Dissertation2007-Asinari

8/13/2019 Dissertation2007-Asinari

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I.U.S.S.

Istituto Universitario

di Studi Superiori

Università degli

Studi di Pavia

EUROPEAN SCHOOL FOR ADVANCED STUDIES IN

REDUCTION OF SEISMIC RISK

ROSE SCHOOL

“BUILDINGS WITH STRUCTURAL MANSONRY WALLS

CONNECTED TO TIE-COLUMNS AND BOND-BEAMS”

A Dissertation Submitted in Partial

Fulfilment of the Requirements for the Master Degree in

EARTHQUAKE ENGINEERING

by

MARIANA ASINARI

Supervisor: Prof. GUIDO MAGENES, Dr. ANDREA PENNA

February, 2007



The dissertation entitled “Buildings with structural masonry walls connected to tie-columns

and bond-beams”, by Mariana Asinari, has been approved in partial fulfilment of the

requirements for the Master Degree in Earthquake Engineering.

Name of Reviewer 1 Prof. Guido Magenes

Name of Reviewer 2 Dr. Andrea Penna



Abstract

i

ABSTRACT

Confined masonry is extensively used in seismic regions around the world. Experimental data about

confined masonry are very scarce and this lack of knowledge affects the seismic safety and the design

practice of masonry structures.

This type of constructions consist basically of masonry panels confined by vertical and horizontal

elements usually of reinforce concrete. This confinement enhances greatly the connection between

structural walls, improves the stability and the strength, provides ductility under earthquake loading

and improves the integrity and containment of earthquake damage in masonry walls.

The present dissertation concerns a general review on confined masonry structures, ranging from

current and past research, taking as a reference the experimental data available in the literature. Failure

and resisting mechanisms are described. Vulnerability and experimental tests in confined masonry are

presented as well. Finally some code recommendations, of different countries, for a proper

construction and resistance verification are given.

Keywords: confined masonry; tie-columns; bond-beams; failures modes; resisting mechanisms



Acknowledgements

ii

ACKNOWLEDGEMENTS

I would like to acknowledge:

• The financial support of the Erasmus Mundus Programme,

• The help, suggestions and support of Dr. Andrea Penna and Professor Guido Magenes,

especially for founding me this research project that may help with the continuities of my studies and

professional life,

• The cooperation with the available experimental data and suggestions of Professor Alfredo

Payer and Professor Carlos Prato,

• The help and cooperation of Saverio,

• The support and sacrifice of my parents José Luis and Beatriz being so far form each other for

so long,

• The support and love of my sisters Cecilia and Florencia,

• The comprehension, patience, support and love of Andrés,

• The great and unforgettable time in Europe with my new friends from all over the world of the

meees programme: Paola, Francisco, Bin, Oil, Jessie, Rena, Nelson, Daniele, Davide x 2, Marco,

Michael and Gopal.



Index

iii

TABLE OF CONTENTS

Page

ABSTRACT ............................................................................................................................................i

ACKNOWLEDGEMENTS....................................................................................................................ii

TABLE OF CONTENTS ......................................................................................................................iii

LIST OF FIGURES ................................................................................................................................v

LIST OF TABLES.................................................................................................................................xi

LIST OF SYMBOLS...........................................................................................................................xiii

1. INTRODUCTION.............................................................................................................................1

1.1 Objectives ..................................................................................................................................1

1.2 Description.................................................................................................................................2

1.3 Construction procedure..............................................................................................................5

1.3.1 Materials for confined masonry construction ..................................................................5

1.3.2 Construction procedure....................................................................................................8

2. FAILURE MECHANISMS AND PERFORMANCE IN PAST EARTHQUAKES ......................12

2.1 Failure mechanisms .................................................................................................................12

2.2 Ductility ...................................................................................................................................15

2.3 Predominant design failures in confined masonry during earthquakes....................................17

2.4 Vulnerability: performance in past earthquakes ......................................................................19

3. RESISTING MECHANISMS .........................................................................................................30

3.1 Resisting mechanism ...............................................................................................................30

4. EXPERIMENTAL TESTS..............................................................................................................34

4.1 Dynamic behaviour of confined masonry buildings through shaking table tests ....................34

4.1.1 Assessment of the response of Mexican confined masonry structure through shaking

table test, Alcocer et al [27] .....................................................................................................34 4.1.2 Seismic behaviour of confined masonry, Tomazevic et al. [28]....................................39



Index

iv

4.1.3 Seismic behaviour of a three-story half scale confined masonry structure, San Bartolomé

et al. [29] 48

4.1.4 Pseudo dynamic tests of confined masonry buildings, Scaletti et al. [30].....................51

4.2 Dynamic behaviour of confined masonry panels under cyclic lateral loads............................55

4.2.1 Experimental behaviour of masonry structural walls used in Argentina, Zabala et al. [5]

55

4.2.2 Behaviour of multi-perforated clay brick walls under earthquake type loading, Alcocer

and Zepeda [31] .......................................................................................................................60

4.2.3 Experimental investigation of the seismic behaviour in full- scale prototypes of confined

masonry walls, Decanini et al. [32]..........................................................................................64

4.2.4 Influence of vertical and horizontal reinforcement: Influence of the tie-column vertical

reinforcement ratio on the seismic behaviour, Irimies [33].....................................................70

4.2.5 Influence of openings in the behaviour of confined masonry: Behaviour of confined

masonry shear walls with large openings, Yáñez et al. [8]......................................................72

4.2.6 Influence of the number and spacing of confining tie-columns: Experimental evaluation

of confined masonry walls with several confining columns, Marinilli and Castilla [34] ........75

4.2.7 Experimental study on effects of height of lateral forces, column reinforcement and wall

reinforcements on seismic behaviour of confined masonry walls, Yoshimura et al. [11] .......77

4.2.8 Effects of vertical and horizontal wall reinforcement on seismic behaviour of confined

masonry walls, Yoshimura et al. [13] ......................................................................................81

4.2.9 Experimental study for developing higher seismic performance of brick masonry walls,

Yoshimura et al. [12] ...............................................................................................................83

4.2.10 Experimental study on earthquake-resistant design of confined masonry structures,

Ishibashi et al. [35]...................................................................................................................86

4.3 Concluding remarks .................................................................................................................89

5. CODE RECOMMENDATIONS.....................................................................................................91

5.1 Quality of masonry ..................................................................................................................91

5.2 Classification of the structural walls........................................................................................94

5.3 Confined masonry....................................................................................................................99

5.4 Resistance verification...........................................................................................................107

5.5 “Simplified method” allowed by the Argentinean code ........................................................111

5.6 Comparison between codes....................................................................................................118

5.7 Conclusions and possible topics to develop...........................................................................125

1. REFERENCES ..............................................................................................................................128

2. ANNEX .............................................................................................................................................1



Index

v

LIST OF FIGURES

Page

Figure 1.2.1 Uses of masonry in Argentina, 1985. 1) Confined masonry and 2) RC frames with

masonry infill; Decanini, Payer and Terzariol [4]...........................................................................2

Figure 1.2.2 Solid brick masonry confined with tie-columns and bond-beams, Kuldeep Virdi [3]........4

Figure 1.3.1 Types of masonry units, Kuldeep Virdi [3]........................................................................5

Figure 1.3.2 Determination of the mortar compressive strength, Bustos [10]........................................7

Figure 1.3.3 Typical anchorages of the reinforcing bars according to EC 6 Kuldeep Virdi [3]............. 8

Figure 1.3.4 Foundations construction and start of the vertical concrete columns.................................8

Figure 1.3.5 Construction of the masonry panel.....................................................................................6

Figure 1.3.6 Arrangement of vertical reinforcement in tie-columns ......................................................6

Figure 1.3.7 Position of the horizontal reinforcement ............................................................................6

Figure 1.3.8 Concrete poured against the boundaries of the masonry panel ..........................................6

Figure 1.3.9 Confined masonry walls under construction Mexico, 1993. Yoshimura et al. [11]........... 8

Figure 1.3.10 Left: confined masonry walls under construction (Jimo, P.R. China, 1999); right:

confined masonry walls under construction (El Salvador, 2001), Yoshimura et al. [12] ...............8

Figure 1.3.11 Confined masonry walls using hollow concrete block masonry units. Las Losas Project

under construction in Villahermosa, Mexico, Yoshimura, Kikuchi, Okamoto and Sanchez [13] ..9 Figure 1.3.12 Confined masonry of clay bricks deposit under construction in Córdoba, Argentina......9

Figure 1.3.13 Confined masonry of clay bricks deposit under construction in Córdoba, Argentina....10

Figure 1.3.14 Construction problems from Blondet et al. [14].............................................................11

Figure 2.1.1 Flexural failure, Zabala et al. [5] .....................................................................................13

Figure 2.1.2 Diagonal cracking under cycling loading, Zabala et al. [5]..............................................13

Figure 2.1.3 Bad connection between horizontal and vertical reinforcement, Universidad Nacional de

Córdoba [16] .................................................................................................................................14

Figure 2.1.4 Compression of the diagonal, Universidad Nacional de Córdoba [16]............................14



Index

vi

Figure 2.1.5 Occurrence of the different failure modes in confined masonry. a) Compression failure; b)

Diagonal crack; c) Flexural failure; Bustos [10]...........................................................................15

Figure 2.2.1 Different ductilities induced by different typologies of masonry, Bustos, Universidad

Nacional de San Juan, Argentina [10]...........................................................................................16

Figure 2.2.2 Ductile response of the confined masonry structures, Decanini and Payer [17] ..............17

Figure 2.3.1 Failure caused by insufficient anchorage of reinforcement in the confinement elements.

Examples of bad disposition of the reinforcement, Decanini and Payer [17]...............................18

Figure 2.3.2 Typical period range in confined masonry constructions. Statistic values for different

types of soils; Decanini and Payer [17].........................................................................................19

Figure 2.4.1 Damage to reinforced concrete column in confined masonry wall due to 1999 Colombia

earthquake, Yoshimura et al. [12].................................................................................................20

Figure 2.4.2 Cracks observed in confined masonry after the Oaxaca earthquake 1999, López Bátiz et

al. [19] ...........................................................................................................................................21

Figure 2.4.3 Damage resulting from an inadequate distribution of the confining elements, López Bátiz

et al. [19] .......................................................................................................................................21

Figure 2.4.4 Damage in a hospital during the Oaxaca earthquake in Mexico, López Bátiz et al. [19] .22

Figure 2.4.5 Failure of a hollow concrete-block masonry wall. The hollow concrete block units are

separated from the RC confining column, Yoshimura and Kuroki [20] .......................................23

Figure 2.4.6 Damage to a confined clay-brick masonry wall in Usulutan, Yoshimura and Kuroki [20]

.......................................................................................................................................................23

Figure 2.4.7 Photograph illustrating typical damage in confined masonry, 1996 Nazca earthquake,

Loaiza and Blondet [21]................................................................................................................24

Figure 2.4.8 Undamaged recently constructed reinforced masonry dwelling in Bermejo, 60km south-

southeast of the epicenter, Rojahn, Brogan and Slemmons [22]...................................................25

Figure 2.4.9 Damage in masonry in Caucete earthquake, 1977 Argentina. Decanini, Payer and

Terzariol [4] ..................................................................................................................................26

Figure 2.4.10 Vertical and horizontal confining elements maintain the stability of the building,

Kooroush Nasrollahzadeh Nesheli [24] ........................................................................................28

Figure 2.4.11 Confined masonry wall in Iran that survive the earthquake, Usam Ghaidan [25]..........28

Figure 2.4.12 Damage to masonry building in the 1998 Mionica earthquake, Nikola Muravljov,

Radovan Dimitrijevic [26] ............................................................................................................29

Figure 3.1.1 Left: Distribution of seismic loads in the building; Right: tension originated by the

gravitational loads before the earthquake, Bustos [10].................................................................30

Figure 3.1.2 Tensions in the confined masonry wall originated by gravitational and seismic loads

during the earthquake, Bustos [10] ...............................................................................................31

Figure 3.1.3 Resisting mechanisms, Universidad Nacional de Córdoba [16] ......................................31



Index

vii

Figure 3.1.4 Resisting mechanisms of confined masonry walls, Universidad Nacional de Córdoba [16]

.......................................................................................................................................................32

Figure 3.1.5 Non-deformable diaphragms and good connection between walls allowing the correct

distribution of the seismic action. Decanini and Payer [17]..........................................................33

Figure 3.1.6 Deformable slab and no capacity of load distribution. Consequence: An important

bending moment in wall M2 is generated. Decanini and Payer [17] ............................................33

Figure 4.1.1 Characteristics of the Specimens, Alcocer et al. [27].......................................................35

Figure 4.1.2 Reinforcement of the Specimens, Alcocer et al. [27].......................................................36

Figure 4.1.3 Final cracks patterns, Alcocer et al. [27]...........................................................................37

Figure 4.1.4 Response envelope for M1 and M3; MCBC: Mexico City Building Code, Alcocer et al.

[27]................................................................................................................................................37

Figure 4.1.5 Typical floor plan of prototype building, used as a basis for the design of 1:5 scale

models, Tomazevic et al. [28].......................................................................................................40

Figure 4.1.6 Typical section prototype building, used as a basis for the design of 1:5 scale models,

Tomazevic et al. [28].....................................................................................................................40

Figure 4.1.7 Reinforcement of floor slabs and vertical and horizontal bonding elements, Tomazevic et

al. [28] ...........................................................................................................................................41

Figure 4.1.8 Earthquake simulator set-up, Tomazevic et al. [28].........................................................42

Figure 4.1.9 Left: instrumentation of models: accelerometers and LVDT-s on model M1; Right: strain

gauges on reinforcing steel of vertical confinement of model M1, Tomazevic et al. [28] ...........42

Figure 4.1.10 Left: instrumentation of models: accelerometers and LVDT-s on model M2; Right:

strain gauges on reinforcing steel of vertical confinement of model M2, Tomazevic et al. [28]..43

Figure 4.1.11 Model M1, northern side-propagation of cracks at the eastern corner, Tomazevic et al.

[28]................................................................................................................................................44

Figure 4.1.12 Left: Model M1: middle pier after test run R200; Right: Model M1: detail of damage to

tie-column after test run R200, Tomazevic et al. [28] ..................................................................45

Figure 4.1.13 Model M1, southern side, cracks after test runs R100, R150 and R200, Tomazevic et al.

[28]................................................................................................................................................45

Figure 4.1.14 Model M2: mechanism of collapse, Tomazevic et al. [28] ............................................47

Figure 4.1.15 Model M2: mechanism of collapse, Tomazevic et al. [28] ...........................................47

Figure 4.1.16 Geometry of the 3-storey confined masonry specimen, San Bartolomé et al. [29]........48

Figure 4.1.17 Specimen after run C, San Bartolomé et al. [29]............................................................50

Figure 4.1.18 Left: Total base shear force vs. displacement at level 1 in run C; Right: Lateral force in

one wall at the time of maximum base shear force at each run (A, B and C), San Bartolomé et al.

[29]................................................................................................................................................51

Figure 4.1.19 Test specimen, Scaletti et al. [30]...................................................................................51



Index

viii

Figure 4.1.20 Left: resonance curves for full scale specimen; Right: natural periods, frequencies,

damping and modal shapes; Scaletti et al. [30].............................................................................52

Figure 4.1.21 Left: input signal for PD test of half scale model; Right: input signal for PD test of full

scale model; Scaletti et al. [30] .....................................................................................................53

Figure 4.1.22 Left: base shear vs first story displacement, pseudo dynamic test of half scale model;

Right: envelopes of base shear vs first story displacement of the half scale model; Scaletti et al.

[30]................................................................................................................................................54

Figure 4.1.23 Left: first story displacement time histories of the full scale specimen; Right: base shear

time histories of the full scale specimen, Scaletti et al. [30].........................................................55

Figure 4.2.1 Model Dimensions, Zabala et al. [5] ................................................................................56

Figure 4.2.2 Outline of the test setup and its instrumentation, Zabala et al. [5] ....................................57

Figure 4.2.3 Shear failure in column of wall 1, Zabala et al. [5]..........................................................58

Figure 4.2.4 Crack pattern developed in the first 4 testing walls, Zabala et al. [5] ..............................59

Figure 4.2.5 Crack pattern developed in walls 5 and 6, Zabala et al. [5] .............................................59

Figure 4.2.6 Characteristics of the specimens, Alcocer and Zepeda [31].............................................61

Figure 4.2.7 Final cracks patterns of the fourth specimens, Alcocer and Zepeda [31].........................63

Figure 4.2.8 Hysteretic curves, Alcocer and Zepeda [31] .....................................................................64

Figure 4.2.9 General dimensions of the prototypes, Decanini et al. [32] .............................................65

Figure 4.2.10 Left: Reinforcement of confined masonry of solid clay bricks; Right: reinforcement of

confined masonry of hollow clay bricks, Decanini et al. [32] ......................................................66

Figure 4.2.11 Test setup and its instrumentation, Decanini et al. [32] ..................................................67

Figure 4.2.12 Initial and ultimate cracks of the testing wall M3, Decanini et al. [32] .........................69

Figure 4.2.13 Experimental models, Irimies [33].................................................................................70

Figure 4.2.14 Damage patterns of walls WC1, Irimies [33] .................................................................71

Figure 4.2.15 Damage patterns of walls WC2, Irimies [33] .................................................................72

Figure 4.2.16 Wall dimensions, Yáñez et al. [8] ..................................................................................73

Figure 4.2.17 Failure Mechanisms, Yáñez et al. [8] .............................................................................74

Figure 4.2.18 Configuration of specimens M1, M2, M3 and M4, Marinilli and Castilla [34] .............75

Figure 4.2.19 Specimens M1 and M2 after testing, Marinilli and Castilla [34] ...................................76

Figure 4.2.20 Specimens M3 and M4 after testing, Marinilli and Castilla [34] ...................................76

Figure 4.2.21 Test setup, Yoshimura et al. [11]....................................................................................79

Figure 4.2.22 a) Specimens with aspect ratio (ho/lo) of 1.51; b) Specimens with aspect ratio (ho/lo)

of 0.84; c) Specimens with aspect ratio (ho/lo) of 0.69, Yoshimura et al. [11] ............................80

Figure 4.2.23 Crack patterns of the specimens, Yoshimura [13]..........................................................83

Figure 4.2.24 Final crack pattern, Yoshimura et al. [12] ......................................................................85

Figure 4.2.25 Specimens details, Ishibashi et al. [35] ..........................................................................87



Index

ix

Figure 4.2.26 Loading history, Ishibashi et al. [35]..............................................................................87

Figure 4.2.27 Response of specimens WW, WBW and WWW, Ishibashi et al. [35] ..........................89

Figure 5.1.1 Tests on brick piers to determine the compressive strength, Bustos [10] ........................92

Figure 5.1.2 Masonry probes under diagonal compression to determine the strength, INPRES-

CIRSOC 103 code [7] and Yáñez et al. [8]...................................................................................93

Figure 5.2.1 Confined masonry. Left: reinforce confined masonry; right: confined masonry.

