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GROUTING IN ROCK TUNNELLING
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GROUTING IN ROCK TUNNELLING
With Selected Case Studies of Post grouting in Kárahnjúkar Hydroelectric Head Race Tunnel
East Iceland.
Joseph Oyeniyi Ajayi
M.Sc. ritgerð í Umhverfis- og audlindafræði
Júní 2007
Umhverfis-og byggingarverkfræðiskor Verkfræðideild Háskóla Íslands
© Joseph Oyeniyi Ajayi, 2007 ISBN: 978-9979-9812-0-6 Printed in Iceland by Háskólaprent, 2007.
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PREFACE This document is a M.Sc. thesis submitted to the Department of Civil and Environmental
Engineering at the University of Iceland (Háskóli Íslands) for the fulfilment of the award
of Master of Science in a Masters program in Environment and Natural Resources.
The study describes the grouting activities carried out at the Tunnel Boring Machine
(TBMs) fullface and the Drill and Blast sections of the Kárahnjukar Head Race Tunnel
(HRT) all the way from the power intake at the Hálslón reservoir to the power house in the
Fljótsdalur valley (45 km). It sheds more light on the need for grouting (pre-and post
grouting) and highlights the procedures adopted to carry out grouting activities, open-up
some of the problems encountered during excavation such as the possibility of
encountering large quantities of high pressure ground water inflow which may delay or
result in a temporary interruption of production. General water treatment along the stretch
of the tunnel will be described.
The effectiveness of both the traditional cementacious grouting, as well as the chemical
counterparts, as the most acceptable methods of sealing the tunnel against water ingress to
prevent any adverse internal environment, prevent unacceptable impact on the external
surrounding environment as well as maintaining hydrodynamic containment of the tunnel
(NTS.Publication No 14. Ch 9). Necessary equipment and materials needed for the
successful operation will be discussed.
In conclusion, this thesis will generally summarise some of the experiences and lessons
learned during the construction of Kárahnjúkar Hydroelectric Head Race Tunnel.
The study was supervised by Birgir Jónsson associate professor at the Department of Civil
and Environmental Engineering, University of Iceland and Björn A. Hardarson, Senior
Geotechnical Engineer at Geotek Consulting Engineering, at present Resident Engineer at
the 11 km Hedinsfjord Road Tunnel, North Iceland. The External examiner was Emeritus
Professor Julius Solnes, Department of Engineering, University of Iceland, former
Minister for the Environment in the Icelandic government.
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Aim / Scope of the project. This research describes the grouting activities carried out at the Tunnel Boring Machine
(TBMs) fullface and the Drill and Blast sections of the Head Race Tunnel (HRT) all the
way from the power intake at the Hálslón reservoir to the power house 45 km away in the
Fljótsdalur valley. This research sheds more light on the need for grouting (Pre- and Post-
grouting), and highlights the procedures to carry out grouting activities in the Head Race
Tunnel (HRT). The research will open up some of the challenges encountered during
excavation such as the possibility of encountering large quantities of unexpected high
pressure ground water inflow which directly or indirectly delay production or may even
bring a temporary interruption to the excavation progress.
The significance and the effectiveness of grouting as one of the ground support methods in
tunnelling will be disscussed and the general performance of Tunnel Boring Machine
(TBM) beeing used for the first time in Icelandic basaltic rock will be overview.
Considering the hydrogeological nature of groundwater in Iceland which is very different
in the geologically younger formations – Late Quartenary and Recent in terms of high
permeability, where active and open fissure swarms enhance the permeability as well as
creating a strong anisotropy as compared to the geologically older formations – Tertiary
and Early Quaternary (Sigurðsson and Einarsson 1988)1. Smaller levels of ground water
ingress can cause problems in the tunnel or in the surroundings. Hence, ground water
ingress can be controlled by drainage, proper pre-excavation grouting and post-excavation
grouting 2(Knut F. Garshol 2003).
The research will analyse the effectiveness of both the traditional cementacious grouting
and chemical counterpart, as the most acceptable methods of sealing the tunnel against
underground water inflow to prevent any adverse internal environment, prevent
unacceptable impact on the external environment as well as maintaining hydrodynamic
containment in order to prevent leakage 3(NTS 2004. Publication No 14). Grouting
activities for stabilization as part of the rock support will also be revealed.
1 Freysteinn Sigurdsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availability and Demand. Jökull, No. 38, 1988. 2 Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. 3 Norwegian Tunnelling Society, NTS (2004). Ch 9. Grouting. Publication Number 14 pp 43-46.
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The proper equipment and materials needed for the successful operation will also be
briefly disscussed.
The Practical application techniques of pressure grouting ahead of the full Tunnel Boring
machine (TBM) and Drill and blast sections of the tunnel as well as already excavated
sections supported by theory are in focus.
In conclusion, it will generally summarise some of the experiences and lessons learned on
the construction of Kárahnjúkar Hydroelectric Head Race Tunnel .
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Acknowledgement. Special thanks to my supervisors Birgir Jónsson, associate professor and Björn Hardarson,
Senior geotechnical engineer at Geotek Consulting Engineering. I am highly grateful for
their great assistance and guidance. My sincere appreciation to all the staff and students of
the Geology and Geography Department, Environment and Natural Resources
Management Department and Civil and Environmental Engineering Department for all the
assistance they rendered during the whole study period. My appreciation to Hnit
Consulting Enginering for supporting the printing of this thesis.
My in-depth thanks to my wife and family for all the support and love from the beginning
of my life.
All honour and glory to God for a successful completion of the study.
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Table of Contents 1 Introduction. ..............................................................................................................13 2 Historical Background and Description of the Project Area. ....................................17 3 Environmental Impacts Assessment..........................................................................24 4 Rock Distribution and Geology of the Project Area. ................................................28 5 Hydrogeology/Groundwater Condition in Iceland. ...................................................31 6 The Tertiary-Early to Middle Quaternary Bedrock...................................................33
6.1 Sedimentary Aquifers. .......................................................................................34 6.2 Late Quaternary and Recent Bedrock................................................................35 6.3 Pyroclastics, lavafields and fissure swarms.......................................................35 6.4 Great springs and springs areas. ........................................................................37
7 Land Use Planning. ...................................................................................................39 8 Design and Construction. ..........................................................................................40
8.1 Design and Construction of the project. ............................................................40 8.2 Excavation Methods. .........................................................................................41
8.2.1 Excavation Class/Rock Support. ...............................................................41 8.3 Bore Classes. .....................................................................................................42
9 Monitoring Instrumentation.......................................................................................46 9.1 Deformation monitoring. ...................................................................................46 9.2 Groundwater Control.........................................................................................46 9.3 Vibration monitoring. ........................................................................................47
10 Grouting into the rock. ..............................................................................................48 10.1 Reasons for grouting in tunnelling. ...................................................................48 10.2 Comments on Pre-grouting and Post-grouting. .................................................48 10.3 Reasons behind increase in the use of pressure grouting. .................................53 10.4 Design of grouting in rock tunnels. ...................................................................54 10.5 Practical basis for injection works in tunneling.................................................55 10.6 Influence of tunneling to the surrounding and the conditions inside the tunnel56 10.7 Condition inside the tunnel................................................................................57 10.8 Issuing Of Site Instruction.................................................................................57
11 Boreholes in Rock. ....................................................................................................59 11.1 Top hammer percussive drilling. .......................................................................59 11.2 Down the hole drilling.......................................................................................60 11.3 Low speed rotary drilling. .................................................................................61 11.4 High speed rotary drilling. (Core drilling). .......................................................61
12 Typical solution example for drill and blast excavation............................................64 12.1 Probing ahead of the face. .................................................................................64 12.2 Drilling of injection holes..................................................................................67 12.3 Cleaning of holes/ Flushing of boreholes for injection. ....................................67 12.4 Placement of packer...........................................................................................68 12.5 Water Testing/Water Pressure Testing. (WPT).................................................68 12.6 Choice of injection materials.............................................................................70 12.7 Injection pressure and procedure.......................................................................70 12.8 Injection records and quality control during injection.......................................72 12.9 Settling of grout, time until next blast. ..............................................................72
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12.10 Drilling of control holes. ...................................................................................73 13 Typical solution example for Tunnel Boring Machine (TBM). ................................74
13.1 Probing a head of the face. ................................................................................74 13.2 Drilling of injection holes..................................................................................74 13.3 Cleaning of holes/ Flushing of boreholes for injection. ....................................75 13.4 Placement of packer...........................................................................................75 13.5 Water Pressure Testing......................................................................................75 13.6 Choice of injection materials.............................................................................75 13.7 Injection pressure and Procedure.......................................................................76 13.8 Injection records and quality control during injection.......................................77 13.9 Setting of grout/ Hardening time. ......................................................................77 13.10 Drilling of control holes. ...................................................................................77
14 Packers.......................................................................................................................79 14.1 Mechanical Packers (expanders). ......................................................................79 14.2 Disposable packers. (Single application). .........................................................80 14.3 Hydraulic packers. .............................................................................................81 14.4 Standpipes techniques. ......................................................................................83 14.5 Tube-a-manchet. ................................................................................................84 14.6 Types used on the Project..................................................................................84
15 Cement-Based Grouts................................................................................................85 15.1 Basic properties of cement grout. ......................................................................85 15.2 Cement particle size, fineness. ..........................................................................85 15.3 Bentonite etc. .....................................................................................................87 15.4 Rheological behaviour of cement grout.............................................................87 15.5 Pressure stability of cement grout and use of high injection pressure...............88 15.6 Grout setting characteristics and durability of cement injection in rock...........95 15.7 Additives and admixtures for cement injection.................................................96 15.8 Equipments for Cement injection. .....................................................................97
15.8.1 Mixing equipment. (Mixer). ......................................................................97 15.8.2 Agitator......................................................................................................98 15.8.3 Grout Pumps..............................................................................................99 15.8.4 Complete systems. ...................................................................................101 15.8.5 Automated mixing and grouting plants ...................................................102 15.8.6 Recording of grouting data. .....................................................................103 15.8.7 HIR ..........................................................................................................103 15.8.8 HFR .........................................................................................................103
16 Chemical Grouts. .....................................................................................................104 16.1 General. ...........................................................................................................104 16.2 Overview of Polyurethane grouts. ...................................................................105 16.3 Adopted Products on Kárahnjúkar Head Race Tunnel....................................111 16.4 Silicate grouts. .................................................................................................114 16.5 Acrylic grouts. .................................................................................................115 16.6 Epoxy resins. ...................................................................................................116 16.7 Bitumen Asphalt. .............................................................................................116
17 High Pressure Ground Water Condition..................................................................121 17.1 Basic Problem..................................................................................................121
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17.2 Practical procedure in high risk areas..............................................................121 17.2.1 Pumping System. .....................................................................................121 17.2.2 Probe drilling /coring...............................................................................122 17.2.3 Injection. ..................................................................................................122 17.2.4 Special cases. ...........................................................................................122
18 Experiences /Summaries of Case Studies................................................................124 18.1 Case study 1. Chainage 135 – 1+ 146. Power Intake. .....................................124
18.1.1 General remarks and recommendation. ...................................................131 18.2 Case study 2. Chainage 19 + 645- 19 + 595. Adit 2 ( TBM Stuck 6-months) 132 18.3 Case study 3. Chainage 1 + 468, 5. Adit 4. (Dry-dyke) ..................................145 18.4 Case study 4. Chainage 002 + 255. Adit 3. ~150- 200 l/sec. ..........................148 18.5 Adit 2. Jökulsa Valve Chamber. Curtain Grouting – Plug 2.2. .......................151
19 Discussions and summaries of lessons learned, results and conclusion..................155 20 References. ..............................................................................................................157
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List of figures and tables. Figure1. Pictures showing the dam area and the vertical Penstock of the project.
Figure 2. Kárahnjúkar Pressure shafts compared to the highest building of the World.
Figure 3. Map Showing reservoirs, tunnels, dams and roads. Description of the Project
Area.
Figure 4.The Project alignment. Showing Description of the Kárahnjukar Head Race
Tunnel (HRT) including Power House, (VIJV- Database 2006).
Figure 5. Map showing the project area, reservoirs, HRT, power house, roads,
transmission lines and the neighbouring communities.
Figure 6. Summary of the Major Environmental Impacts of the project.
Sources. Landsvirkjun website.
Figure 7. Diagram Showing Geology-Longitudinal section.
Figure 8. Map showing the demarcation of Iceland based on the bedrock classifications.
Figure 9. Map of Iceland showing great springs from the Late Quaternary and Recent
Zone, most of them in connection with fissure swarms or lava fields.
Figure 10. Showing differnt Bore Classess, rock types and percentage of face surface.
Source: VIJV Database 2002.
Figure 11. Map showing excavation details of part of head race tunnel, (Power-intake) and
Adit 4 junction. Source, VIJV Database 2004.
Figure 12. Map showing the profile of the Head race tunnel, directions of the three tunnel
boring machines (TBMs) and Jumbo Drilling Machines. (Source. Landsvirkjun Database
2006).
Figure 13a. Multi-boom drilling jumbo in operation for blasting and grouting holes
Figure 13b. Sandvik Tamrock Axera T12 and Atlas copco Rocket Boomer WL3 C drilling
jumbos. At right, a rod changer system for the drilling jumbo (Garshol, 1999).
Figure 14. Pre-grouting drilling system on a Robbins TBM (Garshol, 1999).
Figure 15a. Showing disposable packers (With a distance of 100-150mm between the
rubber sleeves). Source BASF Construction Chemicals.
Figure 15b. Showing inflatable packers, (Source Bimbar).
Figure 15c. Showing Inflatable Packers, (Source Hany AG Equipment).
Figure 16. Dispersing effect of an admixture using Micro-cement.
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Figure 17. Rehological behaviour of Newton and Bingham fluids.
Figure 18. (A1). Curves showing consumption drops and pressure remains constant after
peak.
Figure 18. (A2).Curves showing pressure increases and consumption remain constant after
Peak.
Figure 18 (B1). Curves showing. pressure and consumption remains constant after Peak.
Figure 18 (B2). Curves showing pressure and consumption rise to some value and then fall
rapidly.
Figure 18 (C1). After a rise in pressure to a certain value, the pressure remains more or
less constant, while the rate of consumption goes on rising.
Figure 18. (C2). After reaching a certain value, the pressure quickly drops while the rate of
consumption remains constant.
Figure 19. Mixer. (Photo Hany AG Equipment).
Figure 20. Agitator (Photo Hany AG Equipment).
Figure 21. Grout Pump. (Maximum Pressure 100 bar).
Figure 22. Grout Pump. (Maximum Pressure 200 bar). (Photo Hany AG Equipment).
Figure 23. Complete grout pump. ( Photo Hany AG Equipment).
Figure 24. Automated mixing and grouting plants. (Photo Hany AG Equipment).
Figure 25. Showing Recording Systems. (Photo Hany AG Equipment).
Figure 26a. Showing typical reactive polyurethane prepolymers.
Figure 26b. Water reactive prepolymers.
Figure 27.Core showing Infilling of rock cavities with Polyurethane.
Figure 28. Showing Power intake, tunnel concrete lining outline and grout plan.
Figure 29. Diagram showing Contact, Consolidation and Curtain grouting in the
dismantling chamber (Ch 1 + 135 -1+ 146).
Figure 30. Diagram showing the Chronological advancement of the TBM 2 through fault
#1, #2 and #3.
Figure 31. Diagram showing the length of the Extraordinary Geological Section(EGO)
Figure 32. Diagram showing grouting activities and backfilling of void as part of the rock
support and finishing works.
Figure 33. Geological Mapping showing the geological over break (Dyke).
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Figure 34. Drawing showing grouting/Treatment of dry dyke at Ch 1+ 468. 50
Figure 35. Geological Mapping showing the geological over break (Fault).
Figure 36. Diagram showing the drilling and grouting activities/treatment of faults at
Ch 002 +255.
Figure 37a. Sketches Showing Ring A (15m). Curtain Grouting ( 1,3,5,7,9,11,13,15,17,19,21,23).
Figure 37b. Sketches Showing Ring B (15m). Curtain Grouting ( 2,4,6,8,10,12,14,16,18,20,22,24). Figure 37c. Sketches Showing Ring C (15m). Curtain Grouting ( 1,3,5,7,9,11,13,15,17,19,21,23). Figure37d. Sketches Showing RingD (5m). Curtain Grouting ( 2,4,6,8,10,12,14,16,18,20,22,24). Figure 37e. Curtain drilling and grouting of plug 2.2.
List of Tables. Table 1. Showing the key figures of the project.
Table 2.Ranking of Major Grout Properties.
Table 3. Ranking of Chemical Grouts by Application. (EM 1110-1-3500. 31 Jan 1995).
Table 4. Summary Timeline of chronological advancement of the TBM 2 through fault #1,
#2 and #3.
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1 Introduction.
The Kárahnjúkar Hydro-Electric Project is located in eastern Iceland between the Jökulsa
á Dal and Jökulsa í Fljótsdal rivers. The rivers have their origin from the Vatnajökull
glacier and run towards the northeast. In the Hálsón reservoir the water of the Jökulsá á
Dal will be stored and then conveyed by the headrace tunnel to the pressure shaft and then
to the underground power house in the Fljótsdalur valley about 40 km south of the town of
Egilsstaðir. In the Jökulsá diversion, the discharge of some tributaries to the Jökulsá í
Fljótsdal will be stored in Kelduá reservoir and then conveyed from there by a tunnel to
Ufsarlón pond to be located in the Jökulsá í Fljótsdal river course. The Jökulsá tunnel
conveys water from the Ufsarlón pond to the Jökulsá junction with the headrace tunnel.
Power from the project is transmitted by two 400 kV transmission lines to the port of
Reyðarfjörður where Alcoa a multinational company from the United States of America
has almost completed the construction of a 340 thousand tonnes per year aluminium
smelter. These two interrelated projects are built as a result of the long term governmental
policy to develop Icelandic renewable energy mainly for export revenue and employment
opportunities in the power intensive industries. In the light of this, a 40 year contract to
provide power for the plant was concluded with US Multinational Alcoa in March
2003.4(J. Roby 2006)
Landsvirkjun the owner of this project, founded on 1 July 1965 is an independent legal
entity formerly owned by the Icelandic Government, the City of Reykjavik and in 1983,
the Municipality of Akureyri became a third owner but solely owned by the Icelandic
Government as from 1 January 2007 through taken over the shares of Reykjavik and
Akureyri. Since then, the company is a state-owned partnership.5( http:www.lv.is/EN)
The company with a current installed electricity generation of over 1,200 MW, is the
largest power producers in Iceland. Landsvirkjun has until 2006 owned and operated the
main National Grid which is now owned by a seperate company named Landsnet.
Landsvirkjun generates about 87 % of the country,s electricity.
4 Joe Roby. February (2006). Tunnel Bussiness Magazine. Nothern Exposures. Supplying Iceland,s New hydropower Plant. Tunnelling Overcome Extreme Weather and Geological Conditions. 5 Landsvirkjun Website. http: www.lv.is/EN.
14
The Project area extends from the power house and about 40 km south and southwest, and
onward to the glaciers protruding north from the north-eastern part of the Vatnajökull Ice
cap.
Picture showing the dam area, (Reservoir).
15
Figure1. Pictures showing the dam area and the vertical penstock of the project.
16
Figure 2. Kárahnjúkar Pressure shafts compared to some of the highest buildings of the
World.
17
2 Historical Background and Description of the Project Area.
The establishment of large-scale industry has been on the agenda in East Iceland for the
past three decades and in all this period, the attention has been on the industry site in
Reyðarfjörður. Below are the summary of some disscussions that has came up since the
last three decades:6(http// www.kárahnjúkar.is)
• 1975-1976: Norsk Hydro the Norwegian company examined possibilities of
building an aluminium plant in Reyðarfjörður. The power was scheduled to come
from the 200 MW Fljótsdalur hydroelectric power plant.
• 1980-1985: Rio Tinto Zink the Australian company had plans to build a silicon
metal plant in Reyðarfjörður. The power was also planned to come from the
Fljótsdalur hydroelectric power plant. This project was later abandoned while on
its way.
• 1989-1990: Reydarfjörður was again in mind as a site for large-scale industry
when the multinational firm Atlantal, owned by Hoogovens, Alumax and
Gränges, was searching for a site for an aluminium plant. The final choice was
Keilisnes in South of Iceland and the power was still to come from Fljótsdalur
hydroelectric power plant. Construction was just about to start when the project
was suspended in 1991.
• 1998: Norsk Hydro took up discussions again about an aluminium plant and a
hydroelectric Power plant in East Iceland. Memorandum of understanding was
signed in June 1999, for a 120.000 tonnes smelter and a power plant in
Fljótsdalur with reservoir at Eyjabakkar, known as the Noral Project. In 2000
investors came to the opinion that the smelter needed to be bigger in order to be
profitable then, in May a new memorandum of understanding was signed for the
Noral Project whereby the aluminium plant was now planned to have a
production capacity of about 240.000 tonnes per annum in its first stage and a
second stage with a capacity of about 120.000 tonnes was planned for later. A
project company, Reyðarál, jointly owned by Norsk Hydro and Icelandic
6 Kárahnjúkar – Large Scale Industry in East Iceland. (2002). Landsvirkjun website http//www.Kárahnjúkar.is.
18
investors was to develop the plans for the smelter and Landsvirkjun would
supply the power for both stages with a hydroelectric power plant at Kárahnjúkar
and a diversion in Fljótsdalur Valley.
• In March 2002: It was revealed that Norsk Hydro could not meet deadlines set
for September 1st in the decision process, due to Hydro´s massive investment in
the German aluminium company VAW. However, they still maintained their
interest in the project at a later stage. In the light of this, the Icelandic
Government established a commission to deeply look into other company´s
interest in the project and shortly afterwards talks started with Alcoa, the worlds
biggest aluminium producer.
• April 19th 2002, a joint action plan ( JAP) was signed with Alcoa to explore the
possibility of constructing a state-of-the-art aluminium production plant in
Eastern Iceland. The joint action plan was extended on May 23rd until July 18th,
2002 by which the parties must have made up their mind to sign a formal
Memorandum of Understanding and proceed with the project. In the project,
Alcoa would own and operate a 340,000 metric ton per year aluminium plant
with power from a 500+ MW hydroelectric power station in Eastern Iceland to be
constructed and operated by Landsvirkjun, the National Power Company of
Iceland.
• July 19th 2002 a Memorandum of Understanding (MOU) was signed in
Reykjavik between the Government of Iceland, Landsvirkjun, Iceland´s National
Power Company and Alcoa formalizing their co-operation on a 295,000 metric
ton per-year, low emission aluminium smelter to be built in eastern Iceland. In
the contents of the new Memorandum of Understanding, Landsvirkjun will begin
development of a 630 MW hydropower facilities in eastern Iceland and Alcoa
will embark and carry out environmental and engineering feasibility studies of
the smelter in Reyðarfjordur. The MOU also involved a harbour facility at
Mjóeyri coupled with appropriate infrastructures improvements In eastern
Iceland.
• The Management of Alcoa decided to present the Plant Project to its Board of
Directors for full approval in January 2003 as Landsvirkjun also aimed at
19
submitting the Power Contract, and the Harbour Fund was aiming at submitting
the Harbour agreement, to their respective Boards of Directors for final approval
before the end of December 2002. It was jointly agreed that in the event that
Alcoa, the Government and Landsvirkjun were unable to ultimately agree upon
the basic premises for binding definitive agreements on implementation of the
Plant Project before the end of march 2003, this MOU will terminate, unless the
concerned bodies jointly agreed to extend the term of the MOU or alternatively
replace it by a further Memorandum.
The Kárahnjúkar Hydropower project is located in the highlands of remote northeastern
Iceland. While the project has been discussed for nearly 40 years, it only became a reality
in the last five years. In 2001, a thorough Environmental Impact Assessment (EIA) on the
project was completed. The Iceland Ministry for the Environment gave a final positive
ruling and in 2002, followed by approval from a sizable majority of the Althing, Iceland
Parliament. Later that year, the Ministry for industry issued all of the necessary permits for
the project and preparatory work was able to begin, including road construction. Local
Municipalities issued the necessary construction permits in February 2003 and in March
2004, a 40 year contract to provide power for the new aluminum smelter was concluded
with Alcoa.
The project area is lying generally at an elevation slightly exceeding 600 m above sea
level. It is charactrised by the high plateau which gradually built up and formed over a
time period of some 10 Million years. Into this, valleys have been eroded by various rivers
and the glaciers during numerous glaciation periods resulting in mountains of different
prominence which rise above the plateau.
Kárahnjúkar power plant will have an installed power of 690 MW and the annual
generating capacity will be about 4,600 GWh. To meet this standard, the Jökulsá á Dal
river is dammed by three dams at Fremri Kárahnjúkar. The largest of the dam
Kárahnjúkastífla, is located at the Southern (upper) end of the Hafrahvammar canyon and
is about 730m (2,400 ft) long and 193m (633 ft) high containing 8.5 million cu m of rock.
The structure is concrete-faced rock fill dam (CFRD) type and when completed will be the
20
highest of its kind in Europe, among the highest and fourth largest of its type in the World.
