of the com-plex: Palazzo Lombardia. The building is currently the tallest in Italy, and one of the strongest features ofthe skyline of the city for years to come.The Tower is the new administrative centre for Lombardia’s regional government in Milan; the adminis-trative complex also includes five lower buildings (about 40 m high, called Cores 2, 3, 4, 5 and 6), sur-rounding the high-rise Tower (Core 1), which, 161.30 metres tall, set an Italian record. The building’s sinu-ous interweaving strands recall mountains, valleys, and rivers of the Lombardia region. Their curvilinearforms are adaptable to changing functional requirements and are receptive to the region’s evolving orga-nizational structure. In addition to its headquarter functions, the building accommodates public ameni-ties accessible to all. The winning architectural project for the new regional headquarters was conceived,according to the guidelines set forward by the Adminisration, by the architecture bureaus Pei Cobb Freed& Partners from New York, together with Caputo partnership and Sistema Duemila from Italy in 2003.Given the very strict construction times (October 2006-December 2009), wise and innovative choiceswere enforced as for the design of structures, entirely consisting of reinforced concrete elements. The foun-dation system is a 4m thick reinforced concrete slab resting on soil whose load bearing capacity was pre-liminarily improved by means of the jet grouting technique. The total volume of concrete is about 8.000m3. For the lower layer, 1m thick, including most of the steel rebar, SC 30/37 self compacting concretewas used, whereas for the upper layer, 3m thick, a high performance C 30/37 concrete was employed.The vertical structures of the Tower consist of one 15,5x16,3m inner core hosting stairways and lifts,whose maximum thickness of the walls is 45cm, and 22 circular columns with diameters ranging from120cm at the bottom to 65cm at the top floors. The columns are located along two curved lines anddefine a structural grid of 8,60mx6,50m. For the cast in situ slabs, 35cm thick, C 40/50 concrete wasused. Regarding the construction techniques, both for the vertical load bearing structure (cores andcolumns) and for the horizontal ones (slabs), it was decided to make extensive use of industrialized sys-tems, together with high profile state-of-the-art construction technologies, such as the self-climbing form-work employed to build the core of the Tower, for which high strength concrete (class C 45/55) was used.An hybrid steel encasing/reinforced concrete system was employed to build the columns, allowing thecolumns to be cast at the same time for an height of up to three floors. Slabs consisted in pre-assem-bled panels including the main reinforcement and polyethylene spheres with a diameter of 27cm. In thisway, the total weight of the slab was reduced by about 25% with respect to a solid section and only addi-tional reinforcement had to be placed, thus considerably reducing the construction time. These techniques,coupled to the self-climbing formwork, allowed the construction of each storey of the tower in an aver-age of 5.5 days, as opposed to the 8 days needed with the traditional casting system. The synergic useof these technologies allowed the 38 floors of the Tower, of which those from 1 to 11 have a surface ofabout 2000 m2 and those from 12 to 38 have a surface of 900m2, to be built in about eleven months.This very remarkable reduction in construction times of the structural system of the Tower was reflectedin the global construction ending 60 days ahead of schedule.The roofing of the inner plaza, with a sur-face of about 4200 m2, consists of a steel truss structure made of welded and bolted S355 J2 steel tubeson top of which a double-layer Texlon ETFE (ethylene-co-tetrafluoroethylene) film is installed. The ETFEcushions guarantee a filtering of sunlight of about 50%. EFTE was chosen since it is a transparent plas-tic film, lighter and more resistant than glass, with superior insulating power and easier and less expen-sive to install. The ETFE film is also stable to UV rays and is not altered by environmental pollution andweather conditions. Sustainability, environmental impact, durability and reduced management and main-tenance costs were also consistently pursued, by recurring to the most effective solutions for plants, pho-tovoltaic energy production, air conditioning, façades and ventilation, such as the double-layer façade mak-ing up a ‘thermal wall’ for the internal rooms, with an automated system of sun shades conceived to min-imize heat dispersion and maximize the efficiency of the air conditioning system.