Universidad Nacional de Cordoba [16].........................................................................................95

Figure 5.2.2 Different options of reinforced masonry, Decanini and Payer [17] .................................95

Figure 5.2.3 Minimum dimensions of confined masonry panels with two constraints, Universidad

Nacional de Cordoba [16] .............................................................................................................96

Figure 5.2.4 Minimum dimensions of confined masonry panels with three or more constraints,

Universidad Nacional de Cordoba [16].........................................................................................97

Figure 5.2.5 Structural walls distribution in plan, Kuldeep Virdi [3]...................................................98

Figure 5.2.6 Irregular configurations in plan should be separated in regular potions, Kuldeep Virdi [3]

.......................................................................................................................................................99

Figure 5.3.1 Detail of RC bond-beam showing splicing of re-bars at wall corners, Kuldeep Virdi [3]

.....................................................................................................................................................102

Figure 5.3.2 Dimensions recommended by INPRES-CIRSOC 103 code for tie-columns and bond-

beams, Universidad Nacional de Cordoba [16] ..........................................................................102

Figure 5.3.3 Hoops in critic zones (near corners) and in normal zones, Universidad Nacional de

Cordoba [16] ...............................................................................................................................104

Figure 5.3.4 Recommended details in masonry wall connection in Argentina [16]...........................105

Figure 5.3.5 Recommended details in masonry wall to RC column connection in P.R. China,

Yoshimura et al. [12] ..................................................................................................................105

Figure 5.3.6 Construction of tie-column for confined brick masonry house, Kuldeep Virdi [3] .......107

Figure 5.5.1 Gravity loads, INPRES-CIRSOC 103 [7] ......................................................................111

Figure 5.5.2 Seismic coefficient of design, INPRES-CIRSOC 103 [7] .............................................111

Figure 5.5.3 Determination of torsion moments and shears of each story, INPRES-CIRSOC 103 [7]

.....................................................................................................................................................112

Figure 5.5.4 Determination of the elastic constants, INPRES-CIRSOC 103 [7]................................112

Figure 5.5.5 Geometric characteristics, INPRES-CIRSOC 103 [7] ...................................................113

Figure 5.5.6 Wall stiffness, INPRES-CIRSOC 103 [7]......................................................................113

Figure 5.5.7 Total design shear for each story, INPRES-CIRSOC 103 [7]........................................114

Figure 5.5.8 Design bending moment of each wall for each story, INPRES-CIRSOC 103 [7] .........114

Figure 5.5.9 Design normal resistance for each wall, INPRES-CIRSOC 103 [7]..............................115

Figure 5.5.10 Verification of shear strength, INPRES-CIRSOC 103 [7] ...........................................115



Index

x

Figure 5.5.11 Verification of gravitational loads, INPRES-CIRSOC 103 [7]....................................116

Figure 5.5.12 Reinforcement dimensions of bond-beams, INPRES-CIRSOC 103 [7] ......................116

Figure 5.5.13 Reinforcement dimensions of tie-columns, INPRES-CIRSOC 103 [7].......................117

Figure 5.5.14 Verification of flexion and compression, INPRES-CIRSOC 103 [7] ..........................117



Index

xi

LIST OF TABLES

Page

Table 1.3.1 Strength of the masonry units, INPRES-CIRSOC 103 [7]..................................................6

Table 1.3.2 Typical prescribed composition and strength of general purpose mortars, Kuldeep Virdi

[3]....................................................................................................................................................7

Table 1.3.3 Constituents of the mortar joints given by the INPRES-CIRSOC 103 [7]..........................7

Table 4.1.1 Measured response characteristics, Alcocer et al. [27].......................................................38

Table 4.1.2 Characteristic parameters of shaking-table motion recorded during individual test runs,

Tomazevic et al. [28].....................................................................................................................43

Table 4.1.3 Assumed force distribution in one specimen wall, San Bartolomé et al. [29]................... 49

Table 4.1.4 Shaking table test runs, San Bartolomé et al. [29].............................................................49

Table 4.1.5 First floor displacement, base shear and predominant period as a function of maximum

ground acceleration, Scaletti et al. [30].........................................................................................54

Table 4.2.1 Main features of the six tested walls, Zabala et al. [5] .....................................................58

Table 4.2.2 Measured loads and angular deformations, Decanini et al. [32]........................................68

Table 4.2.3 Properties of the system, Marinilli and Castilla [34] .........................................................77

Table 4.2.4 Test specimens, Yoshimura et al. [11]...............................................................................78

Table 4.2.5 Predicted and observed ultimate lateral strengths and failure modes, Yoshimura et al. [11].......................................................................................................................................................81

Table 4.2.6 List of specimens, Yoshimura et al. [13] ...........................................................................82

Table 4.2.7 Listed of the tested specimens, Yoshimura et al. [12] .......................................................84

Table 5.1.1 Values of compressive strength for different masonry units and mortar joints, INPRES-

CIRSOC 103 code [7] ...................................................................................................................92

Table 5.1.2 Values of shear strength for different masonry units and mortar joints, INPRES-CIRSOC

[7]..................................................................................................................................................94

Table 5.2.1 Maximum heights and number of stories allowed by the INPRES-CIRSOC 103 code [7]

.......................................................................................................................................................97



Index

xii

Table 5.3.1 Maximum area and dimension for confined masonry panels, INPRES-CIRSOC 103 [7]

.....................................................................................................................................................100

Table 5.3.2 Recommended diameters and separation for reinforcement in tie-columns and bond-

beams, Universidad Nacional de Cordoba [16] ..........................................................................103

Table 5.3.3 Recommended reinforcement for tie-columns [37].........................................................106

Table 5.3.4 Recommended reinforcement for bond-beams [37] ........................................................106

Table 5.6.1 Elastic properties of masonry given by the different codes.............................................118

Table 5.6.2 Minimum conditions for walls to be considered as load-bearing walls...........................119

Table 5.6.3 Geometric conditions for confined masonry given by the different codes, Decanini et al.

[38]..............................................................................................................................................120

Table 5.6.4 Resistance verifications in plane given by codes.............................................................121

Table 5.6.5 Resistance verification out-of-plane and verification of the confining elements ............122

Table 5.6.6 Specifications and requirements of the confining elements.............................................123

Table 5.6.7 Specifications and requirements of reinforcement for confining elements .....................124



Index

xiii

LIST OF SYMBOLS

Ac= Cross-sectional area of the longitudinal reinforcement of tie-

columns

Ae = Cross-sectional area of stirrups in one layer

Av= Cross-sectional area of the longitudinal reinforcement of bond-

beams

Bc = Total area of the transverse section of tie-columns

Bm = Cross-section of the masonry wall

C = Design seismic coefficient

D= Projection of the compressive loads in the direction of the

masonry course for the diagonal testing

Em = Elastic modulus of masonryGm = Shear modulus of masonry

H = Height of confined masonry measured between constrains

Ho = Height of confined masonry measured between bond-beams

Ht= Total height of confined masonry measured from the top of the

foundation

Hs = Horizontal seismic loads

L = Length between borders of the confined masonry wall

Lo = Length of confined masonry measured between tie-columns

Le = Length between tie-columns of load-bearing wall

M.R.A.D.= Reinforced masonry with distributed reinforcement (mampostería

reforzada con armadura distribuida)

Mur= Ultimate bending moment and axial loading of confined masonry

wall

Muro = Ultimate bending moment of the confined masonry wall

Muv= Ultimate bending moment in the vertical direction per unit of

length, in the out-of-plane verification of walls

N = Gravitational loads

Nu = Vertical load applied to the masonry wall

Nuo = Compressive strength of the wall

Nur = Ultimate vertical load resisted by the wall



Index

xiv

P = Compressive load for the diagonal testing

RC = Reinforced concrete

URM = Un-reinforced masonry

Vp = Compressive stress acting in the masonry panel

Vur= Maximum shear stress resisted by the masonry in ultimate limit

state

βs = Yield stress of steel

µ = Ductility

σ 'PKm = Mean compressive strength of masonry units

σ 'PK = Characteristic compressive strength of masonry units

σ o =Mean compressive strength generated by vertical loads on the wall

σ 'mo = Compressive strength of masonry

τ mo = Shear strength of masonry

ø = Diameter of the reinforcing steel

Ψ = Reduction factor due to the eccentricity of vertical loads and

slenderness of the confined walls

d = Diagonal length

dc = Transversal dimension of tie-column

ds = Diameter of the reinforcing steel

eo = Thickness of the masonry probes under diagonal compression

ea = Accidental eccentricity due to vertical loads

ec = Complementary eccentricity due to the slenderness effect

et = Eccentricity at the top of the wall

e* = Design eccentricity of load-bearing walls

k = Number of stories above the analysed story

lb = Straight anchorage length

lc = Critical length

q = Lateral gravitational load per unit area of the wall

qs = Seismic load per unit area of the wall

se = Spacing of stirrups

t = Thickness of the masonry wall

List of symbols for chapter 5.6

Italian code

Avi= Cross-sectional area of the i- esima longitudinal reinforcement of

tie-columns



Index

xv

Mr = Resisting moment of tie-columns

di = Length between the reinforcement to the compressive

fc = Compressive strength of masonry fy = Yield stress of steel

l = height of tie-column

t = Thickness

σ tr = Diagonal tensile strength of masonry

σ o = Mean compressive strength

τ ok = Value that variety in function of masonry type

τ m = Medium diagonal tension

Colombian code

Aci = Total area of the transverse section of tie-columns

Ae = Cross-section of the masonry wall

Amd = Effective cross-section of the masonry wall

Amv = Cross-section of the masonry wall to determine shear strength

As= Cross-sectional area of the longitudinal reinforcement of bond-

beams

Ast = Cross-sectional area of the longitudinal reinforcement

Mn = Design bending moment of the confined masonry wall

Mu = Ultimate bending moment of the confined masonry wall

Pnc = Nominal compression load acting in tie-column, positive, N

Pnd = Nominal compression load of the masonry

Pnt = Nominal tension load acting in tie-column, negative, N

Pu = Design compression load of the masonry wall, N

Puc = Design compression load acting in tie-column, positive, N

Put = Design tension load acting in tie-column, negative, N

Re= Coefficient that takes into account the slenderness of the elements

under compression

Vn = Design shear stress resisted by the masonry, N

Vnc = Nominal shear stress acting in the reinforcing concrete

Vu= Maximum shear stress resisted by the masonry in ultimate limit

state, N

Vuc = Maximum shear stress resisted by the tie-column

b = Thickness of the masonry wall, mm

f’c = Compressive strength of the concrete of confining elements

f’m = Compressive strength of masonry

fy = Yield stress of steel, MPa

h’ = Diagonal length, mm



Index

xvi

hp = Height of confined masonry measured between bond-beams

lc = Length of confined masonry measured between tie-columns

lw = Total length of confined masonry

t = Thickness of the masonry wall

ø = Reduction strength coefficient

Peruvian Code

C 1 = Seismic coefficient

F = Axial force in tie-columns produced by the bending moment

L = Total length of confined masonry

Lm = Total length of the highest confined masonry or 0.5L

Me = Bending moment acting in confined masonry Ms = Distributed bending moment per unit length

Nc = Number of tie-columns

Pc = Sum of the gravitational loads

Pg = Service gravitational load

Pm = Maximum gravitational load

Ts = Tension force

U = Importance factor

Ve = Shear force produced by moderate earthquake in the wall

Vet = Shear force in the wall determined in the elastic analysisVm = Shear strength of the confined masonry wall

Z = Zone factor

a = Critic dimension of the confined masonry panel

e = Gross thickness of the wall

f’m = Compressive strength of masonry

h = Height of confined masonry measured between bond-beams

t = Thickness of the masonry panel

v’m = Characteristic shear strength of masonryw = Seismic load uniformly distributed

α = Reduction factor of the shear strength due to slenderness effects

γ = Specific weight

σ m = Maximum axial strength

Mexican Code

As = Cross-sectional area of the longitudinal reinforcement

Asc = Cross-sectional area of stirrups in one layer AT = Cross-section of the confined masonry wall



Index

xvii

FE= Coefficient of reduction that takes into account the slenderness

and the eccentricity

FR = Resistance factor

Mo = Bending moment

MR = Bending moment applied to the plane of the wall

P = Axial compressive load

PR = Design strength of the masonry wall to vertical load

Pu = Design axial load

Q = Factor of seismic behaviour

VmR = Design shear load of masonry, N

d= Length between the reinforcement in tension and the concrete in

compression.

d’= Length of confined masonry measured between longitudinal

reinforcement of tie-columns

f’c = Compressive strength of the concrete of confining elements

fm* = Compressive strength of masonry

fy = Yield stress of steel

hc = Total area of the transverse section of tie-columns

s = Spacing of stirrups

vm = Compressive diagonal strength of masonry



Chapter 1. Introduction

1

1. INTRODUCTION

1.1 Objectives

The issue of the seismic performance and safety of existing mixed masonry and reinforcedconcrete buildings systems is characterized by numerous uncertainties and in some cases by a

real lack of sufficient knowledge. This statement also applies to existing codes, where scarce

or no provisions are given regarding methods of analysis and safety check criteria to be used

in practical applications. Recent code provisions such as Eurocode 8 [1] or the Annex 2 to the

OPCM 3274 [2] provide little or no guidelines to the designer, and are limited most often to

define the general principles for design.

The present dissertation concerns with a general review of the subject, ranging from current

and past research, taking as a reference the experimental data available in the literature, the

characteristics of buildings with structural masonry walls connected to tie-columns and bond- beams.

This type of buildings is characterized by the mutual interaction between masonry and tie-

columns and bond-beams, giving a composite behavior which is essentially similar to what is

now defined in modern construction as “confined masonry”. This type was rather common in

Italy during the first half of the past century, and is still being used nowadays in some regions

of the Italian territory, in spite of the fact that specific national code regulations are not yet

available, as a local building tradition in which the vertical tie-columns are seen as a

confinement for masonry and/or a means to carry concentrated vertical loads.




2

1.2 Description

The construction of confined masonry starts after the Messina earthquake in 1908 and has

become one of the most popular and inexpensive structural construction system used forhousing. With the increased popularity and availability of reinforced concrete and different

types of masonry units, this construction is common for low-rise residential buildings and

individual houses in many areas of Latin America, Indian Subcontinent and Asia as well as in

some parts of Europe. In these buildings masonry shear walls are often the only structural

element assumed to provide resistance to gravitational and seismic lateral loads. It consists

basically of masonry panels confined by vertical and horizontal elements usually of reinforced

concrete.

From the structural and seismic point of view, in Argentina and in many other countries,

masonry can generally be use in two ways for dwelling: confined or reinforced masonry andreinforced concrete frames with masonry infill. Figure 1.2.1 shows the percentage of use of

confined masonry in relation with RC frame with masonry infill in Argentina, 1985. Almost

75% correspond to constructions of confined masonry (1), around 20% to RC frames with

masonry infill (2) and the rest to other typologies like timber and steel structures. In this

document only confined masonry is study.

The main difference between confined masonry and RC frames with masonry infill is that: in

confined masonry the confining elements are not intended or designed to perform as moment-

resisting frames. When such frames are constructed to resist lateral and vertical loads the

purpose of the masonry walls is only for space partitioning, and the construction system iscalled masonry infilled frames. In masonry infilled frames the reinforced concrete frame

structure is constructed first and the masonry is added later between the RC members. In the

case of confined masonry, the masonry walls are load-bearing and are constructed to carry all

of the gravity loads as well as lateral loads, Kuldeep Virdi [3].

Figure 1.2.1 Uses of masonry in Argentina, 1985. 1) Confined masonry and 2) RC frames with masonry

infill; Decanini, Payer and Terzariol [4]

Confined masonry

RC frame with

masonry infill




3

Confined masonry is normally used in buildings up to five stories high. Reinforced concrete

confining elements are horizontal members called bond-beams and vertical members called

tie-columns. Tie-columns have a square section whose dimensions typically correspond to the

wall thickness (15cm in general). Bond-beams width is the wall thickness and the depth is

usually equal to 25cm. Typically, both tie-columns and bond-beams have a longitudinal

reinforcement ratio, based in the gross sectional area, of about 1.2%, this percentage can vary

between the different national codes.

Seismic action, represented by lateral forces applied to each floor and to the roof, is resisted

by a mechanism of walls, coupled by lintels and sills, and interconnected by the floor slabs.

Slabs are assumed to behave as non-deformable diaphragms, being able to distribute the

lateral forces to the walls, Zabala et al. [5]. The floor system generally consists of cast-in-

place reinforced concrete slabs, but very often, prefabricated units are used (such us pre-

stressed concrete joints or planks), Meli et al. [6].

The major improvements in the performance of the confined masonry building over the plain

masonry building are the following ones, Kuldeep Virdi [3]:

• Enhances greatly the connection between structural walls.

• Improves the stability of masonry walls.

• Improves the strength of masonry walls.

• Provides ductility under earthquake loading.

• Improves the integrity and containment of earthquake damaged masonry walls.

Tie-columns and bond-beams confine the masonry walls to give containment after cracking as

the result of the earthquake, avoiding a brittle behavior and allowing the dissipation of energy

under earthquake loading. The confinement must be continuous as is shown in Figure 1.2.2, to

improve the connection among other walls and floor diaphragms.




4

Figure 1.2.2 Solid brick masonry confined with tie-columns and bond-beams, Kuldeep Virdi [3]

Confined masonry walls have limited shear strength and ductility as compared to common

reinforced concrete walls; nevertheless, typical low-cost housing buildings have goodearthquake resistance, because they have large wall densities (ratio of transverse wall areas to

a typical floor area), and because wall layout is symmetric and regular, both in plan and in

elevation.

Depending on the presence, or not, of reinforcement in the masonry panels, different

classifications of confined masonry are currently use. Confined masonry, reinforced confined

masonry and confined masonry without vertical columns is the classification used in the

masonry seismic code of Argentina (INPRES-CIRSOC 103 [7]). The difference between the

confined masonry and the reinforced confined masonry is the presence of horizontal

reinforcement in the cement mortar joints. In this kind of masonry the reinforcement does notgive much more resistance to the panel but gives more ductility and integrity. It must be

mentioned that generally in South America it is not common to use horizontal reinforcement

in confined masonry walls.