(Peter Reina 2006, Joe. Roby 2006).7
Completing the trio are other smaller saddle dams which will be built at Kárahnjúkar,
these are Desjarárstifla dam to the east, and Sauðárdalsstífla dam to the west. Both dams
will be rock fill dams with an earthen core, and together the three will combine the main
57 km2 Hálsón storage reservoir. The three dams will retain the Hálslón reservoir with
about 2.1 × 109 usable storage. When the reservoir is full, its water level will reach a
height of 625 m above sea level, and its shores will reach the edge of Brúarjökull glacier.
It is estimated that the Hálsón reservoir will be filled by late summer in most years.
Surplus water will then be diverted through a spillway chute at the western end of the
Kárahnjúkastífla dam down to the edge of the Hafrahvammar canyon.
On the east side of the mountain Snæfell, the Jökulsá í Fljótsdal river is dammed about 2
km downstream of the Eyjabakkafoss waterfall on the north side of the Eyjarbakkar
wetlands. The intake pond this creates has been named Ufsarlón, and water from three
tributary rivers on the eastern side of the Jökulsá is also diverted into it.
Two steel-lined vertical pressure shafts lead the water from the intake to the underground
powerhouse. Each shaft is 420 m high, and the total head is 599 m. The power house has
six Francis turbines each with a rated output of 115 MW. When water exits the power
house, a tailrace tunnel and canal take it to the course of the Jökulsá í Fljótsdal river.(see
table 1).
The total length of the whole tunnel system is over 70 km, and they vary in depth from
100 to 200 m. Around 40 km of the head race tunnel and parts of the access adit tunnels
has been full faced bored using the three Tunnel Boring Machine (TBMs).
7Peter Reina. (June 2006). Energy: Harsh Climate, Difficult Geology Temper Peace of Work. Joe Roby. ( February 2006). Tunnel Bussiness Magazine. Nothern Exposures. Supplying Iceland,s New Hydropower Plant. Tunnelling Overcome Extreme Weather and Geological Conditions. Focus on Iceland (2005). Extreme tunnelling – Kárahnjúkar. Tunnels and Tunnelling International October 2005. Pg 16 -20. Peter Reina (June 2006). Energy: Iceland Digs Deep To Develop Power in the Wilderness.
21
Figure 3. Map showing the project area, reservoirs, HRT, power house, roads,
transmission lines and the neighbouring communities.
22
Figure 4. Map Showing reservoirs, tunnels, dams and roads. Description of the Project
Area.
23
Figure 5.The Project alignment. Showing Description of the Kárahnjukar Head Race
Tunnel (HRT) including Power House, (VIJV- Database 2006).
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3 Environmental Impacts Assessment.
In accordance with the Icelandic law No. 106/2000, all hydro power project greater than
10MW or with reservoir greater than 3 km2 are to be subjected to Environmental Impact
Assessment. Landsvirkjun, as the developer promoting the project was required to carry
out such evaluation.
The impact area of the project includes highlands by the glacier Vatnajökull as well as
land along the rivers through the valleys of Jökuldalur and Fljotsdalur out to the coast of
Héraðsflói. This expanse of land lies within the following two communities:
• Fljótsdalshreppur
• Fljótsdalshérað.
Hydroelectric power plants always involve changes in the environment when rivers are
dammed and their flow diverted through power-generating stations, their courses below
the dams are altered and land above the dams remains flooded. In addition, channels, new
roads as well as various other visible structures are introduced.
The area affected by Kárahnjúkar Power Plant, particularly through dams, reservoirs and
roads, is in all ramifications very unique and certain parts of it are considered to have high
conservation value.
The impact area of the development stretches from the Brúarjökull glacier down to the sea
at Héraðsfloi. On the one hand, there is a sparsely vegetated highland area with broad open
spaces, glacial rivers and land use limited primarily to sheep grazing and hiking. On the
other hand, there is a lowland area with the main utilisation being agricultural (Sheep,
cattle and forestry farming), along with denser centres of human settlement at Egilsstaðir
and Fellabær. As in many marginal areas, centuries of grazing, especially by sheep, but
later also reindeer, have had impact on highland vegetation.
The Kárahnjúkar hydropower project will leave a lasting impression on its impact area,
drastic changes will occur in the natural terrain and land use in areas where vegetation and
25
fauna are sensitive. Some of the main aspects of nature which will be affected or become
extinct due to the establishment of the plant are as follows:8
• Changes in the landscape occur with the formation of Hálslón reservoir, covering an area of approximately 57 km2 .
• About 32 km2 of established vegetation is submerged by Hálslón and an additional 8 km2 disappear under other structures and storage reservoirs. Valuable habitats of vegetations and small animals are lost under Hálslón.
• Structures impinge upon undisturbed open spaces in the highlands, reducing them by 925 km2 .
• Worth considering changes will occur in the flow of two large glacial rivers due to diversion. Volume increase in Lagarfljót is expected, but average flow in Jökulsá á Dal decreases significantly.
• Important Calving and grazing grounds of reindeer stock are lost and migration routes are disturbed.
• Increased blowing of dust and sand from the Hálslón shores has potential of harming vegetation on Vesturöræfi as the Kringilsárrani reserve is reduced by a quarter.
• Important geological features are inundated and scientific investigation of them becomes difficult.
• Changes will occur in the alluvial flats of Jökulsa á Dal: deposits of sediment will become somewhat vegetated and the river will be restricted to a narrower channel.(Beneficial).
• Many waterfalls, mainly in the catchment of Jökulsá í Fljótdal, will be reduced or become completely extinct.
• Sediment in Lagarfljót will increase, the river becomes darker in colour and conditions for aquatic life deteriorate.
• Some hayfields along Jökulsá í Fljótsdal and Lagarfljót will be dampened (waterlogged) by a raised groundwater table.
• The coastline of Héraðsflói will recede, but natural impact on that area is minimised.
According to Landsvirkjun,s Environmental Policy and Objectives, aimed to minimize
disturbance of the environment during construction and to achieve exemplary project
finish at the end of construction. It has been actively aimed at to show consideration for
the environment. The layout of the project has been the result of co-operation between the
8 Landsvirkjun. (May 2001). Kárahnjúkar Hydroelectric Project. Environmental Impact Assessment. Summary of Environmental Assessment Report.
26
designers and the EIA scientists as well as the conclusion of the Icelandic Government in
its ruling on the environmental impact of the project.
The structures have been designed to appear neatly in their surroundings and to cause
minimum impact to the environment. The entire construction areas, including borrow
areas, disposal areas and temporary camp and working areas, has been planned in light of
the policy9. Main environmental aspects that has been subjected to a significant impact on
the environment and compelled to be adhered to by the concerned parties with the project
are the followings:
• Borrow areas and disposal areas.
• Waste
• Hazardous substances and use of chemicals
• Dangerous substances.
• Noise and vibrations.
• Transportations
• Soil and water
• Flora and fauna. (Ecosystem).
• Air pollution
• Landscape unit
• Cultural heritage
• Education and competence.(Information).
9 Landsvirkjun (Nov 2002). Kárahnjúkar Project – Landsvirkjun’s Environmental Policy and Objectives.
27
Figure 6. Summary of the Major Environmental Impacts of the project.
Sources. Landsvirkjun website.
28
4 Rock Distribution and Geology of the Project Area.
Geologically, Iceland is a young country, built up during the later part of the Cenozoic era,
i.e. Miocene, Pliocene and Quaternary. It is almost entirely composed of lava flows and
eruptive móberg (hyaloclastites/volcanic breccias) while in between are widespread, thin
sedimentary beds. Older geological formations are characterized by igneous intrusions.
The geological formations of Iceland are divided into four main groups according to
stratigraphical age which differ considerably from one another. Oldest is the Tertiary
Basalt Formation formed in the late Tertiary period. Next in sequence of age is the Grey
Basalt Formation which was formed in the very late Pliocene and early to middle
Pleistocene and thirdly the Móberg Formation which was formed in the very late
Pleistocene. These three formations form the bedrock of the country of which the fourth
and youngest formation rests, which is made up of unconsolidated or poorly hardened
beds such as till and glaciofluvial deposits, marine and fluvial sediments and soil, as well
as volcanic tephra and lava flows. This youngest formation was formed at the end of the
Pleistocene and in the Holocene.10(Einarsson 1991)
The bedrock in the project area was formed over the past 6.5 million years, consisting
mainly of thick sequences of basalt flows with intercalated sediments and móberg
formations of various kinds and due to this, mixed conditions arises. This has made
excavation more difficult because of the variation in the strength and elasticity of the
encountered materials at the virgin face.
The basalt is classified into the following three petrographic types namely, tholeite basalt,
olivine basalt and porphyritic basalt. The accumulation rate of lava and the average period
between eruptions in the Fljótsdalur-Jökuldalur area have been determined to be about 500
m (1600 ft) per million years and 20,000 to 30,000 years respectively.11(J. Roby 2006).
Sediments in the area occur as intercalations between lava flows, as well as thick
accumulations filling depressions and old valleys.
10 Þorleifur Einarsson (1991). Geology of Iceland Rocks and Landscape. 11 Joe Roby. ( February 2006). Tunnel Bussiness Magazine. Nothern Exposures. Supplying Iceland,s New Hydropower Plant. Tunnelling Overcome Extreme Weather and Geological Conditions.
29
Móberg formations occur in the upstream part of the project as buried bodies of pillow
lava, pillow breccia, tuff breccia and tuffs.
The nature of the sediments varies with their locations within the lava-pile, in the lowest
part of the pile, most sediments are fine grained and tuffaceous. In the upper part of the
pile, the sediment intercalations indicate cold climate with deposition of conglomerates
and tillites. The thick sediments are of fluvio-glacial origin, mainly consisting of
conglomerates and sandstones. Heavy underground water inflows were encountered
during the tunnel excavation phase with peaks up to 350 l/sec.
Peter Pitts (2006), Tomasz Najder (2006). Personal discussions on site, Joe Roby (2006).12
12 Peter Pitts. Senior Geothecnical Engineer. VIJV- Viseningar Enginering Joint Venture. Kárahnjúkar Hydroelectric Project Supervision Team. Tomasz Najder. Senior Grouting Engineer. VIJV- Viseningar Enginering Joint Venture. Kárahnjúkar Hydroelectric Project Supervision Team.
30
Geology – Longitudinal section
Figure 7. Diagram Showing Geology-Longitudinal section.
31
5 Hydrogeology/Groundwater Condition in Iceland.
The precipitation is very high in Iceland, exceeding 3,000 mm/year in mountainous parts
of southern Iceland, although it may be less than 600 mm/year over wide stretches in the
northern part of the country. 13(Sigfúsdóttir, 1995 & Tómasson, 1982).
The hydrogeological nature of groundwater in Iceland is very different in the geologically
younger formations – Late Quaternary and Recent, as compared to the geologically older
formations – Tertiary and Early Quaternary. In the former ones the bedrock is highly
permeable, while active and open fissure swarms enhance the permeability as well as
creating a strong anisotropy 14(Sigurðsson & Einarsson 1988). These characteristics result
in a concentration of the groundwater flow appearing in springs or groups of springs with
an unusually high discharge and a high seasonal stability. In the older formations, Tertiary
and early to middle Quaternary the bedrock usually has a low permeability, wide open
fissure swarms are absent and the groundwater is restricted to recent sediments of a
limited extension and characterized by strong seasonal fluctuation in the discharge of the
springs. The real conditions are much more diversified beyond this simplified general
classification, but the deviations are varying and are in each case restricted to limited areas 15(Hjartarson et al., 1980). Gíslason and Eugster, 1987, observed a degree of regularity in
the chemical composition of the groundwater. The main factors are the marine component
in the precipitation, the silica increase and cations in the groundwater through reactions
with the rocks as well as the mixing with geothermal water from high-temperature areas.
Local deviations can be traced to organic compounds from peat-bogs in the lowlands and
to recent snowmelt in summer in the highlands. Due to the low population density ~3
13 Paper presented at the Nordic Hydrological Conference, Förde, 28 -30 June 1982. Orkustofnun, mimieographed report No. OS -82059/VOD -10, Reykjavík. 14 Freysteinn Sigurðsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availabilty and Demand. Jökul, No. 38, 1988. 35-54. 15Hjartason Á., Andersen L.J ., Kelsrup N and Rasmussen J. 1980: Explanatory Notes for the International Hydrogeological Map of Europe, Sheet B 2 Island. 55 pp. Gíslason S.R. and Eugster H.P. 1987: Meteoric water –basalt interaction. 11 A field study in N.E. Iceland. Geochimica et Cosmochimica Acta, 51, No. 10, p. 2841 -2855. 15b Freysteinn Sigurðsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availabilty and Demand. Jökul, No. 38, 1988. 35-54.
32
inhabitants/km2, and high precipitation ~ 2000 mm/year, freshwater has been considered
an easily available and inexpensive commodity. Nevertheless, some prime factors have
influenced this development such as the population growth and higher standard of living.
The groundwater in Iceland is generally low in chemical contents and free from pollution. b(Sigurðsson & Einarsson 1988).
33
6 The Tertiary-Early to Middle Quaternary Bedrock.
The bedrock of the Tertiary-Early to Middle Quaternary formations is predominantly built
up of stratified basaltic lava flows, with a number of central volcanoes dispersed in the
strata. The strata are usually slightly tilted in accordance with the tectonic history of
Iceland, but in some areas the tilting even exceeds ten degrees 16(Sæmundsson, 1980). The
past volcanic activity occurred in distinct, elongated volcanic systems, accompanied by
swarms of open fissures compared to the presently active zones. The fissures are now in
most cases tightly closed, but in some regions they have been rejuvenated during later
tectonic events. Some regions are characterized by new fissure zones been formed in the
course of the later tectonic development, though of different intensity as those connected
with the Mid Atlantic rift volcanism. These “young” fissure swarms represent zones of
strongly increased permeability as well as anisotropic elements in the structure of the
bedrock.
The thickness of the lava flows in the basalt sequences is variable, intercalations are
usually much thinner than the lava flows, consisting of ash layers, windblown sand and
sometimes of remnants of soils. The columnar parts of the lava flows have an effective
porosity only in the narrow fissures between the joints, and they may be nearly closed
through alteration, tightening and deformation due to the overburden. The scoriaceous
parts, especially at the contact of lava flows, have a higher effective porosity and
permeability, but they are much thinner than the columnar part. The glassy and vesicular
scoriae are more prone to alteration than the massive columns. The originally higher
permeability in the scoriaceous parts can thus be more strongly reduced than the
permeability in the massive parts. 17(Sigurðsson & Einarsson.1988).
The intercalative layers have originally had a rather high effective porosity, but their
lithological nature makes them highly susceptible to geothermal alteration, which can
reduce their permeability, until it eventually becomes negligible. In principle, the same
applies to sedimentary layers more abundant in the stratigraphically higher parts of the
16 Sæmundsson K. 1980: Outline of the Geology of Iceland. Jökull, 29, p. 7 – 28. 17 Freysteinn Sigurðsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availabilty and Demand. Jökul, No. 38, 1988.
34
sequences in particular in the Early to Middle Quaternary. In the younger formations the
rocks have not been buried as deeply as in the older ones, so the degree of alterations is
much less reduced. Parts of the older (Tertiary) formations have never been buried deep
enough to be subject to any recognized alterations and tightening.
The stratigraphical, lithological and tectonic structure of the central volcanoes is much
diversified, showing inhomogeneities and anisotropies. In these complex systems some
small-scale aquifers with higher permeability may occur, seldom of great extension and
often very irregular. Eventually, only small springs, if any, are to be found under these
hydrogeological conditions. Rejuvenated (Secondarily created fissures swarms are
regionally of importance due to the greater part they play in the geothermal hydrology of
the country).
6.1 Sedimentary Aquifers. During the last glaciation Iceland was to a large extent covered by glaciers, which carried
most of the loose sediments on the surface out to the sea. The present sedimentary cover is
therefore mostly postglacial in age. There is a difference in the most common sedimentary
aquifers between the older geological formation and the younger ones. In the Tertiary-
Early to Middle Quaternary regions the most common sediments are deposits left by the
retreating glaciers, river gravels, rockslides and a thin weathering cover. Principal aquifers
are the rockslides and the river gravels, some of the rockslides have extension of some
km2 and thickness of more than 10 m. Springs issuing from them have a discharge up to
some tens of litres per second, which is the same order of magnitude as that of other great
springs in the Tertiary areas.
18The bedrock in the older formation is classified as impermeable; spring–fed rivers are
rare, surface runoff is high and the rivers are liable to flooding and the gravel plains at the
valley bottoms correspondingly voluminous. In the lower courses of the rivers, as well as
on all the glacial rivers, the sediments are finer, reducing the permeability and making the
construction of wells difficult. It is characterized by strong seasonal variation in rivers as
18 Freysteinn Sigurðsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availabilty and Demand. Jökul, No. 38, 1988. pp 40.
35
well as in the groundwater flow in the river gravels. Seasonal changes in water
temperature are also great which has the tendency of causing some difficulties during the
winter time.
In the Late Quaternary-Recent regions spring fed rivers as well as glacial rivers are
predominant. During the snowmelt in spring – early summer after rain storms most rivers
are in flood and some dry up in between due to lack of steady supply. Potential
sedimentary aquifers may also dry up, due to the high permeability of the bedrock in the
case of this; groundwater extraction is mainly carried out in springs or rocks of high
permeability.
6.2 Late Quaternary and Recent Bedrock. These formations are also of volcanic origin and mostly basaltic in composition.
Recordable numbers of central volcanoes have been active in these younger geological
periods and some are still very active.
6.3 Pyroclastics, lavafields and fissure swarms. A known characteristic of the volcanism during the glacial is the subglacial heaping up of
fragmentary rocks, tuffs, breccias and pillow lavas, due to magma extrusion under an ice
cover and rapid chilling of the lava in the subglacial/glacial meltwater. These rocks do
appear as steep-sided mountains or as extensive layers, mostly of secondary origin. Late
Quaternary and recent bedrock are characterized by high porosity, with strong variation in
permeability. The fine grained tuffs have in most cases narrow pores and low
permeability, which could be further reduced through alteration, post volcanic or due to
later geothermal activity.
The postglacial lava flows have very high permeability (conductivity 0.001-11.0 m/s).
Similarly, to their tertiary counterpart, they have a relatively low storage coefficient (0.01-
1.0). (Sigurðsson & Einarsson.1988). In comparison to the pyroclastic rocks, the
permeability of the interglacial lava flows has often been reduced by razing off of the
scoriaceous top layers, glacial tightening, and effect of deformation or rearrangement of
joints under the glacial ice cover. In general, their permeability is less than in postglacial
lavas.
36
Permeable rocks and swarms of open fissures create strong groundwater currents in the
Late Quaternary Zone and the chemistry of the groundwater may show the influence of
geothermal activity related to the silicic centres.
Figure 8. Map showing the demarcation of Iceland based on the bedrock classifications.
(Sigurðsson and Einarsson 1988).
Explanations.
1. Tertiary and Early to Middle Quaternary basalt regions.
2. Late Quaternary rocks.
3. Quaternary silicic centre.
4. Zone of fissure swarm.
37
6.4 Great springs and springs areas. Quite a number of spring areas have a discharge in the range 5-20 m3/s which is fairly
constant over the year, seasonal fluctuation is often at lesser percentage ~ 10%. Prominent
spring areas are directly or indirectly connected with active fissure zones, recent
volcanism is accompanied by swarms or zones of more or less open fissures. The
secondary permeability forming this fissuring can be very high (hydraulic conductivity
reaching up to 0.1- 1 m/s) which is of similar value with most permeable rocks or even
higher. As the fissure zones extend in the longitudinal direction for tens of km, they have a
very strong, anisotropic “draining effect”. Though to a lesser extent, the volcanic structure
of subglacially formed pyroclastics ridges acts in the same way.
The hydrogeological characteristics of the young rock formations can be summarized as
follows:
• Very strong anisotropic effect of the fissure swarms (anisotropic coefficients 1.5-2 or even more).
• Very high permeabilities, conductivity often in the range 0.001-0.1 m/s.
• A high storage coefficient in the pyroclastic rocks (0.1-0.5), but a low one in the lavaflows (0.01-0.1).
• Very strong concentration of the groundwater flow towards preferred spring areas (yield exceeding 1m3/s).
The central volcanoes in the young, recently active formations play a similar role as their
counterparts in older formations. Often irregular in structure, strongly altered rocks and
the permeability has been drastically lowered over parts of them, especially when
compared to the higher permeabilities found outsides the central volcanoes.
High-temperature geothermal areas, link to these volcanoes served as the most important
source of geothermal heat in Iceland. Fresh water is not abundant in these areas, because
of the geothermal impacts on its chemistry and the reduction in permeability. Most of
them are also situated far away from the areas of major settlements, where other sources of
fresh water are likely more abundant and better accessible.
38
19In general summary, the geologically youngest regions of Iceland, Late Quaternary and
Recent, are abundantly provided with clean and constant groundwater. The oldest regions,
Tertiary and Early to Middle Quaternary, are on the other hand usually deficient in
affluent aquifers and subject to seasonal fluctuations in groundwater flow. Unfortunately,
many settlements with very high demand of clean and ample fresh water are situated in
these geologically older regions. On the other hand, the state of things regarding the
younger formations is encouraging, as more over 70% of the total population are
concentrated on the southwestern part of the country, with an easy access to rich
groundwater basins.
Figure 9. Map
of Iceland showing great springs from the Late Quaternary and Recent Zone, most of them
in connection with fissure swarms or lava fields, (Sigurðsson and Einarsson 1988).
1. Springs or groups of springs with yield > 5 m3/s. 2. Springs or group of springs with yield 1-5 m3/s. 3. No Perennial surface runoff.
19 Freysteinn Sigurðsson and Kristinn Einarsson (1988). Orkustofnun, National Energy Authority, Hydropower Division Reykjavik, Iceland. Groundwater Resources of Iceland- Availabilty and Demand. Jökull, No. 40 1990.
39
7 Land Use Planning.
Landsvirkjun, the owner of the Kárahnjúkar Power Project, is in full support of the land
use plan being studied by the Icelandic Government, to establish a national park or a
protected area on the north side of the Vatnajökull ice cap as a part of an extended
Vatnajökull National Park. 20Landsvirkjun believes that a power project and a national
park in the area can co-exist. Hydroelectric projects and national parks are operated side
by side in many areas of the world. They claim that the project and its operation within a
protected area can establish better conditions for the operation of a protected area.
Preparation works is on-going by the Icelandic Government for the establishment of a
national park north of Vatnajökull, mostly west and southwest of the construction area.
20 Landsvirkjun website: http://www.lv.is/
40
8 Design and Construction.
8.1 Design and Construction of the project. The Head Race Tunne (HRT) is constructed from the main reservoir all the way to the
power house, It will convey the water from the Hálsón reservoir along the 40 km long
pressure tunnel and two 400 m deep vertical shafts to the underground powerhouse in the
western flank of the Fljótsdalur valley. The power house will house 6 vertical Francis
turbines with a total installed capacity of 690 MW and a rated generating flow of 144 m3/s.
The turbine water will be released to the Jökulsá í Fljótsdal river along a 1.1 km long tail
race tunnel and a 2.1 km long canal. Most of the 40 km headrace tunnel, with diameter
between 6.8 and 7.6 m was executed by three full face Tunnel Boring Machines (TBM s).
In the case of Kárahnjúkar Head Race Tunnel, 35,459,2 km and 4,167,3 km 21(VIJV
Database 2006) were excavated by Tunnel Boring Machine and Drill and Blast
respectively. Drill and Blast drives were necessary partly as counter drives to cope with
the rather tight construction schedule, the upper section of the headrace tunnel was
executed by Drill and Blast. During excavation by conventional drill and blast method, the
work face (virgin rock) is perforated by up to 5 m long drill holes. These holes are then
filled with explosives and blasted. Full face excavation means that the excavation is
performed in one operation, on the head of the Tunnel Boring Machine several rolling
cutters are mounted. The high feed pressure of the head against the face of the tunnel,
combined with a rotation of the head, leads to the crushing of the rock at the face, thus the
excavation. Full face excavation is normally a much quicker method than conventional
drill and blast. Futhermore, the mobilization cost of the TBMs are very high, thus
requiring a long tunnel in order to make the method economical. The conventional drill
and blast method provides better flexibility in handling major weakness zones.
Three construction adits was built by the contractor along the route: Adit 1 at Teigsbjarg
(lower), Adit 2 at Axará (Middle) and Adit 3 at Glúmsstaðadalur (upper). The adits are
approximately 1.4 km, 1.7 km and 2.7 km long, these will also provide permanent access
into the tunnel during operation. Based on the schedule, preparatory construction started in
21 VIJV Database December 2006. Viseningar Enginering Joint Ventures. Kárahnjúkar Hydroelectric Project Supervision Team.
41
the fall of 2002 and the main construction in springtime 2003. Impoundment of the
Hálslón reservoir started in September 2006. Electricity from the first generating unit will
begin to flow in late summer 2007, and the project is scheduled to be fully completed in
2009.22(Landsvirkjun Publication 2003)
8.2 Excavation Methods.
8.2.1 Excavation Class/Rock Support. In the TBM excavation section of the tunnel, the rock is divided into excavation classes.