ProjectNew administrative center for Lombardia’s regionalgovernment


Architectural designPei Cobb Freed & Partners Architects, New York,Caputo Partnership & Sistema Duemila, Milan

Architectural Design SupervisionArch. Henry N. Cobb

Structural DesignProf. Franco Mola CE – ECSD S.r.l., Milan

Construction SupervisorInfrastrutture Lombarde S.p.a. – General Manager: Antonio Giulio Rognoni CE

General ContractorConsorzio Torre (Impregilo S.p.A., TechintInfrastrutture, cmb, Cile S.p.A., Montagna CostruzioniS.r.l., Pessina Costruzioni S.p.A., ConsorzioCooperative Costruzioni, Sirti S.p.A.)

Leading ContractorImpregilo S.p.A.

General ManagerGaetano Salonia CE

Site Technical ManagerVinicio Scerri CE

Construction Site General Manager and SafetySupervisorGuglielmo Fariello CE

FormworkDoka italia S.p.A.

Precast vertical elements and beamsCSP Prefabbricati S.p.A.

SlabsCobiax Technologies S.r.l.


“Altra Sede” for the Regione Lombardia


• 1- View of the Tower through the roofing. 2- Construction: facades. 3- Con-struction of Core 1. 4- Cross section of the slab.

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• 5- Ground floor plan. 6- Columns and slab systems. 7- Tower contruction. 8-Slab system. 9- Core 1: reinforcement details.

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The new Sant’Anna Hospital in Como, located in the heart of the Lombardia Region, in north-ern Italy, is a large hospital complex for which a short time for construction, fixed in 900 days,was allowed by the Administration.The complex is made of eight interconnected buildings, with heights up to 22 m, including,respectively, a number of floors ranging between 2 and 3 floors above the ground. The under-ground floors range between one and three, according to the height of each building above theground and the ground level itself varying across the plan. The whole complex is surrounded byreinforced concrete walls with varying height. Two main aspects governed the design choices: the earthquake hazard for the area where theHospital is located and the deep foundations, which caused the foundations themselves and thelower underground floors to be entirely below the water level.As for the seismicity, the Lombardia Region is classified as ‘Zone N.4’ in the Italian Seismic Code,which corresponds to a PGA of about 0.05g. Since the hospital buildings are classified as ‘strate-gic’, an importance coefficient of 1.4 on the PGA is also compulsory, meaning that the designPGA value became 0.07g. Due to their limited height, the buildings are all quite rigid, so thatearthquake effects are strongly prevailing on those of wind loads: earthquake is thus the mainload condition for pre-dimensioning of the structural elements for lateral resistance (global baseshear).The chosen structural elements were reinforced concrete cores, adequately distributed in plan,so as to reduce torque effects due to lateral loads; as for static loading, reinforced concretecolumns were dimensioned to the vertical loads in addition to the flexural effects derived fromthe lateral displacements produced by the cores and also taking into account the interactionwith slab elements under vertical loading.The foundation system includes reinforced concrete basements for the columns, supported byconcrete piles with driven steel formwork, having a maximum length of 26m. The cores are supported by concrete plates resting on reinforced concrete diaphragms. The planconfiguration of the buildings, with a complex distribution of columns, not lined into orthogonalpatterns, gave way to varying spans for the slab structures. The lack of regularity in the distrib-ution of vertical elements is not very strong, though, so, except for some peculiar areas, the meanspans for the slab were 7.50mx7.50m. The ratio between the two spans calls for elements witha two-dimensional behavior. The irregularities in the grids called for the use of construction systems that are totally or most-ly independent on the morphology of the slabs themselves. The most competitive technique atthis regard is the use of in situ casting: continuous prestressed concrete cast-in-situ slabs restingon the columns turn out to be the best choice. An adequate number of prestressed unbondedsteel strands was introduced in the slab, which allowed a reduction of the slab thickness to26cm. The quantity of prestressing strands and additional rebar is globally reduced with respectto ordinary rebar, so that the time to arrange the steel before casting is significantly shortened.Even if the arrangement of prestressing steel must be accurately carried out and supervised,and special construction methods to guarantee the effectiveness of prestressing must beenforced – in particular provisional gaps, sliding supports along the cores and punching shearadditional reinforcement – construction time can still be reduced, because the formworks canbe removed as soon as after only 36 hours from casting, when concrete, whose mix-design wasaccurately studied, reaches a strength of about 30MPa, implying an elastic modulus of about3x104 MPa. In this way, a good productivity level was guaranteed, coupled to high structural per-formance and strong standardization in the construction process.