In the case of confined masonry without vertical elements, there are no tie-columns. This type

of masonry is only used for internal load-bearing walls made of solid clay bricks and in areas

of low seismic activity.

Another classification is based in the different masonry units. Masonry panels are usually

made of clay bricks, clay or concrete blocks, bonded with cement mortar.




5

After examining its good seismic performance, this system became popular in zones of high

seismic hazard. It must be pointed out that confined masonry has evolved essentially through

an informal process based on experience, and that it has been incorporated in formal

construction through code requirements and design procedures that are mostly rationalizations

of the established practice, even after having been validated by structural mechanics

principles and experimental evidence.

In spite of masonry experimental research programs conducted in many countries, Yáñes et al.

[8], the behavior of confined masonry shear walls is still not very well known.

1.3 Construction procedure

1.3.1 Materials for confined masonry construction

Masonry units

Masonry units are classified into the following types: solid, perforated unit, hollow unit,

cellular unit and horizontally perforated unit illustrated in Figure 1.3.1., Kuldeep Virdi [3].

They can be made of clay or concrete. It is forbidden the use of perforated or hollow clay

bricks in the horizontal direction because of their brittle behaviour and the difficulties to build

vertical mortar joints. Also the re-utilization of the masonry units (bricks, blocks, etc) is not

allowed by the codes.

Figure 1.3.1 Types of masonry units, Kuldeep Virdi [3]

Different test are made to each type of masonry brick. These tests consist in tension, axial

compression made with half masonry unit and water absorption. Average of the

measurements is compute and parameters like strength are given for each masonry unit.




6

The strength of the masonry units is given as an example in Table 1.3.1 from the Argentinean

code.

Table 1.3.1 Strength of the masonry units, INPRES-CIRSOC 103 [7]

Type of masonry unit class σ'PKm σ'PK Net section

Kg/cm² Kg/cm²

Solid clay brick A 120 80 >80% gross section

Solid clay brick B 75 45 >80% gross section

Hollow clay block A 120 85 >60% gross section

Hollow clay block B 75 50 >40% gross section

Hollow concrete block I 65 45 >40% gross section

Hollow concrete block II 65 45 >40% gross section

Hollow concrete block III 50 30 >40% gross section

Mortar

Depending on the codes of the different countries there are different specifications for the

mortar joints. According to the specification used in EC 6 [9], several types of mortar can be

used for masonry walls, Kuldeep Virdi [3]:

• General purpose mortar, used in joints with thickness greater than 3mm and produced

with dense aggregate.

• Thin layer mortar, which is designed for use in masonry with nominal thickness of

joints 1-3mm.

• Lightweight mortar, which is made, using perlite, expanded clay, expanded shale etc.

Lightweight mortars typically have a dry hardened density lower than 1500kg/m3.

In the Table 1.3.2 below are shown typical composition of prescribed general purpose mortar

mixes and expected mean compressive strength. This table corresponds to the specificationsgiven in EU 6, where mortars are classified by their compressive strength, expressed as the

letter M followed by the compressive strength in N/mm², for example, M5. Prescribed

masonry mortars, additionally to the M number, will be described by their prescribed

constituents, e. g. 1: 1: 5 cement: lime: sand by volume.




7

Table 1.3.2 Typical prescribed composition and strength of general purpose mortars, Kuldeep Virdi [3]

In the Argentinean code values of the compressive strength of mortar joints is given after 28

days, like is illustrated in Table 1.3.3. Three categories of mortar are distinguished here: high

(H), intermediate (I) and normal (N). Also the proportions of cement, sand and hydrated lime

are given for each type of category. Test to determine the compressive strength for each kindof mortar are made of squares of 7cm of side, Figure 1.3.2.

Figure 1.3.2 Determination of the mortar compressive strength, Bustos [10]

The strength of the mortar is increased with the increase of the cement content and a little

proportion of hydraulic lime enhances the use of the mortar for the joints.

Table 1.3.3 Constituents of the mortar joints given by the INPRES-CIRSOC 103 [7]

Type of mortar joint Cement: hydraulic lime: sand Minimum compressive strength

after 28 days (MN/m²)

High 1:0:3

1:1/4:315

Intermediate 1:1/2:410

Normal 1:1:5

1:1:65




8

Reinforcing steel

Steel bars are used as reinforcement in confined masonry. Reinforcing steel may be assumed

to possess adequate elongation ductility and shall provide with sufficient anchorage length so

that the internal forces are well transmitted between all the members. Anchorage should beachieved by straight anchorage, hooks, bends or loops as shown in Figure 1.3.3. Alternatively

stress transfer may be by means of an appropriate mechanical device proven by tests.

Figure 1.3.3 Typical anchorages of the reinforcing bars according to EC 6 Kuldeep Virdi [3]

1.3.2 Construction procedure

In the case of confined masonry, the masonry walls are considered as load-bearing and are

built to carry all of the gravity loads as well as lateral loads. For this reason load-bearing

masonry walls are constructed first with serrated edges and then the concrete of columns and

beams are poured against the boundaries of the masonry panel. The vertical and horizontal

confining elements are cast simultaneously with the floors, which are constructed as

reinforced concrete slab.

The steps of the construction procedure of the confined masonry are illustrated in Figure 1.3.4

to 1.3.8.

Figure 1.3.4 Foundations construction and start of the vertical concrete columns




6

Figure 1.3.5 Construction of the masonry panel

Figure 1.3.6 Arrangement of vertical

reinforcement in tie-columns

Figure 1.3.7 Position of the horizontal

reinforcement

Figure 1.3.8 Concrete poured against the

boundaries of the masonry panel

In order to achieve effective confinement of walls, vertical confining elements (tie-columns)

should be located at all corners and changes of wall contour, and at all joints, wall

intersections and free ends of structural walls.

The contribution of the tie-columns and bond-beams to the lateral resistance of the masonry

building is normally not taken into account for design. Consequently specific design

calculations for confining elements are not required. The amount of reinforcement in vertical

and horizontal confining elements is determined on an empirical basis. Although the tie-

columns and bond-beams do not provide frame system contribution to the wall, adequate

splicing and anchoring of re-bars is required at all joints.

Confined masonry should be constructed following simple instructions for quality of

workmanship, Kuldeep Virdi [3]:




8

• In dry and hot climate, masonry units should be soaked in water before the

construction in order to prevent quick drying and shrinkage of cement based mortars.

• Same type of masonry units and mortar should be used for structural walls in the same

storey.

• Bracing walls should be constructed in the same time as the load-bearing walls.

• The thickness of individual walls is kept constant from storey to storey.

• In cases where general purpose mortar is going to be used, the mortar joints thickness

should be between 8 and 15mm.

From Figure 1.3.9 to Figure 1.3.13 confined masonry walls, of different types of masonry

units, under construction are showed as is the current practice in various countries.

Figure 1.3.9 Confined masonry walls under construction Mexico, 1993. Yoshimura et al. [11]

Figure 1.3.10 Left: confined masonry walls under construction (Jimo, P.R. China, 1999); right: confined

masonry walls under construction (El Salvador, 2001), Yoshimura et al. [12]




9

Figure 1.3.11 Confined masonry walls using hollow concrete block masonry units. Las Losas Project

under construction in Villahermosa, Mexico, Yoshimura, Kikuchi, Okamoto and Sanchez [13]

Figure 1.3.12 Confined masonry of clay bricks deposit under construction in Córdoba, Argentina




10

Figure 1.3.13 Confined masonry of clay bricks deposit under construction in Córdoba, Argentina

Construction problems

Most surveyed dwellings had important construction problems, mainly due to lack of

knowledge of proper building techniques, poor workmanship, or the use of materials of poor

quality. House owners and builders seem to believe that the reinforced concrete beams andcolumns are the most important structural elements. Accordingly, they pay a lot of attention to

their construction. The results, however, usually tend to be poor. Concrete is prepared with

high water/cement ratio, large aggregate size, and inadequate mixing and vibration.

Aggregates are usually bought from informal quarries, where there is little or no quality

control in its cleanliness and contents of fines or organic material, or of the size of the

material. Furthermore, curing is not considered to be important: beams and columns are

seldom cured, and slabs are sometimes cured by pouring some water on the surface the day

after they are built. As a result of these poor construction practices concrete is often weak,

porous, and full of voids, Figure 1.3.14.

Because there is an understanding of the importance of reinforcement in the strength of the

structure, most elements are overly reinforced, even though steel reinforcement is expensive.

Stirrups, however, are believed to be useful only to maintain the main reinforcement in place,

and in most cases they have open hooks, or are made with small diameter rebar. In areas with

access to welding factories, it is common to find welded steel bars, instead of overlapping

rebar connections. Since safety is always a concern, many owners weld metal doors and

windows to the reinforcement of columns or beams. A common problem observed is

corrosion of the steel reinforcement. This happens because of poor quality of concrete with

small covers to protect from filtration of rain water and other atmospheric agents.




11

The bricks used to build the masonry walls are usually hand made by local artisans, because

they are significantly cheaper than industrial bricks. Furthermore, the quality of the masonry

is generally quite poor due to the mortar joints, Blondet, Dueñas, Loaiza, and Flores [14].

Figure 1.3.14 Construction problems from Blondet et al. [14]



Chapter 2. Failure mechanisms and performance in past earthquakes

12

2. FAILURE MECHANISMS AND PERFORMANCE IN PASTEARTHQUAKES

2.1 Failure mechanisms

In the event of an earthquake, apart from the existing gravity loads, horizontal loads are

imposed on walls. In these conditions, however, the un-reinforced masonry generally behaves

as a brittle material, depending on the intensity of the excitation. Hence if the state of stress

within the wall exceeds masonry strength, brittle failure occurs, followed by possible collapse

of the wall and/or of the building. Therefore one solution to make the un-reinforced masonry

walls vulnerable to earthquakes can be to confined and/or reinforced whenever is possible the

masonry panels, Kuldeep Virdi [3].

The principal failure mechanisms of confined masonry subjected to seismic actions can be

summarized as follows:

• Flexural failure: this king of failure is ductile causing yielding of the vertical

reinforcement. A flexural failure would be desirable because is more ductile than a shear

failure; also the former is more simple to repair, however more research is needed to obtain

this goal. Generally first yielding occurs at the base of the tie-columns as illustrated in Figure

2.1.1.

Tie-columns and bond-beams, according to the specifications set up in EU 6, are not

considered in assessing the flexural resistance of structural walls, Tomazevic [15].




13

Figure 2.1.1 Flexural failure, Zabala et al. [5]

• Shear failure: A brittle failure in the masonry is due to the shear, Figure 2.1.2. Failureoccurs when the applied loads are higher than the shear resistance of the confined masonry.

For this failure to happen the previous failure in flexure must not have occurred. This means

that it can only occur when tie-columns have higher reinforcement, when the masonry panel

has high axial load or when the masonry panel is very long. Research showed that a shear

failure is possible to occur when a strong earthquake hits a confined masonry structure, even

in the case that the structure satisfies the ideal characteristics to obtain flexural failure. The

formation and development of inclined diagonal cracks may follow the path of bed and head-

joints (stepped) or may go through the bricks, depending on the relative strength of mortar

joints, brick mortar interface, and brick units.

According to the requirements of EC 6, the strength of the reinforced masonry members

should be taken into account in the design of confined masonry for seismic load. These is in

disagreement with the existing experimental evidence, that indicates that tie-columns and

bond-beams improve the lateral resistance of a plain masonry wall panel, just as they improve

its energy dissipation capacity and ductility, Tomazevic [15].

Figure 2.1.2 Diagonal cracking under cycling loading, Zabala et al. [5]




14

• Insufficient anchorage of the vertical or horizontal elements: failure that occurs when

the anchorage of the reinforcement is not enough in foundations or when the reinforcement of

the horizontal elements is not well connected to the vertical confinement like is illustrated in

Figure 2.1.3.

Figure 2.1.3 Bad connection between horizontal and vertical reinforcement, Universidad Nacional de

Córdoba [16]

• Crushing of the compressed corners of the diagonal: This kind of failure generallyoccurs in hollow concrete blocks and the crushing zones are situated at the ends of the

diagonal as is showed in Figure 2.1.4

Figure 2.1.4 Compression of the diagonal, Universidad Nacional de Córdoba [16]

Crushing in diagonal

corners

Compressed

diagonal

Bond-beam

Tie-column

Hs




15

It is possible to estimate the theoretical flexural strength of a confined masonry wall in a

simple way, by considering the amount of column reinforcement, vertical load supported by

the wall and yield stress of the reinforcement steel. On the other hand, the cracking shear load

and the maximum shear strength of these walls are more uncertain, since they depend on

several factors like: individual brick strength, mortar and workmanship qualities, vertical

load, amount of columns reinforcement and amount of horizontal reinforcement embedded in

the masonry. In addition, the manufacturing conditions of the bricks and the walls are very

variable, causing high dispersion of the resulting mechanical properties, Kuldeep Virdi [3].

The different failures mechanisms and their occurrence are illustrated in a load vs.

deformation plot Figure 2.1.6, Bustos [10]. In this plot three fields are observed: lineal elastic

field, non linear field until the ultimate load and the last one of failure. As is observed here the

flexural failure is the desirable one, as it presents more ductility.

Figure 2.1.5 Occurrence of the different failure modes in confined masonry. a) Compression failure; b)

Diagonal crack; c) Flexural failure; Bustos [10]

2.2 Ductility

In case of the confined masonry wall system, both of the vertical and horizontal reinforcement

in masonry walls play an important role for expecting higher ultimate lateral strength and

better ductility, as shown in Figure 2.2.1. Ductility is defined as the relation between the

maximum displacement and the displacement when the first crack appears. Different levels of

ductility are achieved for different types of masonry. Un-reinforced masonry has generally

(a)

(b)

(c)

Load

First

crack




16

less ductility in comparison with the rest. When tie-columns and bond-beams are added to the

un-reinforced masonry panel more ductility is expected. In the case of reinforced confined

masonry, the reinforcement does not give much more resistance to the panel but gives more

ductility and integrity. In general the higher ductility is achieved in the case of the reinforced

masonry, where the reinforcement is distributed in the horizontal and vertical direction like is

shown in the third case of Figure 2.2.1. In this figure the graphics are only intending to

represents more less the different ductilities achieved by different configurations of masonry.

Figure 2.2.1 Different ductilities induced by different typologies of masonry, Bustos, Universidad

Nacional de San Juan, Argentina [10]

For masonry structures, it should be provide adequate capacity for energy dissipation in the

inelastic field. This energy dissipation must be comparable with the seismic loads assumed in

the design. This capacity is given with the inclusion of the confinement and the reinforcement

on it. In Figure 2.2.2 is illustrated the effect of the force reduction for the inelastic response.

This reduction is given by the ductility.

Confined masonry

Un-reinforced masonry

Reinforced masonry




17

The INPRES-CIRSOC 103 [7] code gives, for confined masonry with clay solid bricks, a

global ductility of 3 and for masonry with hollow concrete blocks a ductility of 2. From the

comparison of these two ductilities, it is known that confined masonry with clay solid bricks

have major capacity of energy dissipation. This is also verified in the experimental tests

carried out in the National University of Córdoba.

Figure 2.2.2 Ductile response of the confined masonry structures, Decanini and Payer [17]

2.3 Predominant design failures in confined masonry during earthquakes

Common failures in design that produce collapse during earthquakes, in confined masonry,

are listed below, Bustos [10]:

• Missing elements of vertical confinement (tie-column).

• High spaced between tie-columns producing a loss of the effect of confinement.

• Wrong poured of concrete in columns.

• Bad quality of the concrete in tie-columns that can produce propagation of the shear

failure from the masonry panel to the confinement columns.

• Insufficient anchorage of the reinforcement of horizontal or vertical elements, Figure

2.3.1.

• Excessive vertical load that produce more shear resistance in the wall but reduce the

ductility.




18

• Torsion problems due to: bad distribution of the walls, in a given direction not enough

wall density, no vertical continuity in the masonry panels, and differential settlements and big

openings in the slabs.

Figure 2.3.1 Failure caused by insufficient anchorage of reinforcement in the confinement elements.

Examples of bad disposition of the reinforcement, Decanini and Payer [17]

Predominant failures have been found to occur in buildings are due to the shear and not due to

bending as expected. The reasons of the shear failure are:

• The deformation due to shear failure is predominant because masonry panels have

short height and a higher moment of inertia of the transverse section.

• If the effect of the transversal walls is added (including their loads) when the masonry

panel wants to flexure the transverse wall must make off with it and this is difficult to

achieve.

• The bending moment at the base, associated with the static analysis, is reduced due to:

rotation of the foundations, high modes of vibration and by the interaction of slab and wall.

For this reason the conventional design showed an extra capacity in bending.

• Confined masonry is very stiff system due to their large wall densities interconnected

by the floor slab, assumed to behave as non-deformable diaphragms. In addition concentrated

loads (slabs) and distributed loads (walls) are also present enhancing the stiffness of the

system. Due to these effects the accelerations of the structure are very close to the

accelerations at the ground level, Figure 2.3.2.




19

Sa/g

T (seg)1.51.00.5

0.9

0.6

0.3

Typical period range

for masonry constructions

4) Soft clay and sand

3) Deep non-cohesive soils (H>80m)

2) Stiff soils (H<50m)

1) Rock

Figure 2.3.2 Typical period range in confined masonry constructions. Statistic values for different types

of soils; Decanini and Payer [17]

2.4 Vulnerability: performance in past earthquakes

Several studies reveal that masonry construction is the most common solution for housing

construction in Latin America. Mainly two types of masonry are used: adobe (sun-dried mud

blocks) and confined brick masonry. During the recent earthquakes analyzed, adobe

construction and un-reinforced brick masonry had the highest rate of damage or collapse, and

in general good performance was observed in confined or reinforced brick masonry housing.

The seismic behavior of confined masonry buildings has been generally satisfactory and could

be found undamaged even in the most heavily damaged areas. Nevertheless, significant

damages have been observed in near-epicentral regions during strong ground shaking, Meli et

al. [6].

A summary of observed behavior of confined masonry dwellings during past seismic events is

presented next, together with several examples of good and poor housing construction

practices.

Typically damage patterns observed are: 1) Shear diagonal failure of walls, 2) Shear and

bending failure of heads and feet of reinforced columns, 3) Separation of columns from walls,

and 4) Collapse of wood slabs, hollow brick joist slabs and brick jack arch slabs, 5) Another

main structural deficiencies for this construction type lies in the widely different wall

densities in the two orthogonal directions. This deficiency may be eliminated with appropriate




20

architectural design. This construction type is otherwise expected to demonstrate good

seismic performance.