There are basically four excavation classes, namely as follows:
• Excavation Class I
• Excavation Class II.
• Excavation Classs III
• Excavation Class IV.
The excavation class assigned to a particular section of the tunnel depends on the strenght
and stability of the area. It also reveals the deterioration of the rock in the area and hence,
the amount of support to be applied there. The degree of deterioration worsens from
excavation class I to IV (I<IV). In other words, areas classified as excavation class I
requires much less rock support in comparison with area classified as excavation class IV.
Below are the required activities to be carried out under each of the excavation class
according to the design and plan.
Excavation class I requires the following to be carried out:
• Spolt bolting as required.
• Sprayed concrete at least 50 mm thick with wire mesh or steel fibres.
• Drain holes 50 mm, 0.300 mm into rock at 2,0 × 2,0 or 3,0 × 3,0 spacing.
Excavation class II. requires:
• Pattern bolting, bolt lenght 2-5m long, spacing of 1,87m 23(= 1 stroke length of TBM) unless otherwise directed.
• Sprayed concrete of 50-100mm thick with wire mesh or steel fibres.
22 Landsvirkjun Publication 2003. Kárahnjúkar Hydroelectric Project and Transmission Lines FL3 and FL4. 23 Defined as the maximum advancement length of the TBM before re-gripping and advance. (Maximum advance length).
42
Excavation class III. requires:
• Pattern bolting, bolt lenght 2-5m, spacing 1,87m (= TBM stroke length) unless otherwise directed. Bolts drilled by left and right booms may be staggered ½ stroke lengths.
• Sprayed concrete of 100-150mm thick with wire mesh or steel fibres.
• U-channels where directed and rock bolts to be adjusted accordingly.
Excavation class IV requires the following:
• Rock bolting as instructed.
• Sprayed concrete of 200 250mm thick with wire mesh or steel fibres.
• Steel ribs TH 29/48, spacing 1,28m unless otherwise directed.
• Steel lagging as required.
• Rock bolts for anchoring ribs as required or instructed.
• Full circle rib if required and approved by the engineer and designer.
8.3 Bore Classes. Bore classes are used to distinguish the type of rock face been excavated at a particular
section of the tunnel, (see figure 10). Bore classes are basically divided into namely:
• Class HI.
• Class H2
• Class H3
• Class M1
Class M2.
43
Figure 10. Showing different Bore Classes, rock types and percentage of face surface.
Source: VIJV Database 2002.
Figure 11. Map showing excavation details of part of head race tunnel, (Power-intake) and
Adit 4 junction
44
TBM 1TBM 2TBM 3Jumbo 1Jumbo 2Jumbo 3Jumbo 4Jumbo 5Jumbo 6
5 km
10 km
15 km
20 km
25 km
30 km
35 km
Adit 1
Adit 2
Adit 3Adit 4
Power intake
Surge tunnel
Drainagetunnel
Jökulsá diversion tunnelø 7,2 m
Headrace tunnelø 7,6 m
Headrace tunnelø 7,2 m
Tunnel alignment
TBM 3
TBM 2TBM 1
Figure 12. Map showing the profile of the Head race tunnel, directions of the three tunnel
boring machines (TBMs) and Jumbo Drilling Machines, (Source, Landsvirkjun Database
2006).
45
Table 1. Showing the key figures of the project.
Headrace Tunnel (from Hálslon). Cross section Circular and horse shoe. (TBM and Drill and Blast). Diameter 6.8-7.6 m Lenght 39.8 km. Headrace Tunnel (from Ufsar). Cross section circular and horse shoe. (TBM and Drill and Blast). Diameter 5.5-5.8 m Lenght 13.3 km. Varying depth 100-200 m Powerhouse Type underground. Size (L ×W× H) 115 ×14 ×34 m Turbines 6 Francis units Installed capacity 6 ×115 = 690 MW. Design net head 524 m. Discharge 144m/s Power Lines. (Transmission lines FL3 and FL4). Rated voltage 420kV. Operating voltage 220kV FL3 Transm. Capacity at 220kV 915MVA. FL4 Transm. Capacity at 220kV 1,300 MVA. Lenght FL 3 49 km Lenght FL 4 53 km Number of steel towers FL3 159 Number of steel towers FL4 166 Tailrace Tunnel. Type horseshoe ross section Diameter 9 ×9 m Lenght 1,100 m Tailrace Canal. Lenght 2,100 m. Energy poduction. Annual energy 4,450 GWh
Two high tension lines connect the Kárahnjúkar Station in the Fljótsdalur valley to the
Fjarðaál aluminium smelter at Reyðarfjörður, owned by the US Company Alcoa. Each line
is about 50 km long and they go through two separate valleys for safety reasons (Snow
avalanches). The switching station in Fljótsdalur will also be connected to the national
grid,24(Landsvirkjun Publication. 2003).
24 Landsvirkjun Publication 2003. Kárahnjúkar HEP and Transmission Lines FL3 and FL4.
46
9 Monitoring Instrumentation.
Construction of a tunnel will affect the surroundings in different ways. It is of high
importance to monitor these effects to prevent damages to the surroundings as well as the
tunnel. Different types of monitoring instruments are presented below of which most of
them has been used in the construction of Kárahnjúkar Head Race Tunnel, together with a
brief description of how they work.
9.1 Deformation monitoring. Instrumentation. Application.
Borehole extensometer Rock movements around tunnel.
Pressure cell Pressure measurement in tunnel linings
Convergence tape extensometer Movement of tunnel linings.
Increx mobile extensometer Strains and deformations around tunnel
Anchor load cell Monitoring of anchor loads.
Piezo settlement column Water level and settlement control.
Embedment strain-gauge Concrete strain measurements (shotcrete).
Electrical piezometer Control of pore- water pressure.
Settlement system Monitoring of ground settlement
Surface clinometer Structural inclination movement.
Settlement column Soil settlement above tunnel.
Smach accelerograph Dynamic measurements during excavation.
9.2 Groundwater Control. Piezometers.
The measurement of the pore-water pressure in rock mass has been carried out with the
use of Piezometers. They can be installed in boreholes, from ground surface and from the
tunnel. There are several types of Piezometers. The choice of it largely depends on desired
accuracy, cost and user-friendliness. Some of the piezometer types that have been widely
used on this project are listed below:
• Standpipe piezometer.
• Pneumatic piezometer.
• Vibrating wire piezometer.
47
Dipmeters.
Dipmeters is a portable instrument for measuring the water level in wells, standpipes and
boreholes.
Flowmeter. Measures the amount of water inflows in the tunnel.
9.3 Vibration monitoring. Control of vibration is carried out on nearby constructions:
• Heavy constructions.
• Public and Private buildings
• Other structures such as concrete structures in the tunnel, transformers, as well as other structures of importance.
48
10 Grouting into the rock.
10.1 Reasons for grouting in tunnelling. Grouting is the injection of particles suspended in a fluid medium or chemical as a
solution into rock voids mainly for rock reinforcement (support), stabilization of the tunnel
strata and water tightness. Some of the reasons behind grouting are, it support tunnelling
principle such as preventing leakages, protection of groundwater against depletion and
reduces environmental degradation. It also improve stability and durability of the tunnel as
well as leads to smooth excavation.
Excavation of tunnel involves so many risks of which most are related to encountering
unexpected ground conditions. Such risks are the chances of hitting large quantities of
high pressure ground water as well as smaller inflow of ground water which are bound to
cause some problems in the tunnel as well as the surroundings. Ground water inflow
controls remain the most frequent reason for grouting in tunnels, apart from improvement
of ground stability problems. Deteriorating rock condition mostly in weak /shear zones
causing instability problems for the tunnel excavation is another reason for grouting, poor
and unstable ground can be improved by filling discontinuities with a grout material of
sufficient strength and adhesion, i.e. rock reinforcement by grout injection. 25(Garshol
2003, Raymond 1996, J. Warner 2004, C. Kutzner 1996).
In general, ground water ingress can be controlled by drainage, pre-excavation grouting
and post-excavation grouting.
Pressure grouting in rock is executed by drilling boreholes of suitable diameter, length and
direction into the rock material, placing packers near the hole opening or alternatively,
other means of providing a pressure tight connection to the borehole, connecting a grout
conveying hose or pipe between a pump and the packer and pumping a prepared grout by
overpressure into the cracks and joints of the rock surrounding the boreholes.
10.2 Comments on Pre-grouting and Post-grouting. Grouting in tunneling could be classified into two namely:
25 Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. Raymond W. Henn. (1996) Practical Guide to Grouting of Underground Structures. James Warner (2004) Practical Handbook of grouting. Soil, Rock and Structures. Christian Kutzner (1996). Grouting of Rock and Soil.
49
• Pre-excavation grouting or Pre-grouting.
• Post-excavation grouting or Post- grouting.
• Pre-excavation grouting or pre-grouting: Involves grouting ahead of the excavation, where the boreholes are drilled from the tunnel excavation face into the virgin rock in front of the face and the grout is pumped in until refusal pressure is maintained and allows setting, before advancing the tunnel face through the injected and sealed rock volume.
The advantages of pre-grouting are:
• protection against high water inflow during excavation work,
• lowering of the groundwater table can be limited or totally prevented,
• higher grouting pressures can be used,
• strengthening of rock around the tunnel,
• less grout consumption,
• more controlled grout wandering, hence limited grout leakage to the tunnel,
• easier charging of blast or advancement of TBM stroke.
The disadvantages of pre-grouting are:
• interruption of excavation work
• blasting and deformations around the tunnel might open joints and fracture again,
• expensive if done from the surface into the tunnel,
• orienting of grouting holes might be difficult since the leaking spot is not directly seen.
Post-excavation grouting, or Post grouting: where the drilling for grout holes and
pumping in of the grout material to refusal pressure take place somewhere along the
already excavated part of the tunnel due to unacceptable water ingress beyond minimum
requirement. It is worth knowing that, post-grouting is a supplement to pre-grouting, to
seal off the remaining spot leakages if necessary most especially when the pre-grouting
has not satisfied the required average tightness within a given section of the tunnel i.e. 2
l/100 m. It is interesting to observe that post grouting is much more effective when the
same area has already been pre-injected, as the usual problem of leakage points shifting
from one tunnel location to another, without really sealing them off are in most cases
avoided. This is due to the facts that the latter process seals the open joints in the rock
50
before the water starts to flow through the dyke, discontinuities or faults, whereas with
post-grouting the water has started to flow into the tunnel and the joints have to be blocked
with the water flowing through them. One of the possible problems that have to be faced
with post-grouting is “grout wash-out”. Based on experience, it has been indicated that the
time and cost of reaching a specified result by post-grouting may be 30-60 times higher
than by pre-grouting hence, its highly recommended that in cases where large water
inrushes are expected and most especially at high ground water head, to carry out probe
drilling ahead of the face and to embark on pre-grouting if large water in flow is detected. 26(P. Tolppanen & P. Syrjanen 2003)
A few advantages of post grouting are as follows:
• does not disturb excavation work,
• easier to plan, since rock surface are visible,
• grouting can be limited to a certain area where leakage exists,
• grouting of tunnel bottom is a problem,
• lower pressure can be used,
• water usually finds a new route, and,
• Water inflows from the tunnel floor are difficult to localize.
Based on experience, it’s generally agreed that post-grouting is extremely expensive and
time consuming as well as less effective than pre-grouting. In difficult situations, it can be
close to impossible to be successful (complicated).
Other grouting activities that could be classified under post grouting activities are the
following: (Raymond W. Henn. 1996).27
• Contact grouting.
• Consolidation grouting.
• Curtain grouting.
26 P. Tolppanen and P. Syrjanen (2003). Hard Rock Tunnel Grouting Practice in Finland, Sweden and Norway – Literature Study. 27 Raymond W. Henn. (1996). Practical Guide to Grouting of Underground structures.
51
• Contact Grouting: Contact grouting involves the filling of voids between concrete
linings, cast-in-place or pre-cast, and the host geologic material. It also includes the
filling of voids behind steel and cast iron liner segments. It is also used to fill
similar voids between steel penstock lining backfill concrete and the host rock.
Some excavations in rock require an initial support using pre-cast concrete
segments; steel ribs and lagging, or shotcrete followed by a cast-in-place concrete
liner. The voids left between these two lining systems often require contact
grouting. Contact grouting is otherwise known as “backfill” or “backpack”
grouting.
Cast- in- Place Concrete Lining:
The occurrence of voids between cast-in-place concrete linings and rock is not uncommon,
particularly when concrete is placed overhead, as in tunnel crowns and chamber roofs.
Voids can also form in the lower quarter-arcs in circular tunnels containing reinforcing
steel and around curved structures such as elbows that form the transition from shafts into
tunnels.
Voids in concrete placed overhead usually occur because concrete behaves like a fluid
during placement and consolidation by vibration before it takes an initial set. In its fluid
state, concrete tends to maintain a horizontal surface; therefore, a void will form at the
high point of a pour. Voids also develop due to the presence of trapped air, a poor concrete
placement procedure, insufficient concrete slump, or unstable concrete. Obstructions to
concrete flow during placement from items such as rock bolts, mine straps, embedded
conduit and piping, or steel set and lagging, could also cause voids. A predetermined
pattern is given for hole layout, spacing, depth and refusal pressure and is shown on the
contract drawing. 28(KEJV,VIJV, Raymond. Henn 1996). Two and one hole patterns on
3m centres were used in the power intake of the Head Race Tunnel. The other technical
requirements such as mix design, injection pressure and refusal pressure (3 bars) were also
specified.
28 KEJV: Kárahnjúkar Enginering Joint Ventures. Kárahnjúkar Hydroelectric Power Project Designer. VIJV: Viseningar Joint Ventures. Kárahnjúkar Hydroelectric Power Project Supervision Team.
52
• Consolidation Grouting. Consolidation grouting involves the filling of open joints, separated bedding planes,
faulted zones, cavities, and other defects in the rock up to some distance (e.g 15 m),
usually a minimum of one tunnel diameter, beyond the excavation limits. Consolidation
grouting strengthens the foundation material and reduces the flow of groundwater into the
structure. In the case of high-pressure water tunnels, consolidation grouting minimizes the
flow of water outward through the structure’s lining into the surrounding rock after the
facility has been put into service. Moreover, water and associated increase in water
pressure in the foundation material adjacent to the structure caused by the outward flow
may be unacceptable for structural, aesthetic, and even environmental reasons.
29(Raymond H.1996, Arvind Shroff & Dhananjay L. Shah.1999).
Defects in the rock surrounding the excavation may be naturally occurring, been in
existence prior to the excavation, or the defects may have developed as a result of ground
vibrations during excavation. It is also possible for existing defects to have been worsened
by the excavation process. All excavated surfaces “relax” or move, into the opening after
the rock is removed. Movement of this nature may cause once tight joints to open and
bedding planes to move a part. Additionally, the rock beyond the excavation limits is
disturbed by forces created by the act of excavating. In general, vibrations and expanding
gas pressures caused by blasting as in the case of drill and blast will disturb the rock
surrounding the excavation to a greater degree than mechanical excavation methods do
such as tunnel-boring machine (TBM).
The contract specification enclose the requirement to perform consolidation grouting, and
the need for consolidation grouting based on the decision taken by the designer will be
based upon geotechnical data that is collected during the geological site investigation
phase of the project as well as the construction methods used. A predetermined pattern,
shown on the contract drawings, is adopted for hole layout, spacing, and depth. The
specification should also contain other technical requirements, such as mix design,
injection pressure, and refusal criteria.
29 Raymond W. Henn. (1996). Practical Guide to Grouting of Underground structures. Arvind Shroff & Dhananjay L. Shah.(1999). Grouting Technology in Tunnelling and Dam Construction. 2nd Edition.
53
In general, unit pricing is used to offset pay for consolidation grouting and items such as
the depth of holes drilled (15m), the number of set-up, and the quantity of grout mixed and
placed by volume are common pay items.
• Curtain Grouting.
Curtain grouting of tunnel is made as post-grouting. It is carried out for stabilization
purposes or as sealing of visible leaks or cracks. The grouting is done as a radially fan
pattern, either perpendicular to the centre line or in a direction crossing as many cracks
and fissure planes as possible. Since curtain grouting is done as post-grouting the utilized
grout pressure is limited (low), contrary to pre-grouting where high sealing pressure can
be adopted.
Curtain grouting is also done to prevent the tunnel becoming a drainage pipe in the rock
after a few years, when temporary filled cracks will be flushed clean again. 30(Petterson
and H. Molin 1999).
Pressure grouting into the rock mass surrounding a tunnel, is a technique that has existed
for more than 5 decades, and it has developed rapidly during the last 2 decades. Much of
the development into a high-efficiency economic procedure has taken place in
Scandinavia. Pressure grouting has been successfully carried out in arrange of rock types
ranging from weak sedimentary rocks to granitic gneisses and has proofed effective
against very high hydrostatic head as well as in shallow tunnels.
Contact, consolidation and curtain grouting were intensively carried out in the power
intake stretch of the tunnel which covers the first 1-km of the Kárahnjukar Head Race
Tunnel, these section of the tunnel are with cast in place concrete lining.
10.3 Reasons behind increase in the use of pressure grouting. Some of the rationale behind the increase in the use of pressure grouting ahead of the face
in modern tunnelling works within the last two decades are the following:
• Limits on permitted ground water drainage into tunnels are now frequently imposed by concerned authorities; this is mainly due to environmental protection and sometimes to avoid settlement above the tunnel. A typical
30 Sten-Ake Petterson and Hans Molin (1999).Grouting and Drilling for Grouting: Purposes, applications, methods and equipment with emphasis on dam and tunnel Projects. Atlas Copco Crealius AB Publication.
54
effect of settlement is damage on the surface, e.g. to infrastructure like buildings, roads, drainage pipes, supply lines and cable ducts.
• The risk of unexpectedly major water inrushes can be virtually eliminated as a result of the systematic probe drilling ahead of the face which prematurely detects the possibility of hitting major water features.
• Poor ground ahead/behind the face can be substantially improved and stabilized before or after exposure by excavation. This goes a long way to improve the stability, thus reducing the risk of uncontrolled collapse in shear zone.
• Permanent sprayed concrete tunnel linings are increasingly being adopted since the saving potential in projects cost and time is of importance, this remain one of the main reason for the increased interest in permanent lining shotcrete technology. This operation can not be effective under wet conditions hence; grouting operations might become a possible solution.
• The risk of pollution into and from the tunnel can be avoided or limited. This is because once the ground has been treated with pre-grouting coupled with post-grouting if required; it becomes less permeable for stored materials to freely ingress or egress from the tunnel.
10.4 Design of grouting in rock tunnels. Design of grouting in rock tunnels encompass the development and specification of
drilling patterns, the mix design, the grout materials to be used and the methods and
procedures to be applied during execution. These variables can be controlled by grouting
engineers, geologists or specialists; these are varied depending on the actual local
conditions in the tunnel, all aiming at achieving reliable results. The success of grouting
activities could not be accurately predicted because of the nature of the technique and the
lack of details regarding ground conditions. The indirect signs and effects on water ingress
after operations are used to asess the effectiveness of the grouting operations.
The design of tunnel grouting operations depends upon the best estimates of the average
permeability of the rock through which the tunnel is to be driven. These include
calculation of the likely water ingress, drawings showing procedure such as the depth,
angle and pattern of the intended drilling, execution procedure covering all aspects of the
operation and the materials specification targeted at satisfying the required water tightness
of the tunnel.
Some basic facts to work with in the design of grouting in rock tunnels are as follows:
55
• Once the water tighteness requirements are defined, project data and all available information about rock conditions and hydrogeology can be analyzed and compared with those requirements. This involves calculation of potential ground water ingress under different typical situations. This helps in determining the sequence of steps to take to meet the required tightness of the excavated tunnel.
• Resulting tightness in terms of water ingress achieved can be measured quite accurately. This means that it is possible to move to a reasonable comparison between targeted water ingress and the actual result and pinpoint if the situation is satisfactory or not. If the results are positive, the work will continue without changes but with close monitoring of the ingress.
• In case the result is negative, i.e. too high rate of water ingress, the information should be used to decide how to modify the situation to ensure satisfactory results in comparison to the requirements for the remaining tunnel excavation. Execution in stages may be options until a satisfactory result are achieved. Post-grouting will be embarked upon for any excavated sections of the tunnel which do not meet the requirements of the specification until the overall results are acceptable. 31(K.F. Garshol 2003).
10.5 Practical basis for injection works in tunneling. Reasons for pre-injection in tunnelling are different, and the operations can be carried out
under quite variable geological- and hydro-geological conditions. These factors contribute
to the execution of pre-injection in a given case. Some common practical facts exist when
carrying out pre-injection planning and execution.
Limited working space and logistics problems are typical in tunnelling work. In order to
keep the rate of tunnel face advance high, it is highly important that all work sequences are
as rapid as possible, with as small disturbance and variation as possible, as well as with a
smooth change from one operation to the next. This is an important factor for the cost of
the tunnel, since the time associated expenses are running whether there is face advance,
or not.
Voluminous information generated during drilling of holes and during execution of the
injection itself must be treated with care in order to avoid very complicated and time
consuming decision procedures.32(K.F Garshol 1999).
31 Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. 32 Knut. F Garshol (1999). Use of Pre-injection and Spilling in Front of hard rock TBM Excavation. Paper presented at Tenth Australian Tunnelling Conference 1999. Melbourne, Victoria 21-24 March 1999.
56
10.6 Influence of tunneling to the surrounding and the conditions inside the tunnel.
Tunnel excavation affects the immediate surrounding to some extent. This depends on the
location of the tunnel, its design and purpose, ground conditions and hydro-geological
conditions. Such influence could cause problems.
Some of the issues of importance that need evaluation are the followings:
• Location of the tunnel, especially in relation to other infrastructure, other excavation, lakes, rivers and ground water level. Most of the tunnels are below the local ground water level.
• Rock and soil cover, type and characteristics of tunnelling ground, water conductivity of the ground.
• Effects of in and out leakages on economy, environment, safety and health. It has been known that, out leakages can cause as much of damages as the other way round. The hydropower pressure conduits will loose water and electricity production and sewage may cause contamination/pollution. 33(K. F Garshol 2003).
Possible impacts of tunnel excavation on the surroundings are as follows:
• Lowering of the ground water level can cause number of effects among which are, ingress of oxygen to wood foundations which causes rotting. Some rocks, like alum shale may swell due to the creation of gypsum, causing damage to foundations and other structures. (Norwegian cases).
• Ground water resources like springs and wells may be influenced or lost and vegetation may be drastically affected and farming activities damaged.
• Out-leakages effects largely depend on the type of liquid and components in the liquid that is leaking out as well as the hydrostatic head. Some of the effects of water are splitting, jacking or washing out effects at high head and water influx at unwanted locations also at lower head. In the case of contaminated water such as sewage, hydrocarbon liquids, poisonous liquids, gases and others could lead to severe environmental problems in the surrounding.
• In general, Inflow of groundwater may cause settlement of soil deposits above the tunnel; it’s characterized with clay deposits which loose their pore pressure easily. In the case of buildings and other structures founded on clay, serious damages may result. Cases of this nature may arise at water ingress level of 1 to 5 l/min. per 100 m tunnel. (K.F. Garshol 2003).
33 Knut F. Garshol (2003). Pre- Excavation Grouting in Rock Tunnelling.
57
10.7 Condition inside the tunnel. The consequence of water ingress inside the tunnel is visible; these effects are different in
the construction phase in comparison to the operation stage.
• In the case of excavation down a decline, water tends to flow to the face and has to be pumped out. The level of acceptable quantities are usually smaller for a TBM excavation, approximately 0.5 m3/min at the face will cause problems in comparison with drill and blast (D&B) where 0.2 – 2.5 m3/min may be handled with determinant based on a number of other factors. Constant pumping of water may be required which may become an important cost factor at high pumping head/high volumes.
• Concentrated high pressure inrush, may lead to flooding and severe problems as well as delay (time loss). Conditions such as distributed water ingress and generally wet conditions will cause problems such as poor conditions for shotcrete application, concrete works, construction road works, derailing, construction phase dewatering and drainage. Water may have a high or low temperature, leading to a very poor working environment and it may also contain salt common in the case of sub-sea tunnel. Salt water produces corrosion and problems with all electrical equipment underground.
• Largely depending on rock type and quality, water can create instability, rock decomposition, and rock swelling as well as washing out.
• In the operational face of the tunnel, technical installation of different kinds will be installed such as the permanent ventilation system, electrical supply and operation systems in the case of metro tunnels. Humid conditions will over time cause corrosion and the likes of which the maintenance and repair cost may become high.
In cold climate and ventilated tunnels, water ingress can cause ice build up. This should not be allowed and has to be taken care of if it occurs. In the case of traffic tunnel, even local drips (less than 1 l/min. per 100m tunnel) of minor or no concern above the freezing point, can turn into serious problems when the frost volume is high enough.(Garshol 2003).