ProjectNew Sant’Anna Hospital

LocationComo, Italy

ClientInfrastrutture Lombarde and Azienda OspedalieraSant’Anna

Architectural and plants designBortolazzi Consulting

Structural designProf. Franco Mola CE – ECSD S.r.l., Milan

On-site SupervisorFrancesco de Probizer CE

Construction SupervisorInfrastrutture Lombarde S.p.A. - General Manager:Antonio Giulio Rognoni CE

General ContractorS.A.N.C.O. SCarl (Altair, GDM Costruzioni, Aster)


New “Sant’Anna” Hospital


• 1- Overview of the completed Hospital. 2- Plan view of the complex (architect -ural drawing).

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• 3- Column-slab joint. 4- Floor plan. 5- Vertical section. 6- Facades assembling.

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EAST PARKINGHOSPITAL

NORTH PARKING


The theatre stands on a long narrow lot bounded by the orthogonal street grid, and is in aplace of transition between the historic downtown and an urban outskirts almost a largecity’s. The position of the architectural complex has not compromised the language of conti-nuity of the sur-rounding buildings, indeed, it has relaunched their vitality through the greatappeal it enjoys within the city. The theatre main entrance opens on two broad streets: the viaBattisti and the via Martel-li, which meet at an angle, right at the building entrance and onthe entrance to the plaza. Volume-wise, three elements identify the theatre’s principal use assi-gnments: the halls and the foyer, the stage tower, and the dressing rooms. The project calledfor three halls: a principal hall seating 998 between the orchestra and three orders of bal-conies; a smaller one containing 160 seats, and a multi-use hall seating one hundred at thethird level, useable both for small entertainments and for rehearsals, since its dimensions arethose of the stage. The choice of theatre-hall conformation fell on the traditional horseshoe,to guarantee good visibility and acoustics. The public accesses it through the building-highfoyer: two grand stairways serve the three balconies and a route along the perimeter of thefoyer joins the three balconies with three rings. The dressing rooms are divided into four sin-gles, three doubles, three quadruples and six group dressing rooms. In the basement there isonly the wardrobe. The first floor of the backstage zone is taken up by some of the dressingrooms and by the dressmaker’s. The second floor holds more dressing rooms, while the thirdfloor, besides other dressing rooms, also holds offices, including the press office, and in the cen-tral part the choir rehearsals hall, useable too as a conference room. The fourth floor holdsthe final dressing rooms, the rest of the offices, a workshop, the carpenter’s shop (directly con-nected to the stage tower), and the rehearsal room.


ProjectNew “Verdi” Theatre

LocationPordenone, Italy

ClientMunicipality of Pordenone

DesignIng. Carlo Filipuzzi, arch. Paola Moretti – Interstudios.r.l., Udine

Structural engineer consultant Ing. Carlo Filipuzzi

Yard managementArch. Ermanno Dell’Agnolo

General contractorMazzi Impresa generale SpA, Verona

PhotographsInterstudio s.r.l.

“Verdi” Theatre


• 1- Plan of the project, right downtown. 2- The reinforced-concrete sculpture pla-ced in the foyer. 3- The foyer, characterized by the sculpture and the fair-face bea-ring structures of the roof and of the galleries accessing the three higher balconies.

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• 4- Building plan at the level of the main hall orchestra seating. 5- The building’stop floor (level of the boxes, 12.90 m). 6- Longitudinal section through building. 7-Facade, on the via Martelli.