The 25 January 1999 Colombia Earthquake

The magnitude Mw=6.2 earthquake occurred in an epicentral area near the cities of Armenia

and Pereira, with populations of 270.000 and 380.000, respectively. These were the largest

cities affected by the earthquake although other smaller cities were also severely damaged.

The total number of deaths in Armenia alone was about 1.000, and about 5.000 people were

injured in this city. Armenia was the city that suffered the highest rate of deaths and damage

in dwellings, Rodriguez and Blondet [18]. Extensive structural damage occurred in some of

the newly constructed confined masonry walls, in which adjacent masonry walls were

separated from the reinforced concrete confining columns, as shown in Figure 2.4.1.

Figure 2.4.1 Damage to reinforced concrete column in confined masonry wall due to 1999 Colombia

earthquake, Yoshimura et al. [12]




21

The 30 September 1999 Oaxaca, Mexico Earthquake

On September 30, 1999 at 11:31:00 h (local time), a magnitude Mw=7.5 earthquake occurred

with its epicenter located southeast form the city of Puerto Escondido, Oaxaca. Observations

made during a visit to the affected region on the behaviour of confined masonry showed nostructural damage, only cracks is walls as is illustrated in Figure 2.4.2 and 2.4.3. It is

estimated that 40% of the walls present diagonal cracks this is a direct consequence of

construction problems observed, like over-reinforcement and insufficient anchorage between

beams, columns, slabs and foundations, López Bátiz et al. [19].

Figure 2.4.2 Cracks observed in confined masonry after the Oaxaca earthquake 1999, López Bátiz et al.

[19]

Figure 2.4.3 Damage resulting from an inadequate distribution of the confining elements, López Bátiz et

al. [19]




22

Also a hospital under construction suffered damage during the 30 of September earthquake.

This building was built of a mix between confined masonry and frames of reinforced concrete

localized in the frontal and lateral parts of the construction. Both structures suffer damage as

is showed in Figure 2.4.4.

Figure 2.4.4 Damage in a hospital during the Oaxaca earthquake in Mexico, López Bátiz et al. [19]

The 13 January 2001 El Salvador Earthquake

The main shock of this earthquake occurred at 17:33, Saturday the 13th of January 2001. Its

Magnitude was Mw=7.6. The epicenter, 100km southwest of the city of San Miguel, El

Salvador, was located off coast of Central America, Yoshimura and Kuroki [20].

A large number of the buildings of confined masonry wall construction exist in El Salvador,

but most were not severely damaged during the earthquake. A few rare cases of damage to

confined masonry buildings show where hollow concrete block walls and clay masonry brick

walls were severely damaged. In the building shown in Figure 2.4.5, a concrete block

masonry wall has separated into parts due to shear cracking, and part of the wall has separated

from its adjacent RC confining column and overturned in the out-of-plane direction. Also,

shear cracks formed in the clay brick masonry walls of the building in Figure 2.4.6 have

penetrated the RC confining columns. This damage seems to have been caused by pounding

from a collapsed building rather than by ground shaking.




23

Figure 2.4.5 Failure of a hollow concrete-block masonry wall. The hollow concrete block units are

separated from the RC confining column, Yoshimura and Kuroki [20]

Figure 2.4.6 Damage to a confined clay-brick masonry wall in Usulutan, Yoshimura and Kuroki [20]

The 21 January 2003 Colima, Mexico Earthquake

The earthquake occurred in the coastal region of the state of Colima and had a magnitude of

about 7.6 Mw. From 13.500 dwellings reported damaged, about 2.700 collapsed. Confined

masonry dwellings suffered mostly minor damage and most damage was concentrated in

dwellings of un-reinforced masonry or adobe, Rodriguez and Blondet [18]. Only some

fissures appear in the confined masonry, this was a direct product of the bad construction ofthe tie-columns and bond-beams.




24

Past earthquakes reported in Perú

Damage to confined masonry in Perú was reported in the following years, Loaiza and Blondet

[21]:

• The 12 November 1996 Nazca, Perú Earthquake: the 12 November 1996 Mw =7.7

Perú subduction zone earthquake occurred off the coast of southern Peru, near the intersection

of the South American trench and the highest topographical point of the subducting Nazca

Ridge. Some damage was found in confined masonry after the earthquake. Figure 2.4.7 shows

a slender wall in the right.

• The 23 June 2001 Atico, Perú Earthquake: in the late afternoon of June 23, 2001, a

colossal earthquake with a magnitude of Mw=8.4 took place in the coastal waters off the

District of Arequipa and the town of Atico, Perú. The magnitude of the event makes it the

largest in the world in the last 25 years. This earthquake caused nearly 2000 deaths, 3.000

injuries, 26.000 homes destroyed and 34.000 damaged homes and left 190.000 people

homeless. Confined masonry walls have generally shown a good seismic performance, and no

significant damage was found during this earthquake in Perú.

Figure 2.4.7 Photograph illustrating typical damage in confined masonry, 1996 Nazca earthquake, Loaiza

and Blondet [21]

The 3 March 1985 Chile Earthquake

This was an interplate event in the subducted Nazca plate, with Mw = 7.8. Its epicenter was

located 20km from the Pacific coast of central Chile. This earthquake is considered one of the

most important experienced in Chile in the 20th century and has been compared to the great1906 Valparaiso earthquake. Most of the severe damage occurred in adobe dwellings in rural




25

areas, particularly near Llolleo, where an acceleration record had a peak ground acceleration

of 0.67g, Rodriguez and Blondet [18].

The 23 November 1977 Caucete, Argentina Earthquake

The epicentre of the destructive San Juan, Argentina earthquake of November 23, 1977 was

located near the eastern slopes of the Andes Mountains approximately 80km northeast of the

city of San Juan, capital of San Juan province. On the basis of teleseismic and local

seismograph data, the earthquake has been assigned a magnitude of 7.4 (Ms), a depth of 30km

or less. The main shock was followed by a large aftershock sequence including at least one

magnitude 6 event and was felt throughout much of southern South America, including

Buenos Aires 950km to the southeast and Sao Paulo 2.100km to the northeast. The earthquake

caused extensive damage in the Province of San Juan, particularly in the towns of Bermejoand Caucete, respectively located approximately 60km south southeast and 70km southwest

of the epicentre. Most notable effects of the earthquake included vast areas of liquefaction

(hundreds of square kilometers), complete or partial collapse of hundreds of adobe dwellings,

and damage to numerous cylindrical wine storage tanks. Approximately 65 persons were

killed, 284 injured and 20.000 to 40.000 left homeless, Rojahn, Brogan and Slemmons [22].

Most of the adobe dwellings and un-reinforced masonry wall buildings collapse, including the

building which housed the town’s electric power generating plant. Recently constructed

reinforced and confined masonry walls buildings that where designed to resist earthquakes, on

the other hand, where not damage, Figure 2.4.8. As indicated by the bed in the center of this photograph taken eight days after the earthquake, some residents of Bermejo chose to sleep

outdoors after the earthquake. In Figure 2.4.9 proportions of damage for different dwellings

are given.

Figure 2.4.8 Undamaged recently constructed reinforced masonry dwelling in Bermejo, 60km south-

southeast of the epicenter, Rojahn, Brogan and Slemmons [22]




26

Figure 2.4.9 Damage in masonry in Caucete earthquake, 1977 Argentina. Decanini, Payer and Terzariol

[4]

Collapse or strong damage

Damage of consideration No significant damage

Recoverable

Un-reinforced masonry

Confined masonry

Adobe dwellings

Soil failure




27

The 1985 Mexico Earthquake

The 1985 Michoacan earthquake was the result of the subduction of the Cocos Plate under the

continent and has been classified as of the interplate type. Several reports have been published

on observed damage in structures during this earthquake. Most of these reports describedamage in Mexico City. Little information can be found on the behavior of masonry

dwellings near the epicentral location the Pacific coast. However, it is accepted that most of

the damage occurred in Mexico City, with little or no damage in masonry dwellings in either

Mexico City or in the area near the epicenter, Rodriguez and Blondet [18].

No important damage or collapse was found in buildings of confined masonry in the city of

Mexico. In particular, no damage was found in dwellings of resent construction that follow

the code regulations. In the epicentral zone some damage was found.

The 26 December 2003 Bam, Iran Earthquake

A devastating earthquake hit the city of Bam in the south of Iran at 5:26 a.m. local time,

Friday, 26 December 2003. Based on the government of Iran's February estimate, the

earthquake caused more than 43.000 deaths, 30.000 injuries, and left 70.000 homeless. It

caused extensive damage to residential and commercial buildings and emergency response

facilities. Essential buildings usually play a very important role in emergency response, but

this was not the case in the Bam earthquake. Damage to the fire station, hospitals, and

municipal and communications buildings caused serious problems in emergency response

soon after the earthquake, Sassan Eshghi and Kiarash Naserasadi [23].

Confined masonry demonstrated good seismic performance. All structural walls of all

masonry buildings, one or two stories, irrespective of whether they are constructed with

bricks, cement blocks or stone, confining elements must be constructed. Vertical and

horizontal confining tie-columns and bond-beams provide integrity for the building and make

a seismic resistant structure. By constructing tie-columns in the main corners of the buildings,

the connection of walls at the intersections can be maintained. It should be noted that good

seismic performance of confining ties could be expected only if the ties are well executed. In

other words, the ties with poor quality of concrete are not able to develop a seismic resistant

mechanism.

In order to have a three-dimensional resisting system, tie-columns should be properly

connected at all intersection points to tie-beams. If there is no suitable detailing for

reinforcing bars in the concrete joints, the building can not stand against earthquakes.

Moreover, the distance between axes of two successive tie-columns should be limited to 5

meters. The confined masonry buildings, which did not observe the abovementioned points,

failed during the Bam Earthquake, Kooroush Nasrollahzadeh Nesheli [24].




28

Figure 2.4.10 Vertical and horizontal confining elements maintain the stability of the building, Kooroush

Nasrollahzadeh Nesheli [24]

Another’s earthquake cause: technical school at Avaj, Qasvin province is designed by

ORDENS and constructed of confined masonry walls, Figure 2.4.11. It has survived the

earthquake while a neighbouring building was totally destroyed. It has brick walls anchored to

the foundations and to the rigid floors and roof by means of tie-columns and ring beams,

Usam Ghaidan [25].

Figure 2.4.11 Confined masonry wall in Iran that survive the earthquake, Usam Ghaidan [25]

The 1998 Mionica and the 1999 Trstenik, Yugoslavia Earthquake

In the1998 Mionica (magnitude Mw=5.7) and the 1999 Trstenik (magnitude Mw=5.1),

damage to confined masonry buildings was not extensive. Figure 2.4.12 shows damage to

masonry buildings in the 1998 Mionica earthquake. A number of older un-reinforced masonry

buildings were damaged in the earthquake however confined masonry buildings performed




29

well and did not suffer any significant damage, as illustrated in the figure, Nikola Muravljov,

Radovan Dimitrijevic [26].

Figure 2.4.12 Damage to masonry building in the 1998 Mionica earthquake, Nikola Muravljov, Radovan

Dimitrijevic [26]



Chapter 3. Models

30

3. RESISTING MECHANISMS

3.1 Resisting mechanism

Under an earthquake, the walls are subjected to gravitational and seismic loads as is shown inFigure 3.1.1. Tensions are generated in the masonry panel by the combination of the

gravitational loads and the oscillating earthquake loads. The slab acts like a rigid diaphragm

and transfer the loads directly to the walls parallels to the seismic action, M2. The inertial

forces originated in the walls perpendicular to the seismic action (M1) are also transfer to the

wall M2 in part by the slab and by the connections between the two walls.

It is observed that the resistance to the seismic loads is provided by the wall M2, parallel to

the seismic action.

Figure 3.1.1 and 3.1.2, before and during the earthquake respectively, illustrate the loads andthe tensions originated in the resisting wall M2. This masonry panel will be subjected to

flexion with compression and shear. In constructions of low height, the shear will be

predominant effort.

Figure 3.1.1 Left: Distribution of seismic loads in the building; Right: tension originated by the

gravitational loads before the earthquake, Bustos [10]

Tensions originated

by gravitational

loads

Rigid diaphragm Vertical loads



Chapter 3. Models

31

Figure 3.1.2 Tensions in the confined masonry wall originated by gravitational and seismic loads during

the earthquake, Bustos [10]

Under these loads, confined masonry can be modeled to resist the actions in: two confining

columns (a1), two confining beams (a2) and the masonry panel can behave like two

diagonals, one in compression (a4) and the other one in tension (a3), Figure 3.1.3.

a2

a1a1

a2

a4a3

Hs

Figure 3.1.3 Resisting mechanisms, Universidad Nacional de Córdoba [16]

In this way two resisting mechanisms are generated: A and B, Figure 3.1.4. In the resisting

mechanism A, the diagonal is under compression. Under the action of low amplitude forces

the diagonal a3 may fail in tension, and mechanism A is the only one left to resist the seismic

loads.

Gravitational loads

Seismic action

ΔHs

Tensions originated by

gravitational and seismic loads

Approximated distribution of

shear tensions



Chapter 3. Models

32

RESISTING MECHANISMS

Hs/2 Hs/2

A B

d

Figure 3.1.4 Resisting mechanisms of confined masonry walls, Universidad Nacional de Córdoba [16]

It is very important to remark the fact of having links between the walls panel for a good

behaviour under an earthquake action. They must perform a rigid box for an efficient load

transmission.

To materialize the previous condition it is necessary to have slabs that behave as non-

deformable diaphragms and sufficient anchorage of reinforcement in the confinementelements, Figure 3.1.5. In this way the mechanism of walls and interconnected floor slabs are

able to distribute the lateral forces to the walls parallels to the seismic action.

Figure 3.1.6 presents the consequences of a bad connection between slabs-walls and walls-

walls. In this case, due to a wrong distribution of forces the panel M2 has an important

bending moment in the perpendicular direction.

a3



Chapter 3. Models

33

Figure 3.1.5 Non-deformable diaphragms and good connection between walls allowing the correctdistribution of the seismic action. Decanini and Payer [17]

Figure 3.1.6 Deformable slab and no capacity of load distribution. Consequence: An important bending

moment in wall M2 is generated. Decanini and Payer [17]



Chapter 4. Experimental tests

34

4. EXPERIMENTAL TESTS

The results of experimental tests concerning confined masonry walls are summarized and

discussed in this chapter. Dynamic behaviour of confined masonry walls test using shaking

tables or cycling loading are described with the correspond results of cracks patterns,

hysteretic behaviour, energy dissipation, stiffness degradation and other mechanical

characteristics. Experimental tests concerning the study of the influence of openings and

distance between tie-columns in the strength of the confined masonry are given as well.

4.1 Dynamic behaviour of confined masonry buildings through shaking table tests

4.1.1 Assessment of the response of Mexican confined masonry structure through

shaking table test, Alcocer et al [27]

Little information is presently available on the response of three-dimensional confined

masonry structures subjected to controlled dynamic excitations, like those applied through

shaking tables tests.

The dynamic behaviour of two small-scale confined masonry buildings tested in shaking table

is discussed. Specimens were half-scale models of typical low-cost housing buildings of one

and three stories constructed in Mexico, hereafter referred to as M1 and M3 specimens

respectively. Models were subjected to a series of seismic motions characteristic of Mexican

subduction events recorded in the epicentral region.

Walls were made of hand-made solid clay bricks confined by reinforced concrete tie-columns

and bond-beams. In the direction of the earthquake-simulator motion (E-W), three wall axes

were built, Figure 4.1.1. The facade walls had door and window openings, whereas the middle

walls were solid. In the prototype, the middle wall axis divides two adjacent dwellings. In the

transverse direction (N-S), four walls were built to improve the gravity load distribution

among walls, and to control possible torsional deformations. Models were symmetrical and

the wall distribution was uniform over the specimen height.




35

Figure 4.1.1 Characteristics of the Specimens, Alcocer et al. [27]

Tie-column and bond-beam reinforcement was made of four longitudinal wires and hoop

reinforcements spaced at 100mm. In M3, aimed at increasing wall shear strength, controlling

damage and achieving a more stable behaviour, hoop spacing was reduced to 30mm at tie-

columns ends. Floor systems were cast-in-place reinforced concrete solid slabs supported on

bond-beams. Slabs were reinforced with 4,76mm diameter deformed wires, spaced each

150mm in both directions. The models were built on a steel platform, Figure 4.1.2.




36

Figure 4.1.2 Reinforcement of the Specimens, Alcocer et al. [27]

Two earthquake motions recorded in epicentral regions in Mexico were used as basis for the

testing program. One was the motion recorded in Acapulco, Guerrero, in April 25, 1989,

during Mw=6.8 earthquake with PGA=0.34g. The other was that recorded in Manzanillo,

Colima, in October 10, 1995, during Mw=8.0 earthquake with PGA=0.40g.

TEST RESULTS

Analysis of data confirmed that shear deformations controlled the response like is showed in

Figure 4.1.3. In M1, damage was mainly characterized by horizontal and inclined cracks. The

first inclined cracks formed near the wall center, and propagated towards the corners of tie-

columns ends, except for walls MS4 and MN4, where behavior was dominated by a shear-

sliding mechanism (horizontal cracks at the walls base) and inclined cracks in the lower part.

First diagonal cracking occurred at a drift ratio to 0.36%. Crack propagation into the tie-

columns ends, thus shearing off these elements, was recorded at a drift ratio to 0.67%. At the

end of the test runs, maximum recorded drift ratio was 1.83%.

In M3, damage was mainly concentrated in the first story, ground floor. In general, walls

exhibited one or two large inclined cracks at 45-deg (X-shaped). First diagonal cracks formed

at a drift ratio to 0.25%. Penetration of inclined cracking to tie-columns ends was recorded at

a drift ratio to 0.43%. A full soft-story mechanism was readily observed during test runs for

which the maximum recorded drift ratio was 1.75%.

In the second story, few horizontal cracks at the base of the walls were observed, whereas inthe third story no cracking was observed.




37

Based on the failure mode observed, the analytical model for design and assessment could be

simplified by assuming that all inelastic deformations would take place at the first story and

would be controlled by shear.

Figure 4.1.3 Final cracks patterns, Alcocer et al. [27]

Hysteretic loops were typical of confined masonry structures. The elastic limit was defined by

the occurrence of the first inclined cracking in the masonry wall; strength was achieved whenthe maximum base shear was attained; and the ultimate limit state was considered at a lateral

drift ratio when 20 percent reduction in strength was recorded, Figure 4.1.4.