10.8 Issuing Of Site Instruction. Immediately after excavation, the geology of the new section is mapped and analyzed,
stating the characteristics of the rock mass and the features associated with the area. Also
considered are the results of the probe holes or core samples taken from the virgin rock
ahead of the cutter head in the case of TBM, or the blast face in drill and blast (D&B).
Observations such as the amount of water ingress and rate of penetration during drilling of
the probe hole goes a long way in guiding the engineers or specialists on the next stage of
action. In case of enormous amount of water inflow with high rate of drill rod penetration,
58
pre-grouting, if feasible cementacious type, are instructed in conjunction with rock support
depending on the excavation class. In the case of only enormous water inflow with low
rate of rod penetration/ strong bedrock, only pre-grouting often with chemicals are
instructed.
The decision is normally issued out in the form of site instruction to the contractor, which
describes and formally instructs the procedure and precautions to be followed in achieving
desired results. Information such as the orientation, diameter and depth of drilling holes,
chainages, diagram/drawing, sequence of injections, types of grout materials to be used
and equipments, general comments/precautions and closing criteria including refusal
pressure are instructed.
The results of grouting action performed are submitted for assessments including the
amount of grout intake per hole as well the effects on the amount of water ingress.
This goes a long way to determine the next line of action to be taken after settling time of
the grout perhaps to advance by few strokes in the case of TBM or one round of blast as
the case with drill and blast. 34(Tomasz Najder 2006, Peter Pitts 2006).
34 Tomasz Najder (2006). Personal Discussions. VIJV: Viseningar Joint Ventures Senior Grouting Engineer. Kárahnjúkar Hydroelectric project Supervision Team. Peter Pitts (2006). Personal Discussions. VIJV: Viseningar Joint Ventures Senior Geotechnical Engineer. Kárahnjúkar Hydroelectric project Supervision Team.
59
11 Boreholes in Rock.
Basically drilling methods can be divided into two main groups:
A. Percussive Drilling: Here the rotation of the drill bit is done during the rebound from
the rock in the bottom of the hole.
B. Rotary Drilling: Here, the rotation of the drill bit takes place while it rests against the
rock in the bottom of the hole. 35(H. Frank Eggington et al; 1996. ADITC, Petterson and
H. Molin 1999).
11.1 Top hammer percussive drilling. This is the most common drilling method in hard rock – as well as medium hard rock
tunnelling. The drill rods are attached to the drilling machine using coarse threads and the
energy from the hammer blows travel through the drill rod to the drill bit at the end.
Torque is being constantly delivered by the drilling machine for the drill rod and drill bit
rotation at certain speed which could be in the range of 80 to 160 rev. / min. with hole
diameters often above 102 mm (4 inch.) and depth limited to approximately 60 metres.
The drill rods are coupled in the case of holes length greater than the rod length (5 m.).
The commonly used borehole diameters are 51 mm, 64 mm as well as 76 mm.
Within the last two decades, the hydraulic drilling machine has completely replaced
pneumatic machines. Most of the modern hydraulic machines can penetrate at 1.5 to 2.0
m/min. even in hard granitic rock. In the case of longer grouting holes, the directional
deviation depends on a number of factors, primarily the chosen equipment and practical
procedures and secondarily on the condition of the rock. Holes drilled near horizontal have
been known with higher deviation than vertically drilled holes. By a careful and slow start
of the hole, preferably until the first drill rod length has entered into the rock and by a
slightly reduced feeder pressure, the deviation can be drastically reduced. Stiffeners could
also be applied to the first drill rod, thus further reducing the deviation. The disadvantage
with stiffeners on the drill string is the problem of ground seizing in poor ground. There is
high risk of getting the drill string stuck in the hole. 35 H. Frank Eggington et al; (1996). The Mannual of Methods, Applications and Management. 4th Edition. 1996. Australian Drilling Industry Training Committe Limited. (ADITC). Sten-Ake Petterson and Hans Molin (1999).Grouting and Drilling for Grouting: Purposes, applications, methods and equipment with emphasis on dam and tunnel Projects. Atlas Copco Crealius AB Publication
60
During the drilling of injection holes, it is important that the borehole is as circular as
possible and with the correct diameter. The packer will then have the best possible chance
to seal the hole without any problems. In order to achieve high productivity and good
economy, drilling of probe holes and injections of more than 5 m length, will require
hydraulic equipment for the handling of drill rods, including coupling and decoupling.
High productive percussive drilling requires constant water flushing for removal of the
drill cuttings, proper flushing remains very important to reduce the risk that fine materials
that may be blocking joints and cracks which are meant for effective grout penetration.
The remaining rock cuttings may also interfere with the packer seal. A secondary grinding
of particles which arises from the rotation of the couplers and the drill rods and friction
against the bore holes walls. This secondary grinding action and the risk of squeezing fines
into joints and cracks are highly reduced by sufficient water flushing. (Petterson and H.
Molin 1999).
Advantages associated with percussion drilled grout holes are:
• Much better penetration in rock than for other methods.
• Smaller and light drill rigs can be used easily moved from one hole to other on surface.
• Low drilling cost compared to rotary drilling.
• Optimization of the equipment when drilling through layers of different hardness and thickness.
Top hammers percussive drilling is the most common and generally also the least
expensive method, but it limits the hole depth and causes the greatest hole deviations
which results to increased numbers of holes and costs as well as lower quality.
11.2 Down the hole drilling. This technique is also a percusive drilling method except that, the drilling hammer works
directly on the drill bit and follows the drill bit into the borehole The drill rods serve as
feeder pressure, rotational torque and to convey the flushing medium. Due to the facts that,
the hammer blows are always directly on the drill bit, the rate of drilling is then not much
influencd by the hole lenght. The typical rotation speed is 10 to 60 rev./min. With hole
diameters of 85 mm and upward.
61
Down the hole drilling methods is not often used for drilling of injection holes in
underground works. Though, it may be considered in special cases such as if the greater
hole diameter is of benefit and a long hole with small deviation is required or if it is
necessary to use a casing for hole stabilization.
Down the hole drilling gives straighter and deeper holes, but at the same time requires a
larger hole diameter. When running at high pressures, compressors are needed.
11.3 Low speed rotary drilling. This methods works by point crushing under the rotary drill bit, due to the rotation and
axial feeding. The methods is not efficient in hard rock and the minimum diameter
necessary makes it unsuitable for injection drilling. Rotation speed will decrease with
increasing bit diameter.
Rotary cutting drilling. Rotation speed is 50-600 rev/min.
Rotary crushing drilling. Rotation speed is 50-120 rev/min
Auger drilling. (H. Frank Eggington et al; 1996).
11.4 High speed rotary drilling. (Core drilling). Core drilling is also a rotary drilling method but in this case, the drill bit is a cutting tool
and not crushing. The drill rods are steel pipes and the drill bit is a ring shaped bit with
diamonds as the cutting material. The feeding pressure and rotation torque is produced by
the drilling machine at the hole opening. The equipments are normally hydraulic and the
machine is typically powered by electricity.
Core drilling is not used for normal injection drilling, but for investigations ahead of the
tunnel face and for special case injection at greater depth. Cores of rock material that is
retrieved from the borehole are produced for inspection, laboratory testing and geological
logging. Normal hole diameters ranges from 45-56-66-76 mm as well as 86 mm.
Hole lenghts in the range of 100 to 300 m are possible, depending largely on rock
condition and equipment. Drilling capacity is usually up to 5 m/h and the deviations in the
range of 2-3% for short holes (<15 m) and around 5 % for long holes.
62
Core drilling produces round and smooth holes and typically the clogging of cracks and
joints in the rock is less (reduced), compared to percussive drilling. The cost and time
needed for core drilling is still much higher than for percussive drilling hence, it is only
used in special cases.
Some advantages of high speed rotary drilling are:
• The same equipment can be used for both investigatory and grout hole drilling.
• Continuous or intermittent exploration of the rock is possible over the whole lenght of the hole.
• Drilling can be done to greater depths.
• Straight and greater depth holes are of minimum deviation.
• Limited clogging of the rock fissures, flushing removes all cuttings from the hole.
• Wide applications in all kinds of rocks
• It is possible to use most of the power alternatives to drive the equipment.
• Good penetration speed in soft formations
• Easy packer installation into the hole
High flush water speed in the annulus does not cause cuttings to block the fissures which are to be grouted later. (Petterson and H. Molin 1999).
Requirements for grouting drillholes.
The drilling of grout holes is carried out in many different ways, depending on the
demands from the ground being dealt with, the type of grouting to be carried out, the
availability of equipment, etc.
In general, the method selected must:
• Drill a straight hole,
• Protect the hole walls from caving in,
• Produce drill cuttings of such a size, that they can be flushed out without closing the fissures in the ground or blocking the subsequent grouting,
63
• The drilling method shall be as economical as possible.36( H. Frank Eggington et al; 1996, Petterson and H. Molin 1999)
Other requirements are:
• Hole directions
• Drillhole diameters
• Hole depths
• Cleaning of grout holes
• Casing ( Rotating casing and Non-rotating casing).
Hole straightness: The straightness of a drillhole depends on several factors which include;
• the drilling equipment
• the drill feed positioning
• collaring
• cracks and fissures in the rock
• hardness of the rock
• feed thrust and bit impact
• drilling angle
• drillhole inclination
• rotation speed
• the annulus gap (area difference between drillbit and drill string).
• Unstable set up of the drill rig
• Choice of drill bit.
36 H. Frank Eggington et al; (1996). The Mannual of Methods, Applications and Management. 4th Edition. 1996. Australian Drilling Industry Training Committe Limited. (ADITC). Sten-Ake Petterson and Hans Molin (1999).Grouting and Drilling for Grouting: Purposes, applications, methods and equipment with emphasis on dam and tunnel Projects. Atlas Copco Crealius AB Publication
64
12 Typical solution example for drill and blast excavation.
Drilling for grouting (20-30m) - Grouting – Drilling for blasting (3-6m) – Blasting –
Scaling – Mucking – Bolting – Shotcreting – Control.
12.1 Probing ahead of the face. In the excavation of tunnel, all information about the virgin rock condition in the face may
be scanty though, the general average conditions may fairly be well known but cases of
local features (dyke, fault, joints or cracks) that may suddenly exposed yielding enormous
litres of high water pressure. In addition to this, the striking differences between normal
hard rock tunnelling conditions and the sudden occurrence of a major shear zone
containing swelling clay and crushed rock can be catastrophes if exposed without any
warning. In most cases, this leads to delay of production.
Probing ahead of the tunnel face by percussive drilling is one of the ways of reducing
these afore-mentioned risks. Log information gathered during drilling such as the rate of
drilling rod penetration, types of drill cuttings, colour of sludge, changes in fragmentation
of the sludge, loss or reduction of flushing water, amount of water inflows among others,
all linked to depth from the tunnel and goes a long way to predict the rock conditions as
well as the hydro-geological conditions a head of the face. Based on the observations and
the resulting interpretation, decision can be made on possible action regarding additional
drilling, execution of pre-injection, embark on few strokes of the cutter head or start of
drill and blast as the case may be.
This method remains the best for investigation of water within a reasonable time and cost
frame operating from a tunnel face. In most kind of full face Tunnel Boring Machine,
custom designed percussive drilling equipment is attached for practical use. In the case
whereby additional investigation is suggested, a core hole will produce a lot more
information and more accurate data, but it takes too much time to be used as a routine tool
and off course, more expensive. 37(Knut F. Garshol 1999, Petersson and H. Molin 1999, H.
O. Honnestad & K. F Garshol 2002, P. Tolppanen, P. Syrjanen 2003).
37Sten-Ake Petterson and Hans Molin (1999).Grouting and Drilling for Grouting: Purposes, applications, methods and equipment with emphasis on dam and tunnel Projects. Atlas Copco Crealius AB Publication.
65
In drill and blast excavation, the equipment is already there and the additional effort of
drilling some of the blasting holes as overlapping to a greater depth for probing ahead is
very common. In general, Overlapping holes are drilled (20-30m) from the face covering
the perimeter of the tunnel and extending some distance beyond the blasting holes are
used. The number of probe holes required will largely depend on the size of the tunnel,
the rock type and ground water regime as well as the potential consequences of not
detected problems. The probability of problem detection increases proportionally to the
number of holes drilled. 38(H. O. Honnestad et al; 2002, Tomasz Najder 2006, Personal
Communication on site.
In the case of sub-sea tunnels, which normally lie below rivers or lakes or somewhere with
a high risk if the rock cover is shallow, the probe drilling must be targeted at more than
just detecting water but as well as water tightening or grouting holes.
The advantages of percussive probe drilling are the low cost, the speed of execution and a
fairly high probability of detecting major serious features. Disadvantages lies in the
difficulty of interpreting variation observed except in high contrast features. (Pettersson
and H. Molin 1999, H. Frank Eggington et al; 1996).
In the case of possibility of serious problem within the probing depth and more specific
information is required, a combination including core drilling is often used though, it is
more expensive and time consuming but produce rock sample for detail inspection, where
the exact location of all features can be logged and analyze.
It is also possible most especially at the preliminary stage to use borehole radar systems,
seismic tomography, electrical resistance investigations and similar sophisticated
techniques.
In conclusion, supplies of packers will be maintained in the tunnel throughout all probing
and injection operations for immediate use to seal holes should in case high water inflows
are encountered at any time while drilling. Hans O. Honnestad, K. Garshol and R. Dimmock (2002). Pre-injection in hard tunnels using Degussa Rheocem Microcement. Degussa International UGC Publication. 38 Hans O. Honnestad, K. Garshol and R. Dimmock (2002). Pre-injection in hard tunnels using Degussa Rheocem Microcement. Degussa International UGC Publication Tomasz Najder 2006. VIJV Senior Geotechnical Enginner. Personal Communication on site.
66
Figure 13a. Multi-boom drilling jumbo in operation for blasting and grouting holes.
Figure 13b. Sandvik Tamrock Axera T12 and Atlas copco Rocket Boomer WL3 C drilling
jumbos. At right, a rod changer system for the drilling jumbo (Garshol, 1999).
67
12.2 Drilling of injection holes. Length, Orientation and Number of Holes: All injection holes generally have a length of
about 20 to 30m for cement grouting or 15m for polyurethane grouting but the length may
be adjusted as necessary to meet specific requirements. All holes shall be drilled at an
angle to the tunnel direction to permit a treated grout curtain that fulfils the requirements.
There are situations with very dominating joint orientations that may call for an adapted
borehole direction.
Generally, the spacing of injection holes will be between 1 and 1.5m around the perimeter
of the tunnel. Additional holes may be drilled out from the centre of the tunnel if
considered necessary to achieve either waterproofing or face stability in zones of
weakness. All injection holes are reference numbered.
12.3 Cleaning of holes/ Flushing of boreholes for injection. The first requirement is to provide efficient water flushing during drilling of the hole. The
water pressure shall be at the maximum specified by the drilling equipment manufacturer
and shall be ensured by a special pressure booster on the drilling jumbo. In areas of rocks
with clay filled fissures/joints, carefully supervised flushing operations remain very
important.
Further cleaning of the injection holes may be prescribed as either a combination of water
and compressed air or by high pressure water cleaning. If there are zones in the borehole
that may collapse if soaked in water or if the water ingress from the hole is larger than 10
l/min then, the flushing may be omitted.
Flushing by water and compressed air should be done by a stiff plastic hose using water at
10 bar pressure, combined with some compressed air. Push the hose to the bottom of the
hole, open up the water and air and withdraw the hose while flushing is on. Flushing of
boreholes for grouting should be done as a routine matter and any necessary deviations
must be decided by the supervisor based on the borehole records. (Tomasz Najder 2006
Pers. Comm., Knut F. Garshol 1999).
68
12.4 Placement of packer. The packer is normally placed near to the borehole opening and the hole is injected over
its entire length in one single step. The packer placement depth shall be between 1.5 and
3.0m depth into the borehole, adjusted to the ground conditions and locations providing a
good sealing. However, allowance must be made for a number of possible situations that
may require a different packer placement. It is clearly important that packers are installed
with great care in order to provide effective seals within the drill holes sufficient to
withstand the subsequent grout pressure.
High ground water pressure and very poor rock may speed-up a face failure and the
appropriate action is to place the packer at greater depth (5 m). It happens sometimes that
a fissures causes water and grout backflow to the face and that the packer must be placed
at a depth larger than the depth of intersection between the borehole and this fissure. In
some cases, the borehole is locally disturbed by weak rock material and local wedge
fallout causing the packer to slide or to leak. Placing it deeper is normally the corrective
measure to solve the problem. In principle, there should be an overlap of tight rock (a
buffer either from sound rock or grouted rock from the previous injection round) of
between 5- 7 m in front of the face. Packer placement lies in this zone. 39(Hotter, K G et al,
1996, H. O. Honnestad & K. F Garshol 2002).
12.5 Water Testing/Water Pressure Testing. (WPT) Some of the rationales behind conducting water tests in connection with grouting are:
• To test the ground for stability and the need for strengthening.
• To test the leakage of water and locate the leakage route, to ascertain consequences that this may have on the surrounding structures as well as to evaluate the influence distance.
• The outcome of the afore-mentioned goes a long way to investigate how to stop the leakage of water.
39 Holter K.G et al, 1996. Tunnelling through a sand zone: Ground Treatment experiences from the Bjoroy sub sea road tunnel, Proceeding of North American Tunnelling 1996. Ed: Ozdemir, pp 249-256, Vol 1 (AA Balkema Rotterdam). Hans O. Honnestad, K. Garshol and R. Dimmock (2002). Pre-injection in hard tunnels using Degussa Rheocem Microcement. Degussa International UGC Publication
69
• Another important reason is to decide which type of grout material, method and the refusal pressure to be used.
It is advisable to combined water tests with geological and geotechnical investigations,
such as MWD (measurement while drilling) and core drilling.
Different methods to execute water tests are:
1. Natural flow from an area (Seepage test).
2. Natural flow into or out from a hole (Seepage or permeability test).
3. Pump out test (Permeability test,)
4. Pump in tests (Permeability test, WPT)
Test 1 and 2 involve free running water, test 3 and 4 are done under pressure.
Water Pressure Test (WPT) is otherwise known as “Water Loss Test” in some countries.
WPT´s are done in situ by sealing off the opening to a drill hole and pumping in a
measured volume of water under constant pressure. The flow is a measurement of the
permeability of the ground, but does not directly give a K-value for the permeability. The
result is presented as Lu (for Lugeon, after Maurice Lugeon). One lugeon is the water
intake of 1 litre per metre of drillhole length per minute at a pressure of 1 MPa (10 bars)
above the existing hydrostatic pressure head. 40(Heuer, R.E. 1995, E. Friedrich- Karl 1994,
Water Resources Commission N.S.W 1980. Grouting Manual 3rd edition, Houlsby, A.C.
1976.)
Test hole diameter is mostly in the 40 -100 mm range, but is not decisive for the
permeability test. Before starting a WPT, the hole should always be flushed clean.
Interconnection of adjacent holes with test hole should be monitored and the flow through
of water. For identification purposes, paint is used as an additive to the water pumped in.
40 Heuer, R.E. (1995). Estimating Rock Tunnel water inflow. Proceedings, Rapid Excavation and Tunnelling Conference, Society for Mining, Metallurgy, and Exploration, Inc., Cushing-Malloy, Inc., Ann Arbor, Mich., 41-60. Ewart, Friedrich-Karl, (1994). Rock Permeability and Groutability Related to Dams and Reservoirs, lecture notes, University of Paderborn, Germany, September 1994. Water Resources Commission N.S.W Australia., (1980). Grouting Manual 3rd edition, 1980. Houlsby, A.C. (1976)“ Routine Interpretation of the lugeon Water Test”, Quarterly Journal of Engineering Geology, Vol. 9, The Geological Society, 1976.
70
Water pressure testing of boreholes should not be considered as a routine activity. The
long time spent in comparison to the information value produced is the main reason for
this. There is no good reason to invest time and money in such measurement.
12.6 Choice of injection materials. In drill and blast excavation, cement grouting is the original and in general the most
common method in controlling the water inflow. In the case of Kárahnjúkar Head race
tunnel excavation, Ordinary Portland Cement (OPC) of different mix design and largest
granule size of about 125 µm were used. The thickness of the mixes varies from one
another based on the water cement content (W/C) and the typical settling time (bleeding
period) for these mixes is 2 hours under 5o-20o C, allowing work to proceed without any
delay. The range of commonly used and adopted mix design are Mix 211-05 W/C 0,5,
Mix 208-05 W/C 0,67, Mix 203-05 W/C 0,45. (VIJV- Database 2006).
Depending on the ground conditions and the required level of maximum ground water
leakage into the underground openings, chemical grout may be considered. Inflow of
ground water through joints and cracks in the face or elsewhere may cause problem of
grout washout and backflow, such problem can be solved by injection of quick foaming
polyurethane which can be used as a temporary flow blockage
12.7 Injection pressure and procedure. The decision criteria for injection to be carried out must be specified. This is in most cases
based on measured water in-leakages from the probe holes and can be a given number of
l/min from a single hole or a maximum sum leakage from all the probe holes. In general,
the balance between these criteria and the target tunnel tightness must be based on
experience and local rock conditions, with the option of feedback from actual results
during operation.
Maximum injection pressure has to be evaluated on a running basis and especially it has to
be checked against local conditions in the tunnel. Very poor rock conditions in the face
area, high hydrostatic water pressure and existing backflow will be indicators that
maximum pressure must be limited, even if the rock cover is hundreds of meters away.
71
In pre-injection, the maximum allowed pressure should be used from the beginning of
injection or alternatively until one of the following occurs:
1) No more grout is accepted by the ground at maximum allowed pumping pressure.
11) The maximum specified grout quantity for the hole has been reached, regardless of
pressure but re-grouting of the hole after settling time until refusal pressure is achieved
becomes paramount.
Allowed maximum grouting pressure should be at least 50 bars above the static ground
water head; unless there are special reasons identified requiring a lower maximum
pressure. In pressure sensitive situations, it should be recognized that the danger of
causing damage by lifting, splitting or other deformation is linked to effect of pressure and
quantity rather than to pressure alone. 41(Barton, N and Quadros, E., 2004, Arvind Shroff
& Dhananjay L. Shah.1999, Knut F. Garshol 2003, Sjostrom O.A 2003).
Injection procedure:
1) Start injection of the lowest hole in the face and work upwards. (Bottom-up approach).
11) A hole is finished when the maximum specified grout quantity per hole is reached and
the refusal pressure attained.
111) if backflow of grout and water into the tunnel is detected, this should be minimize by
reducing the pump output and adding accelerator to create a blockage of the backflow.
1V) if during the injection process two or more holes become connected as indicated by
grout backflow through the hole, close the packer in the connected hole and continue
grouting the current hole. The maximum volume of grout to be pumped before the stop
criteria are reached shall be multiplied by the number of connected holes. If the maximum
pressure is reached before the maximum quantity, then the connected holes shall be
injected as well, if they take any grout.
V) All holes that do not require injection shall be filled with a stable cement grout.
41 Barton, N and Quadros, E., (2004) “ Improved understanding of high-pressure grouting effects for tunnels in hard rock” ISRM 2003 – Technology road map for rock mechanics, South African Institute for Mining and Metallurgy. Arvind Shroff & Dhananjay L. Shah. (1999). Grouting Technology in Tunneling and Dam Construction. 2nd Edition. Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. Sjostrom O.A. (2003). Principles and Ground water Control through Pregrouting in Rock Tunnels.
72
12.8 Injection records and quality control during injection. Records of the injections data have to be taken as a matter of routine; most of these may
be automatic by computerized recording or otherwise, there must be well prepared forms
to be used in the tunnel during work progress. It must be well defined who is responsible
for the record keeping. As a minimum, the following information must be recorded:
1) General data like tunnel chainage, date, time and shift, person who does the recording,
identification and location of all holes, measured water flow from the holes.
11) Per hole: packer placement location, type of packer, length of hole, grout mix design,
pressure at start and end, time at start and end, total grout quantity, any leakages
(backflow) and any connections to other holes.
111) Results of quality control holes.
The following quality control tests shall be carried out:
• Cup tests to ascertain setting time and bleeding.
• Marsh cone tests to indicate viscosity
The frequency of the tests will be determined by the site supervisor.
12.9 Settling of grout, time until next blast. The typical setting time for Ordinary Portland Cement (OPC) grout with additives is 2-3
hours ensuring smooth running of work. As the water pressure increase and the fissure
openings gets larger the risk of grout material failure and wash out will increase. It is
difficult to give general rules on how this situation is evaluated, but the consequences of a
failure, time required for setting, the water pressure and fissure size in the ground have to
be considered.
In the case of accelerator been used to shorten setting time, this off course is a good help
for the actually accelerated grout, but based on experience, it has been proofed that only a
part of the grout in one injection stage is normally accelerated. If the next planned activity
is drilling of boreholes for control of the injection results or for a next round of grout
holes, always start drilling in the area where the previous injection was terminated (To
allow maximum setting time). (Knut F. Garshol 2003)
73
12.10 Drilling of control holes. Control holes must be evaluated to assess the efficiency of any injection stage. Control
holes are drilled on both sides of all holes that yielded water flow above the injection
criteria. Re-injection may be necessary depending on the result of testing in the control
holes and compliance with the specified criteria.
