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slope stability
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ORIGINAL PAPER
Slope Stability Problems and Back Analysis in Heavily JointedRock Mass: A Case Study from Manisa, Turkey
Mutluhan Akin
Received: 13 February 2012 / Accepted: 1 May 2012 / Published online: 24 May 2012
� Springer-Verlag 2012
Abstract This paper presents a case study regarding
slope stability problems and the remedial slope stabiliza-
tion work executed during the construction of two rein-
forced concrete water storage tanks on a steep hill in
Manisa, Turkey. Water storage tanks of different capacities
were planned to be constructed, one under the other, on
closely jointed and deformed shale and sandstone units.
The tank on the upper elevation was constructed first and
an approximately 20-m cut slope with two benches was
excavated in front of this upper tank before the construc-
tion of the lower tank. The cut slope failed after a week and
the failure threatened the stability of the upper water tank.
In addition to re-sloping, a 15.6-m deep contiguous
retaining pile wall without anchoring was built to support
both the cut slope and the upper tank. Despite the con-
struction of a retaining pile wall, a maximum of 10 mm of
displacement was observed by inclinometer measurements
due to the re-failure of the slope on the existing slip sur-
face. Permanent stability was achieved after the placement
of a granular fill buttress on the slope. Back analysis based
on the non-linear (Hoek–Brown) failure criterion indicated
that the geological strength index (GSI) value of the slope-
forming material is around 21 and is compatible with the in
situ-determined GSI value (24). The calculated normal–
shear stress plots are also consistent with the Hoek–Brown
failure envelope of the rock mass, indicating that the
location of the sliding surface, GSI value estimated by back
analysis, and the rock mass parameters are well defined.
The long-term stability analysis illustrates a safe slope
design after the placement of a permanent toe buttress.
Keywords Slope stability � Back analysis � GSI �Non-linear failure criterion � Water storage tank �Heavily jointed rock mass � Retaining pile wall �Block punch index test
1 Introduction
The potable water supply of a settlement is usually stored in
water tanks of different capacities. The dimensions of a tank
are related to the water demand, calculated with respect to
the population. A water tank both regulates the water
pressure in the network and reserves a water supply trans-
mitted from the source location. Furthermore, a water tank
should be adequately elevated in order to fully maintain the
hydraulic pressures required for potable water network
distribution. Thus, water storage tanks are mostly located on
hills or uneven terrain, and a cut slope is usually excavated
so as to construct the concrete tank on a flat surface.
Although slope stability problems concerning water storage
tanks are not very common during the construction or post-
construction period, fatal events may occur after such
incidents (Calderon et al. 2009). Water leakage from the
tank may considerably reduce the shear strength of the slope
material, leading to slope failures and catastrophic acci-
dents. Moreover, excavations to create cut slopes during
construction may trigger slope instabilities, which may also
affect the safety of nearby structures. The failure of slopes
and the substantial costs of remedial measures are mostly a
consequence of unsatisfactory geological and geotechnical
investigations and inadequate interpretation of acquired
data during preliminary design (Lee and Hencher 2009).
The construction of two reinforced concrete (RC) water
storage tanks started at the end of 2005 in Manisa, Turkey,
to store a portion of the potable water demand. The
M. Akin (&)
Department of Mining Engineering, Yuzuncu Yil University,
Zeve Campus, 65080 Van, Turkey
e-mail: [email protected]
123
Rock Mech Rock Eng (2013) 46:359–371
DOI 10.1007/s00603-012-0262-x
location map of the study area is shown in Fig. 1. The
latitude and longitude of the construction site are
533338.27 E and 4273272.91 N. The capacities of the tanks
are 3,000 and 7,500 m3, respectively. Both tanks were
constructed, one under the other, on steep terrain (slope
inclination C30�). The horizontal distance between the two
tanks is approximately 30 m. In addition, the elevation of
the 3,000-m3 tank (WT1) is 153.3 m, whereas the 7,500-m3
tank (WT2) is situated at 133.8 m. A general cross-section
of the construction area with the location of the two RC
water storage tanks is provided in Fig. 2. It should be noted
that the construction area was restricted by an expropria-
tion boundary and an unusually steep cut slope had to be
excavated to place the two structures in a tight area, due to
space limitations. After the excavation, the cut slope failed
and the failure affected the stability of WT1. Several
effective and ineffective remedial works were carried out
to retain the failed slope and WT1.