Figure 4.1.4 Response envelope for M1 and M3; MCBC: Mexico City Building Code, Alcocer et al. [27]




38

The measured response characteristics are shown in Table 4.1.1. Cycles within the elastic

limit experienced some hysteretic attributed to wall flexural cracking at initial stages. As it is

common in confined masonry structures, specimens attained their maximum strength at loads

higher than those associated to first inclined cracking. Specimen M1 showed stable and

symmetric loops up to large drift ratios, whereas in M3, hysteretic curves were stable and

symmetric up to the strength limit state, after which a severe strength and stiffness decay,

because of damage over the panels and at tie-columns ends, was developed. As it is customary

in shear-governed members subjected to inelastic deformations, response curves exhibited

severe pinching, especially at very large lateral drift ratios associated to failure of the

structure. In M3, at the ultimate limit state, a fast degrading process, involving sliding along

the first story inclined cracking and crushing of masonry and concrete, was clearly observed.

It was apparent that stories 2 and 3 laterally deformed very slightly, suggesting a rigid body

motion over the first story. This phenomenon led to a concentration of deformations and

damage at the first story which performed as a soft-story with shear governed mechanism.

Table 4.1.1 Measured response characteristics, Alcocer et al. [27]

Stiffness decay was observed at low drift ratios, even before first inclined cracking became

apparent. This phenomenon is attributed to incipient wall flexural cracking, and perhaps, to

some micro-cracking in masonry materials, local loss of mortar bond and adjustment of brick

position. After first inclined cracking, but before reaching strength, the decay increased with

drift ratio. At larger drift ratios, decay remained nearly constant. At this stage, stiffness decay

is associated to cracking and crushing in masonry walls and RC confinement members.

The energy dissipated during the tests was computed as the area within the hysteretic loops

from the base shear–drift relations. M1 dissipated, in absolute terms, more energy that M3;

moreover, at same drift ratios, M1 also dissipated more energy. At present, it is contended that

the failure mode of M1, characterized by shear and sliding mechanisms, contributed to the

difference in behaviour in both specimens.




39

4.1.2 Seismic behaviour of confined masonry, Tomazevic et al. [28]

Two models of a three-storey confined masonry buildings have been tested on a simple uni-

axial earthquake simulator. The models have been made at 1:5 scale according to the

assumptions of the theory of complete models. Both models where identical and represented

a typical three- to four-storey house with structural walls at 5.78m in one and 4.65m distance

in the other direction, designed according to Chilean engineering practice. Since the

distribution of structural walls of the model was not symmetric in both directions, model M1

was tested in the longitudinal, whereas model M2 was tested by subjecting it to the simulated

earthquake ground motion in the transverse direction. By testing the model transversally, the

possible torsional effects have been studied.

The building was a three-storey structure, composed of ground floor and two typical storeys

with storey height 2.47m. According to the design, structural walls were built with hollow blocks units: in the ground floor, the thickness of the walls was 17.5cm, in the upper storeys;

however, the thickness of the walls was reduced to 14cm. Lime-cement mortar in the

proportion of 1:0.25:4 (cement: lime: sand) was used to construct the walls. Vertical tie-

columns were reinforced with 4ø8mm bars and were grouted with concrete with a

characteristic compressive strength 16MPa.

Bearing walls were supported by a continuous reinforced concrete strip foundation. Floors

were cast in situ monolithic reinforced-concrete slabs 12cm thick, roof structure was wooden.

Typical plan and vertical section of smaller, square unit, which has been used as a basis for

the design of the models, are shown in Figure 4.1.5 and Figure 4.1.6. The distribution ofconfining vertical element, tie-columns can be also seen in these figures. Reinforcement of

floor slabs, tie-columns and bond-beams are illustrated in Figure 4.1.7.




40

Figure 4.1.5 Typical floor plan of prototype building, used as a basis for the design of 1:5 scale models,

Tomazevic et al. [28]

Figure 4.1.6 Typical section prototype building, used as a basis for the design of 1:5 scale models,





41

Figure 4.1.7 Reinforcement of floor slabs and vertical and horizontal bonding elements, Tomazevic et al.

[28]

Both models have been tested by subjecting them to a sequence of simulated earthquake

ground motion with increased intensity of motion during each subsequent test run, Figures

4.1.8 to 4.1.10. During the shaking-table tests, the displacement and acceleration response of

the models has been measured at three points at each storey level. The changes of strain in

vertical reinforcement of typical tie-columns have been also followed. Similar behaviour of both models has been observed, with symmetrical amplitudes of vibration at both sides of the




42

models, despite the expected torsional behaviour in the case of the model tested transversally.

As a result of relatively high wall/floor area ratio in both directions of the tested structures,

the observed seismic resistance was very high, in both cases. However, significant strength

degradation has been observed after the attained maximum value, with increased damage to

the masonry wall and subsequent falling off of the masonry.

Simulation of seismic loads was made with the first 24 seconds of ground acceleration record

of Montenegro earthquake of April 15, 1979, N-S component of the Petrovac record, with

peak ground acceleration of 0.43g has been used for simulation of earthquake ground motion.

Several individual test runs in the shaking-table were made, the characteristic of parameters

use are described in Table 4.1.2.

Figure 4.1.8 Earthquake simulator set-up, Tomazevic et al. [28]


gauges on reinforcing steel of vertical confinement of model M1, Tomazevic et al. [28]




43


gauges on reinforcing steel of vertical confinement of model M2, Tomazevic et al. [28]

Table 4.1.2 Characteristic parameters of shaking-table motion recorded during individual test runs,


TEST RESULTS

Development of cracks and damaged propagation in the structural elements of both models

during the shaking-table tests has been inspected visually.




44

In model M1 no damage was observed after the initial phases of tests, test run R5 and R25.

The first tiny crack was at the first floor after the test run R50, the real initiation of diagonally

oriented cracks in the walls in the first storey was test run R75. The cracks were not all

oriented in the same direction. No cracks could be seen in the walls orthogonal to the

direction of seismic motion.

During the test run R100, the existing cracks propagated, Figure 4.1.11(a). In some walls, new

diagonal cracks developed, oriented in the other diagonal direction. Some horizontal cracks

have been also observed in the parapets, passing through the mortar joints.

The damage was serious during run R150, Figure 4.1.11(b). Most of the cracks passed

through mortar joints and it start the crushing of masonry units in the middle. Severe stiffness

degradation was observed as a result of damage to wall, occurred during test run R150, and,

consequently, large displacements amplitudes of vibration have been measured, no damage

has been observed to the walls, orthogonal to seismic excitation. Also, no damage has beenobserved to confining elements.

Heavy damage occurred during test run R200, Figure 4.1.12 and Figure 4.1.13. In the first

floor, all the walls oriented in the direction of excitation disintegrated and fall out of the

confinement: in the middle sections of the walls, masonry units crushed, and at vertical

borders the walls separated from the confining elements. This indicates that tie-columns and

bond-beams are only active until a certain level of lateral displacements; afterwards they

cannot prevent the disintegration of the masonry, unless it is reinforced with horizontal,

mortar bed-joint reinforcement.

During test run R200, the central wall partly collapsed. Parts of the wall failed in shear, in

some parts; however, sliding shear failure was the reason of collapse.

Figure 4.1.11 Model M1, northern side-propagation of cracks at the eastern corner, Tomazevic et al. [28]




45

Figure 4.1.12 Left: Model M1: middle pier after test run R200; Right: Model M1: detail of damage to tie-

column after test run R200, Tomazevic et al. [28]

Figure 4.1.13 Model M1, southern side, cracks after test runs R100, R150 and R200, Tomazevic et al. [28]




46

On the other hand, the behaviour of model M2 during shaking tests was basically similar as

the behaviour of model M1. Obviously, different position of the model on the platform during

testing and non-symmetry of structural system with regard to the transversal axis caused

slightly different damage propagation.

First cracks in the walls of model M2 developed during moderate excitation in the beginning

of shaking test (R25). Diagonally oriented cracks in the walls were not located symmetrically:

on the southern side of the model, cracks have occurred at the eastern part of the shear-wall.

After the run R50, the crack pattern became symmetric, since cracks in the other; previously

not damage parts of the shear-walls have been also observed. In both walls, diagonal cracks

have been oriented from the corners at the bottom to the corners at the top of the model. No

cracks have been observed in the middle part of the shear-walls.

During test run R75, the cracks propagated along the whole height of the model. In the walls

where diagonal cracks in one direction have been observed after test run R25, cracks in theother diagonal direction have occurred. In the central shear-wall hardly visible cracks

developed in both diagonal directions.

Model M2 was seriously damaged during test run R100. A system of cracks, oriented in both

diagonal directions, developed in all elements of all shear-walls in the direction of seismic

motion. Most of the cracks passed through mortar joints and the first signs of crushing of

masonry unites have been observed in the middle.

During test run R150 the damage to model walls increased. The walls of the first and second

storey stated to falling of. Initialization of micro-concrete at the joints between vertical andhorizontal tying elements has been also observed.

The extent of damage to model M2 during test run R200 significantly changed the dynamic

characteristics of the model. Practically all walls in the first storey failed: the masonry units

crushed and the broken parts of the walls simply disintegrated and fell out, so that the model

was left standing mainly due to confining elements, without any masonry infill. The damage

in the upper storeys did not increase.

The model M2 was submitted to a repeated strong excitation R200/1. It did not cause further

increase of damage to the structure, so the model was subjected to a series of sinusoidal

motion which followed the decayed natural frequency of the masonry in the first storey,

increased immensely, the model started pounding with the rigid steel supporting structure,

fixed at a distance of about 10cm from the model to the foundation slab. As the model

consequently leaned to the steel structure, and did not fall on the opposite side, the testing was

terminated. The collapse mechanism is illustrated in Figure 4.1.14 and Figure 4.1.15.




47

Figure 4.1.14 Model M2: mechanism of collapse, Tomazevic et al. [28]

Figure 4.1.15 Model M2: mechanism of collapse, Tomazevic et al. [28]




48

4.1.3 Seismic behaviour of a three-story half scale confined masonry structure, San

Bartolomé et al. [29]

In this research it was study, analytically and experimentally, the seismic behaviour of a

reduced scale model (1:2.5) through shaking table test. The walls represent one perimetric

wall of a 3-story building, made of clay masonry confined by reinforced concrete elements.

The geometry of the specimen is given in Figure 4.1.16. Reinforced concrete slabs with added

load were used in the model. The specimen weight was 57.78KN; therefore the axial stress in

the first-story walls was 0.33MPa.

Figure 4.1.16 Geometry of the 3-storey confined masonry specimen, San Bartolomé et al. [29]

The masonry units were solid clay bricks of 11MPa of compressive strength. The mortar was

1:4 (Portland cement: sand) with a compressive strength of 6MPa. The concrete of the tie-

columns and bond-beams had a compressive strength of 15MPa and elastic modulus

Ec=13700MPa. The axial compression was tested with four masonry prisms and it results of

6MPa and the elastic modulus was E=1510MPa. To obtain the shear strength, four squaremasonry prisms were tested to diagonal compression, giving a shear strength of 0.8MPa and a

shear modulus G=450MPa. Vertical reinforcement in each column was 4#5.5mm wire steel.

The yield stress was 220MPa and the ultimate stress was 316MPa. Horizontal reinforcement

was added in the first-story in a small ratio of 0.016% (1#1.8 every 3 layers anchorage in the

columns), even that in Perú it is not common to use horizontal reinforcement in confined

masonry walls.

The specimen was design according to the Peruvian code (ININVI). Flexural and shear

capacity of the walls were computed, Table 4.1.3. The shear capacity of each wall was

evaluated using the formula (1) proposed by San Bartolomé, 1990. The results of the shear




49

capacity were greater than the values associated to the yielding flexural capacity and also to

the maximum flexural capacity; therefore, a flexural failure was expected.

Table 4.1.3 Assumed force distribution in one specimen wall, San Bartolomé et al. [29]

In the dynamic test the input wave was the L component of the May 31, 1970 earthquake,

recorded in Lima. The horizontal excitation was in the wall direction and the peak platform

displacement for each run is shown in Table 4.1.4.

Table 4.1.4 Shaking table test runs, San Bartolomé et al. [29]

TEST RESULTS

In run A no cracking occurred. At run B a flexural crack appeared at the walls base, causing

the yield of the vertical reinforcement. The shear failure in both first-story walls occurred in

run C, at the end of the test the specimen was in an irreparable condition as is shown in Figure

4.1.17. During this run the horizontal reinforcement broke, showing that it effectively worked

under dynamic conditions, including the sliding of the upper stories across the first-story

diagonal cracks.




50

Figure 4.1.17 Specimen after run C, San Bartolomé et al. [29]

First flexural crack prediction: a moment M= 31.4KN-m was obtained at the base for one

wall, while experimental result in run B was M= 32.9KN-m.

Shear strength prediction: the formula 1 was applied for the wall and the shear strength

obtained was of 22KN. This prediction is 13% less than the experimental value obtained inrun C (24.9KN), so the correlation is acceptable.

The following comments can be made:

• The failure of the specimen was concentrated only at the first story, while the upper

stories the actual shear force never surpassed the theoretical shear strength, so their failure

was avoided.

• Referring to the ductility factor the obtained value was 1.8, this calculated with the

shear forces obtained, Figure 4.1.18. This value is less than the Peruvian Code specification

for confined masonry Rd= 2.5, so the Code does not appear to be conservative.

• The platform acceleration was 0.54g at the instant the shear failure occurred. This

value has never been recorded in Perú.

• The research has shown that a shear failure is possible to occur when a strong

earthquake hits a confined masonry structure, even in the case that the structure satisfies the

ideal characteristics to obtain a flexural failure. Therefore, the design process of a confined

masonry buildings should include the possibility of a shear type of failure to avoid structural

collapse.




51

Figure 4.1.18 Left: Total base shear force vs. displacement at level 1 in run C; Right: Lateral force in one

wall at the time of maximum base shear force at each run (A, B and C), San Bartolomé et al. [29]

4.1.4 Pseudo dynamic tests of confined masonry buildings, Scaletti et al. [30]

Pseudo dynamic tests were carried out to investigate the behaviour of confined masonry

structures. Two-story one-bay specimens, with two parallel walls connected by stiff horizontal

slabs, were used for the tests. One full scale specimen and one half scale model were built based on Peruvian standards for confined masonry. Figure 4.1.19 show the dimensions of the

specimens. Scale factors were 2 for displacement, 1 for acceleration as for strain, angular

distortion, stress and elastic modulus, square root of 2 for time and 4 for mass and force. The

mass of the full scale specimen was 15.26tn (including added masses of 2.54tn on each slab).

Figure 4.1.19 Test specimen, Scaletti et al. [30]




52

The materials used for the specimens were: clay bricks units laid with a 1:4 (cement: sand)

mortar. The concrete used for footings, columns and slabs had a nominal strength of 20MPa.

Columns were reinforced with 4#3 longitudinal bars and stirrups #2 at 14cm, except near the

joints, were 10cm spacing was used. The yielding stress of the steel was 410MPa.

The test program included: static test of half scale model under monotonic loading, shaking

table test of half scale model and pseudo dynamic test of one full scale specimen and one half

scale model.

Steady-state resonance test were performed using a small rotating eccentric weight exciter,

producing a horizontal sinusoidal force parallel to the walls. Figure 4.1.20 shows typical

resonance curves, obtained for the full scale model. Damping was estimated by considering

the specimen as a one degree of freedom system and using the bandwidth method. Modal

shapes were obtained from the ratios of acceleration amplitudes at resonance. Natural periods,

frequencies and percentage of critical damping of the specimens tested are also showed inFigure 4.1.20.

Figure 4.1.20 Left: resonance curves for full scale specimen; Right: natural periods, frequencies, damping

and modal shapes; Scaletti et al. [30]

For the PD tests of the half scale model the input signal was the same used for the shaking

table test. It consisted of a series of 5Hz sine waves with different amplitudes, Figure 4.1.21.

During the first stage of the test, while the specimen had little damage, this 5Hz base motion

was equivalent to a static loading. The maximum acceleration in the input signal is 1.3g,

although the specimen failed during the stage with maximum acceleration of 1.06g. The

integration time interval was 0.004sec.




53

The full scale specimen was tested with a ground acceleration corresponding to the NO8E

component of the Lima earthquake of October 10, 1966. The test was repeated four times,

scaling the record as required to have a maximum ground acceleration of 293.6gal (original

record), 400, 800 and 1200gal. the integration time interval was 0.004sec, Figure 4.1.21.

Figure 4.1.21 Left: input signal for PD test of half scale model; Right: input signal for PD test of full scale

model; Scaletti et al. [30]

TEST RESULTS

For the half scale model thin cracks were observed at the base of the walls from the beginning

of the test. Diagonal cracks developed during the second stage and became increasingly

important after 4 seconds. A large strain increment in the longitudinal steel reinforcement of

the columns occurred at the same time. The failure mode was shear, involving both masonryunits and mortar joints. Diagonal cracks were also observed in the second level, this damage

can be related with defects on the construction in one of the walls. A plot of base shear versus

first story displacement is shown in figure 4.1. The behaviour was almost linear during the

first two stages of test, while the story drift angle was less than 1/1000. Stiffness degradation

and hysteretic were important from the third stage. The right part of Figure 4.1.22 compares

envelope curves from static and dynamic tests reported by San Bartolomé et al., 1991, with

those from the PD tests. Good agreement was found between results of static and shaking

table tests. Lower values obtained in the PD tests may be due to strain rate.




54

Figure 4.1.22 Left: base shear vs first story displacement, pseudo dynamic test of half scale model; Right:

envelopes of base shear vs first story displacement of the half scale model; Scaletti et al. [30]

For the full scale model the failure mode was by shear. Cracks were noticeable in the first

story walls after the 400gal earthquake. The second story walls remained practically

undamaged. Table 4.1.5 lists maximum first floor displacement (u), maximum shear base

shear (V), and predominant response period (T) for different levels of ground acceleration (a).

Although maximum displacements and base shears correspond to only one point of each

record, their relative magnitudes and the period elongation reflect the importance of

nonlinearities in the response. The specimen failed at an average shear stress in the first level

of 0.32MPa, considerably lower than that reached by the half scale model. The allowable

design stress in the current Peruvian Code is 0.16MPa. First story displacement time histories

and base shear time histories of the full scale model are shown in Figure 4.1.23

Table 4.1.5 First floor displacement, base shear and predominant period as a function of maximum

ground acceleration, Scaletti et al. [30]




55

Figure 4.1.23 Left: first story displacement time histories of the full scale specimen; Right: base shear

time histories of the full scale specimen, Scaletti et al. [30]

4.2 Dynamic behaviour of confined masonry panels under cyclic lateral loads

4.2.1 Experimental behaviour of masonry structural walls used in Argentina, Zabala et

al. [5]

Confined masonry is extensively used in seismic regions of Argentina. Experimental data

about confined masonry built using local practice are very scarce and this lack of knowledge

affects the seismic safety and the design practice of masonry structures.