Re-injection is carried out using a pattern of holes alternating with the previously injected
holes. This is followed by further control holes and further injection stages as necessary,
until the specified criteria are achieved.
Limited post excavation grouting may be embarked on later if necessary to achieve the
specified inflow requirements in local areas of the tunnel if it is not possible to achieve the
criteria in particular areas by pre-injection alone.
74
13 Typical solution example for Tunnel Boring Machine (TBM).
13.1 Probing a head of the face. Most of the full face Tunnel Boring Machine is equipped with custom designed
percussive drilling equipment practically used for drilling grouting holes and probe hole a
head of the cutter head. One of the visible disadvantages of this attached percussive
drilling equipment is the in-ability to drill holes close to the invert i.e. below 8 o, clock
and 5 o, clock. Overlapping probe holes are maintained during excavation in order to
ascertain the state of the virgin rock ahead of the cutter head and also to guide in taking
precautionary approach against the chances of encountering high water pressure or shear
zones which may lead to temporary delay in production.
Figure 14. Pre-grouting drilling system on a Robbins TBM (Garshol, 1999).
13.2 Drilling of injection holes. Drilling a head of a hard rock Tunnel boring machine is difficult because of the very
limited available space close to the tunnel face. The TBM occupies almost all the volume
for the first roughly 15 minutes. Typically the angle drilling is between 15o and 20o
depending on the arrangement of the drill rigs on the TBM. The length, orientation and
number of injection holes are specified and reference numbered. 42(H. O. Honnestad & K.
F. Garshol 2002, K. F. Garshol 1999, Schunnesson H. 1996).
42 Knut F. Garshol (1999). Use of Pre-injection and Spiling in Front of Hard Rock TBM Excavation. Paper Presented at Tenth Australian Tunnelling Conference 1999. Melbourne, Victoria 21-24 march 1999.
75
13.3 Cleaning of holes/ Flushing of boreholes for injection. Constant flushing of the injection holes while drilling must be ensured. The effect of
neglecting this can lead to rapid blockage of the intersected water bearing channels meant
for grout. Sludge and rock cuttings from the drilling process are forced into the openings
by the injection material as well as the pumping pressure.
13.4 Placement of packer. The packer placement depth shall be between 1.5 and 2m ahead of the TBM face. The
depth will depend on the angle of the hole and length of the cutter head. For example for a
hole collared 4m behind the cutter head face in an open TBM the packer should be at least
6m into the hole. However, allowance must be made for a number of possible situations
that may require a different packer placement. It is clearly important that packers are
installed with great care in order to provide effective seals within the drill holes sufficient
to withstand the subsequent grout pressure.
13.5 Water Pressure Testing. Water pressure test is not required in TBM excavation and not normally been carried out.
It found more application in consolidation and curtain grouting to determine the lugeon
and the procedure as well as the stages of grouting that may be required.
13.6 Choice of injection materials. The choice of injection materials largely depends on the amount of inflow from the probe
holes or the total amount of leakages from the face. In general, high inflow of water
requires a very fast reaction grout which makes chemical grout more effective. On the
other hand, cement grout with accelerator could also perform the same function by
catalyzing the setting rate as well as enhancing the stability of the zones in the case of
shear zones. In conclusion, inflow per hole greater than 5 l/sec or 15 l/sec from 2 holes
requires very fast reaction grout such as polyurethane.
Hans O. Honnestad, K. Garshol and R. Dimmock (2002). Pre-injection in hard tunnels using Degussa Rheocem Microcement. Degussa International UGC Publication Schunnesson, H., (1996). “ Probing ahead of the face with percussive drilling” Tunnels and Tunnelling, January 1996, pp 22-23.
76
13.7 Injection pressure and Procedure. In pre-injection, the maximum allowed pressure (50-110 bars) should be used from the
beginning of injection or alternatively until one of the following occurs:
I ) No more grout is accepted by the ground at maximum allowed pumping pressure.
11) The maximum specified grout quantity for the hole has been reached.
Other closing criteria are that injection should be interrupted in case of:
• Intensive leaking of non- reacted polyurethane (chemical) into the tunnel.
• The refusal pressure of approximately 110 bars is reached.
• The expanded polyurethane grout fractures the rock close to the surface.
• The packer is not set tightly in its position, but moving towards the mouth of the hole.43(Degussa UGC Europe, Tomasz Najder Personal site discussions, VIJV Database 2006)
Injection procedure:
1) Start injection of the lowest hole in the face and work upwards. (Bottom-up approach).
II) A hole is finished when the maximum specified grout quantity per hole is reached and
the refusal pressure attained.
III) if backflow of grout and water into the tunnel is detected, this should be minimize by
reducing the pump output and adding accelerator to create a blockage of the backflow.
IV) if during the injection process two or more holes become connected as indicated by
grout backflow through the hole, close the packer in the connected hole and continue
grouting the current hole. The maximum volume of grout to be pumped before the stop
criteria are reached shall be multiplied by the number of connected holes. If the maximum
pressure is reached before the maximum quantity, then the connected holes shall be
injected as well, if they take any grout.
V) All holes that do not require injection shall be filled with a stable cement grout.
43 Degussa (BASF) UGC Europe (2006). Practical injection course UGC Europe Region Hagerbach, Switzerland, 4-5. May 2006. Tomasz Najder (2006). Senior Grouting Engineer. Viseningar Joint Ventures. Kárahnjúkar Hydroelectric Power Supervision Team. VIJV Database (2006). Viseningar Joint Ventures. Kárahnjúkar Hydroelectric Power Supervision Team.
77
Alternatively, two holes could be drilled consecutively, often opposite to each other and
grout until the required number of holes is completed.
13.8 Injection records and quality control during injection. Adequate injection data should be undertaken as a matter of routine; General information
such as:
• Tunnel chainage, date, time and shift, person who does the recording, identification and location of all holes, measured water flow from the holes.
• Per hole: packer placement location, type of packer, length of hole, grout mix design, pressure at start and end, time at start and end, total grout quantity, any leakages (backflow) and any connections to other holes.
• Results of quality control holes.
13.9 Setting of grout/ Hardening time. The setting time of Ordinary Portland Cement (OPC) grout is normally 2-3 hours and the
hardening time could be between 8-10 or even 2 days partly depending on the local
condition. Addition of additives and admixtures (accelerator) such as cemsil, sodium
silicates and industrial salts drastically reduced the setting time in order to avoid delay to
work.
In the case of chemical grouting such as polyurethane, its fast reaction time of 10-15
minutes and hardening time of maximum 5-8 hours further reduces the time waste and
hence, ensure smooth production and advancement of the Tunnel Boring Machine.
13.10 Drilling of control holes. Control holes must be evaluated to assess the efficiency of any injection stage. Control
holes are drilled on both sides of all holes that yielded water flow above the injection
criteria. Re-injection may be necessary depending on the result of testing in the control
holes and compliance with the specified criteria.
Limited post excavation grouting may be embarked on later if necessary to achieve the
specified inflow requirements in local areas of the tunnel if it is not possible to achieve the
criteria in particular areas by pre-injection alone. Further decision on additional grouting
78
or the advancement of the Tunnel Boring Machine could then be taken. 44(Knut F. Garshol
2003, H. O. Honnestad & K. F. Garshol 2002).
44 Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. Hans O. Honnestad, K. Garshol and R. Dimmock (2002). Pre-injection in hard tunnels using Degussa Rheocem Microcement. Degussa International UGC Publication
79
14 Packers.
Packers are mechanical tools used to ensure a tight connection (seal) between the pumping
hose and the borehole drilled into the rock formation or soil for the purpose of injecting a
grout at high pressure. They are devices inserted into a grout hole that expands
mechanically or by inflation to restrict the flow of grout to a specific part of the grout hole.
A typical packer consists of a pipe with a coupling in the end as well as an elastic
expander that can be inserted into the hole and expanded against the borehole wall. The
expander will anchor the packer in place so that the injection pressure is not forcing it out
of the hole and also seal off the pressurized section of the borehole from the tunnel side.
The injection pump hose is hooked up to the pipe the pump can be put on.
Different packer types exists, most manufacturers produce in working principle the same
types of packers though, the quality, sizes and technical details may be different. The
ground condition, availability, price and a number of other factors help in selecting
packers for individual project.45(Raymond H.1996, P. Tolppanen & P. Syrjanen 2003,
James Warner 2004, Knut F. Garshol 2003, C. Kutzner 1996).
14.1 Mechanical Packers (expanders). Mechanical packers are re-usable types which are commonly used on construction sites.
The diameter of the rubber expander has to be in a certain relation to the borehole
diameter and the maximum expansion available. Mechanical packers are available in
different standard lengths, commonly from 1.0 m to 5.0 m in steps of 0, 5 m. Users can
also produce their own pipes locally and any suitable length. Connectors are used to join
standard pipe lengths for deep packer placements.
A ball valve or similar parts is normally fitted to the end of the packer-pipe, after the
completion of injection the ball valve is closed and the pump hose disconnected.
45 Raymond W. Henn. (1996) Practical Guide to Grouting of Underground Structures. James Warner (2004) Practical Handbook of grouting. Soil, Rock and Structures. Christian Kutzner (1996). Grouting of Rock and Soil. P. Tolppanen and P. Syrjanen (2003). Hard Rock Tunnel Grouting Practice in Finland, Sweden and Norway – Literature Study.
80
The valve remain closed with the packer in place until the completion of setting time to
keep the ground water pressure then, the packer may be removed and cleaned for re-use in
a different hole. Mechanical packer is applicable for both cement and polyurethane grout.
Pressurized grout would flow back into the tunnel without closing the valve. (Raymond
W.Henn. (1996).
14.2 Disposable packers. (Single application). The working principles of disposable packers are the same to re-usable counterpart
though, they are constructed so that when expanded, the expansion is automatically locked
to allow removing the inner- and outer pipes used to place the packer and expand it. The
packer has a one-way valve to keep pressurized grout in place without backflow when
releasing the pump pressure and removing the pipes.
The Mechanical packers and the disposable ones are not very effective in weak ground,
proper sealing without backflow may be difficult, they may start sliding in the hole and
sometimes they get stuck in the wrong position. 46(BASF (Degussa) Construction
Chemicals,)
46 BASF (Degussa) (2007) Construction Chemicals Europe AG.
81
Figure 15a. Showing disposable packers (With a distance of 100-150mm between the
rubber sleeves). Source BASF Construction Chemicals.
14.3 Hydraulic packers. Inflated packers are otherwise known as the hydraulic packers, the inflation is carried out
through high water pressure supplied through a separate thin line from the tunnel to the
packer location. The packer is handled through a single pipe, which is also the grouting
pipe after expansion. Hydraulic packers have a much wider expansion range and they will
seal better (lengthwise). In other words, hydraulic packers are more expensive. They are
very practical for Water pressure testing in long holes because they are very quick to
expand, deflate and remove and the low risk of backflow around them as well as the good
sealing properties in poor ground are very favourable. (Bimbar 2004, Knut F. Garshol
2003,)
A combination of a hydraulic packer and a disposable self-locking packer can be very
useful and economical under very weak conditions. The hydraulic packer allows
expansion of the disposable packer mounted in the front of it, and the hydraulic packer
82
helps to prevent backflow and sliding during injection. After completion of grout setting
time, the hydraulic packer is removed and the disposable one remains in place.
Figure 15b. Showing inflatable packers. (Source Bimbar 2004).
83
Figure 15c. Showing Inflatable Packers (Source Hany AG Equipment).
14.4 Standpipes techniques. Under poor ground condition, the placement of packer may become very difficult and the
borehole stability may also be at risk, this is very common in shear zones and highly
broken ground in hard rock tunneling. Combination of this condition with high water
pressure often leads to loss of stability and collapse. The drilling of long holes in this kind
of condition becomes a problem as the drill string could easily get stuck and will sometime
break in the bore hole.
One of the best ways of dealing with serious problems of this nature is to use the so called
stand-pipe technique. Using an oversize drill bit of 76 mm diameter to drill to a depth of 3-
4 m, then a steel pipe of suitable diameter (55 mm, 66 mm) could be easily inserted into
the hole, the pipe is then grouted with a fast setting cement grout. Once the grout has set
the pipe can be used to extend the hole by a smaller diameter hole (51 mm) diameter drill
bit. In case of any serious drilling problems of any kind, a packer is placed safely and
tightly in the steel pipe and the drill part can be injected. After completion of grout setting,
the drilling may be resumed for another step of bore hole deepening. The process is
repeated as required until a desire or instructed depth is reached. Plastic pipes could be
84
used to replace the steel pipes as described above as it does not cause any problems for
Tunnel boring machine or Road header excavators.
14.5 Tube-a-manchet. This type of methods is not commonly used in rock injection underground but found its
wide application in soil injection (mostly vertical holes). The principle is based on a
hydraulic double packer been inserted into a pre-grouted sleeve pipe which is surrounded
by a weak mortar often called a mantel grout, which is a simple cement grout with
relatively high content of Bentonite clay. The sleeve pipe has rubber sleeves (non-return
valves) at fixed distance and these valves can be activated individually by injection
pressure between the double packers. The mantel grout is designed so weak that it will
split from the injection pressure and grout can flow into the ground without any leakages
along the bore hole. The packer can then be moved as needed and a given valve can be
grouted many times. Due to the sub-horizontal orientations of most injection holes in the
tunnel, the use of mantel grout around the sleeve pipe for proper filling could be non-
effective as it results to un-wanted grout leakage along the hole. 47( Jaroslav Verfel,
1989, Raymond W. Henn,1996).
14.6 Types used on the Project. • Mechanical packers of different lengths and diameters.
• Disposable packers.
• Inflatable packers of different lengths and diameters. (Bimbar packers ½ m/ 1 m
long.)
47 Jaroslav Verfel (1989). Rock Grouting and Diaphragm Wall Construction. Development in Geotechnical Engineering 55. Elsevier Science Publishers. Raymond W. Henn. (1996). Practical Guide to Grouting of Underground structures.
85
15 Cement-Based Grouts.
15.1 Basic properties of cement grout.
15.2 Cement particle size, fineness. The two important parameters governing the permeation capability of cement are the
particle size and particle size distribution. The average particle size can be expressed as
the specific surface of all cement particles in a given quantity. The finer the grinding, the
higher is the specific surface, or Blaine value (m2/kg).
In determining Blaine value, the particle size distribution may vary and the important
factor is the maximum particle size, or as commonly expressed the d95. The d95 gives the
sieve dimension where 95% of the cement particles will pass through and the remaining
5% of the particle population is larger than this dimension. The maximum particle size
should be small in order to avoid premature blockage of fine openings, caused by jamming
of the coarsest particles and filter creation in narrow spots. 48(Houlsby, A.C. 1990, James
W. 2004, K. Bowen R. 1980, C. Kutzner 1996, K.F. Garshol 2003, Houlsby A.C 1983).
In general, cement with the highest Blaine value will normally be the most expensive, due
to finer grinding and will also give better penetration into fine cracks and openings. From
an injection perspective, this cement is characterized with the following basic properties:
• A highly ground cement with small particle size, will bind more water than a coarse cement. The risk of bleeding (water separation) in a suspension created from a fine cement counterpart is lower and a filled opening will remain more completely filled.
• The finer cements have a quicker hydration and a higher final strength but cause the disadvantages of shorter open time in the equipment. A high temperature increases the potential problems of clogging of lines and valves. Mixing of fine cements must be closely controlled, to prevent heat
48 Houlsby, A.C., (1990). Construction and design of cement grouting, a guide to grouting in rock foundations, John Wiley and Sons, New York, 1990. James Warner (2004). Practical Handbook of Grouting. Soil, Rock and Structures. John Wiley & Sons, Inc. Robert Bowen (1981). Grouting in Engineering Practice. 2nd edition. Applied Science Publishers LTD. London. Christian Kutzner (1996). Grouting of Rock and Soil. A.A Balkema /Rotterdam/ Bookfield/ 1996. Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. MBT International Underground Construction Group. Division of MBT (Switzerland) Ltd., 2003. Houlsby A.C. (1983). Cement Grouting - A Compilation of Recent Papers. Water Resources Commission., NSW, Australia.
86
development caused by the friction in the high shear mixer, and hence even quicker setting.
The effectiveness of these characteristics largely depends on the efficiency of the mixing
process enough to separate the individual particles and properly wet them. In pure and
finer cement with the addition of water suspension, the tendency of particle flocculation is
high after mixing which is against effective production.
The effects of the water reducing admixture (or dispersing admixture) when mixing
cement could be seen on catalyzing the specific surface or Blaine value (m2/kg). Other
great effect of water reducing admixtures is the lowering of viscosity at a fixed w/c-ratio.
The importance of lowering water content is to improve final strength of grout, as well as
lowering permeability and ensure better chemical stability. More important in the cases
with higher water content is that the permeability is so high and the strength is low that
any flow of water can lead to mechanical erosion and chemical leaching out of hydroxides
(hydration products from cement reacting with water). (K.F. Garshol 2003).
Figure 16. Dispersing effect of an admixture using Micro-cement.
87
15.3 Bentonite etc. Bentonites are used to reduce the bleeding of grouts and a dosage of 3 to 5% of the cement
weight has a strong stabilizing effect. Its composition is clay containing some percentage
of clay minerals of the smectite group (usually montmorillonite) characterized by its large
volume increase on wetting.
Because of strong tendency of standard cement to separate when suspended in water,
enhanced by the normal use of water cement ratio greater than 1.0, bentonite has been
traditionally used constantly to reduce this effect and in combination with cement for
grouting of rock and soil.
Bentonite is natural clay from volcanic ashes and the main mineral is montmorillonite. It is
hydrophilic or water swelling (Brady and Clauser 1986, K.F. Garshol 2003).
Basically, two types of bentonite exist:
Sodium- bentonite (Na -)
Calcium bentonite (Ca -)
Sodium- bentonite (Na -): Because of its high stabilizing and swelling characteristic
(between 10 and 25) times the original dry volume when mixed in water, it has been
commonly used as an additive in cement grouts.
15.4 Rheological behaviour of cement grout. Cement mixed in water as a stable paste or an unstable suspension in terms of water
separation behaves according to Bingham’s Law. On the other hand, water and true liquids
have flow behaviour according to Newton’s Law. These laws are illustrated as follows:
Bingham,s Law: T = c + n dv/dx
Newton,s Law: T= n dv/dx.
Where: T= flow shear resistance (Pa).
n = Viscosity (Pa s)
dv/dx = shear velocity (s-1)
c = cohesion (Pa)
88
Figure 17. Rehological behaviour of Newton and Bingham fluids.
A cement suspension demonstrates some cohesion which has to be overcome for any flow
to be initiated as compared to the liquid counterpart. If the internal friction is negligible,
the paste will then behave in a similar manner as liquid. The rheological parameters of
cement suspensions can be affected by w/c- ratio, by chemical admixtures, by bentonite
clay and by other mineral fillers.
15.5 Pressure stability of cement grout and use of high injection pressure. The efficiency of a grouting operation is very difficult to realize since the whole operation
is carried out below ground level. In other words, it can be very well controlled by
(I) Controlling the properties of the grout,
(II) Controlling grout pressure and
(III) Controlling the rate of consumption of the grout.
Some of the precautions to be taken before starting grouting are:
(I) Ensure that the correct borehole and circulation line is connected.
(II) Grout ingredients such as clay, bentonite, additives etc are available in ample quantity
near the mixing plant and ready for use.
(III) Pump, agitator, mixers etc are working properly and have sufficient fuel.
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(IV) Standpipe fittings and pressure gauges are operating properly
(V) Spare standby equipment is available and anything foreseeable which might interrupt
grouting. (Arvind Shroff & Dhananjay L. Shah.1999).
Refusal Criteria.
Each grout should continue until refusal is reached except
(I) When severe leakages is observed on the surface away from the hole or along the
grout pipe.
(II) When the consumption reaches a predetermined limit
(III) When there is no grout absorption at the limiting pressure specified.
(IV) When an undesirable trend develops indicating bulb formation.
The grouting operation may then be resumed after 24 hours and pressure regulated
carefully to avoid repetition of the phenomena leading to the stoppage of the grouting.
Refusal is considered to have been reached when the intact grout at the desired limiting
pressure is less than 2 litre/minute averaged over a period of 10 minutes and the pressure
greater than 3.5 kg/cm2, or 1 litre/minute when the pressure is below 3.5 kg/cm2.
Time – Pressure – Consumption Curves.
During injection the flow rate of a properly mixed grout should be constantly monitored
and plotted against the grout pressure to understand what is happening below ground. The
pressure at the grout pipe is often much less because of friction losses that occur in long
lengths of hose leading to the pipe, coupled with the existence of a large pressure drop at
all fittings and manifolds.
Typical response curves of flow rate versus pressure and flow rate versus time has been
studied and by means of these curves, one can follow closely how the rock has responded
to the grout being injected 49(Koerner, 1985, Mistry 1988).
49 Koerner, R.M. et al. (1985). Acoustic emission monitoring of grout movement issues in dam grouting. Proceeding session Geotechnical Engineering. Division of ASCE. American Society of Civil Engineer., New York, pp. 202-206. Mistry, J.F. (1988). Important Aspect of River Valley Projects. Mahajan Book Distributors, Amdavad, Gujarat, India, Vol. 2. Arvind Shroff & Dhananjay L. Shah. (1999). Grouting Technology in Tunneling and Dam Construction. 2nd Edition.
90
Based on close observation of the grouting procedure in the project, the behaviour of holes
could be classified into six different categories depending on time- consumption
(litre/minute), and time pressure data. (kg/cm2). (Arvind Shroff & Dhananjay L.
Shah.1999, Jefferies M.G et al., 1982, Baker, C. et al., 1982). These six pairs of standard
curves (A1, A2, B1, B2 and C1, C2) of which each pair indicates certain conclusions are
defined and interpreted below:
Interpretation of curves.
(1) Figure 18. (A1). Consumption drops and pressure remains constant after peak.
This pair of curves shows filling of cracks with the grout material. After reaching the
maximum pressure, a few minor cracks might open out which are again filled up. This is
clear from the curves beyond the peak, i.e. the pressure curve is approximately parallel to
the time axis while the consumption at the nearly constant pumping energy goes on
decreasing. This is the ideal curve and in such case, the procedure should be continued and
grouting completed at the specified pressure until refusal.
Jefferies M.G et al. (1982). Electronic monitoring of grouting. Proceeding. Conference. Grouting in Geotechnical. Engineering., New Orleans. W.H. Baker (ed). American Society of Civil Enginner. New York, pp 769 -780. Baker, C. et al., (1982). Use of grouting in caisson construction. Proceeding Conference. Grouting for Geotechnical Engineering. New Orleans. W.H Baker (ed). American Society of Civil Enginnering. New York, pp. 874-891.
91
(11). Figure 18 (A2). Pressure increases and consumption remain constant after Peak.
This pair of curves shows that the pressure slightly falls due to opening of the cracks.
After filling the crack, the pressure rises and the rate of consumption remain constant. If
the rate of consumption is within permissible limits, the grouting could be stopped. If the
pressure achieved is more than specified, the operation may be continued at a suitably
reduced pump speed. This is also an ideal curve (Scenario) and, in such cases, the grouting
should be continued at the specified pressure until refusal.
92
(111). Figure 18 (B1). Pressure and consumption remains constant after Peak.
This pair of curves indicates that at the peak pressure, the grout continues to travel in the
cracks unchecked. Thus, there is no control on the extent of the curtain. In this case the
grouting operation may be stopped after injecting a certain quantity of grout. The hole
should be left for a while to allow injected grout to set; grouting may then be resumed
after a lapse of about 48 hours, when the cracks will have been partly sealed by the setting
grout.
93
(1V). Figure 18 (B2). Pressure and consumption rise to some value and then fall rapidly.
The pair of curves indicates the opening out of new cracks at peak pressure. On filling of
these new cracks, the pressure and consumptions remain constant throughout, indicating
that at the second peak pressure the grout travel is continuing uncontrolled. In this case
also, the grouting operation may be stopped after injecting a certain quantity of grout (200-
300 kg of cement depending on the length). The hole should be washed thoroughly and the
injected grout allowed setting. Grouting may then be resumed after a lapse of about 48
hours.
94
(V). Figure 18 (C1). After a rise in pressure to a certain value, the pressure remains
more or less constant, while the rate of consumption goes on rising.
This pairs of curves reveals abnormal behaviour in the hole. They indicate that there may
be leakage of grout through natural strata or along the hole or some upheaval in the rock
strata. Grouting should therefore be carried out under strict supervision. Possible spots of
leakage should be located by close inspection around the area and interconnection of grout
holes, if any should also be looked for. If leakage is noticed or if the grout consumption
exceeds maximum grout take (75 kg of grout per metre depth of hole), the grouting should
be stopped and resumed only after a lapse of about 48 hours. Immediately on locating the
leakage point, it should be plugged by excavating a small pit around it and filling this with
lean cement concrete. If there is any interconnected hole through which the grout leaks,
the hole should be immediately washed by pumping water under adequate pressure so that
it does not get chocked before regular grouting is taken up in that hole. Upheaval gauges
should be closely observed to see if there has been any upheaval; if an upheaval is noticed,
grouting should be stopped forthwith. The grout hole should be thoroughly washed and
grouting resumed after about 48 hours with pressure sufficiently reduced so as not to cause
any upheaval. Sometimes grouting can be continued at a lower pressure if upheaval
remains within allowable limits. (Mistry 1988).