In this paper, the repeated failure of a cut slope and a series
of remedial works for slope stabilization are explained. Fur-
thermore, the slope failure which occurred after the con-
struction of the upper tank is back-analyzed to assess the shear
strength parameters of the slope-forming material. In addition,
the movements monitored after the construction of a retaining
pile wall is evaluated and, finally, the efficiency of the
placement of a granular fill buttress on the slope is analyzed.
2 Geology of the Study Area
The study area is situated on highly fractured and deformed
rocks of the Bornova Flysch Zone. The flysch zone com-
prises large blocks of Mesozoic limestone, basalt, serpen-
tine, and radiolarian chert with a highly disturbed clastic
matrix of Cretaceous to Paleocene age (Okay and Altiner
2007). Moreover, Neogene-aged yellowish brown marl
layers crop out in the same zone. The foundations of the two
water storage tanks as well as the cut slope exist in the
clastic matrix of the Bornova Flysch Zone. It consists of
gray graphitic shale and alternating beds of sandstone and
shale units (Fig. 3a, b). These units are closely jointed,
sheared, and folded with a chaotic structure (Fig. 3c). There
is no observed groundwater table in the study area, except
local wet zones after heavy rains. As the above-mentioned
geological units exhibit similar geotechnical properties, it is
quite difficult to differentiate the exact unit boundaries in
the construction area. The discontinuity spacing of the slope
material is mostly between 5 and 10 cm. As a particular
note, the discontinuity surfaces are usually smooth and
soapy, which drastically decreases their shear strength,
especially during heavy rains. There is no certain discon-
tinuity orientation, as the rock mass is heavily jointed. In
this highly deformed material, a circular slope failure is
more probable than a structurally controlled instability,
since there is no distinct discontinuity surface on which
failure can occur (Anderson and Richards 1987; Ozdemir
and Delikanli 2009; Sharifzadeh et al. 2010). The sliding
surface in heavily jointed rock masses involves both natural
discontinuities aligned on the sliding surface and some
shear failure through intact rock (Wyllie and Mah 2004).Fig. 1 Location map of the study area
Fig. 2 A general cross-section of the construction area (note: the
vertical scale is exaggerated)
360 M. Akin
123
3 History of the Slope Instabilities
A detailed geotechnical survey was not carried out on the
construction site during preliminary investigations in this
project. The physico-mechanical properties of the litho-
logical units were estimated by observational studies and
literature data without any laboratory or in situ tests.
In the preliminary stage of the project, the construction
of the 3,000-m3 tank (WT1) at the higher elevation was
started. When the tank construction was about to be com-
pleted, an approximately 20-m high cut slope with two
benches was excavated in front of WT1 to make room for
the foundation of the 7,500-m3 water storage tank (WT2) at
the lower elevation (Fig. 4). Having completed the slope
excavation, the foundation of WT2 and a drainage ditch
along the tank perimeter were excavated. One week after
excavation of the drainage ditch, the cut slope in front of
WT1 instantly failed. Major and progressive tension cracks
at the top bench and a small-scale horizontal movement at
the toe were observed after the failure (Fig. 4). The slip
surface shown in Fig. 4 is estimated considering the main
tension crack and the horizontal movement at the slope toe.
Eventually, the slope failure threatened the stability of
WT1. Also, a separation of several millimeters in scale
between the main tank and the maneuver room sections of
WT1 occurred and was monitored (Fig. 5).
As an immediate remedial measure to prevent slope
failure and to protect the stability of WT1, a granular toe
buttress was constructed and re-sloping was performed by
removing slope material from the crown to lower the
sliding forces (Fig. 6). Further slope movement was pre-
vented by the above-mentioned temporary remedial mea-
sures. However, the toe buttress covered a large portion of
the WT2 foundation and the temporary support had to be
removed in order to construct WT2.