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The construction of the Lombarda bank’s new headquarters in Brescia, in the area of urbanexpansion, is the result of a design search for an alternative to the tower typology, often usedfor the design of buildings having a high services concentration. The complex is conceived asa sort of virtual cube, 50 metres on a side, consisting of two lateral wings connected on thenorth by a suspended volume and on the south by a low body having truncated-pyramidalroof and by a glazed walkway at the tenth floor. The two lateral wings are assigned to ope-rations offices; the block to the north is reserved to halls and offices for meeting the public,and the volume that concludes in a truncated pyramid houses the two great congresses halls,seating 500 and 150 persons. They may be accessed independently for any public use. Thecomplex features a sharp contrast between the white of the marble cladding of the stairbodies’ four corner volumes and the transparent surfaces of the continuous glazed facades.In the east and west wings completely transparent bands of glazing alternate with glassblocks, while for the suspended building body a “dual skin” solution was adopted with a brise-soleil towards the south, and a wholly glazed façade to the north. The corporate offices buil-ding runs up twelve floors above ground, reaching a height of 54 metres, and has two base-ment floors, one outfitted for services and the other for parking, the total net area being27,000 square metres. To this is added the 25,000 square metres of underground parkingadjacent. The office building structure features two trios of steel latticework beams placed atelevations 23.35 and 44.15 metres. Their span is 36 m and their function is to sustain thenine-floor suspended body. The beams bear on reinforced-concrete columns stabilized by twoof the four stairwells. For the foundations, considering the soil’s low compressibility and its bea-ring capacity, separate direct footings were chosen.


ProjectBanca Lombarda Center

LocationBrescia, Italy

ClientSBIM – Società Immobiliare Mobiliare S.p.A. , Brescia

ArchitectGregotti Associati International s.r.l., Milano

Structural engineerSajni e Zambetti s.r.l., Milano

System design Amman Progetti S.p.A., Milano

General contractorColombo Costruzioni S.p.A., Lecco

PhotographsDonato Di Bello – Gregotti Associati InternationalS.p.A.

Banca Lombarda Center


• 1- Plan. 2-3 Phases in the construction of the structures: all reinforced-concretecomponents were in situ poured.

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• 4- Ground-floor plan. 5- Second-level plan.

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The building, located on the southwest outskirts of the city of Biella, is essentially characterizedby the juxtaposition of two truncated pyramids: one, upright, takes in and configures the basicstructural system, i.e. the great reinforced-concrete spherical caps representative of the entirespatial order. The other, inverted and differently oriented relative to the axes of the system ofcaps themselves, houses the main stairwell, making possible the natural statics equilibrium ofthe whole. The construction develops on the whole on four levels above ground, besides thebasement floor and the roof terrace, for a total net area of 1500 square metres. Between theground floor, housing the accesses and reception desk, and the first floor is a mezzanine floorhaving as its basic function to act as an intermediate between the spaces, all wholly devotedto housing, small and large events, such as showings, reviews and exhibitions and the videoprojections they involve. The first floor develops within the great spherical caps. On the second and top floors, directly connected with the overlying terrace, space is found fora cafeteria and a small multimedia library, while on the basement floor is the great exhibitionarea, naturally lighted by special large skylights “excavated” in the structure of the hanging gar-den. This permits its conversion, as needed, into a sizeable conference hall. The building is stron-gly characterized by the massive use, not only of r.c., but also of marble and stone. The wholemain construction above ground bears on five columns, placed at the sides of an equilateraltriangle, that is in respect of the primary design of a truncated pyramid, which is what, sub-stantially, the building configures. The two columns near the top of this triangle, towards theeast, are 0.60 m in diameter, while the remaining three, placed at the corners and in midspanof the west side, are 0.70 m in diameter. Starting from the upper level (ground floor) thecolumns change typology, being converted into characteristic oblique pilasters, two of which –along the west axis – display a variable rhomboidal section owing to their tapering towards thetop.