95
VI. Figure 18. (C2). After reaching a certain value, the pressure quickly drops while the
rate of consumption remains constant.
This pair of curves also shows abnormal behaviour of the hole. They indicate that there
may be leakages of grout through natural strata or along the hole or some upheaval in the
rock strata. Grouting should therefore be carried out under strict supervision. The
procedure as outlined for case (V) above should be followed for this case (VI) also.
15.6 Grout setting characteristics and durability of cement injection in rock. Important for the quality and durability of cement grouts is the w/c-ratio and perhaps the
grout are stable or segregating. Stable grouts and also w/c- ratio below a certain limit has
been achievable with the aid of modern grouting technology.
Stable or almost stable suspensions contain less excess water than unstable one. Hence,
grouts with low water content are advantageous in the following ways:
-During grouting.
• the risk is minimized that expelled water will damage the partially set grout.
• grouting time is shortened because little excess water has to be expelled.
• the reach and the volume of grout can be closely delineated.
96
• effective filling of joints, including branches.
• higher density, leads to better removal of joint water and less mixing at the grouting front. 50(Houlsby, A.C. 1990, K.F. Garshol 2003).
-After hardening.
• greater strength.
• Better durability.
• Better adhesion to joint wall
• Low permeability.
15.7 Additives and admixtures for cement injection. There are cases where unexpected backflow can occur through the face or even further
back in the tunnel, possible indications are that a borehole is in contact with extremely
large channels with a lot of high pressure water. In these situations, it is beneficial to
accelerate the cement setting and hardening by stopping the backflow and allowing further
injection of the ground without loss of material as well as, by stopping unnecessary spread
of grout outside of a reasonable distance from the tunnel. Additives and admixtures are
used in cement and non-cement-based grouts to modify their fluid and set characteristics 51(Naudts et al; 2003, Naudts 2000, P. Tolppanen & P. Syrjanen 2003).
Cement-based grout additives can be admixtures, bentonite, mineral additives, or
pozzolan, such as blast furnance slag or silica fume. Admixtures used for cementitious
grouts are are, for example:
• Plasticizers and superplasticizer to reduce the water-cement ratio.
• Accelerators to prevent grouts from leaking into the tunnel or ground surface
• Additives to reduce bleeding and shrinkage. e.g. bentonite or silica based products.
50 Houlsby, A.C., (1990). Construction and design of cement grouting, a guide to grouting in rock foundations, John Wiley and Sons, New York, 1990. Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling. MBT International Underground Construction Group. Division of MBT (Switzerland) Ltd., 2003. 51 Naudts et al., (2003). Additives and Admixtures in Cement-based Grouts. Naudts et al., (2000). “New On –site Wet milling Technology for the preparation of Ultrafine Cement-based Grout” Canadian Geotechnical Conference. Montreal. October. P. Tolppanen and P. Syrjanen (2003). Hard Rock Tunnel Grouting Practice in Finland, Sweden and Norway – Literature Study.
97
• expanding additives.
• Retarders to slow hydration.
Some of the additives used are:
• Slag
• Fly-ash
• Natural Pozzolan
• Trass: Clay-phyllosilicate
• Silica fume
• Bentonite.
• Locally available fillers.
15.8 Equipments for Cement injection.
15.8.1 Mixing equipment. (Mixer). Cement and water together with additives are fed into a mixer. Several mixer types exist;
• Normal Paddle mixers which are simple to use and rather cheap but the mixing results are not good enough for high quality grouting.
• A Turbo mixer is more suitable than a paddle mixer. A centrifugal pump circulates the grout at a high speed (1300-1400 rpm, max. 1435 rpm) in the turbo-mixing container and creates a sharing action between the fractions for good quality mixing.
• Colloidal mixer: This type of mixer produces the best result. In the colloidal type mixer shear forces are also created in the mixer housing. The shear forces may be created by high turbulence in the casing as in the case of Hany system or the shear forces are created by close tolerance between the impeller and casing 52(Pettersson & Molin, 1999, Houlsby, A.C. 1990).
The mixing time and mixing speed are important factors influencing the grout quality. A
typical mixing time for OPC is 4-5 min. The finer cements require more intense mixing.
The maximum batch size is normally 80 % of the container volume. In a colloidal mixer,
the temperature might increase several degrees due to the energy release of the shear force
52 Sten-Ake Petterson and Hans Molin (1999).Grouting and Drilling for Grouting: Purposes, applications, methods and equipment with emphasis on dam and tunnel Projects. Atlas Copco Crealius AB Publication. Houlsby, A.C., (1990). “Construction and design of cement grouting, a guide to grouting in rock foundations, John Wiley and Sons, New York, 1990.
98
braking. This might induce early hardening of the grout and should be controlled by the
agitator (Pettersson & Molin, 1999). Wearing of the mixer should as well be put in
control.
Figure 19. Mixer. 53(Photo Hany AG Equipment).
15.8.2 Agitator. In order to keep the grout at a low viscosity and to prevent sedimentation, the grout should
always be agitated. The agitator acts as a holding tank with grout ready for grouting.
Inside the slowly revolving agitator, the grout suspensions are homogenized and possible
air bubbles removed. It is normally twice the size of the mixer and rotates at
approximately 60 rpm. It is highly recommended to attach one agitator per pump and per
grout mixture.
53 Hany AG Equipment. http://www.haeny.com/geraete.html.
99
Figure 20. Agitator (Photo Hany AG Equipment).
15.8.3 Grout Pumps Market wise, two types of pumps for grouting are available;
• The progressive cavity pump (pump without valves).
• Piston pumps (the valve type pump).
The pump flow and pressure capacity must be sufficient to perform a satisfying grout
operation; these parameters should also be controllable and individually adjustable during
grouting work. The maximum practical pressures needed in tunnel grouting is 10 MPa
(100 bar). The grout flow should be high enough to avoid separation of the grout. The
maximum grain size for the pump is usually varies between 3-8 mm, and should be
checked prior to grouting work.
100
Figure 21. Grout Pump. (Maximum Pressure 100 bar).
Figure 22. Grout Pump. ( Maximum Pressure 200 bar). (Photo Hany AG Equipment).
101
15.8.4 Complete systems. This is a complete unit, a mixer, agitator, pump and even a data logger. Here, standardized
components are individually mounted onto the container or stationary plants.
Figure 23. Complete grout pump. (Photo Hany AG Equipment).
102
15.8.5 Automated mixing and grouting plants • Excellent mixing quality
• High capacity
• Accurate dosing
• Consumption record
Figure 24. Automated mixing and grouting plants. (Photo Hany AG Equipment).
103
15.8.6 Recording of grouting data. These are an automatic / computerized logging tool. An auto logger is easily use with
experience of an Excel and PC.
The logged parameters are typically flow, pressure, volume, time, real time and hole
number. All parameters are shown in real time and display, and stored on a PC card. The
logged data are store and import into standard programs for viewing and printouts.
15.8.7 HIR • Simple Recording System with chart
• Recorder for pressure and flow.
• Totalizer for accumulated flow.
• Automatic switch off of grout pump when
• Reaching the pre-adjusted pressure or
quantity
• Solid stainless steel housing (IP 65).
Grout recorder.
15.8.8 HFR • Recording System with PC software
• Recording of data of up to 8 grout lines
Simultaneously.
• Recording of data on PCMCIA Memory
Card.
• Real-Time recording on PC
• Graphical display of P+Q curves
• Flow recording by inductive flow meters or
Pump strokes.
Figure 25. Showing Recording Systems. (Photo Hany AG Equipment).
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16 Chemical Grouts.
16.1 General. Chemical grouts are injected into voids as solutions, in contrast to cementitious grouts,
which are suspensions of particles in a fluid medium. Chemical grouts react after a
predetermined time to form a solid, semisolid, or gel. The process of injecting a
chemically reactive solution that behaves as a fluid but reacts after a predetermined time to
form either a solid, semisolid, or gel is known as chemical grouting.
Chemical grouts were developed in response to a need to develop strength and control
water flow in geologic units where the pore sizes in the rock or soil units were too small to
allow the introduction of conventional Portland-cement suspensions.
Chemical grouting technology has expanded with the addition of organic polymer
solutions and additives that can control the strength and setting characteristics of the
injected liquid. 54(EM 1110-3500. 1995, Karol R.H 1990, Karol R.H. 1983).
Commonly used chemical grouts in construction industry today include the following:
• Water reactive polyurethane.
• Silicates grouts.
• Acrylic grouts.
• Epoxy Resins.
• Bitumen (Asphalt).
Some chemical and mechanical properties that need to be evaluated in the selection of a
grout for a particular application are the following:
• Viscosity.
• Durability.
• Gel time.
• Sensitivity.
54 EM 1110-3500. (1995). EM 1110-1-3500. Engineer Manual. The U.S Army Corps of Engineers. Engineering and Design –Grouting Technology (1995). Geotechnical and Materials Branch, Engineering Division, Directorate of Civil Works (CECW-EG). Karol R.H (1990). Chemical Grouts and their properties. Marcel Dekker. Inc. New York. Karol R.H. (1983). Chemical Grouting. Marcel Decker Inc., New York.
105
• Toxicity.
• Durability.
(EM 1110-1-3500. 95).
16.2 Overview of Polyurethane grouts. Water reactive polyurethane grouts were introduced into the grouting industry during the
late sixties by the Takenaka Company in Japan under the trade name TACSS. In the 1980,
exclusive rights for TACSS was taken over as the successes in stopping major leaks in
tunnels democratize the traditional sodium silicate cement combinations as well as the
seepage control grouting (acrylamide grouting). By the mid-eighties there were more than
10 manufactureres of polyurethane grouts, coupled with era of custom-made formulations
tailored to specific project 55(Naudts, 2003). Due to the longevity problems faced by one
component water reactive hydrophilic polyurethanes, the two- component polyurethane
foams were gradually introduced into geotechnical industry and later became popular.
Polyurethane grouts are probably the most popular type of solution grouts (Chemical
grouts). It has been known to be used for close to four decades and have contributed to the
democratization of the grouting industry. Solution grouts pertains to grouts that behave
like Newtonian fluids, contrary to suspension grouts which behave like Binghamian fluids.
Solution grouts are injectable into very fine apertures, not accessible to suspension grouts
(even microfine).
Polyurethane grouts are made up of a large family of solution grouts hence, it is very
difficult to make generic statements for the whole group. Most polyurethane grouts are
non-evolutive or true solution grouts. A true solution grouts is characterised by a flat
viscosity curve, followed by a sudden increase in viscosity immediately prior to gelation
or curing. Others known example of typical true solutions are acrylamide, acrylate, epoxy
resins and silicates grouts.(Naudts, 2003).
55 Naudts A. (2003). Irreversible Changes in the Grouting Industry Caused by Polyurethane Grouting: An overview of 30 years of polyurethane grouting. Geotechnical Special Publication 2003 Issue 120; Vol 2, page 1266 – 1280.
106
Basically, polyurethane grouts are categorised into three, based on the recognition by the
construction industry:
• One component foaming grouts (Water reactive polyurethane grouts).
• Two component foaming grouts (Polyol – isocyanides combination)
• Two component polyurethane elastomers.
Water reactive polyurethane grouts is further sub-divided to two main sub-categories-
• Hydrophobic polyurethane prepolymer grouts (resins): They react with water but
repel it after the final (cured) product has been formed.
• Hydrophilic polyurethane prepolymer grouts (resins): They react with water but
continue to physically absorb it after the chemical reaction has been completed.
The aforementioned sub-categories and the two other categories are further divided into
sub-families of polyurethane grouts.
The sensitivity which defines the ability to control factors influencing the reaction or
curing pattern during the reaction of most polyurethane grouts is typically lower than the
sensitivity of cement- based suspension grouts. The longevity of the end product and its
chemical resistance are typically far superior to the cement-based counterpart (Naudts,
1990). The toxicity of most of the polyurethane grouts has been reduced to an acceptable
level as an approval for use in potable water application has been obtained for several
polyurethane categories of grouts despite some ill-conceived applications in the past. 56(Naudts, 2003, EM 1110-1-3500. 1995, Skanska Environmental Report 1997).
56Naudts A. (2003). Irreversible Changes in the Grouting Industry Caused by Polyurethane Grouting: An overview of 30 years of polyurethane grouting. Geotechnical Special Publication 2003 Issue 120; Vol 2, page 1266 – 1280. EM 1110-1-3500. (1995). EM 1110-1-3500. Engineer Manual. The U.S Army Corps of Engineers. Engineering and Design –Grouting Technology (1995). Geotechnical and Materials Branch, Engineering Division, Directorate of Civil Works (CECW-EG). Skanska AB (1997). Environmental Report. Sweden.
107
One component foaming grouts (Water reactive polyurethane grouts).
Polyurethane prepolymer grouts are characterised with the ability to react with the in-situ
groundwater to create a foam or gel that is either hydrophobic or hydrophilic. They are
component products using the underground water as a reaction partner to create the end-
product. The catalyst such as a tertiary amine is excluded as a components since it only
affects the rate and the direction of the polymer forming process 57(Hepburn, 1992,
Anderson 1998). Additional catalyst only speeds up the gelation process.
During this exothermic reaction, the hydrophobic polyurethanes expand and penetrate
pervious media: fine cracks as narrow as 8 micron as well as soil with a permeability
coefficient as low as 10-4 cm/s. The reaction and penetration is enhanced by the formation
of CO2 . Their penetrability is determined by their viscosity and reaction time, It has been
revealed that, the penetration of fine aperture is a very slow process hence, requires
multiple hole grouting to obtain an economically justifiable and technically sound
grouting operation. ( Anderson 1998). Before gelation, the viscosity of the grout
decreases due to the formation of of C02 and during gelation, the viscosity increases
substantially to prevents futher penetration. The formation of either a gel or foam during
the reaction depending largely on the amount of water they are being mixed with is the
end product.
Water-reactive prepolymer are high molecular grouting materials, primarily produced by
mixing a polyol with an excessive amount of poly-isocyanides to form a low prepolymeric
compound containing some free OCN groups. The injected resin is made up of this
prepolymer, plasticizer, diluting agents, surfactants and the amine catalyst.
The mechanism of reaction among the isocyanides, polyol and other components remain
complicated. In simple terms classification, the following happens:
• The reaction between the isocyanides and the polyol yields a pre-polyurethane.
57 Hepburn, C. (1992) Polyurethane elastomers. Elsevier Applied Science. London, New-York. Anderson, H. (1998) Chemical Rock Grouting. An experimental study on polyurethane Foams: Chalmers University of Technology, Goteborg. S-41296.
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• The reaction of poly-isocyanides with water liberates carbon dioxide and urea
derivatives
• The reaction of poly-isocyanides with ureido develops molecular links and high
molecular formation.
These reactions occur because of the existence of the free OCN groups in the grout, which
react with the compounds containing active hydrogen atoms, such as hydroxyl, water,
amino and ureido. The hydrogen atoms move to link up with the nitrogen atoms of the
poly-isocyanides and form high molecular polymers.
Two-component Polyurethane Foam Grouts.
The result of the reaction between a polyol (R-OH) and an isocyanate (RI-NCO) known to
be toxic is the formation of polyurethane. Largely depending on the type of polyol,
blowing agents, catalysts, a wide variety of foams with different characteristics can be
formed, only differing in: density, cellular structure, compressive strength, reaction
pattern, water absorption and fire resistance.
The isocyanate has a high affinity for water and thus has a tendency to steal the
isocyanate, leaving not enough isocyanate for the polyol to form a complete reaction.
These products react with the in-situ water and expand during the exothermic chemical
reaction to release carbon dioxide. They are totally stable after reaction with limited
flexibility. The reaction time can typically be adjusted between 45 seconds up to one hour
by adding a tertiary amine-based catalyst, (Accelerator).
Figure 26a. illustrates a typical reaction curve for water reactive polyurethane
prepolymers. First there is an induction time during which the viscosity of the water and
polyurethane mix remains constants followed by the reaction time, during which the
foaming starts. The latter is associated with a decrease in viscosity followed by a very
rapid increase in viscosity before final gelation. 58(Naudts 2003, Karol R.H. 2003).
58 Naudts A. (2003). Irreversible Changes in the Grouting Industry Caused by Polyurethane Grouting: An overview of 30 years of polyurethane grouting. Geotechnical Special Publication 2003 Issue 120; Vol 2, page 1266 – 1280. Karol R.H. (2003). Chemical Grouting and Soil Stabilization. 3rd Edition. Revised and Expanded. Marcel Dekker – Taylor and Francis CRE.
109
Figure 26a. Showing typical reactive polyurethane prepolymers. (Naudts 2003).
Figure 26b. Water reactive prepolymers. (Naudts 2003).
In order to start the reaction, there is a minimum enthalpy required, which is higher, as the
pressure is higher. This indicates that, at a given pressure, if the enthalpy is too low
(temperature too low) the reaction does not start unless the products are mixed thoroughly
with the water. (Naudts 2003). Heating the grout (maximum 60oC but never directly
applying the heat to the grout containers) speeds up the rate of reaction and hence,
eliminates the possibility of grout wash-out under most circumstances, (Figure 26 b).
Above the reaction line the reaction takes place, and below this line reaction does not
occur unless mechanical agitation of in-situ water and polyurethane grout exists. The
110
turbulent flow of the urethane-water mixture through cracks and channels often provides
for enough in-situ mixing. Some of the recently developed polyurethane resins react
regardless of pressure and groundwater temperature. (e.g. Minova Carbo-Tech products).
During grouting, the carbon dioxide generated during the chemical reaction will generate
additional pressure, as the grout flows through the cracks and pore channels, pushing the
grouts into very fine cracks and crevices.
The long term durability of grouts has been deeply looked into 59(Jakubowicz, 1992;
Naudts, 1990) which revealed that in highly acidic environments; polyurethane is
anticipated not to break down more than 10% over a 100 year time period. In an alkaline
environment (such as in concrete or in limestone formations), a weight loss of 35% has
been projected over a 45 years period due to hydrolytic degradation of the cured
polyurethane. Soils grouted with polyurethane do not display signs of long- term
degradation. 60(Oshita, et al.1991; Takenaka 1976,; Landry, et al. 2000). Hence, water
reactive polyurethane grouts do not last “for ever” in highly alkaline environments but are
very durable in neutral and acidic environments.
Typical applications of two components polyurethane grouts include: stabilization of
unstable rock in mines; sealing previous formation in front of flood bulkheads; post
grouting; sealing gaps and joints around ventilation doors, etc.
59 Jakubowicz. I. (1992). Determination of long term properties for polymer based grouts. Conference proceedings. Nordiskt Symposium I berg injektering. Ed. Lundblom. Chalmers University of Technology, Goteborg. Pp 50-56. Naudts, A. (1990). “Research Project Fort Cady Minerals: Chemical resistance of modified cement-based grout and polyurethane grouts Accelerated ageing test” Unpublished. 60Oshita, T., Kitano, M., and Terashima, K. (1991) “ Long Term Durability of Sails Solidified with Hydrophobic Polyisccyanate type grout “Sociey of Materials Science. Japan. December. Pp.1552-1557. Takenaka (1976). Taccs Know-how books. Internal Publications. Landry, E, Lees, D, and Naudts A. (2000): “New Developments in Rock and Soil Grouting: Design and Evaluation” Geotechnical News. September 2000. pp 38-44.
111
Figure 27. Core showing infilling of rock cavities with polyurethane.
16.3 Adopted Products on Kárahnjúkar Head Race Tunnel. Cementitious grouts, which were the first to be injected historically continue to be the
most frequently used on the project. The two chemical grouts adopted for the grouting
activities in the Kárahnjúkar project for control of high water ingress as well as tunnel
stabilization are the following:
(a) BASF Construction Chemicals Products. (MEYCO MP 355 A3).
MEYCO MP 355 A3 is a robust two-component (A and B), solvent-free polyurethane
injection resin specifically designed for filling big voids and jointed rock, stabilization of
loose rock, and cutting of running water. The foaming reaction time is significantly
dependent on the temperature of the Polyurethane resin, the rock and the ground water.
Fields of application are:
• Stabilization of fractured rock, sands and gravels and land-fill materials.
• Void filling
• Repair of concrete structures.
• Control of high volume water structure.
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Features and benefits of MEYCO MP 355 A3 according to manufacturers are:
• When in contact with water, the product forms rigid foam. Without the presence of
water, it forms a stiff, rubber- like material.
• Ability to react also in dry condition to cured. (With or without water).
• Sealing of running water.
• Robust rigid foam.
• Modification of the reaction can be achieved by addition of accelerator and
thixotropic agent to Component A. 61(Technical Data Sheet MEYCO MP 355/A3).
Packaging.
MEYCO MP 355/A3 is available in the following packaging:
Component A: 25kg cans or 200kg drums
Component B: 30kg drums or 240 kg drums.
Application and procedure.
Components A and B are delivered ready- to use. They are injected in the proportion of
1:1 by volume using a two component injection pump equipped with a static in-line mixer
nozzle. MEYCO MP 355 A3 Accelerator 10 and 15 are the catalyst often used.
(b) Minova CarboTech Chemical products. (CARBOPUR WFA).
CARBOPUR WFA is an extremely fast-reacting two- component- injection resin(A and
B), CFC-free and halogen free, characterize with foaming action when water added,
without water, hardens without foaming. It ensured permanent sealing against high levels
of water inflows in pits, shafts and tunnel sections.
The resin is design for:
• Sealing and consolidation in water-bearing strata.
• Sealing against strong water ingress (also seawater)
• Sealing against water under pressure e.g. from strata, dams or shaft walls.
• Stabilization and sealing works in tunnels
• Repair of old shafts and tunnels
61 Technical Data Sheet MEYCO MP 355/A3. Degussa (BASF) Construction Chemicals (Europe) AG.
113
• Stabilization of crown abutments in tunneling.
• Sealing of anchoring of sheet pilings etc in ground water.
Features and benefits of CARBOPUR WFA over others are:
• Very fast setting.
• Used for wide ranging injections.
• Stabilizing effect.
• Compatible with groundwater. B-Component slightly hazardous to water.
• Applicable at temperature between -25 and 30˚C, recommended for sealing water
at low temperature.62( Technical Data Sheet CarboPur WFA)
Packaging.
Component A 20 kg in a tin can (blue cover) or 200 kg in a drum.
Component B 24 kg in a tin can (black cover) or 240 kg in a drum.
(c ) Composition and Properties.
CarboPur WFA, Component A honey in colour is a mixture of various polyols and
additives which reacts with the B-component dark brown in colour to form a tough /hard
polyurethane resin. CraboPur, component B is a polyisocyanate on the basis of 4,4-
diphenylmethane diisocyanate (MDI).
The mixed resin penetrates the structures to be sealed. Most part of water in there is
displaced due to hydrophobicity and the viscosity of the resin. Traces of water make the
resin foam. Based on its contact with water, the resin foams up more or less. Thus, the
mechanical properties vary a lot. The cured resin is resistant against many acids, alkali and
salt brines as well as organic solvents.
According to the manufacturers, there are no objections against the use of CarboPur WFA
in respect of groundwater and drinking water hygiene and very compatible with concrete
and steel.
62 Technical Data Sheet CarboPur WFA. Minova CarboTech GmbH. Tunnel and Civil Engineering.
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(d) Processing.
The two components are pumped by a two component pump at the volumetric ratio of 1:1,
they are mixed thoroughly in a static mixer unit prior to injection into the strata through a
packer installed in a previously drilled borehole. In contact with water, the resin then
foams up. When the reaction does not meet any more water it hardens without foaming to
form a pore-free material thus, a water-tight shell is formed which in turn is surrounded by
a zone consolidated by foamed-up polyurethane. This leads to a permanent sealing and
consolidation. (Minova Carbo Tech GmbH).
(e) CarboAdd Fast.
This is an accelerator for the reaction of polyurethane and silicates resins. It is a solution
of catalyst which accelerates the polyol-isocyanate-reaction. Prior to use CarboAdd Fast is
added in small amounts (0-2 %) to the bright polyol component and mixed thoroughly.
16.4 Silicate grouts. Because of silicates grouts low cost and low viscosity, sodium silicates have found its
application in soil and rock grouting for stabilization or for ground water control for
decades. It has PH of 10.5 to 11.5 making it to be quite aggressive. The working safety
and health of using silicates has been of major concern.
Liquid silicates (water glass) are produced by dissolving vitreous silicate in water at high
temperature (900o C) and high pressure. The liquid is later diluted by water to reach a
viscosity level that can be used for injection purposes in soil and fine cracks in rock. A
normal injection grout will have a viscosity of about 5 cP and the gel produced is water
rich which is weak and to some extent unstable.