4 Back Analysis of the Initial Slope Failure
The estimation of shear strength parameters along the
sliding surfaces is quite difficult in slope engineering
(Sonmez et al. 1998). The limit equilibrium back analysis
of a failed slope is one of the most reliable approaches to
determine the shear strength of slope material at the time of
failure (Sancio 1981; US Army Corps of Engineers 2003;
Topal and Akin 2009). The shear strength parameters
obtained by the back analysis of slopes are accepted as
being more consistent than those obtained by laboratory or
in situ testing during remedial measure design (Popescu
and Schaefer 2008). In conventional back analysis, the
internal friction angle or cohesion is assumed in order to
calculate the other parameter, considering a factor of safety
of 1.0. Although back analysis based on linear failure cri-
terion is mostly applied in soil slopes, the same procedure
can be followed on very weak rock mass, which is trans-
formed into a soil-like material as a consequence of
chemical weathering or alteration (Cai et al. 2007; Sha-
rifzadeh et al. 2010). On the other hand, in recent years, the
geotechnical characterization of homogeneous and isotro-
pic rock masses has mostly been performed using theFig. 3 Gray graphitic shale (a), the alternation of sandstone and shale
(b), a close-up view of heavily jointed slope material (c)
Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 361
123
geological strength index (GSI) system (Morales et al.
2004; Marinos et al. 2006). The GSI system proposed by
Hoek et al. (1995) allows the determination of rock mass
strength and deformation parameters for both hard and
weak rock masses.
The back calculation of shear strength parameters of
sliding surfaces using the linear Mohr–Coulomb criterion is
independent from normal stress. However, the failure
envelope of a closely jointed rock mass is non-linear and is
sensitive to normal stresses (Sonmez et al. 1998; Yang and
Yin 2004). The Hoek–Brown non-linear failure criterion
(Hoek and Brown 1980; Hoek et al. 2002) has been com-
monly employed for the back analysis of slope failures in
heavily jointed rock masses (Sonmez et al. 1998; Sonmez
and Ulusay 1999; Cai et al. 2007; Sharifzadeh et al. 2010).
The shear strength parameters of a failure surface in such
rock masses can be determined for a specific normal stress
using the material constants of the Hoek–Brown failure
criterion (m and s) as a function of the rock mass rating
(RMR) system or the GSI system.
Fig. 4 The 20-m high cut slope
with two benches and first slope
failure after excavation
(modified after GDBP 2006)
Fig. 5 Separation between the
main tank and the maneuver
room after slope failure (a side
view, b overhead view)
Fig. 6 Immediate slope
remediation after the first slope
failure (modified after GDBP
2006)
362 M. Akin
123
The non-linear Hoek–Brown failure criterion for
homogeneous and isotropic rock masses is defined by the
equation below:
r01 ¼ r03 þ rci� ½mb � ðr03=rci
Þ þ s�0:5 ð1Þ
where r01 and r03 are the maximum and minimum principal
effective stresses acting upon the sliding surface, rciis the
intact rock strength, and mb and s are the material
constants, which are determined by the following
formulas in accordance with the GSI:
mb ¼ mi � exp½ðGSI� 100Þ=ð28� 14DÞ� ð2Þs ¼ exp½ðGSI� 100Þ=ð9� 3DÞ� ð3Þ
where mi is the intact material constant and D is the dis-
turbance factor of rock mass due to blasting or excavation.
The GSI value can be directly determined in the field
based on site conditions, although sampling for laboratory
testing is extremely difficult in heavily jointed sedimentary
and metamorphic rock masses such as shale, flysch, and
schist. In addition, alternative procedures may be imple-
mented in order not to overestimate the mb and s values, as
overrated input parameters may lead to unrealistic results
in the slope stability back analysis using the non-linear
approach (Unal et al. 1992; Sonmez et al. 1998). The back
analysis of failed slopes using the GSI system can be
performed with a trial and error approach following the
procedure first presented by Sonmez et al. (1998). The
calculation steps are as follows:
(a) A GSI value called GSI(s) is assessed and the material
constant s is determined using Eq. 3.
(b) The material constant mb is calculated considering the
existing slope geometry and slip surface in limit
equilibrium software using the Hoek–Brown failure
criterion given in Eq. 1, which satisfies the limit
equilibrium condition (FOS = 1.0).
(c) The calculated material constant mb in the previous
step is employed in Eq. 2 and discloses the second
GSI value, named GSI(m).
(d) The calculation step is carried out for different values
of GSI(s) to obtain a variety of GSI(s) and GSI(m) pairs.
(e) The results are presented in a GSI(s)-GSI(m) graph
and a straight line passing from the origin with an
inclination of 45� is drawn. The inserted line inter-
sects the GSI(s)-GSI(m) curve at a certain point
identifying the GSI value of the investigated rock
mass (GSIRM).