In order to obtain better knowledge about the seismic behaviour of confined masonry wallsused in the seismic region of Argentina tests were performed on six full-scale model walls at

the Earthquake Research Institute of the National University of San Juan (IDIA). The walls

were built with handmade solid ceramic bricks, 18 cm wide. Wall confinement was provided

by reinforced concrete columns, with nearly square sections, 20cm wide by the thickness of

the wall. The design of the tested models was based on the typical building layout used by the

San Juan Provincial Institute of Housing (IPV), and built with the recommendations given by

the INPRES-CIRSOC 103 code. Model dimensions are showed in Figure 4.2.1.




56

Figure 4.2.1 Model Dimensions, Zabala et al. [5]

The walls were tested under a prescribed constant vertical load and allowing free rotation ofthe upper end. The vertical load was applied through a stiff steel beam by means of two

vertical servo-controlled actuators, Figure 4.2.2. The tests were performed by applying cycles

of lateral displacements at the wall head. The instrumentation consisted of one displacement

transducer controlling the horizontal displacement of the wall head, two vertical displacement

transducers at both sides of the model, two diagonal displacement transducers, three load cells

mounted in series with the hydraulic jacks and a number of strain gages applied to some

columns reinforcement bars.




57

Figure 4.2.2 Outline of the test setup and its instrumentation, Zabala et al. [5]

TEST RESULTS

Table 4.2.1 summarizes the main features of the six tested walls. The models 1 to 4 developed

the crack pattern presented in Figure 4.2.4. This pattern includes diagonal cracking of themasonry panel and partial separation of the confinement columns. These walls clearly show a

shear failure, but sustained their strength for a displacement up to 20mm. None of these walls

reached their theoretical flexural capacity and the final state was controlled by the columns

shear strength. This is due to the fact that, under large displacements, diagonal cracking of

masonry extended to the columns, Figure 4.2.3. Compression failure never occurred.




58

Table 4.2.1 Main features of the six tested walls, Zabala et al. [5]

Notes:

(1) Considering the horizontal load applied at the horizontal actuator level, the applied vertical load and σs= 420MN/ m2 (yield stress of the steel)

(2) Vur= (0.3 σ +0.6 τmo)Bm. Where σ = compressive stress, τmo = diagonal shear strength of small masonry

probes. τmo= 0.3 MN/ m2

(3) Additional strength due to horizontal masonry reinforcement.

Figure 4.2.3 Shear failure in column of wall 1, Zabala et al. [5]




59

Figure 4.2.4 Crack pattern developed in the first 4 testing walls, Zabala et al. [5]

Walls 5 and 6, having a shear capacity clearly larger than the flexural capacity, reached, by

hardening of the vertical reinforcement bars, strength values substantially larger than the

theoretical flexural capacity.

Under the applied displacement cycles with increasing amplitude, it was observed that these

walls (Figure 4.2.5) maintain their strength and their energy dissipation ability for larger

displacement amplitudes than walls 1 to 4. Bending-induced horizontal cracking was

observed and the separation between column and panel did not occur. The final state is

controlled again by the shear strength of the column at the joints with the confinement beams.

Figure 4.2.5 Crack pattern developed in walls 5 and 6, Zabala et al. [5]




60

The INPRES-CIRSOC building code allows a reasonable estimation of the wall strength,

based on measured strength in diagonal shear test of small masonry probes, only in the case of

lightly reinforced columns, not providing larger flexural capacity than shear capacity (See

Table 4.2.1). For larger reinforcement ratios of columns, the wall strength is controlled by the

shear strength of the confinement columns and beam joints.

The code should require the capacity design of columns and joints reinforcement, considering

the maximum expected shear force induced by the compressed masonry strut, arising from the

cracking pattern of the panel. For the used brick type, a compression failure of this strut is not

likely to occur and therefore the wall strength becomes controlled by the vertical

reinforcement of the columns. The amount of transverse reinforcement in critical zones of the

confinement columns and beams normally used in practice is insufficient in order to sustain

this shear force.

4.2.2 Behaviour of multi-perforated clay brick walls under earthquake type loading,

Alcocer and Zepeda [31]

To evaluate the behaviour and to develop analysis, design and construction guidelines of this

type of brick walls, four large-scale isolated load-bearing walls were built and tested under

constant vertical axial load and cyclic lateral loads.

Previous research conducted on this issue made clear that the mode of failure of these bricksis quite brittle. On the other hand, its economic advantages compared to the traditional hand-

made bricks have made multi-perforated bricks an increasingly popular construction system

for low-cost housing.

The control specimen, N1, consisted of an un-reinforced wall panel, made of multi-perforated

bricks, confined in its vertical edges with tie-columns built within hollow clay bricks, Figure

4.2.6. In specimens N2 and N3, the minimum horizontal reinforcement ratio as required by

the Mexico City Building Code, was provided. N2 was confined with similar internal tie-

columns as in N1, whereas in N3 external reinforced concrete tie-columns were used. In

specimen N4, the horizontal reinforcement ratio was almost four times the minimum value;

internal tie-columns were built using special hollow pieces fabricated to achieve a larger tie-

column cross sectional area.




61

Figure 4.2.6 Characteristics of the specimens, Alcocer and Zepeda [31]

TEST RESULTS

Final crack patterns and hysteretic loops are shown in Figure 4.2.7 and Figure 4.2.8

respectively. In N1, damage was mainly concentrated in two large inclined cracks that

extended into the lower ends of the internal tie-columns. After the “x” crack pattern wasformed, the wall lost its capacity for carrying vertical and lateral loads.




62

A more uniform distribution of cracks was observed in specimens with horizontal

reinforcement; the larger the amount, the more uniformly distributed the cracking was. Main

cracks were inclined at 45° and extended through bricks and mortar joints. At same drift

levels, crack widths in horizontally reinforced specimens (i.e. N2 to N4) were smaller than

those recorded in N1. Typically, flexural cracks formed at a drift angle of 0.09%; the x pattern

of inclined cracking was formed at 0.15%.

Model N2 abruptly failed after fracture of four horizontal wires, which led to a shear

compression failure of few multi-perforated bricks along the crack (and along the internal

compression strut), and to shearing off the lower ends of the tie-columns.

Damage in specimen N3 was concentrated in the panel with extension of some fine cracks

into the tie-column. Similarly to N2, some bricks exhibited spalling of their exterior walls

after crushing or fracture of interior walls. Specimen N4 exhibited a very uniform distribution

of fine cracks over the wall. Crushing and spalling of exterior brick walls in the seconduppermost brick course triggered the failure. Analysis of strain gages on the horizontal

reinforcement of N2 to N4 indicated that wires remained elastic only in N4.

The most severe damage in tie-columns was observed in N1, although closely spaced

crossties and hoops were placed at their ends. In the other specimens, horizontal

reinforcement better controlled the wall shear deformations and clearly improved the stability

of the behaviour after cracking, thus delaying the crack extension into the tie-columns.

The inter-story drift angle R is defined as the ratio of the applied horizontal displacement

measured at the slab level to the specimen height.




63

Figure 4.2.7 Final cracks patterns of the fourth specimens, Alcocer and Zepeda [31]

Based on the observations made during the tests, and on the analysis of the instrumentation,

the following conclusions were developed:

• Masonry diagonal compressive strengths, related to design shear stresses, varied with

the amount of mortar penetration into the multi-perforated bricks. Larger strengths were

obtained with fluid mortars.

• First inclined cracking occurred at a drift angle of 0.1%, disregarding the amount ofhorizontal reinforcement, as well as type and detailing of the tie-columns.

• Shear deformations governed wall behaviour.

• As compared to walls without panel reinforcement, walls reinforced horizontally with

deformed cold-drawn small-diameter wires exhibited a superior behaviour in terms of lateral

strength, deformation and energy dissipation capacities, strength degradation, damage

distribution, and crack widths.




64

• The increase in lateral strength was not linearly proportional to the amount of

horizontal reinforcement. Moreover, the mode of failure is strongly dependent on the

horizontal reinforcement ratio “ ph” and its yield stress “ fyh”.

• The contribution of the horizontal reinforcement to the wall lateral strength was afunction of the lateral displacement and the type of tie-column.

• As compared to walls with internal tie-columns, the specimen with external RC tie-

columns exhibited higher lateral strength, stiffness, energy dissipation and deformation

capacities and a more stable behaviour.

Figure 4.2.8 Hysteretic curves, Alcocer and Zepeda [31]

4.2.3 Experimental investigation of the seismic behaviour in full- scale prototypes of

confined masonry walls, Decanini et al. [32]

In this investigation test results obtained in the laboratory of the National University of

Cordoba, Argentina are presented and discussed. In this test series, 8 confined masonry panels

are subjected to the effects of horizontal loads simulating the seismic action.

Four of the masonry panels were made of solid clay brick and the rest of hollow clay bricks.

Confined masonry panel M1 was made of solid clay bricks and has not openings, M2 to M4

were made also of solid clay bricks but they present an opening in the middle of the panel. M5

and M6 were confined masonry panels made of hollow clay bricks without openings and

finally walls M7 and M8 were also of hollow clay brick but with an opening in the centre ofthe panel. General dimensions of the prototypes are shown in Figure 4.2.9. They all were built




65

with the recommendations given by the INPRES-CIRSOC 103 code. The design of the tested

models was based on the typical building layout used in Córdoba.

The reinforcements used for both typologies are illustrated in Figure 4.2.10.

Figure 4.2.9 General dimensions of the prototypes, Decanini et al. [32]




66

Figure 4.2.10 Left: Reinforcement of confined masonry of solid clay bricks; Right: reinforcement of

confined masonry of hollow clay bricks, Decanini et al. [32]

The walls were tested allowing free rotation of the upper end. No vertical load was applied,

Figure 4.2.11. The tests were performed by applying cycles of lateral displacements at the

wall head. The loading device consisted in two “Amsler” jacks of 10tn of maximum capacity.




67

Figure 4.2.11 Test setup and its instrumentation, Decanini et al. [32]

TEST RESULTS

In Table 4.2.2, results are showed for the different models, also showing the maximum

angular deformation, first cracking, ultimate cracking and maximum load achieved in the test.

The following comments can be made:

• There is not substantial difference between the levels of loads reached by the forceapplied in one sense and the other of the different steps considered, although some differences

are found in the values of maximum angular deformation.

• In the walls made of solid clay bricks, the loads that correspond to the ultimate

cracking are approximately twice the one that produces the initial cracking.

• On the other hand, for walls made of hollow clay bricks the load that produced the

ultimate cracking is only 20% higher than the one that produced the initial cracking.

• The loads that produced initial cracking of the walls with openings are half of the load

that produced the same effects in the wall without the openings.

• Ultimate loads for masonry walls of solid clay bricks are 50% more than the

maximum loads for masonry walls made of hollow clay bricks, but the load that produce the

initial cracking is higher for the hollow clay bricks than for the solid clay bricks.




68

Table 4.2.2 Measured loads and angular deformations, Decanini et al. [32]

In Figure 4.2.12 final cracking pattern is showed for walls of solid clay bricks M3. The failure

mechanisms in all the cases were due to shear. The maximum shear stresses for the panels of

solid bricks were between 1.20kg/cm² and 2.20 kg/cm². On the other hand for hollow bricks

this maximum shear stress was around 1.00kg/cm² and 1.30kg/cm². These values are less that

the ones given by the INPRES-CIRSOC 103 code may be attributed to the poor resistance of

the mortar joints utilized in the models.




69

Figure 4.2.12 Initial and ultimate cracks of the testing wall M3, Decanini et al. [32]

The reinforcement utilized in the tie-columns was 4 φ8mm for the walls conformed by solid

bricks, that correspond two the minimum reinforcement recommended by the Argentinean

code for seismic zone 3 and 4. And for the hollow bricks the reinforcement of the tie-columns

was 4 φ6mm corresponding to seismic zones 1 and 2.

All the tests showed that the bending strength of the walls was higher than the required in this

experience. No flexural failure was registered.

The stiffness reduction consequence of the non linear behaviour of the masonry panels

observed is as follow:

• Stiffness reduction in relation to the initial one:

Initial crack 50 to 70% of the initial stiffness

Complete crack 20 to 30% of the initial stiffness

Maximum load 5 to 20% of the initial stiffness




70

4.2.4 Influence of vertical and horizontal reinforcement: Influence of the tie-column

vertical reinforcement ratio on the seismic behaviour, Irimies [33]

The influence of both confining of un-reinforced masonry walls and vertical reinforcementration in tie-column on the seismic behaviour of the masonry walls is investigated. Three half

scale, two-story masonry walls were tested. One of them was an un-reinforced masonry wall;

and two others were confined masonry walls. The wall models were tested under reversed

lateral cyclic loading statically applied, in presence of a constant vertical force. The

experimental results indicated that the confining of an un-reinforced wall by RC tie-columns

led to the increase of the lateral resistance and to the change of failure mechanism.

Figure 4.2.13 shows the two models, W1 was an URM wall and two models were confined

masonry walls (WC1 and WC2). The difference between these last two was the amount of

vertical reinforcement. The model WC1 had 4ø8mm steel bars and the model WC2 had4ø10mm steel bars. The cross section of the vertical bars in tie-columns of the WC2 model

was 1.8 times greeted than the WC1 model.

Figure 4.2.13 Experimental models, Irimies [33]

The load application and instrumentation were under reversed lateral cyclic loading statically

applied, in presence of a constant vertical force. The vertical load was applied on the top

beam by means of hydraulic jacks.

The load was distributed along the RC beam by a rigid steel beam. The level of vertical

loading provided a constant compressive stress equal to 0.4MPa, including the wall weight, at

the wall’s base, which is considered typical for 3 and 4 stories housing buildings in Romania.




71

TEST RESULTS

The URM specimen, W1, revealed a flexural behaviour, as expected. One single horizontal

crack in the bed joint at the model’s base occurred for the two loading’s directions. At the

limit drift ration of 0.35%, the crack’s length was about 75% of model’s length. The crushingof masonry units occurred at a drift ratio of 1.0%. The test was continued up to a drift ratio of

2.8% without a severe diminution of the lateral load.

The confined specimens presented, at small displacements, only flexural cracks in tie-columns

and in first story masonry panel, Figure 4.2.14 and 4.2.15. The first major damage observed in

these specimens was a diagonal crack in the first level masonry panels developed at a lateral

load equal to about 93% maximum load in both directions. The diagonal cracks were,

generally, stair-step in mortar joints. In the WC2 model, the diagonal cracks were followed

immediately by shear cracks in the column-beam joints. In the model WC1 model, shear

cracks in joints occurred at large drift ratios, before the maximum lateral load. As theamplitude of displacements cycles increased the shear cracks in tie-columns widened

considerably followed by large slips along the bed joints, which resulted in crushing of the

concrete and the masonry at the corners and inside the panel. There was a good bond between

the tie-column concrete and the adjacent masonry.

Figure 4.2.14 Damage patterns of walls WC1, Irimies [33]




72

Figure 4.2.15 Damage patterns of walls WC2, Irimies [33]

In the URM model the measured lateral resistance was about 54KN and was attainted after the

beginning of the masonry crushing at the corners. In the confined models the lateral resistance

was 2.2 times (WC1) and 2.7 times (WC2) higher than that of URM.

The increase of 1.8 times of vertical reinforcement amount in tie-columns led to enhance of

the lateral strength of about 20%. The lateral load reduced after the considerable opening of

shear cracks in tie-columns.

The finals conclusion leads that confining of URM wall by RC tie-columns determined both

the later strength and stiffness increase and the change of failure mechanism.

The lateral strength of confined walls increased with the increase of vertical reinforcement

amount in tie-columns. However, this increase was not proportionally. All the specimens

showed good seismic behaviour at large displacements.

4.2.5 Influence of openings in the behaviour of confined masonry: Behaviour of confined

masonry shear walls with large openings, Yáñez et al. [8]

Significant research has been carried out in different countries to study the behavior of

confined masonry walls. In order to study the behavior of lightly reinforced confined masonry

shear walls with openings, sixteen full-scale specimens were tested. Eight specimens were of

concrete masonry units and eight of hollow clay brick masonry units. The specimens were

designed to have shear failure with diagonal cracks in the masonry panel. The test parameters

were the masonry unit type (concrete and clay) and the size of openings (four cases).




73

All the specimens had a panel with openings of different sizes, two columns and a beam on

top. The bond pattern was of the running bond type. The openings had no special concrete

confinement elements around its borders. The hollow areas of the masonry units were not

filled with mortar or grout, except close to the openings where a vertical reinforcement bar

was placed. Figure 4.2.16 shows the dimensions of the four patterns of the specimens (there

were 2 specimens for each pattern).

Figure 4.2.16 Wall dimensions, Yáñez et al. [8]

The testing set up is conformed by a horizontal cyclic load applied along the axis of the top

beam, and controlled by displacement. There was no vertical load applied. Two cycles at each

deformation level were applied. During testing, the development of cracks and damage were

registered. Five levels of damage were defined: (a) first visible cracking in the columns, (b)

first visible cracking in the masonry panel, (c) beginning of diagonal cracking, (d) primary

and secondary diagonal cracking in the wall segments in both sides of the openings, and (e)

formation of the final cracking pattern.

TEST RESULS

All the specimens failed in shear. Two failure mechanisms, shown in Figure 4.2.17, appeared

with diagonal cracking and mixed cracking. The first mechanism corresponds to a diagonal

crack spanning at least half of the width of the specimen. The second mechanism corresponds

to a crack that develops horizontally and then diagonally, or vice versa, with similar spans in

each case. In both mechanisms, the cracks propagate though the mortar joint due to a low

adherence between the mortar and the masonry units. This situation appears more often in

concrete masonry unit specimens.

The first cracks appeared horizontally in the confinement columns, and in the lower coursesof the masonry panels. While the horizontal reinforcement under the openings was not




74

broken, the damage concentrated in the wall segments in both sides of the openings. In the

specimens with no horizontal reinforcement under the openings (pattern 4), the strength

degradation and the width of the cracks was notorious once the diagonal cracks reached the

vertical bar reinforcement close to the openings.

Figure 4.2.17 Failure Mechanisms, Yáñez et al. [8]

The load-deformation behaviour of the specimens with opening is non linear, having an initial

linear elastic zone. The load-deformation capacity depends on the inclination of the diagonal

struts that can be developed in the specimens with openings, and on the tensile capacity of the

confinement column or vertical reinforcement in the border of the opening that work as ties ofvirtual strut and tie models.