The formation of gel in the ground is normally associated with release of water from the
gel and shrinkage. Due to the low gel strength, it will have limited resistance to ground
water pressure, most especially in relatively large cracks and joints.
The liquid silicates need hardener to create a gel. Acids and acidic salts (like sodium
bicarbonate, sodium aluminates) will cause such gel-formation. In recent time, chemical
systems such as methyl and ethyl di-esters are mostly used as hardener which exhibits
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much better practical properties by improving the quality of the final grout. (Knut F.
Garshol 2003).
Because of the effects that characterize gel formation of silicates grouts in the ground,
such as release of water from the gel and some shrinkage, silicates grouting as a
permanent ground water control could not be effective as the effects can lead to new
leakage channels over time. The chemical stability also remained questionable in many
cases. Silicates grouting for temporary ground water control will mostly be the best
practical application. In rock injection, it is best advice, to ensure that cement injection to
fill up the larger channels precede silicates grout application.
The durability of silicate grouts in low PH cement- environment also remains at stake as it
is unfavorable. 63(Arvind Shroff & Dhananjay L. Shah.1999).
16.5 Acrylic grouts. Acrylates were introduced as less toxic alternatives to the toxic acrylamide compounds
that are no longer available as grout.
The acrylic grouts came into application about 5 decades ago, but for cost implications,
these were based on acrylamide. The toxic nature of acrylamide products over the years
has drastically prevented them from being used. In the Hallandsåsen tunnel excavation
Southern Sweden, addition of acrylamide and n-methylolacrylamide used as acrylic grout
in ground water control caused ground water pollution and poisoning of livestock, 64(Skanska AB Environmental Report 1997). In other words, this component could be
excluded from an acrylic grout.
Polyacrylates are gels formed in a polymerization reaction after mixing acrylic monomers
with an accelerator in aqueous solution. Polymerized polyacrylates are not dangerous for
human health and the environment. In contrast, the primary substances (monomers) can be
very dangerous before their complete polymerization. Injection materials polymerize very
quickly, sometimes before the monomers completely polymerize, a considerable amount
can be diluted by the ground water and depending on the porosity, permeability and local 63 Arvind Shroff & Dhananjay L. Shah. (1999). Grouting Technology in Tunneling and Dam Construction. 2nd Edition. 64 Environmental Report (1997). Skanska AB. Sweden.
116
geology of the area, and the action may lead to large scale contamination and pollution.
The working safety of personnel is at risk and the use of products containing acrylamide is
carcinogenic with cumulative effect in human body. (K. Garshol 2003).
Acrylics gel materials are very useful for injection into soil and rock with predominantly
fine cracks. It is usually injected with less than 20% monomer concentration in water and
the product viscosity is therefore as low as 4 to 5cP. The gel-time can typically be chosen
between seconds and up to an hour.
The strength of the gel will primarily depend on the concentration of monomer dissolved
in water, but also which catalyst system and catalyst dosage being used. Acrylic is
typically not used in areas where it is subject to wetting and drying or freezing and
thawing. 65(EM 11101-3500. 1995, Karol, R.H. 1990). In general, the chemical stability
and durability of acrylic gels are very good.
16.6 Epoxy resins. Epoxy products has found very limited use in rock injection under ground, this is due to
the cost of epoxy and the difficulty of handling and application during operation.
Epoxy resin and hardener must be mixed in the right proportions for a complete
polymerization to take place if not; the quality of the product will be negatively affected.
The reaction is exothermic and it’s not effective in filling very large cavern. Epoxy is
characterized with high viscosity except special solvent are used. The working safety and
associated environmental risk has reduced the adoption of epoxy for rock injection
underground. 66( Bruce. D. et al; 1998).
16.7 Bitumen Asphalt. Hot bitumen grouting technology has continually evolved since its early applications
almost a century ago in France, Germany and the USA to seal persistent leaks in tunnels,
65 EM 11101-3500. (1995). EM 1110-1-3500. Engineer Manual. The U.S Army Corps of Engineers. Engineering and Design –Grouting Technology (1995). Geotechnical and Materials Branch, Engineering Division, Directorate of Civil Works (CECW-EG). Karol R.H 1990. Chemical Grouts and their properties. Marcel Dekker. Inc. New York. 66 Bruce. D. et al; (1989). “ High Flow Reduction in Major Structures: Matreials, Principals, and Case Histories” Grouts and Grouting Proceedings: Geo-Congress 98. Boston, MA. Pp. 156-175.
117
below dams and for erosion protection along canals in conjunction with cement based
suspension grout.
In scenarios where others types of grout has proved less effective, heated liquid bitumen
(asphalt) could be an alternative. Selected quality of bitumen are heated to a sufficiently
high temperature (200o C to 230o C) characterized with low viscosity allowing easy
pumping. The softening point should be around 95 to 100oC. The output must be adapted
to the water head. 67(Naudts 2003, Naudts 2003, K. Garshol 2003, Jaroslav Verfel 1989).
Examples of successful field applications of hot bitumen grouting in projects such as
Lower Baker Dam, U.S.A., 1920s, 1950s, 1964 and 1982, Stewartville Dam, Ontario,
Canada – 1980s, (Deans, et al.1985), Kraghammer Sattel, Germany, 1963 (Schönian,
1999), Potash Mine, Canada -1997, Quarry in Eastern United States, 1998, Milwaukee
Tunnel, Wisconsin, U.S.A – March, 2001, and New Yung Chung Tunnel, Ilan, Taiwan
2002, has demonstrated that the application of bitumen technology is an efficient,
economical and powerful tool to prevent or stop seepage and major leaks, (Naudts 2003).
The nature of most bitumen grouting application involved emergency situations in which
very serious water inflow problems needed to be solved. Ideal bitumen quality will change
from an easily pumped fluid material to sticky, high viscous and non-fluid asphalt at the
water temperature.
As the bitumen grout comes in contact with the water, the viscosity of the grout increases
rapidly resulting in lava like flow forming a hard insulating crust at the interface between
water and bitumen and shelters the low viscosity, hot bitumen behind it. The crust or skin
is re-melted from within when hot bitumen continues to be injected. 68(Schönian, 1999).
After injection into the water stream, bitumen will rapidly loose its high temperature and
rapidly and dramatically change its rheological properties. The bitumen gets sticky and
thus gradually blocks the flow. Additional cementitious grout ensure permanent and very
stable barrier.
67 Naudts et al; (2003). Hot Bitumen Grouting: The antidote for catastrophic inflows. Jaroslav Verfel (1989). Rock Grouting and Diaphragm Wall Construction. Development in Geotechnical Engineering 55. Elsevier Science Publishers. Naudts et al., (2003). Additives and Admixtures in Cement-based Grouts. K. Garshol. (2003). Pre-Excavation Grouting in Rock Tunnelling. 68 Schönian, Erich. (1999) The Shell Bitumen Handbook. Shell International Petroleum Company Limited.
118
Some of the distinct advantages of bitumen as a grout materials to stop or control water
flow, especially under high pressure at high flow rates is that, it can be injected for a very
long time (days-even weeks) into the same grout hole without the risk of either premature
blockage due to its good insulating characteristics and absolutely no wash-out (resist
dilution) because the faster the water flows, the faster the bitumen cools off. Cement–
based suspension grout is often injected in conjunction with hot bitumen to compensate for
the associated thermal shrinkage of the bitumen; making the bitumen less susceptible to
creep; and to increase the mechanical strength of the end product. On the other hands, the
equipment and set-up are generally more complex for bitumen grouting than for the
applications involving regular cement based grouts or solution grouts. The operating
temperature of the surface pipe system needs to be in the range of 180-225˚C (356-
437˚Fahrenheit). The piping system used for grout delivery from the bitumen pumps to the
sleeve pipe “stinger” located at the end of the bitumen grout hole, must either be pre-
heated with oil, heat trace, or steam, potentially through a re-circulation system as well as
the grout been insulated and equipped with temperature sensors and pressure gauges.
(Naudts 2003).
It is noteworthy that, the apparent Lugeon value which defines the permeability coefficient
of the formation using grout as a test fluid typically decreases with time during the
execution of hot bitumen grouting contrary to cement grouting operations due mainly to
the excellent penetration properties of the bitumen into the formation. 69(Landry, et al.,
2000).
Environmental wise, “hard” oxidized blown bitumen with a high solidification point are
environmentally friendly with as well as a long history of successful use for lining
portable water reservoirs and fish hatcheries ponds in United States of America. America
Water Works Association (AWWA) has adopted oxidized bitumen proven to be in
compliance with her standards for leachate resistance of materials use in portable water
applications. In summary, if planned with sound engineered by competent contractor with
suitable equipment using appropriate type of bitumen the “hot bitumen in conjunction with
cement grouting” technique has never fails to stop high inflows. The grouting can be
69 Landry, et al., (2000). New Developments in Rock and Soil Grouting: Design and Evaluation” Geotechnical News. September 2000. pp 38-44.
119
applied safely, economically, despite adverse conditions and last but not the least, blown
oxidized bitumen grouts could be considered the most environmentally friendly grout
presently available in the market (Naudts 2003,).
Advantages and Limitations of Chemical Grouts.
Chemical grouts contain no solid particles hence; they can be injected into rock materials
containing voids that are too small to be penetrated by cementitious or other grouts
containing suspended solid particles. Chemical grouts can therefore be used to control
water movement in and to increase the strength of materials that could not otherwise be
treated by grouting. It has been used in filling voids in fine granular materials, very
effective in sealing fine fissures in fractured a rock or concrete, commonly used in
stabilization or for increasing the load-bearing capacity of fine grained materials in
foundations and for the control of water in mine shafts, tunnels, trenches and other
excavations.(see table 2 and 3).
Chemical grouts suffer from the disadvantages they are often more expensive than
particulates grouts. Large voids are often grouted with cementitious grout, and chemical
grouting is carried out as required. Chemical grouts are also restricted or totally banned in
some circumstances due to potentially toxic effects that have been observed with some of
the unreacted grout components. Potential groundwater is a major consideration in the
selection of the type of grouts to be used in many cases. (EM-1110-1-3500. 1995, Skanska
Environmental Report 1997).
120
Table 2. Ranking of Major Grout Properties.
Table 3. Ranking of Chemical Grouts by Application. (EM 1110-1-3500. 31 Jan
1995).
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17 High Pressure Ground Water Condition.
17.1 Basic Problem. High pressure local water features creating a very high-flow water in-rush leads to
problem in the tunnel excavation by temporary delaying production. In hard rock
excavation such as granite and granitic gneiss with moderately high overburden, only few
stretch of the tunnel stands the possibility of intersecting highly jointed areas producing
large water in-rush at very high pressure. 70(Knut F. Garshol 1999, H O. Honnestad et al;
2002, Holter, K.G. 2005, Holter K.G et al, 1996, Sjostrom O.A 2003, Schunnesson, H.,
1996. K. F Garshol 2003).
Some common features that may be added to magnify the problems are but not limited to
the following:
• Weak rocks or rock zones, or heavy jointing and crushing, (shear zone).
• Too low pumping capacity for dewatering or poor drainage capacity.
• Tunnelling down slope or the access tunnel is on a down slope.
• Frequent intersection of water bearing channels.
17.2 Practical procedure in high risk areas. In practice, there are a few methods of handling water inflow problems in tunnelling
excavation, most especially in shear zones.
17.2.1 Pumping System. The capacity of the pump chosen must be based on predicted and actual project conditions,
the reserve capacity should be minimum 100%. Excavation along down gradient requires a
70 Knut F. Garshol (1999). Use of Pre-injection and Spiling in Front of Hard Rock TBM Excavation. Paper presented at Tenth Australian Tunnelling Conference 1999. Melbourne, Victoria 21- 24 March 1999. H.O. Honnestad et al; (2002). Hans O. Honnestad, Knut Garshol and Ross Dimmock, (2002). Pre-injection in hard tunnels using Degussa Rheocem microcement. Degussa International UGC Publication Holter, G.K (2005). Modern Pre-Injections in Underground Construction; Saving Time and Costs in Difficult Ground. Degussa (BASF) UGC Europe, 8048 Zuerich, Switzerland. Holter K.G et al; (1996). Tunnelling through a sand zone: Ground Treatment experiences from the Bjoroy sub sea road tunnel, Proceeding of North American Tunnelling 1996. Ed: Ozdemir, pp 249-256, Vol 1 (AA Balkema Rotterdam). K. F Garshol (2003). Knut F. Garshol (2003). Pre-Excavation Grouting in Rock Tunnelling.
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stepwise pumping system with buffer tanks and decreasing capacity down slope has been
effective.
17.2.2 Probe drilling /coring. The effectiveness and safety effect of probe drilling is proportional with the number of
appropriately oriented holes; there are always the probabilities of executed probe drilling
not penetrating the water bearing local features/channels. In high risks zones, it is
recommended that a minimum of 4-5 drillings of 25-30 m long holes should be carried
out. (H O. Honnestad et al; 2002, K. F. Garshol 2003).
17.2.3 Injection. In the case of high water pressure found at a drilled depth smaller than the planned probe
length, additional 2-3 meters further drilling could be carried out. More holes are then
drilled into the same area and grouted. New holes to check the effectiveness of the
grouting are made from the same face position, if the contact depth is less than about 15
m. In the case of the first contact and injection made at a depth greater than 15 m, further
excavation until 5 to 10 m remains can be executed to shorten the drilling for control-and
grouting purposes.(Holter, G.K. 2005).
17.2.4 Special cases. Despite probe drilling, pre-grouting and core drillings, the probability of still intersecting
local high pressure water still remains when drilling a blast round (D&B) or control holes
(TBM). In this case, an extra round of grouting may be required. Necessary precautions
must be taken to avoid allowing high pressure water too close to the face.
In the case of a poor and jointed rock in the face area, the risks may be high that local
ruptures in the face occur, grouting may then become difficult due to the flushing out of
pumped grout materials. Time lost and cost is usually the result while trying to established
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favorable condition for controlled grouting to be executed. 71(Holter K.G. et al; 1996, Knut
F. Garshol 1999, Skeggedal, T. 1983, Schunnesson, H. 1996).
71 Skeggedal, T. (1983). “The use of tunnel boring machines (TBM)”, Norwegian Tunnelling Society, Publication No 2, Oslo 1983. Schunnesson H. (1996). Probing ahead of the face with percussive drilling”, Tunnels and Tunnelling, January 1996, pp 22-23.
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18 Experiences /Summaries of Case Studies.
18.1 Case study 1. Chainage 135 – 1+ 146. Power Intake. This case covers the concrete lining of the Power Intake tunnel, which is about 1 km long
from the portal at Hálslón reservoir. It was excavated by the drill and blast method and
extends to the dismantling chamber for the third Tunnel Boring Machine (TBM).
Extensive post grouting activities were carried out covering the entire section ranging
from contact grouting to seal off the membrane between the host rock and the concrete
lining, while consolidation and curtain grouting were specifically carried out to
consolidate the niche (TBM chamber) meant for the Tunnel Boring Machine 3
breakthrough which never came to reality due to a number of problems for the machine
boring the tunnel upstream.
Works were performed according to the designers 72(KEJV) drawing 14-C-2.04.012 and
Site Instructions TN-58 and TN-59 as well as additional instructions issued during the
operation. Drilling and grouting was extended to consolidate the rock close to a fault
located between chainages 1+108 and 1+111 according to Site Instruction TN-62
(upstream from the chamber).
The only target was (apart from curtain grouting) to consolidate the rock around the lined
tunnel periphery, especially close to the fault. The water leakages through the construction
joints and drainage 6” pipes connected to the fault were only observed as minor seepages
through fine cracks in the lining.
The sequence of the works was planed as follows:
1. Drilling and consolidation grouting of 5 rings (8 holes in an umbrella shape per
ring) between chainages 1+105 and 1+117 close to the fault with 15 m long holes
into the rock Ø 76 mm. The rings were located upstream from the chamber based
on the geological mapping.
2. Curtain grouting with 3 rings (12 holes per ring) between chainages 1+137 and
1+143 with 15 m long holes into the rock Ø 76 mm within the chamber.
72 KEJV. Project designer drawing 14-C-2.04.012. Headrace tunnel Ch 1 + 126 to 1 + 146. Concrete linning, plan and profile. TN denotes Site instruction serial number.
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3. Consolidation grouting with 3 rings (12 holes per ring) between 1+125 and 1+133
with 4 m long holes in the rock Ø 76 mm within the chamber between sections as
above, (see figure 29).
In reality all bore holes drilled in the chamber were longer than 3-4 m (1 to 2 m in the fault
section) passing the reinforced lining and backfilled concrete.
In each consolidation grouting ring close to the fault (according to point 1) 8 holes were
drilled and in the consolidation and curtain grouting rings in the chamber (according to
point 2 and 3) 12 holes were drilled. The grouting holes were drilled with an alternating
pattern of 30 degrees (staggered) to prevent damages to the reinforcement.
In each section rings no. 1, 3 (or 5) were drilled and grouted first. After those rings were
completed rings in between were open. As a rule all holes in one ring were drilled first
before the grouting start and the grouting was performed from the invert up to the crown
to refusal pressure, (bottom up approach).
Grouting according to the site instructions should have been performed with cement based
grouts and thickening procedure (w/c content from 0,80 down to 0,45 to achieve the
refusal pressure of 10 bars corresponding to a grout flow of 40 l/min).
Lugeon water tests were performed at the first ring at Ch 1+105 only. There was no reason
to do further water testing in other rings due to huge water inflow from bore holes. Even in
the dry holes the pressure of 10 bars could not be achieved at the highest pump capacity of
50 l/min.
Consolidation grouting close to the fault (1)
The huge water inflows with high pressure were observed under the spring line and mostly
on the left hand side (LHS) looking downstream. In ring C (chainage 1+111) the drilling
and grouting program was revised not to open more than 2 holes at the same time from the
first appearance of water. These holes were grouted immediately one by one to stop the
huge water inflow. This inflow could risk the transportation of crushed material, sand and
gravel into the tunnel.
When the grout take exceeded 3000 l the grouting was suspended to the next day. In 3
holes with directions between 06.00 and 09.00 (using clock position), polyurethane
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grouting was performed to stop the water and to limit the range of grout penetration within
the open feature which was probably reaching the surface under the dam and above the
Power intake. Such holes had to be re-drilled close to their previous position. 3 holes
mentioned above were re-drilled 5-9 times.
Ring B at chainage 1+108 was left to be completed as the last ring in the grouting
program.
The curtain grouting (3) and consolidation grouting (2) in the chamber.
The curtain grouting between Chainage1+137 and 1+143 was performed as the second
section. The huge water inflow and grout-take was observed in general above the spring
line and close to the crown. The restriction of number of holes drilled at the same time was
the same as in the already performed fault section but to a smaller extent.
After completion of the curtain grouting, consolidation grouting in 1+125-1+133 with
short 4 m holes in the rock was started. The main water inflow and grout-take were
observed on the RHS, mainly concentrated between 12.00 and 03.00 o’clock (opposite as
in the fault section).
The general grout consumption in this section was considerably lower than in the fault
section and curtain grouting section, (see figure 29).
During grouting for consolidation at Ch 1+129 (ring B) suddenly a huge rock fall occurred
close to the adjacent ring C at Ch 1+133 and the tunnel was evacuated (07.01.05). A new
site instruction for an additional ring was written (TN-71) and the additional 12 holes were
opened 4 m in to the rock.
Huge leakages were located between clock position 12.00 and 02.00. Considerably high
cement grout consumption was noticed above the spring line in general. Evidently a new
crack system in form of a chimney was reopened between the already grouted curtain and
the consolidation rings.
The absolutely worst drillhole had direction 02.00 (clock position). This hole was grouted
in stages for 5 days – the total cement grout consumption was close to 25.000 l without
obtaining the refusal pressure of 10 bars. The hole was finally plugged with 460 kg of
polyurethane.
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Test holes
After completion of ring B at Ch 1+108 close to the fault 2 new vertical bore holes for
testing at chainages 1+131 and 1+121 were done (TN-72). In the first one after completion
of drilling, a huge water inflow was observed and was evidently followed by two rock
falls on the lining which was heard. The water stream changed rapidly from negligible to a
maximum of ca 80 l/s during these events.
The water inflow in the hole at Ch 1+121 was moderate and stable. This hole was grouted
with standard cement grout. The hole at Ch 1+131 was finally plugged with polyurethane.
For estimation of the rock mass stability at Ch 1+131 additional 5 holes between clock
position 11.00 and 01.00 at chainages 1+127 and 1+131 were instructed (TN-74). In only
one hole at Ch 1+127 new water was observed and ca 0,50 m of empty room above the
lining while drilling. This hole was grouted with polyurethane. All the remaining 4 holes
were almost dry and plugged with standard cement grout.
Drainage pipe systems
Finally the grouting of drainage pipes located close to the invert (i.e. tunnel floor) on both
sides of the tunnel was performed with cement mixtures. The pipes were mounted to
prevent the washing out of the fresh concrete mass during the pouring of the backfill
concrete.
Two plastic 1” pipes on the left hand side (looking upstream) drained water from small
leakages from the rock and concrete backfilling along the wall. These pipes were not
connected to the pumping pit located on the right hand side and were discovered by
digging. These pipes were refilled with a moderate cement grout volume (3120 l).
The 3 steel 6” pipes on the left hand were responsible for water coming from small
leakages like on the left hand side and collecting a huge amount of water coming directly
from the fault. To minimize the cement grout-take while all pipes were indirectly
connected through the rock mass the following procedure was employed:
- each individual pipe grouting was suspended after 2000 l of thick 0,45 mixture and
next pipe was taken for grouting,
- The other ungrouted pipes were opened to flush away water, air and dissolute
grout.
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- To decrease the range of grout penetration in the rock mass, wooden dust and
accelerator based on industrial salt were applied.
The total grout consumption to refill the drainage pipes on the right hand side was 40.470
l.
Complementary testing and grouting
During the grouting operation all leaking joints were sealed simultaneously. Only one ring
close to Ch 1+137 near the crown was still leaking in this section longitudinal cracks in
the concrete appeared probably after the rock falls.
To test the concrete’s condition and its conductivity, 2 additional core holes 4 m in to the
rock were instructed at Ch 1+137, 3 close to the leaking joint mentioned above.
The hole at 01.00 o’clock provided a small amount of water (1-2 l/s) coming only from the
backfilling concrete. The total thickness of the lining and backfilling concrete was 3, 30 m.
In general the concrete quality seamed to be poor. The 4 meter long hole in the rock was
dry – the sandstone was fractured but not affected by any grout.
The hole at direction 12.00 o’clock was dry and the quality of the concrete was estimated
as normal (no cracks). Above the backfilling concrete a 0, 45 m empty room refilled with
cement grout applied during contact grouting or curtain grouting was noticed. This hole
was later refilled with cement grout – the amount was close to the theoretical volume of
the bore hole.
The cement grouting of the leaking hole at 01.00 o’clock was suspended due to grout
losses of cement 0, 45 at a low pressure though existing cracks in the concrete lining. This
hole was completed the day after with a low pressure Polyurethane injection (ca 15 l only).
A total of 291.440 l of cement grout and 5215 l of polyurethane resin was used in this
grouting.
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Figure 28. Showing plan of power intake, tunnel concrete lining outline and grout plan up to the gate shaft, (Ch 135- Ch 282.813). This further extends to the TBM dismantling chamber. (Ch 1+ 146)
Figure 28 shows the plan view of the power intake from the portal (Ch 135) up to the gate
shaft (Ch 282.813). Contact grouting (of two to one pattern) were carried out and four
consolidation rings of 15 m long grout holes (8 holes per ring) were made close to the gate
shaft between Ch 0+ 254.313 and Ch 0+ 272.313. This concrete lining section of the
tunnel further extends downstream (east) to the TBM dismantling chamber (Ch 1+ 146).
See figure 29.
130
Figure 29. Diagram showing extension of the power intake plan seen from above (see
figure 28), and Contact, Consolidation and Curtain grouting in the in the presently cement-
filled dismantling chamber, (Ch 1 + 105 -1+ 146).
Contact, Consolidation and Curtain grouting in TBM Chamber.Contact, Consolidation and Curtain grouting in TBM Chamber.
131
18.1.1 General remarks and recommendation.
1. The distance between the rings should not exceed 2-2, 5 m.
2. The drilling of holes with huge water leakages and material transport should be
done in ascending stages from the collar.
3. To avoid local empty rooms and “air pillows” above the backfilling concrete close
to the crown, the grout hoses with simple valves type Tube-á-manchette (sleeve
port pipes) should be mounted.
4. From the same reason as an option the conventional contact grouting through drill
holes should be performed to a refusal pressure not lower than 10 bars at the collar
related to the grout flow 40-50 l/min.
5. The core drilling through the lining should be performed in sleeves left in the
structure to prevent damages to the reinforcement and extra sleeves for additional
holes should be applied.
6. In the case of open cracks and voids thick cement grouts with lowest w/c ratio of 0,
45 are recommended with additives like sand ballast, wooden dust or chips and
accelerators like sodium silicate.