Following the back analysis, the instantaneous cohesion
and internal friction angle along the existing failure surface
can be calculated by application of the non-linear Hoek–
Brown failure criterion, considering the normal stress and
the GSIRM value.
For the actual slope failure, the shear strength parame-
ters of the sliding surface mobilized at the time of failure
were estimated by means of back analysis using the non-
linear (GSI) approach. The slope stability back analyses
were conducted using the Slide v.5.0 software (Rocscience
Inc. 2002) and the slope geometry before the failure was
considered in the analyses (Fig. 7). In addition, the slope
was kept in dry conditions in the back analysis, since no
groundwater table was observed in the field and in bore-
holes drilled after the construction of the pile wall support.
Due to the impossibility of sampling in heavily jointed
rock mass, the uniaxial compressive strength (UCS) of the
slope material was determined through block punch index
(BPI) tests (Ulusay and Hudson 2007) using thin cylin-
drical slices of rock pieces from the slope material. The
calculated BPIc (corrected BPI) was then converted to the
UCS in accordance with the equations presented by Ulusay
and Hudson (2007). In the BPI test, thin cylindrical disc-
shaped specimens prepared from cores or blocks are put
into an apparatus which is designed to fit the well-known
point load device (Ulusay et al. 2001). The specimens are
loaded and forced to break by a rectangular rigid punching
block. In this study, disc slices used in the BPI tests were
drilled from rock blocks obtained from the investigation
area. The unit weight was also determined on the same
discs. Consequently, the average unit weight and UCS of
Fig. 7 Slope geometry before
the first failure considered in the
back analyses
Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 363
123
the intact slope material are 17.3 kN/m3 and 15.3 MPa,
respectively.
The GSI value of the rock mass studied was directly
determined in the field in accordance with the latest
quantitative GSI chart recommended by Sonmez and Ul-
usay (2002). More realistic GSI values can be estimated in
this chart by means of structure rating (SR) and surface
condition rating (SCR). The SR value is assigned based on
the volumetric joint count (Jv), whereas the SCR of the
discontinuities is calculated considering roughness,
weathering, and infilling parameters. The volumetric joint
count (Jv) of the slope material in the study area is around
21 with respect to in situ measurements. On the other hand,
discontinuity surfaces are generally smooth, highly
weathered, and contain soft clay infillings with a thickness
of\5 mm. Then, the SR and SCR values were found to be
27 and 4, respectively. The GSI value of the slope material
is 24, as seen in Fig. 8, indicating a blocky and disturbed
rock mass. It should be noted that the material constant mi
was selected as 4 in the back analysis with regards to the
recommended mi values for clastic rocks by Hoek et al.
(1995), because the triaxial test is almost impossible to
carry out on such rock types. Additionally, a disturbance
factor (D) value of 0.8 was employed in the back analysis
in accordance with Hoek et al. (2002), as the slope was
excavated mechanically and was subjected to a minor
disturbance due to stress relief from overburden removal.
The GSI(s)-GSI(m) graph obtained from the back anal-
ysis of the failed slope following the procedure proposed
by Sonmez et al. (1998) is illustrated in Fig. 9. The GSIRM
value of the failed slope was found to be 21, as seen in
Fig. 9. As the surface characteristics of discontinuities
were very poor and the slope material was tectonically
deformed, sheared, and jointed with a chaotic structure, the
GSIRM value of the rock mass assessed by back analysis is
reasonable and is compatible with the GSI value of 24
determined in the field.
The Hoek–Brown failure envelope of the slope-forming
rock mass was constructed using the GSI value of 24
determined in the field and the related material constants
(mb: 0.034, s: 8.5e-6, a: 0.5) calculated in accordance with
Eqs. 2 and 3 (Fig. 10). Based on the back analysis con-
sidering the pre-failure slope geometry and the location of
the sliding surface in Fig. 7, the normal and shear stresses
acting at the bottom of each slice of the observed failure
surface was calculated. These data pairs were plotted onto
the non-linear Hoek–Brown failure envelope of the
investigated rock mass, as depicted in Fig. 10. These
normal–shear stress plots mostly fall on the Hoek–Brown
failure envelope, indicating that the location of the sliding
surface, the estimated GSI value via back analysis, and the
rock mass parameters accurately represent the studied
failure.