The stiffness of specimens with an opening ratio of 11% of the total wall area is close to that

of the specimens without openings. The stiffness degrades markedly with the subsequent

cycles due to cracking.

The rate of the stiffness degradation is smaller as the opening size increases, especially in the

concrete masonry walls.

The shear capacity of the specimens was reached for the inter-story drift range of 2.0‰ to

4.0‰. For walls with larger openings, the maximum strength decreases and is reached for

inter-story drift larger than 4‰. It is conservative to consider the shear capacity proportional

to the net transverse area of walls with window openings.

The tests indicate that for these specimens with small horizontal reinforcement in the masonry

panel, the crack widths are quite large for small inter-story drift. In order to keep the crack

widths under 1.5mm, the inter-story drift ratio should be no larger than 1.0‰, whereas to

keep the crack widths under 3.0mm, the inter-story drift should be no larger than 2.0‰.

The tests also indicate that the confinement concrete frame keeps the integrity of the

specimens under inter-story drift ratios as large as 13‰ in spite of large damage (large crack




75

widths and large strength and stiffness degradation). It is interesting to note that this large

deformation level cannot be reached in lightly reinforced partially grouted masonry walls.

4.2.6 Influence of the number and spacing of confining tie-columns: Experimental

evaluation of confined masonry walls with several confining columns, Marinilli and

Castilla [34]

The effect of the number and spacing of confining-columns in the seismic behavior of

confined masonry walls was evaluated experimentally. A set of four full-scale confined

concrete masonry walls of the same nominal area was constructed at the IMME, to be tested

against constant vertical load and reversed cyclic lateral load. The first specimen “M1”

consisted of one panel and two confining-columns. The second specimen “M2” consisted of

two panels and three equally spaced confining-columns. The third specimen “M3” also

consisted of two panels, but the central confining-column was located at ⅓ of the specimen

length. Finally, the fourth specimen “M4” contained three panels and four equally spaced

confining-columns. Figure 4.2.18 shows the configuration of the four specimens. The length

and the height of the specimen were 300 and 230cm, respectively.

The basic components of the masonry walls were hollow concrete blocks, mortar and

confining elements.

Figure 4.2.18 Configuration of specimens M1, M2, M3 and M4, Marinilli and Castilla [34]




76

Each one of the walls was tested against lateral loads applied at the top of the wall. A steel

box was placed around the top beam and fastened to it with bolts, so as to ensure an adequate

distribution of the lateral loads along the wall. The loads were applied with alternating and

increasing displacement-controlled cycles until the limit state of the walls was reached. Each

cycle was repeated as many times as necessary to achieve stability in the corresponding

hysteretic loop. The lateral loads were applied using hydraulic jacks. In addition each wall

was subjected to a constant vertical load to simulate gravity effects. The vertical load was

applied with a stiff steel girder and three dead weights, weighting in total 13.8tn. To ensure a

uniform distribution of the vertical load along the wall a sand bed was placed between the

steel girder and the top of the wall.

TEST RESULTS

In Figure 4.2.19 and 4.2.20 can be seen the cracking produced during testing in Specimens

M1, M2, M3 and M4, respectively. Graded 45º cracking was found in all the masonry panels.

The cracks even propagated to all the confining-columns.

Figure 4.2.19 Specimens M1 and M2 after testing, Marinilli and Castilla [34]

Figure 4.2.20 Specimens M3 and M4 after testing, Marinilli and Castilla [34]




77

However, there are some facts to highlight: Specimen M1 suffered a horizontal crack along a

mortar joint; Specimens M2 and M4 showed a widespread cracking distribution and

Specimen M3 showed in the largest masonry panel a cracking pattern similar to that observed

in M1, while the cracking at the shortest panel was similar to that observed in M2 and M4.

Based on the results obtained, it can be said that the presence of more confining-columns at a

smaller spacing seems to spread the cracking along the masonry panels, thus improving the

damage distribution. The values contained in Table 4.2.3 show that the inclusion of confining-

columns tends to increase the strength of the walls. It is important to remember that all the

tested walls had the same nominal transverse area and were tested against the same vertical

load.

Table 4.2.3 Properties of the system, Marinilli and Castilla [34]

From these results it may be concluded that the inclusion of more confining-columns

improves the ability of the walls to make larger incursions in the inelastic range. This can be

explained considering that less spaced confining-columns is able to perform a better

confinement of the masonry panels.

The analysis of the deformations obtained during the tests shows that the general behaviour of

the walls was governed by shear deformations, even for the specimens which deformations

are not shown herein.

The results show that the inclusion of confining-columns in walls of the same nominal

transverse area increases the initial stiffness, increases the system ductility, allows a better

damage distribution in the masonry panels in conjunction with a lesser spacing of the

confining-columns, and tends to increase the strength of the walls. Otherwise, the inclusion of

confining columns does not seem to improve the energy dissipation capacity or the equivalentdamping ratio, and decreases the equivalent ductility of the walls.

4.2.7 Experimental study on effects of height of lateral forces, column reinforcement and

wall reinforcements on seismic behaviour of confined masonry walls, Yoshimura et al. [11]

In order to investigate the effect of the height of application point of lateral loads and

reinforcing steel bars in walls and columns in improving the seismic behaviour of confined

concrete block masonry walls, an experimental research program was conducted. A total oftwelve one-half scale specimens are tested under repeated lateral loads. Specimens are tested




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to failure with increasing maximum lateral drifts while a vertical axial load was applied and

maintained constant. The specimens adopted are two-dimensional (2D) hollow concrete block

masonry walls with different parameters such as shear span ratio, inflection point and percent

of reinforcement in confining columns and walls. All the specimens are approximately one-

half scale models of one-bay-one-story masonry walls using hollow concrete block masonry

units. The thickness of all the walls is 100mm and that confined by cast-in-place RC columns

with 100mmx100mm cross-sections along their extreme edges and T-shaped RC collar beams

along their tops, Table 4.2.4.

Table 4.2.4 Test specimens, Yoshimura et al. [11]

These specimens were tested under repeated lateral forces, and constant axial compressive

stress of 0, 0.48 and 0.84MPa respectively. The test setup adopted in the present study is

illustrated in Figure 4.2.21 Test setup consisted of steel reaction frames and two hydraulic

actuators, fixed to the frame in order to simulate the constant vertical loads and in plane

lateral repeated forces.




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Figure 4.2.21 Test setup, Yoshimura et al. [11]

TEST RESULTS

Cracks were concentrated along the diagonals, Figure 4.2.22. However, the cracks in

specimen (1), which failed in shear, were observed to be converging towards the centreextending through the blocks whereas in case of the specimen (2), which failed in flexural

failure mode, the cracks were observed mostly along the horizontal joint mortar.

Depending upon the modes of failure, similar crack patterns were developed in the specimens

(5) and (6) in with aspect ratio 0.84, which failed in flexure and shear failure modes

respectively. The specimens (9) and (10) (c) both failed in shear and thus showed a much

more uniform inclined cracking. At failure, the cracks penetrated into the confining columns

showing a rapid reduction in the lateral load carrying capacity of the specimens.




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Figure 4.2.22 a) Specimens with aspect ratio (ho/lo) of 1.51; b) Specimens with aspect ratio (ho/lo) of

0.84; c) Specimens with aspect ratio (ho/lo) of 0.69, Yoshimura et al. [11]

The experimental study was conducted to investigate the seismic performance of the confined

concrete hollow block masonry walls considering the parameters such as height of inflection

point (0.67h0, 1.08h0 and 1.11 h0), shear span ratio ( M/Qd =0.58~1.77) for aspect ratios (1.51,

0.84 and 0.67), tensile reinforcement ratio ( pt =0.04~0.29%), horizontal wall reinforcement

ratios ( ph=0%, 0.08% and 0.18%) and vertical axial stress (σo= 0.84 and 1.8MPa). The

present test results were also compared with the test results of the past in order to investigate

the accuracy of the terms or factors in the existing equation. Based on the observations during

tests and analysis of data, the following conclusions were arrived at: irrespective of the height

(a)

(b)

(c)




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of point of application of lateral forces to the specimens, that is, whether the inflection point is

low or high, it may be concluded that the vertical axial load has positive effect on the value of

ultimate shear stress of the specimens. Test results are showed in Table 4.2.5.

Table 4.2.5 Predicted and observed ultimate lateral strengths and failure modes, Yoshimura et al. [11]

4.2.8 Effects of vertical and horizontal wall reinforcement on seismic behaviour of

confined masonry walls, Yoshimura et al. [13]

In order to investigate the effect of vertical and horizontal wall reinforcing methods on

seismic behaviour of confined masonry wall, eight confined masonry walls specimens with

different details in wall reinforcement, Table 4.2.6, are tested under a constant gravity loadand alternately repeated lateral forces. The test results indicate that the vertical and horizontal

wall reinforcing bars provide in confined masonry walls play an important role for developing

higher strengths and better deformability.




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Table 4.2.6 List of specimens, Yoshimura et al. [13]

The setup adopted consists in constant gravity load applied by a hydraulic jack with a capacity

of 343KN, and alternately repeated lateral forces were applied by a double-acting hydraulic

jack with 980KN capacity. Important displacements and strains in reinforcing bars were

measured by transducers and wire strain gages, and processed simultaneously by a personal

computer.

TEST RESULTS

Summarizing the test results; in the early stage of loading up to the story drift of 0.1%, hair

cracks occurred along through the horizontal joint between bottom of masonry wall and top-

surface of foundation beam for all the specimens. These cracks are caused by tensile stress

due to flexure of the masonry wall. The crack pattern is shown in Figure 4.2.23.




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Figure 4.2.23 Crack patterns of the specimens, Yoshimura [13]

In case of the confined concrete masonry wall system, both of the vertical and horizontal

reinforcement in masonry walls play an important role for expecting higher ultimate lateral

strength and better ductility. This is because high shear stresses are induced in confinedmasonry wall panels especially when the masonry walls are subjected to lateral forces and

expected to fail in brittle shear failure mode.

4.2.9 Experimental study for developing higher seismic performance of brick masonry

walls, Yoshimura et al. [12]

To investigate the effective seismic strengthening methods for masonry walls in developingcountries, a total of twenty eight un-reinforced masonry (URM) and confined masonry (CM)

walls were constructed and tested. The specimens include two-dimensional (2D) and three-

dimensional (3D) masonry walls with and without wall reinforcing bars or U-shaped

connecting bars, which were tested under constant gravity load and repeated lateral forces.

A total of twenty-eight un-reinforced and confined masonry wall specimens with different

wall-to-wall connection details are listed in Table 4.2.7.




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Table 4.2.7 Listed of the tested specimens, Yoshimura et al. [12]

The test setup adopted in the present study consisted of steel testing frame and two hydraulic

actuators, fixed to the frame in order to simulate constant vertical and in plane lateral repeated

loads. The vertical load was applied to the specimens by a hydraulic jack with 2.000kN

capacity, connected to servo-impulse controller system in order to keep constant vertical load

during the test.

TEST RESULTS

The crack and crack propagation during the tests were monitored and recorded by marking the

cracks at the end of the half cycle of loading while the specimen was held at the maximum

displacement, although crack widths were not filed. The cracks were partially closed with

load reversal. The final crack patterns developed in the selected specimens are shown in

Figure 4.2.24. Almost all the confined masonry specimens in L-series (whose infliction point

is 0.67 times wall height) failed in shear mode and also sliding was recorded in some of the




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specimens. However, the clear separation of wall from the RC confining column was not seen

in un-reinforced confined masonry specimens although the cracks were developed along the

vertical joint. On the other hand, un-reinforced confined masonry specimens in H-series

(whose inflection point is 1.1 times wall height) showed a distinct separation of wall from the

RC confining column. This type of wall separation was not seen in confined masonry

specimens provided with horizontal wall reinforcement and U-shaped connecting bars, though

few cracks were developed. All H-series specimens failed in flexure mode at first and

ultimately in either sliding or shear failure mode in some cases. Further, the un-reinforced

specimens failed in flexure, i.e. cracks developed in horizontal bed joint at the bottom course

and also in other courses on the tension side.

Figure 4.2.24 Final crack pattern, Yoshimura et al. [12]

The reinforcing steel i.e. horizontal wall reinforcement and U-shaped connecting bars used in

the specimens mentioned in this report can be seen as one of the critical parameters affectingwall ductility. However, it was impossible to evaluate the ductility factor of the tested wall

specimens correctly because these specimens did not show the perfect elasto-plastic

behaviour. Therefore, it may be concluded that horizontal wall reinforcement and U-shaped

connecting bars improve the deformation behaviour of wall after attaining the maximum

ultimate lateral load.

Based on the observations during tests and analysis of data, the following conclusions were

obtained.

• The confined masonry wall system is effective to improve the poor seismic

performance of the ordinary URM, by enhancing the lateral load carrying capacity.




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• The confined masonry wall system with connecting bars at the vertical wall-to-wall

connections as well as the horizontal wall reinforcing bars developed reasonably higher

ultimate lateral strength with the increase of vertical axial load and showed better ductility as

compared to the un-reinforced wall specimens.

• The wall separation effect from the RC confining columns can be avoided by

providing the U-shaped connecting bars at the wall-to-wall or wall-to-column joints as

recommended in China.

• The increase in axial stress tends to increase the lateral load carrying capacity of the

masonry walls and the observed values showed that the ultimate flexural strength could be

well predicted by the existing equation.

In brief, it can be concluded that the horizontal wall reinforcement and/or connecting bars

provided between masonry walls and RC columns play an important role to improve the poorseismic performance in the ordinary URM and CM walls, by enhancing the ductile behaviour

to some extent and lateral load carrying capacity.

4.2.10 Experimental study on earthquake-resistant design of confined masonry structures,

Ishibashi et al. [35]

One story specimens were designed and constructed following the requirements of the

Mexico City Building Code (DDF 1987 & 1989). The experimental variable was the flexuralcoupling between two wall panels having different openings between them.

Specimens are shown in Figure 4.2.25. The first specimen W-W was practically lacked of

flexural coupling, the walls were only connected through high-strength dywidag bars, that

transfer the lateral forces between them. In the second model, WBW, the walls were linked by a

cast-in-place reinforced concrete bond-beam and slab, forming a door opening. Finally the

last model WWW had a parapet between the walls, this time forming a window opening.

Specimens were 5m long and walls were 2.4m and 1.6m long separated by 1m opening. The

height of the walls was 2.5m. Walls were built with hand-made solid clay bricks. Concretehad a compressive strength of 200kg/cm². The cement mortar joint had a ratio of 1:3 (cement:

sand), with a compressive strength of 125kg/cm². The steel reinforcement had nominal yield

strength of 4200kg/cm². Tie-columns were reinforced with 4#3 longitudinal bars and #2

hoops spaced 20cm, this space was reduce at the end of the columns were it was of 7cm.

Bond-beams were reinforced with 4#4 longitudinal bars and #2 hoops spaced 20cm.




87

Figure 4.2.25 Specimens details, Ishibashi et al. [35]

The lateral shear was applied through a static type hydraulic jack. A vertical force of 5 kg/cm²

was applied to simulate the gravitational loads during the test. The alternated lateral loads

used for the test are shown in Figure 4.2.26. First test were load controlled with maximum

shears equal to 5, 10 and 18tns. In the second stage displacement controlled cycles were

applied up to 0.012.

Figure 4.2.26 Loading history, Ishibashi et al. [35]




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TEST RESULTS

Final cracks patters are showed in Figure 4.2.27. In all the specimens damage was governed

by diagonal cracks. First cracks appeared near the corners and propagated fast to the wall

center with increasing deformation levels during the test. Almost not damage was observed inthe rest of the masonry panel. The specimen ww presented a more uniform crack distribution

tran the other models. Brick crushing occurred but not damage in the mortar. By the end of

the test few vertical cracks were observed between the wall panel edge and the tie-column.

In the models WBW and WWW diagonal cracks extended through the walls from the corners of

the opening to the diagonally opposite corner. No cracks were observed in the parapets of the

model WWW. Tie-columns present flexural cracking uniformly distributed along the height.

After diagonal cracks occurred strength and stiffness decay were observed, particularly at

large drift ratios (0.012).

Maximum measured strengths were 75% higher than the calculated capacities using code

recommended masonry strengths. Stable hysteretic loops were observed up to 0.005 drift

ratio, this limit deformation is small in comparison with well detailed RC frame or wall

structures. At larger deformations than 0.005, severe degradation was noted. Therefore, large

reductions on the elastic spectral ordinates for this type of structures cannot be justified.

Although the type of opening affected the initial stiffness of the specimens, the stiffness decay

was similar and follows a parabolic curve. A wide-column model can be used to predict the

initial stiffness of a structure.




89

Figure 4.2.27 Response of specimens WW, WBW and WWW, Ishibashi et al. [35]

4.3 Concluding remarks

Considering all the experimental tests in confined masonry listed previously some

conclusions can be pointed out:

• Most of the tests presented in this chapter correspond to masonry panels subjected to

cyclic loading and few correspond to buildings subjected to shaking table tests.

• The behaviour of hollow clay bricks and hollow concrete blocks is brittle in

comparison with the solid clay bricks. And in most of the test that used the two first ones

spalling was produce.

• Shear failure was the predominant failure mode.

• Based on the failure mode observed, the analytical model for design and assessment

could be simplified by assuming that all inelastic deformations would take place at the first

story and would be controlled by shear.




90

• None of the walls reached their theoretical flexural capacity and the final state was

controlled by the columns shear strength. This is due to the fact that, under large

displacements, diagonal cracking of masonry extended to the columns.

• Compression failure never occurred.

• A more uniform distribution of cracks was observed in specimens with horizontal

reinforcement; the larger the amount, the more uniformly distributed the cracking was.

• The mode of failure is strongly dependent on the horizontal reinforcement ratio.

• The loads that produced initial cracking of the walls with openings are half of the load

that produced the same effects in the wall without the openings.

• The experimental results indicated that the confining of an un-reinforced wall by RC

tie-columns led to the increase of the lateral resistance and to the change of failure

mechanism.

• The inclusion of more confining-columns improves the ability of the walls to make

larger incursions in the inelastic range. This can be explained considering that less spaced

confining-columns is able to perform a better confinement of the masonry panels.

• The test results indicate that the vertical and horizontal wall reinforcing bars provide

in confined masonry walls play an important role for developing higher strengths and better

deformability.