ProjectBoglietti Palace – Cultural Center

LocationBiella, Italy

ClientGiovanni Boglietti – “Obiettivo Domani” CulturalCentre

ArchitectsArch. Alberto Rizzi, Biella

AssistantsLuca Gibello, Filippo Chiocchetti, Paolo Strobino,Francesca Frigato

Structural engineer Orio Delpiano, A.I.R.E., Biella

General contractor Lasimon S.a.s., Biella

PhotographsArchivio Studio Rizzi, Costantino Merlini, DavideLovatti, Roberto Marchisotti

Boglietti Palace


• 1- The building in plan and volume. 2- Structural system of the large sphericalvaults. 3- Basement-floor plan. 4- Ground-floor plan.

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• 5- First-floor plan. 6- View of building from the south during final finish phases.7- Section through building.

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The Immaculate Heart of Maria parish complex stands in an area donated by the borderingSalesian Sacred Heart hospice. During design it seemed fitting to exploit the lot’s orographicconformation in relation to the main street, as did the need to lower the foundation plane toreach a soil of consistency adequate to permit direct foundations. Thus was permitted: the shif-ting of the catechism classrooms and the services below the sacristy into a band as long asthe lot, lighted by a light shaft with plantings; the location below the church of the multi-usecommunity salon seating 450; and the construction of the weekday chapel for seventy faith-ful. On the upper floor were located the assembly hall, with access from the sacristy, the sacri-sty itself with parish office, and hygienic facilities for the handicapped. Over the sacristy, suspended from the roof, are the priest’s quarters. The bell tower, placedpseudobaricentrically, completes the whole; it is accessed from the sacristy. The entire oeuvrewas built of fair-face white lightweight structural cement concrete, having as well an excellentheat-transfer coefficient, which meant the adoption of walls only 250 mm thick that metenergy-consumption standards. The lightness of the statics scheme was made possible by thespecial technologies applied for construction. One such was that used for the ribbed triangular lacunar plates with fine triangular mesh thatpermitted very low structural weights and were aesthetically pleasing as well as acousticallysuitable. Stimulating too was the design and construction of the five bombé “sail” walls of tra-pezoidal form, all different the one from the other. Extremely complex was the construction ofthe belltower, especially at the points where it was grafted onto the arches and at the floorstructures where the in-depth beams arrive with strong traction forces. The sacristy floor struc-ture is a dual slab lightened with polystyrol pads, poured in situ without break. The three St.Andrew’s crosses, placed on each end of the frames connecting the walls to take up horizon-tal forces, were built on site, positioned and solidly joined in the pour of the ends of the rec-tangular section walls.


Project“Cuore immacolato di Maria” parish complex

LocationFormia, Latina, Italy

ClientParroco Maccioni Don Gesuino – Arcivescovo diGaeta, Mons. Vincenzo Farano

ProjectArch. Bernardo Re

Geothecnical advisoryIng. Giovanni Gambacorta

PhotographsArch. Bernardo Re

“Cuore immacolato di Maria” parish complex


• 1- Church plan.

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• 2- Church front facade. 3- The soffit of the special roof created with ribbed trian-gular plates. 4- The wooden intermediate floor suspended from the concrete frame.