The relationship between shear strength and normal
stress is more accurately represented by a non-linear
model. Furthermore, in the non-linear failure approach, the
shear strength parameters mobilizing on the failure surface
is normal stress-dependent. The instantaneous shear
strength parameters are obtained by the intercept and the
inclination, respectively, of the tangent to the non-linear
relationship between the shear strength and normal stress
(Hoek et al. 2002). In other words, the term ‘instantaneous’
indicates the shear strength parameters at a certain normal
stress level on the non-linear failure envelope.
Therefore, the instantaneous shear strength parameters
along the existing failure surface (ci and /i pairs) were
determined regarding the actual normal stress at the bottom
of each slice. In the analyzed slope, the variation of ci and
/i with different normal stresses is illustrated in Fig. 11.
The normal stress level on the actual sliding surface attains
a maximum value of 130 kPa. On the other hand, the
instantaneous cohesion varies between 6 and 28 kPa,
whereas the instantaneous internal friction angle changes
between 21� and 50�.
5 First Remedial Measure: Application
of the Retaining Pile Wall
After the first failure, it was planned to construct retaining
piles in front of WT1 in order to stabilize the constructed
tank. Therefore, both the safety of WT1 would be provided
and the toe buttress on the foundation of WT2 would be
removed. In addition to retaining piles, a new re-sloping was
also performed by lowering the slope angle of benches, to
decrease sliding forces. The shear strength parameters were
assessed for the pile design due to the presence of insuffi-
cient data for the slope material. Two different material
zones were distinguished during the design phase. The first
zone on the upper level of the slope was represented by
disturbed material which was affected from the first slide.
The second zone under the first subdivision was the undis-
turbed section of the rest of the slope. The slope material
parameters used for the design of the retaining piles are
presented in Table 1 (GDBP 2006). It should be kept in
mind that the shear strength values in Table 1 are not related
to the back analyses performed in this study. The new slope
model with the retaining pile wall is presented in Fig. 12.
As shown in Fig. 12, the new cut slope between the two
water storage tanks has three benches with lower inclina-
tions (54�–57�). Furthermore, 15.6-m long RC piles
(diameter 80 cm) without any anchors were proposed to
support WT1. The axial distance(s) between each pile is
1.60 m. After the preliminary design, a contiguous bored
pile wall was constructed in accordance with the submitted
support model. The slope was then re-excavated with three
364 M. Akin
123
benches. Subsequent to concrete pile wall construction and
re-sloping, the buttress at the slope toe was removed.
Additionally, a total of three 22-m deep boreholes were
drilled between WT1 and the retaining pile wall. Each hole
was cased with an inclinometer casing to monitor probable
lateral movements in the slope and the retaining pile wall.
Fig. 8 Determination of the GSI value of the slope material using the proposed chart by Sonmez and Ulusay (2002)
Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 365
123
6 Second Slope Failure after Pile Wall Construction
One week after finishing the retaining pile wall construc-
tion and the removal of the toe buttress (22.04.2006), a
lateral movement along the longitudinal slope axis (parallel
to the failure direction) was noticed by inclinometer mea-
surements. Fifteen days following the first inclinometer
measurement (07.05.2006), the lateral movement attained a
maximum value of 10 mm. The inclinometer data indi-
cated that the slope in front of the retaining pile wall was
still unstable and that the slope was still moving along the
same failure surface (Fig. 13). A large-scale tension crack
was also observed on the slope benches as an obvious sign
of slope instability (Fig. 14a). In addition to the tension
crack, progressive small-scale cracks occurred adjacent to
WT1 (Fig. 14b). A granular toe buttress was once again
placed immediately to stabilize the slope. The lateral
movement in the slope was prevented after this toe buttress
application (11.05.2006), according to the inclinometer
measurements shown in Fig. 13.
The second failure after the retaining pile wall con-
struction indicated that the support was not sufficient to
provide stability for the cut slope and WT1. The slope
material facing the retaining pile wall was the only
resisting force for the bending moments on the anchorless
piles. The release of the resisting slope material after the
failure resulted in a lateral movement towards the longi-
tudinal slope axis in the retaining piles, due to the lateral
earth pressure on the backs of the piles. Besides, when the
cumulative displacement graph in Fig. 13 is observed, it
can be clearly seen that the first lateral movement started
almost from the bottom of the piles (around 12 m). Finally,
it can be concluded that the decrease of resisting forces
acting on the pile wall after the second slide caused sig-
nificant pile displacements in the contiguous retaining piles
without any anchorage.