7. In huge leaking holes with lengths up to 7 m, the consolidation effect in the rock
can be achieved with Polyurethane grouting but the grout piston Polyurethane
pumps with capacity no less than 10 l/min must be applied.
8. Cold weather conditions and water temperatures below 10° C in the rock mass or
used at the mixing plant cause an uneven and ineffective grout penetration due to
very long cement grout setting time.
Accelerators like industrial salt and sodium silicate as well as water heaters are
strongly recommended.
9. In the case of Polyurethane grouting the frost sensitive components should be kept
within the temperature range of +5° C - +25° C.
Polyurethane materials from Minova – CarboTech are suitable for grouting down up to -5°
C without the need of heating as in the local Kárahnjúkar weather conditions. Therefore
they are strongly recommended instead of the commonly used Degussa-BASF PU
products. (VIJV Grouting team. VIJV- Database 2006, Tomasz Najder 2006).
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18.2 Case study 2. Chainage 19 + 645- 19 + 595. Adit 2. ( TBM Stuck for 6-months). Extraordinary Geological Occurrence, (EGO) Section. Ch 19 +640-19+597.
This 50 m section of the tunnel is 19 km downstream the Head Race Tunnel. It
lies within the middle section called Head Race Tunnel two. This 50 m section is
characterised with three big faults, with very high leakages through the faults, and
blocky rock fall during excavation. Backfilling of voids up to 600 m3 were
experienced and the Tunnel Boring Machine (TBM) got stuck for six months
within this section before it could further advance.
Geological Description Ch19+640 to 19+597. Adit 2.
The general stratigraphical sequence between Ch 19+640 to Ch19+597 follows
that of the proceeding 20m of tunnel i.e., olivine basalt overlying tillite.
However where this section of tunnel differs is both in the quantities of water
ingress and in the nature of the fractures.
The total water ingress in this area was in the order of 100 l/sec coming from two
main areas; namely from a fracture at Ch19+620 and a second fracture at
Ch19+600. It is this water that has proved the major hindrance to backfilling and
re-boring through the faulted zones.
Structurally the area consists of several larger fractures in association with very
closely to closely spaced minor fractures. These “fractures” consist of both joints
and very small scale faulting along with faults with throws of 1 to 3m. The main
structures have a strike and dip of 220° / 85° East. There are however three other
fracture sets giving in total four fracture sets, three sub-vertical fracture sets
(striking approximately at 220° , 250° and 160°) and one fracture set with a dip of
between 0° to 30°.
There were three “problem” areas between Ch19+640 to 19+597.
The first area occurred between Ch19+640 and Ch19+630 where two faults
intersected causing a large overbreak of around 80 m3.
The second area caused more problems mainly due to large volumes of water
ingress along the fracure at Ch19+620. The highly fractured blocky nature of the
ground in this area in combination with the large water ingresses caused a void of
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some 145 m3 to form above the tunnel crown between approximately Ch19+626
and Ch19+613.
The third area was the worst geological conditions thus far intersected on the
Project. Here we have the four joint sets mentioned earlier. These joints can be
described as very closely spaced (60 to 200mm) to closely spaced (200mm to
600mm) with a range of persistence from the minor joints of low (1 to 3m) to very
high (>20m) for the major fractures, often planar striated, sometimes open with
clay filling and with a total water ingress of up to 30 to 50 l/sec. All of this has led
to a void opening up above the tunnel crown between Ch19+610 to Ch19+597,
(the main part of the void is between Ch19+604 to Ch19+597) with a volume of
around 500 m3.
Expected geology ahead of the face.
Borehole HRT2-BH7 was drilled from headrace tunnel 2 and extends ahead of the
face to around Ch19+562. Very briefly this borehole indicates the following:
From Ch19+597 to Ch19+580 the fracture spacing is generally very close (60mm
to 200mm) to closely spaced (200mm to 600mm) and the fractures are generally
striated. The sub-horizontal joints are spaced at between 0.3m to 1.3m. A water
bearing joint thought to strike at 220° will intersect the right hand side of the tunnel
at around Ch19+591.
From Ch19+580 to Ch19+562 the geology appears to improve with the fracture
spacing increasing to widely spaced (600mm to 2000mm) and the spacing of the
sub-horizontal joint set inceases to around 6m. The joint planes are generally
planar rough.
In summary the zone from Ch19+597 to Ch19+580 must be treated with extreme
caution. Hopefully the conditions are better than the core indicates, but the
likelyhood is that the ground will want to form a void as previously. Sealing off the
water by pre-grouting will be carried out to at least try to minimise the influence of
water on the tunnel stability.
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From the recently carried out geophysics survey we can expect to be in variable
ground conditions until Ch19+400. From Ch19+400 to Ch19+090 good ground
conditions can be expected before we encounter what may be a major lineament at
Ch19+090 to Ch19+000. Further good ground conditions are indicated from
Ch19+000 to Ch18+600. 73(VIJV- Database 2005, Peter Pitts 2005).
TBM 2 Chronology of Measures for passing Ch 19+645 to 19+595.
Due to continuing conveyor problems,TBM2 was stopped by the contractor on 24th
February 2005, in order to break out for the already delayed Drill and Blast (D&B)
Downstream Access & Jokulsa access tunnels at the Adit2 Junction area, and to re-aligned
the main tunnel conveyor.
TBM2 resumed boring on 26 March 2005, and immediately returned weekly production
figures of 218 and 274 metres/week in generally Class 1 rock conditions between Ch.
20+337 to 19+650.
By Ch.19+645, (20/04/05) the previously good Class I rock conditions started to
deteriorate rapidly, and Class IV Rock Support was being instructed from Ch. 19+640
onwards. The ground was particularly blocky with clay partings, that needed immediate
shotcreting behind the cutterhead. As a first measure the Class IV rib spacing was reduced
to 0.4m centres with full steel lagging, and cautious advance continued up to Ch. 19+622.
Following an inspection by The Engineer, the steel spiling measures described in the
Design documents were instructed in order to reduce the ravelling of the ground in front of
and above the cutterhead.
In the first week of May 2005, very slow advance of the TBM followed up with 0.4m
centred ribs, and spiling pre-reinforcement continued, the first void was now visible
behind the cutterhead, and ground loss at the cutterhead crown was continuing. The spiling
measures as described in the design were proving inadequate. Discussion on the use of
Polyurethane foam to stabilise the area were iniated by the Owners representative (VIJV)
73 VIJV (2005) Database: Viseningar Joint Ventures, Database. Construction Supervision Team. Kárahnjúkar Project Hydrelectric Power Project. Peter Pitts (2005). Personal Discussions and Internal Reports on Advancement of TBM 2. VIJV: Visen-ingar Joint Ventures. Senior Geotechnical Enginneer. Kárahnjúkar Hydroelectric project Supervision Team.
135
but not implemented until Robbins confirmed no damage to the TBM could be expected
through contact with such foams.
In the second week of May 2005 the TBM continued to be advanced slowly with full
Class IV support, however a massive 150 l/sec ingress of water was also encountered
which only exacerpated the situation and caused more ground loss above the cutterhead,
and temporary loss of rotation
On 11/05/05 a joint decision was made between Owners Representative/Landsvirkjun and
Impregilo S.P.A Iceland Branch (Contractor) to stop further advance, remove the
previously installed steel ribs and reverse the TBM 10 metres back to 19+630 to a safe
working area, and to construct a concrete bulkhead ahead of the TBM by protecting the
cutterhead with timber and polythene and then simply pumping concrete ahead into the
fault zone. Backfill concreting works continued in the third week of May 2005, however
by 17/05/05 the wash out of the concrete by the ingressing water was so severe it was
impractical to continue. The concrete backfill (621m³) had filled 80% of the tunnel
approximately, but the uppermost and critical roof section could not be filled due to the
ingress of water. Rapid reacting polyurethane foam was mobilised to site in order to seal
off the water and resume concrete backfill, drainage relief holes were also drilled. The
small amount of foam already on site was seen to be effective in sealing off the water
locally, but supplies were soon expended, by 25/05/05. Further supplies of foam did not
arrive until 31/05/05.
In the first week of June, further supplies of foam arrived and were used to seal off the
main areas of water ingress, and further concrete backfill/grouting operations were
attempted, in order to consolidate an area above the cutterhead, from which further pre-
grouting measures for the ground ahead could be collared from. On 07/06/05, the TBM
started re-excavating through the concrete plug, which continued for the next few days
again followed up with 120 steel ribs on 0.4m centres. The TBM was advanced 15m to Ch
19+614 and through the first fault by 15/06/05.
However advanced core drilling ahead of the cutterhead indicated a second and wider fault
zone lay ahead between Ch 19+610 and 19+600
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At Ch. 19+614, drilling for pre-grouting and consolidation ahead of the face was
attempted using self-drilling injection pipes previously instructed to be brought to site by
the Owners representative. Attempts were limited by the poor rock conditions and grout
loss due to wash out. The TBM was advanced a further 2 metres and a second ring of pre-
grouting attempted, pre-grouting operations continued up until 11/07/05, and up to Ch
19+610. In this period the pre-grouting operations were hindered by the Contractors lack
of grout packers and experienced grouting operatives. The TBM resumed boring through
the pre-consolidated area on 11/07/05 again with HEB120 steel rib support and full steel
lagging support, advancing approximately 1 metre per day. By 21/07/05, the TBM had
advanced a further 13m to Ch. 19+597, and was already out of the main fault and in better
rock on the RHS of the cutterhead. At this juncture the TBM main conveyor broke down
and was not repaired until the Saturday nightshift of 23/07/05, whereupon the continuing
collapse of material during the stoppage had been found to have blocked the cutterhead
and rotation had been lost.
Further attempts to restore the rotation of the TBM head failed, and it became neccessary
to remove the previously installed ribs in order to reverse the TBM. However these ribs
where supporting loose boulders from the void above. It was necessary to consolidate
grout this area to a depth of 1.5 to 3 metres before it was safe to remove any ribs.
Consolidation grouting operations, removal of short sections of ribs, and attempts to
restore rotation continued throughout July and August. However the voided are above the
cutterhead continued to extend and ravel further, adding to the collapsed material pressing
against the cutterhead and preventing rotation. Again at the OR’s instigation, 12 m long
90mm steel pipes were installed above the cutterhead to limit further ground loss during
TBM reversal operations, but with little effect.
Into September further attempts at consolidating, removal of ribs, reversing of the TBM
continued back as far as 19+609 where rotation of the cutterhead was eventually restored
on 15/09/05. On restoration of rotation the TBM was moved forward under an overlapping
umbrella of self drilling bolts (spiles) and pre-grouting up to Ch 19+604 by 27/09/05.
However beyond this chainage the roof was again observed to be open void with
continuing loosening of the surrounding rock mass. To proceed further could have resulted
137
in a major de-stabilisation of the whole area. It was therefore jointly decided to pull back 5
m, and construct a watertight concrete bulkhead and backfill the void up to 2m above the
tunnel crown and to re-excavate through this second concreted plug.
The Bulkhead was constructed and the void backfilled up to crown level between
28/09/05 to 10/10/05. Re-boring through this second concrete plug continued to Ch.
19+605 by 12/10/05, where further void consolidation was carried out. The TBM then
advanced back to the originally achieved face at Ch 19+597, where pre-consolidation
grouting of the undisturbed ground ahead was carried out and normal TBM production
resumed. The stability of the section was restored to a large extent at this period to ensured
the stability of the tunnel and further advancement of the TBM. Additional rock support
were postponed till the finishing work period where intensive reinforcement, further
backfilling of voids and grouting of fifteen long rings of consolidation holes (12 holes per
ring) were carried out, (see figure 32) Both cement and polyurethane grouts were used.
Table 4. Summary Timeline of the operation described in Case 2.
21/04/05: Encountered first fault zone Ch.19+640
07/05/05: Advance by full class IV support measures to Ch.19+620
11/05/05: Encountered major water ingress which overcame normal pre-
grouting/consolidation techniques. cutterhead rotation lost at Ch.19+620.
13/05/05: TBM reversed to safe working area Ch 19+620 to 19+630
06/06/05: First Concrete bulkhead/plug constructed at 19+630
15/06/05: Re-excavate through concrete plug19+630 to 19+615
09/07/05: Pre-grout and consolidate second and third fault zones 19+615 to 19+600
21/07/05: Slow but steady advance by full class IV measures through collapsing
ground 19+615 to 19+597
21/07/05: Conveyor breakdowns, - 48 hour downtime
23/07/05: TBM Rotation lost at Ch 19+597
15/09/05: Consolidation grouting, removal of ribs and reversing of TBM to restore
rotation 19+597 to 19+610
27/09/05: Re-excavate through collapsing material 19+610 to 19+605
138
27/09/05: Further major collapse prevents further advance 19+605
04/10/05: Second bulkhead and plug constructed at 19+610
20/10/05: Re-excavation through plug 19+610 to 19+597
30/10/05: Consolidation grouting ahead of Ch19+597
Nov 2005: Boring Resumed.
(See figure 30).
139
Figure 30a. Diagram showing the Chronological advancement of the TBM 2 through the
faults #1, #2 and #3, (See Table 4. Summary timeline).
140
Figure 30b. Diagram showing the Chronological advancement of the TBM 2 through fault
#1, #2 and #3, (see table 4. Summary timeline).
141
Could the faulted zone have been predicted ?
According to the Site Investigation (SI), the particular area of tunnel is situated in the LA
middle suite of rocks which are predicted to occur between Ch19+200 to 21+800. The
nearest SI borehole to the fault zone is 1,5 km west at Ch 18+100 (#JB40), the nearest SI
borehole to the east is 4 km distant at Ch 23+600 (#FV47). There is a 5.5 km interval
between SI boreholes in this area, and the SI indicates an “information gap” between Ch
19+700 to 20+700. However this information gap section of tunnel was excavated mainly
in Class I rock without major geological incident.
The SI goes on to predict the likely rock class distribution to be 70% Class I, 30% Class II,
9% Class III and just 1% ( 26m) as being Class IV. Furthermore the likely risk of
encountering loose blocky ground, fault zones and sudden and major ingress of water to be
low or moderate risk.
According to this information no systematic probe drilling ahead of the TBM was being
implemented at Ch 19+640, and it is unreasonable to have expected a 45 metre wide fault
zone to have been predicted from the SI information provided. No systematic core drilling
was being made from the TBM either mainly due to the Contractors lack of suitable
equipment and experienced drillers.
With the major benefit of hindsight, surface lineations are apparent on aerial photos that
are coincident with lineations and faults encountered at Ch 19+600. Similarly, it could be
said that a 5.5 km interval between investigation boreholes was too wide for accurate
geological predictions, especially in view of using a hard rock TBM ill equipped for
dealing with soft-ground conditions. Angled boreholes on much closer spacings (say 1
km) would in retrospect have been a much more accurate way of detecting and predicting
steeply dipping fracture zones.
Since encountering the fault zone, the Owners Represeentative(OR) has suggested and
Landsvirkjun (LV) has instigated surface geophysical investigation, from which it can
now be seen that surface magnetometry is a particularly useful tool in identifying
lineations such as dykes or faults. (VIJV- Database 2005).
Other recommendations by the Owner’s Representative (OR) to improve the geological
predictions ahead of tunnel faces are:
142
• Continuous percussion probe holes ahead of the face.
• Underground core drilling, (Diamec drills).
• Desk Study, (Study topography maps).
• Surface Mapping, (Locating major faults and dykes i.e. through GPS).
• Surface Geophysics, (ISOR).
• Surface core drilling, (Angled holes).
Finishing Works.
Extraordinary Geological Occurrence, (EGO) Section. Ch 19 +640-19+597.
As part of the on-going finishing works, before filling the tunnel with water, further
reinforcement and concrete lining of this section (see figure 31) is been carried out. This is
the only section of the TBM excavation zone that is been concrete lined and thus reveals
the extent of the deterioration of the zone.
Drilling and grouting for backfilling of the voids, contact grouting to seal the membrane
between the host rock and the concrete lining and fifteen rings of consolidation holes (ring
A-N see figure 32), 12 holes per ring was instructed as part of the efforts to ensure
stabilisation of the tunnel strata.
Both cement grouts and chemical counterparts are used most especially in high leaking
holes through the concrete.
143
Figure 31. Diagram showing the length of the Extraordinary Geological Section (EGO).
NB, the tunnel should be circular in the cross-section.
144
Figure 32. Diagram showing consolidation grouting rings (A-N) and backfilling of voids
as part of the rock support. NB, the tulle should be circular in the cross-section.
145
18.3 Case study 3. Chainage 1 + 468, 5. Adit 4. (Dry-dyke) Geology Mapping.
Few meters before the actual chainage of this case, the geological mapping revealed very
strong gray olivine basalt and strong scoria. Then we get strong brown conglomerate (CG)
and sandstone in the lower wall. Then moderately altered rock with coating of yellow clay
on joint surfaces with dripping of water. Then columnar dyke with columns approximately
10 cm in diameter with water seeping and dripping from several places. Altered zone with
clay filled contact on both sides of the dyke. Poor blasting results and the face is un-even.
The boundary between the dyke and host rock is clay filled.
Geological overbreak. (Ch 1+ 460 – 1+472).
Figure 33. A cross section of the tunnel showing geological mapping at the dyke.
146
Drilling of grouting holes.
Figure 34. Drawing showing drilling and grouting/plan view seen from above (lower picture).
147
Drilling and Grouting.
Drilling of series of 24 holes (diameter Ø 51 mm) spaced at 1 m intervals from the lowest
point at the invert (floor) to the crown (ceiling) on either side. The holes were at 45˚out
from the tunnel axis and collared 7 m back from the face and all holes were drilled long
enough to intersect the required feature (the dyke). A technical log kept from drilling
includes location of cavities, cracks, drill bit rate of penetration and uniformity of rotation.
Directly after drilling of each individual hole, the automatic packer BVS 40 Ø 42 – 55 mm
(Minova) with steel extension pipe were inserted and installed close to the dyke.
Grouting/ Treatment of dyke.
The rationale behind the post grouting of the dyke is to prevent leakages in and out of the
tunnel which can drastically affect the production of electricity and could as well gradually
lead to environmental degradation. Hence, consolidation of the area was jointly instructed
by the Supervising Engineer and the Designer. Grouting was performed with Polyurethane
pump type CT-GX 45-11 with the min grout delivery 25 l/min requiring approximately 7
m3/min of compressed air. Polyurethane foam CarboPur Minova (two component system)
with accelerator CarboAdd fast (added in amounts of 2% to component A) was used.
The sequence of grouting was from the lowest position (in the invert) up to the crown. All
holes were grouted to refusal pressure (110 bars) measured on the manometer at the pump.
Three hundred liters of Polyurethane were taken as the maximum injected volume per
individual hole. If refusal pressure is not reached, then an interruption of 2 minutes is
made before continuing grouting again with polyurethane (two components and
accelerator).
In the case of any individual hole with high grout-take (above total 600 liters), the
procedure above would consequently be repeated after every further 300 liters.
Closing Criteria: Some adopted criteria for interrupting the injection e.g. in case of:
• Intensive leaking of non-reacted polyurethane into the tunnel.
• The refusal pressure of approximately 110 bars is reached.
• The expanded Polyurethane grout fractures the rock or shortcrete close to the
surface.
148
• The packer is not set tightly in its position, but moving towards the mouth of the
hole.
18.4 Case study 4. Chainage 002 + 255. Adit 3. ~150- 200 l/sec. This case is located about 2 km into the tunnel from the Halslon reservoir and lies within
the Drill and Blast section of the Head Race Tunnel (Power Intake). The geological
mapping revealed a sequence of scoria above olivine basalt, discontinuity by a fault which
is grey to dark grey and very strong tholeiite basalt, grading into highly vecular/scoracious
basalt, strongly reddish to purple locally. Joints are generally widely spaced and of low to
medium persistence. Joint infill of soft brown clay is common.
Despite the information from the probe holes which revealed water bearing features, the
grouting holes were not grouted to the refusal pressure during pre-injection as instructed
by the Owners Representative. Secondly, the invert (floor) were neglected during
grouting. This resulted in enormous water inflow after blasting (150-200 l/sec) which then
required additional watertightening by post grouting. This was targeted to prevent
leakages in and out of the tunnel which goes against good tunnelling principles.
149
Figure 35. Geological Mapping showing the geological overbreak in case 4.
150
Figure 36. Diagram showing the drilling and grouting activities/plan viewed from above
(lower picture).
Drilling and grouting activities were carried out in a similar way with the above (figure 34.
Ch 1 + 468.5), and the closing criteria was the same. Coffer dam was built downstream of
the grouting area to ensure that unreacted polyurethane does not flow out of the tunnel
with the run-off water there by preventing possible pollution.
151
18.5 Case Study 5. Adit 2. Jökulsa Valve Chamber. Curtain Grouting – Plug 2.2. The Plug is a concrete structure used to permanently close the Head Race Tunnel from the
adit (Adit 2). Curtain and consolidation grouting were carried out for stabilisation reasons.
This involved 4 rings (12 holes per ring), 15 m long in an umbrella shape, spaced at 3-4 m.
Cement grouting was carried out until refusal pressure of 10 bar was reached. Figure 37a
shows the pattern of ring A, 37b shows the pattern of ring B, and 37c shows the pattern for
ring C, while ring D shows the pattern for the consolidation holes (5m).
Plug 2.2 which lies at the Jökulsa í Fljótsdal right branch, cuts across a fault (see figure
37e) of which the grouting of the area further consolidates the zone. Cement grouts are
used for the plugs in general and Plug 2.2 was observed with very low intake during
grouting. All holes were grouted to the refusal pressure of 10 bar.
Figure 37a. Sketches Showing Ring A (15m). Curtain Grouting ( 1,3,5,7,9,11,13,15,17,19,21,23).
152
Figure37b. Sketches Showing Ring B (15m). Curtain Grouting ( 2,4,6,8,10,12,14,16,18,20,22,24).
Figure 37c. Sketches Showing Ring C (15m). Curtain Grouting ( 1,3,5,7,9,11,13,15,17,19,21,23).
153
Figure37d. Sketches Showing RingD (5m). Consolidation Grouting ( 2,4,6,8,10,12,14,16,18,20,22,24).
154
Figure 37e. Curtain drilling and grouting of plug 2.2.
155
19 Discussions and summaries of lessons learned, results and conclusion.
“Grout”is defined 74(ASCE) as “a material injected into a soil or rock formation to change
the physical characteristics of the formation. Grouts can be divided into two categories, the
suspended- solid grouts, such as cement and bentonite, and those materials which are true
solutions, or chemical grouts.
Rock grouting is probably the best known of grouting practices to engineers working in
tunnel design and construction. Its application is found on most permanent dams, tunnels
and numerous other structures such as oil rigs, river barges and in mines to avoid sudden
ingress of water, and to stabilize and seal the ground ahead of excavation operations.
The purpose of grouting in rock tunnelling varies on projects but would mostly be:
• Reducing the permeability against water inflow.
• Enhancing the rock strengths.
• Filling of voids.
The importance and necessity of grouting in rock tunnelling cannot be over-estimated;
pre-grouting activities enhances the smooth running and advancement of the excavation
face in order to ensure that projects are completed within assumed time frame. This goes a
long way to safe time, money and any un-necessary claims that may come up.
Proper pre-grouting activities could almost eliminate post-grouting activities which has
been known to be time consuming, very expensive and in practice difficult in achieving
the desired and expected water tightness of the tunnel.
In shear zones, 360˚grouting pattern should be considered as this eliminates movement of
water through the other side of the feature.
A manageable grouting team is highly recommended from the beginning of any rock
tunnelling which could be increased in number with time if required. This alleviates
additional efforts that may need to be put on post-grouting thereby saving cost.
74 ASCE: American Society of Civil Enginners.
156
Grouting objectives in rock tunnelling go beyond sealing the tunnel against underground
water inflow, as to prevent any adverse environmental effect inside the tunnel, prevent un-
acceptable impact on the external environment, as well as maintaining hydrodynamic
containment but also for overall stabilization of the tunnel strata.
Post-grouting treatment in tunnelling are often necessary to consolidate the rock mass,
minimise water ingress or provide intimate contact between the tunnel lining and the rock
mass (contact grouting).
Chemical grouting has brought an irreversible development into the grouting industry
within the last five decades, with the wider application of fast reactive resins in
underground constructions such as mines and tunnels.
Effective and continuous monitoring of surrounding aquatic bodies is highly
recommended even beyond the construction period of the project.
Proper materials, equipment and technical know-how from competent and experienced
grouting practitioners remain adequate ammunition in successful rock grouting.
Multiple hole grouting is highly recommended to many problems connected with
impermeabilization and stabilization of rock tunnelling.
Up to date injection records and quality control during injection remains an important part
of successful grouting in rock tunnelling.
The three Tunnel Boring Machines (TBMs) used in the excavation of the Kárahnjúkar
Head Race Tunnel (HRT) had successful breakthroughs, despite high water inflow that
characterised parts of the tunne,l through the help of planned and successful grouting
activities (pre-grouting) ahead of the excavation face and afterwards, shear zones being
successfully consolidated (post-grouting). This has in no small measures, increased
confidence regarding the durability and overall stabilization of the tunnel strata.
157
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