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The St. John the Baptist parish complex was built in Lecce, in the Stadio district, a zone onthe outskirts featuring numerous residential buildings and economic and subsidized publicbuilding construction. The design of the centre was thus made up with the aim of creating anurban pole of attraction, a true “community house”, able finally to attribute recognizability tothe entire area. The complex comprises a series of volumes that de-limit and define a foot plaza, an internalcourt and a closed space, the Walled Garden, conceived for meditation. The church’s assem-bly hall is square-plan, 24 m on a side. Next to it is a rectangular wing holding the sacristyand the weekday chapel.The assembly hall’s bearing structure is sustained by just four columns, which identify a full-height central basin, trapezoidal in plan, connoted by a strong feeling of lightness. On the hallperimeter are the service spaces, lower in height. The four columns are con-nected by isola-ted beams, at a height of eight metres. One of the beams of this internal frame extendstowards the entrance wall, piercing it and projecting beyond cantilever-wise. From it hangs thelarge cross that connotes the building façade. The hall roof bearing structure, which reachesa peak elevation of 15.65 m, is seen as a lacunar complex formed of square-section beams1.60 m on a side. In one section, the roof cantilevers out by 7 m. In the lateral areas thebeams and the columns create a square-mesh modular grid 6 m on a side. The frame structure arranged on the perimeter is clad with plastered tuff block walls. Theparish works building has an r.c. frame bearing structure. Of special interest is the stairway,created by a cantilevered slab. The campanile is shaped like a portal, expanded in height. Thetwo vertical outside septums are 0.30 m thick, as is the septum supporting the stairway, andhave a total height of 28 m. The foundations are a one metre deep footing. The cladding forthe Parish hall and for the entrance security lock, both inside and outside, is of Lecce stone.


ProjectNew parish complex of San Giovanni Battista

LocationLecce, Italy

ClientArchdiocese of Lecce, Italy

ArchitectArch. Franco Purini, prof. Laura Thermes – Rome

AssistantsLuigi Paglialunga, Massimiliano De Meo

Structural engineeerIng. Enzo Pierri (church), ing. Andrea Cinuzi (belltower)

Supervision of construction Raffaele Parlangèli

General Contractor Fratelli Marullo, Calimera (Lecce)

PhotographsStudio Purini Thermes

“San Giovanni Battista” parish complex


• 1- Ground-floor plan and facade (line of section 4). 2- First-floor plan and in-terior facades (lines of sections 2 and 3).

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• 3- Sezione trasversale.

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The project concerns a corporate complex including offices and shops set down in and com-pleting the residential and business center of a subdivision called Somada. The triangular-planbuilding features a large internal patio-garden. Lining the patio’s three sides are the shop spaces,on the raised ground floor and the first floor, while the spaces assigned to offices are on the 2nd

and 3th floors. At the building’s heart, besides its planted area, is the cylindrical core. Within it,starting from the raised ground floor, the helicoidal stairs and a glazed-wall elevator develop. Itculminates in a conical-roof skylight. The shops, all enjoying extensive glazing, and the offices areserved, and accessed, on all levels by broad porticoed foot ways. On the raised ground floor, thearea fronting on the court is furnished around the whole perimeter with chairs in which to sitand enjoy the outdoor area. Formally, the complex is characterized by a number of elements,which spring from the search for an appropriate architectural system, filled out with the neces -sary functional elements. The building comprises a bearing frame structure of semi-precast type-K reinforced concrete. Being conceptually connected with the traditional reinforced-concreteframe, it is composed of reinforced-concrete precastings (beams, stair flights) light-weightpredalles-type floor structures, prestressed-concrete honeycomb slabs and in situ-poured mem-bers (foundations, columns, perimetral and dividing walls, and the stairwells). Characteristic ofthe system is the wet-type node, where beam-column structure continuity is created by the sim-ple overlapping of reinforcings and a successive solidizing pour, made at the same time as thefloor structure is poured. The structures subsystem consists of v.r.c. semi-precastings, whose joins,effected by reinforcings and integrating pours, create a statically indeterminate structural com-plex. Its performance characteristics can be traced back to an equivalent in situ-poured struc-ture, as regards the absorption of both vertical and horizontal forces.


ProjectBusiness Center

LocationResidential and shopping centre “Somada”, Italy

DesignArch. Enzo Zacchiroli, Bologna

Precast structure designIng. Mauro Ferrari, Reggio Emilia

General contractor for structureEdilquattro spa, Montecavolo di Quattro Castella (RE)

General contractor for precast structuresAPE spa, Montecchio Emilia (RE)

PhotographsPatriziaVirginia Belli, Bologna

“Somada” Business Center


• 1-2 Sections through the building. 3- Second-level plan: structural details.

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Documents

Italian National Report – Research and Construction