7 Final Solution for Stability: Permanent Toe Buttress
It was of great importance to maintain the permanent sta-
bility of WT1 on the upper elevation after the first slope
failure. However, the constructed pile wall support was
unsatisfactory for slope stabilization. Therefore, an
improved solution that would result in a factor of safety
sufficient to resist additional slope movements was
implemented. Toe counterweights and buttresses are gen-
erally efficient for the mitigation of slope instability (Rowe
2001). The application of a temporary toe buttress after the
first and second slides prevented additional slope dis-
placements. Therefore, a larger buttress was constructed
Fig. 9 GSI(s)-GSI(m) graph obtained from the back analysis of the
failed slope using the non-linear approach
Fig. 10 Hoek–Brown failure envelope of the studied rock mass and
the normal–shear stress pairs acting on the observed failure surface
calculated by means of back analysis
Fig. 11 Instantaneous shear strength parameters along the existing
failure surface (ci and /i) graph
366 M. Akin
123
for efficient stabilization. However, the water storage tank
at the lower elevation (WT2) had to be shifted about 10 m
in the direction opposite to the longitudinal slope axis to
make room for the toe buttress. As previously mentioned,
the construction area was restricted by an expropriation
boundary which made the shifting quite impossible. Hence,
the expropriation boundary was officially enlarged by the
municipal council to create extra space. Consequently, the
site was expanded, which permitted the construction of
both the toe buttress and WT2.
The final slope geometry with a granular buttress is
depicted in Fig. 15. It is important to notice that the new
buttress entirely covers the slope benches and applies a
higher resisting force. No lateral displacements were
observed in the ongoing inclinometer measurements after
the installation of the new buttress. Having completed the
installation of the new buttress on the slope, the larger RC
water storage tank at the lower elevation (WT2) was con-
structed in front of the new support (Fig. 16).
8 Long-Term Stability Assessment
8.1 Estimation of Peak Ground Acceleration
The long-term stability of the final slope geometry was also
analyzed by the slope stability analysis considering the
seismic effect in this study. The project area is located in a
seismically active zone in the Western Anatolia Region. A
Table 1 Specific slope material parameters used for the retaining pile wall design (GDBP 2006)
Unit weight (cn)
(kN/m3)
Cohesion
(c) (kPa)
Internal friction
angle (/) (�)
Modulus of elasticity
(E) (MPa)
Poisson’s
ratio (t)
Zone 1 (disturbed
material)
19 24 27 200 0.35
Zone 2 (undisturbed
material)
19 35 28 500 0.35
Fig. 12 New slope model with
retaining piles and re-sloping
(modified after GDBP 2006)
Fig. 13 Cumulative displacement graph from inclinometer 1 (paral-
lel to the longitudinal slope axis) (modified after GDBP 2006)
Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 367
123
significant extensional regime in this region resulted in
numerous normal faults and graben systems (Bozkurt
2001). The Manisa Fault exists in the very close vicinity of
the project area. This normal fault is about 40 km in length
and lies in the southern margin of Manisa city (Fig. 17).
Although a moment magnitude (Mw) of 5.2 was recorded in
1994 in this fault segment (Emre et al. 2005), Kayabali and
Akin (2003) and Ulusay et al. (2004) assigned values of 7.2
and 7.4, respectively, to the Gediz Graben which is 150 km
long in total and is the main tectonic feature around the
study area. Therefore, the maximum expected earthquake
with a moment magnitude of 7.4 was considered in the
long-term stability assessment in this study. Additionally,
the epicentral distance (Re) to the main segment of the
Gediz Graben is around 25 km.
The peak ground acceleration in the project area was
evaluated by two different regional attenuation relation-
ships of Ulusay et al. (2004) and Kayabali and Beyaz
(2011), given in Eqs. 4 and 5, respectively. In these
equations, PGA is the peak ground acceleration (cm/s2),
Fig. 14 Slope failure-related problems after the construction of the retaining pile wall (a tension crack on slope benches, b small-scale crack
near the water tank, c retaining pile wall and slope, d inclinometer casing between WT1 and retaining pile wall)
Fig. 15 Final slope geometry
with a granular toe buttress
(modified after GDBP 2006)
368 M. Akin
123
Fig. 16 Final slope design
(a WT1 and upper buttress,
b WT2 and lower buttress,
c complete view of WT2)
Fig. 17 Simplified map of
graben systems around the study
area (modified from Bozkurt
2001)
Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 369
123
Mw is the moment magnitude, and Re is the epicentral
distance (km):
PGA ¼ 2:18e0:0218ð33:3Mw�ReÞ ð4Þ
log PGA ¼ 2:08þ ð0:0254M2wÞ þ ð�1:001 logðRe þ 1ÞÞ
ð5Þ
The attenuation relationship proposed by Ulusay et al.
(2004) resulted in a peak ground acceleration of 272 cm/s2,
whereas the relationship of Kayabali and Beyaz (2011)
resulted in a PGA of 113 cm/s2. Therefore, the maximum
peak ground acceleration (272 cm/s2) determined by
Ulusay et al. (2004) was accepted for the project area.
8.2 Determination of the Seismic Coefficient
In the seismic slope stability analysis, the determination of
the seismic load acting on the analyzed slope is of great
importance. A pseudostatic approach is mostly used in
seismic slope stability analysis, where the effects of an
earthquake are represented by constant vertical and/or
horizontal accelerations (Kramer 1996). Appropriate
pseudostatic coefficients should be selected, as the seismic
coefficient is a measure of the pseudostatic force on the
slope. However, there are no certain rules for the deter-
mination of the pseudostatic coefficient in the literature
(Kramer 1996). Hynes-Griffin and Franklin (1984) sug-
gested that appropriate pseudostatic coefficients for earth
slopes should be one-half of the peak ground acceleration.
For this reason, a maximum of 136 cm/s2 horizontal seis-
mic load (one-half of 272 cm/s2) on the analyzed slope is
taken into consideration in this study.
8.3 Long-Term Stability
Long-term stability of the analyzed rock mass should be
maintained, as the close vicinity of the project area is
surrounded by residential places and a slope failure
may lead to both economic and human loss due to a
significant overflow from water storage tanks. Therefore,
the long-term stability of the supported slope design was
investigated using the rock mass parameters determined
by back analysis. The Slide v.5.0 software (Rocscience
Inc. 2002) was employed during analysis. It should be
noted that the non-linear failure criterion was taken into
consideration. The factor of safety of the final slope
design in static conditions is 1.95. In seismic conditions,
considering a maximum of 136 cm/s2 horizontal seismic
load, the factor of safety decreases to 1.52 (Fig. 18). The
calculated safety factor is acceptable even in seismic
conditions, considering the degree of risk in the project
area.
9 Conclusions
In this paper, the repeated failure of an excavated slope in
heavily jointed shale and sandstone units with a chaotic
structure was evaluated via back analysis considering the
non-linear approach. When compared with field estima-
tions, the geological strength index (GSI) value obtained by
back analysis yields satisfactory results. Furthermore, the
estimated failure surface of the analyzed slope was verified
by comparing normal–shear stress plots versus the Hoek–
Brown failure envelope derived from the field-based GSI
value. It should be kept in mind that the shear strength
parameters are normal stress-dependent in such closely
jointed rock masses and the non-linear failure approach
gives more realistic results. Therefore, assigning specific
shear strength parameters during the design phase may lead
to excessive work and insufficient remedial measures in
such slope stability problems. Finally, the slope design
with permanent granular counterweight seems to be quite
stable in accordance with the limit equilibrium analysis
performed using the non-linear approach in this study. The
most important message derived from this case study is that
proper engineering is important to avoid failure of engi-
neering structures.
Fig. 18 Stability analysis of the
final slope supported by a
permanent granular buttress
370 M. Akin
123
Acknowledgments The author is grateful to the General Directorate
of the Bank of Provinces (GDBP) for providing the necessary tech-
nical information about the project. The author also thanks Dr. Samad
Joshani-Shirvan and Dr. Margaret Sonmez for their comments on the
use of language. The author would like to express his sincerest
gratitude to Prof. Dr. Resat Ulusay for his valuable comments and
assistance during the block punch index (BPI) tests. Akademi Soil and
the Rock Mechanics Laboratory deserve thanks for the sample
preparation before BPI testing. Thanks are due to the anonymous
reviewers for their valuable and constructive comments.
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