Chapter 5. Code recommendations

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5. CODE RECOMMENDATIONS

In this chapter some recommendations given by the Argentinean (INPRES-CIRSOC 103) and

the European (EU 6 and EU 8) codes are presented and discussed. It must be pointed out that

confined masonry has evolved essentially through an informal process based on experience,

and that it has been incorporated in formal construction through code requirements and design

procedures that are mostly rationalizations of the established practice, even after having been

validated by structural mechanics principles and experimental evidence.

Finally a comparison between codes from EU, Italy, Peru, Mexico, Argentina and Colombia

are added in order to have a general view of all the recommendations existing nowadays in

confined masonry.

5.1 Quality of masonry

The strength qualities of the masonry are characterized by the following parameters:

• Compressive strength, σ’mo

• Shear strength, τmo

The tensile strength of the masonry due to bending in the plane of the wall is not considered.

The deformability of the masonry is defined by the followings parameters:

• Elastic modulus, Em

• Shear modulus, Gm

The materials used for the masonry panels were characterized by testing brick piers under

simple compression and small masonry probes under diagonal compression, according to the

INPRES-CIRSOC 103 Code [7].




92

Compressive strength

The compressive strength, measured in relation to the gross area, is a strength index of the

masonry in compression, and is useful for design and control. This index is determined by

tests at 28 days; codes also give some minimum values as indicated in Table 5.1.1.

This procedure consists in tests on bricks piers, Figure 5.1.1. In these tests the value of the

compressive strength σ’mo is equal to the characteristic compressive strength. This value is

determined as that achieved by the 95% of the testing.

The piers must be fabricated following the procedures used in the zone where the construction

is to be located, taking into account the conditions and qualities of the masonry units used for

the construction.

Figure 5.1.1 Tests on brick piers to determine the compressive strength, Bustos [10]

Table 5.1.1 Values of compressive strength for different masonry units and mortar joints, INPRES-

CIRSOC 103 code [7]

Values of σ'mo (MN/m²)

Type of mortar jointType of masonry unit

High Intermediate Normal

solid clay brick class A 4 3.5 3

solid clay brick class B 2.5 2 1.5

Hollow clay block class A 3 2.5 2

Hollow clay block class B 2 1.5 1.2

Hollow concrete block class I and II 3 2.5 1.5

Hollow concrete block class III 2 1.5 1.2

Shear strength

The shear strength, measured with relation to the gross section is another strength index of themasonry and is useful for design and control.




93

The tests, in this case, consist in small masonry probes under diagonal compression, Figure

5.1.2. This trial mix is tested after 28 days of curing.

Figure 5.1.2 Masonry probes under diagonal compression to determine the strength, INPRES-CIRSOC

103 code [7] and Yáñez et al. [8]

In these tests the value of the shear strength (τmo) is taken as the characteristic shear strength,

which is determined as the value achieved by the 95% of the tests. Minimum values are

suggested code in Table 5.1.2. The shear strength is obtained by dividing the projection of the

load parallel to the rows of masonry by the respective gross section.

The “r” value must be at least equal to 20cm. The relation between “r/d” should be equal or

higher than 0.3.

D = 0.7 P

τm = D / d . eo




94

Table 5.1.2 Values of shear strength for different masonry units and mortar joints, INPRES-CIRSOC [7]

Values of τmo (MN/m²)

Type of mortar jointType of masonry unit

High Intermediate Normal

solid clay brick class A 0.4 0.35 0.3

solid clay brick class B 0.35 0.3 0.25

Hollow clay block class A 0.35 0.3 0.25

Hollow clay block class B 0.3 0.25 0.2

Hollow concrete block class I and II 0.35 0.3 0.25Hollow concrete block class III 0.3 0.25 0.2

Masonry deformability: elastic modulus and shear modulus

The approximated values given by INPRES-CIRSOC 103 are:

• Elastic modulus:

• Shear modulus:

5.2 Classification of the structural walls

Structural walls are classified by the INPRES-CIRSOC 103 code in:

• Confined masonry, Figure 5.2.1.

- Confined masonry

- Confined masonry with horizontal reinforcement

- Confined masonry without vertical confining elements

• Reinforced masonry, Figure 5.2.2.

Em = 800 σ'mo for dynamic loads

Em = 300 σ'mo for loads of large duration

Gm = 0.3 Em




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Also each of these typologies can be made of different masonry units, like: solid clay bricks,

hollow concrete blocks or hollow clay blocks.

Figure 5.2.1 Confined masonry. Left: reinforce confined masonry; right: confined masonry. Universidad

Nacional de Cordoba [16]

Figure 5.2.2 Different options of reinforced masonry, Decanini and Payer [17]

Structural walls for good seismic performance

The performance of building subject to earthquake motions is governed by the inter-

connectivity of structural components as well as the individual component's strength, stiffness

and ductility. Thus the details to provide good seismic resistance can be classified in two

categories:

Details for complete load path:

• Provide wall to wall connection i.e. tying of walls.

• Provide means for walls to foundations connection.




96

• Provide connection of bond-beams to roof.

• Provide connection of walls to bond-beams.

• Provide stiff in their plane floors/roofs.

Details to improve structural components strength and ductility:

• Improve the compressive strength of structural components.

• Improve the bending strength of structural components.

• Improve the shear strength of structural components.

• Improve the ductility, μ of the structural components.

Codes give a minimum length for the masonry panel to achieve the compressive strength on

the diagonal. For this reason, to establish this condition, walls are classified into: walls with

two supports, and walls with three or more supports.

• Masonry panels supported in two sides must have the minimum lengths given in

Figure 5.2.3.

Figure 5.2.3 Minimum dimensions of confined masonry panels with two constraints, Universidad

Nacional de Cordoba [16]

• Masonry panels supported in three or more sides must reach the minimum lengths

given in Figure 5.2.4.

constraint

constraint

La

H1

Wall with two horizontal constraints

LaH1

2.2≥ 1.5≥

La m( ) H1 m( )

For M.R.A.D. La 1.20≥




97

Figure 5.2.4 Minimum dimensions of confined masonry panels with three or more constraints,

Universidad Nacional de Cordoba [16]

Another condition for a good seismic performance is given by the maximum heights and

number of stories for the different classification of walls and seismic zones, Table 5.2.1.

Table 5.2.1 Maximum heights and number of stories allowed by the INPRES-CIRSOC 103 code [7]

Maximum height maximum number Maximum height maximum number

hn (m) of stories N hn (m) of stories N

Confined masonry 12.5 4 9.5 3

Confined reinforced masonry 15.5 5 12.5 4

Reinforced masonry 15.5 5 12.5 4

Confined masonry 6.5 2 4 1



Confined masonry 6.5 2 4 1



Unconfined masonry 3.5 1 _________ _________

Seismic zones 1 and 2 Seismic zones 3 and 4

Solid clay bricks

Structural walls

solid clay brick

Hollow clay block

Hollow concrete block

Type of masonry unit Type of wall

(1) Only allowed in case of internal load-bearing walls and low seismic activity.

H1

La

onstraint

onstraint

constraint

Wall with three or more constraints

LaH1

2.6≥ 0.90≥

La m( ) H1 m( )

For M.R.A.D. La 0.80m≥

(1)




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Recommendations and general principles of the structural composition according to INPRES-

CIRSOC 103 [7] and EU 8 [36]:

• Load bearing walls are required in two orthogonal directions of the building, Figure

5.2.5.

• Both directions need a minimum density of walls to resist the seismic loads.

• Regular configuration, both in plan and elevation, i.e. uniform and symmetrical, as

shown in Figure 5.2.6.

• It is suggested that no variations of resistance, stiffness or mass in plan and in

elevation are to be introduced.

• Rigid floors should interconnect walls to ensure diaphragm action.

• In constructions of two or more stories, the walls must be vertically aligned from

storey to storey.

• If possible, masonry panels should be supported in all the sides.

• The openings in walls, slabs and roof must be located in such way that they generate

the least possible tensions.

• Stable foundation should be able to transmit the maximum seismic loads from the

superstructure to the foundation soil.

Figure 5.2.5 Structural walls distribution in plan, Kuldeep Virdi [3]

Masonry buildings with horizontal irregularities and lack of symmetry may have considerable

eccentricity between the mass centre and stiffness centre giving rise to damaging coupled

lateral/torsional response. Horizontal irregularities in the form of extensions, projections etc.

may cause stress concentration and local failures since these extensions are prone to vibrate

separately from the rest of the structure. On the other hand vertical irregularity in masonry

building may cause stress concentration at a horizontal plane that can lead to total collapse. In

order to achieve satisfactory redundancy at least to lines of load bearing walls are required ineach principal direction of the building. Lack of rigid floors will prevent proportionate load




99

transfer onto walls at each floor level as well as will not provide out of plane restraint.

Unsupported masonry walls at floor level tend to separate at corners and/or fail out of their

plane, causing collapse of floor or roof, Kuldeep Virdi [3].

Figure 5.2.6 Irregular configurations in plan should be separated in regular potions, Kuldeep Virdi [3]

• Vertical regularity is achieved by uniform distribution along the height of the building

of stiffness and masses. Lack of vertical regularity may lead to horizontal plane of

weakness/stress concentration and collapse.

• Mixed structural systems, such as a combination of masonry structural walls in one

level and RC frame in the next, are not allowed. For the purpose of planning flexibility it is possible a combined system consisting of RC columns and masonry shear walls. For such

configurations the masonry bearing walls should be reinforced and the RC members should

be connected into RC floors forming frames. The vertical reinforcement of the masonry shear

wall should be anchored into the floor to ensure load transfer.

• The floors are rigid in their plane providing diaphragm action and interconnected with

masonry walls. To this end the floors should be constructed in a single plane. In cases where

large openings are present in the floor, such as for stairways the contour of the opening should

be strengthened with a bond-beam. Also two-way slabs are preferred to one-way slabs, as

they distribute the vertical gravity loads more uniformly onto the masonry walls.

5.3 Confined masonry

In this section some recommendations are given for confined masonry. In these

recommendations, specifications about the maximum area of the masonry panels and the

maximum distance of panels allowed are listed according to the INPRES-CIRSOC 103.

The dimensions and the area are given depending on the thickness and the height of the paneland the seismic zone as well, Table 5.3.1.




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Table 5.3.1 Maximum area and dimension for confined masonry panels, INPRES-CIRSOC 103 [7]

Maximum area of the Maximum dimension of the panel (m)

masonry panel wall thickness wall thicknessSeismic zones

(m²) ≥ 17cm < 17cm and ≥ 13cm

1 30 7 4.5

2 25 6 4

3 and 4 20 5 4

The following recommendations regarding the configuration and size of openings should be

observed according to EU 8, Kuldeep Virdi [3]:

• Openings should be vertically aligned from storey to story.

• The top ends of openings in the storey should be horizontally aligned.

• Openings should not stop continuous RC bond-beams (at lintel and/or roof level).

• Openings should be located symmetrically in the plan of the building so as not to get

in the way of the uniform distribution of strength and stiffness in two orthogonal directions.

Tie-columns and bond-beams

On the other hand, some recommendations are given for the confining elements. Thus, in case

of tie-columns, general prescriptions are as follows (INPRES-CIRSOC 103):

- In external walls: should be located at all corners and changes of wall contour, and

at all joints, wall intersections and free ends of structural walls.

- In internal walls: should be located at all corners and changes of wall contour, and

at all joints, wall intersections and free ends of structural walls.

- When the masonry panels area is bigger than the maximum dimensions given in

Table 5.3.1.

- Vertical confining members are also necessary at both sides of any opening.

Position of bond-beams:

- Level of foundations.

- At every floor level.




101

- In roof level.

- In intermediate levels when the area, the maximum dimensions or the relation between

sides is higher than the ones given in Table 5.3.1

- In the case of sloping roof.

Recommendation for the confining elements

In the case of confined masonry construction bond-beams are constructed as part of the

vertical and horizontal masonry confining elements. Bond-beams should be constructed in-

situ from reinforced concrete and cast simultaneously with the floor slab. Bond-beams should

be cast on top of all structural walls at every floor level, Kuldeep Virdi [3].

Bond-beams are constructed because:

• They conform confined masonry shear walls in combination with tie-columns.

• They improve the in-plane stiffness of floors to provide diaphragm action.

• They transfer the horizontal load from the diaphragm to the structural walls.

• They connect the structural walls together and provide out-of-plane support.

• They connect the RC tie-columns.

EC8 specifies the following minimum requirements, Figure 5.3.1:

• Concrete of class 15 should be used.

• Cross section size should be not less than 150x150mm.

• Four mild steel re-bars with total area 240mm2.

• To ensure integrity of the bond-beam the longitudinal re-bars at corners and wall

intersections should be spliced a length of 60φ.

• Transverse reinforcement-stirrups re-bar φ6 @ 200mm intervals.




102

Figure 5.3.1 Detail of RC bond-beam showing splicing of re-bars at wall corners, Kuldeep Virdi [3]

On the other hand, INPRES-CIRSOC 103 code specifies that the concrete used in the

confining elements have minimum compressive strength of 11MN/m², and the minimum

recommended content of cement in the concrete is 250kg/m³.

The dimensions of tie-columns and bond-beams should have the proportions given in Figure

5.3.2. The minimum section of tie-columns can be computed with the following expression:

Figure 5.3.2 Dimensions recommended by INPRES-CIRSOC 103 code for tie-columns and bond-beams,


Bc (cm²) = 0.025*Vp (kg)




103

For the determination of the reinforcement, INPRES-CIRSOC 103 gives an approximated

procedure.

• Tie-columns: the total reinforcement section (Ac) of the tie-column of a given story

can be computed with the next expression:

• Bond-beams: the total reinforcement section (Av) of the bond-beams of a determined

story can be computed with the following expression:

• Minimum reinforcement section of tie-columns and bond-beams determined with the

expressions listed above can not be less that the minimum reinforcement section computed

with the following expressions (first one for seismic zones 1 and 2, and second expression for

seismic zone 3 and 4):

Minimum values are given in the Argentinean code for different types of reinforcement and

seismic zones, Table 5.3.2.

Table 5.3.2 Recommended diameters and separation for reinforcement in tie-columns and bond-beams,


Reinforcement type ADN 420 and ADM

420 Reinforcement type Al 220

Longitudinal Longitudinal

Seismic

zones

ReinforcementHoops

ReinforcementHoops

4 bars diameter 4.2mm 4 bars diameter 6mm

1 and 2 diameter = 6mm each 20cm diameter = 8mm each 20cm

4 bars diameter 4.2mm 4 bars diameter 6mm

3 and 4 diameter = 8mm each 20cm diameter = 10mm each 20cm

• The diameter of hoops (ds), for tie-columns, can be computed as:

Ac = (1 + 0.25k)*Vp*(Ho/Lo)*(1/βs)

Av = Vp*(1/βs)

Amin = (0.25 + 0.13k)*t*(1/βs) Amin = (0.35 + 0.18k)*t*(1/ βs)

ds = (0.20 + 0.1k)*se




104

• In critical zones the total section of hoops (Ae), for tie-columns, in one layer is

determined by:

The separation of hoops for tie-columns can not be more than half of the transversal

dimension of the column or not more than 10cm, Figure 5.3.3.

• Hoops in critical zones of bond-beams are recommended to be twice the reinforcement

that corresponds to the normal zones, and the separation between them not more than 10cm.

In normal zones maximum separation can be 20cm.

Figure 5.3.3 Hoops in critic zones (near corners) and in normal zones, Universidad Nacional de Cordoba

[16]

Recommendations are also given to provide a good anchorage in the corners, Figure 5.3.4 and

connections between masonry walls panels and RC columns, Figure 5.3.5.

Ae = (0.5*Vp)/(dc*βs)*se




105

Figure 5.3.4 Recommended details in masonry wall connection in Argentina [16]

Figure 5.3.5 Recommended details in masonry wall to RC column connection in P.R. China, Yoshimura

et al. [12]




106

According to EC 8 the resistance of the RC bond-beam should not be taken into consideration

in the design calculations. Consequently there is no mandatory design through calculation for

the bond-beams. The design parameters are determined on empirical basis. In Table 5.3.3 the

reinforcement can be determined based on the seismicity of the location the number of storeys

and position for vertical confining elements and in Table 5.3.4 for bond-beams. Also Figure

5.3.6 illustrates the design and position of the reinforcement for tie-columns.

Table 5.3.3 Recommended reinforcement for tie-columns [37]

Table 5.3.4 Recommended reinforcement for bond-beams [37]




107

Figure 5.3.6 Construction of tie-column for confined brick masonry house, Kuldeep Virdi [3]

5.4 Resistance verification

The verifications in the confined masonry are the following two according to INPRES-

CIRSOC 103 code: one is to verify the structure when it is subjected to seismic action, and

the other one is when the masonry structure is under gravitational loads. Verification under

seismic loads must be computed for ultimate stress condition.




108

Confined masonry in-plane verification given by INPRES-CIRSOC 103

In this respect two verifications are needed:

• Failure mechanism in shear

• Failure mechanism in flexion with compression

The maximum shear stress resisted by the confined masonry wall (Vur) can be determined in

terms of the basic shear strength τmo of the masonry and the compressive strength σo

generated by the vertical loads, according to the following expression:

On the other hand, the shear stress resisted by the confined masonry wall (Vur) should reachthe following condition:

For the verification in bending, the ultimate resisting moment can be determined by:

When gravitational loads are acting on the masonry wall, the ultimate resisting moment can

be determined by one of the following expression, depending on the case:

or

In confined masonry panels, whose height is not more than 9m or 3 stories in 1 or 2 seismic

zone, or 7m and 2 stories in seismic zones 3 and 4, and also fulfil the condition given below,

it is assumed that the confined masonry wall satisfies the verification in bending and

compression:

If the confined masonry panels are subjected to gravitational loads the following effects have

to be taken into account:

• Normal stress due to vertical loads.

Vur = (0.6* τmo + 0.3*σo) Bm

Vur ≤ 1.5* τmo*Bm

Muro = Ac * βs * Le

If Nu ≤ (Nuo/3) : Mur = Muro + 0.3 Nu L

If Nu > (Nuo/3) : Mur = (1.5*Muro + 0.15* Nuo* L)*[1-(Nu/Nu0)]

(Ht / L) ≤ 2.5




109

• Bending moments due to the eccentricity of the load transmission, from the inter story

or roof, that supports the masonry wall in consideration.

• Bending moments due to the lack of coincidence of the axis of the wall in the storey

above with that of the wall under consideration.

• Slenderness effects.

• Accidental eccentricity of the load, originated by construction imperfection of the wall

under consideration.

Each of the topics listed above are a required verification in the INPRES-CIRSOC 103.

The last in-plane verification is related to the vertical loads. The ultimate strength of the

confined masonry walls subjected to vertical load is determined with the following

expression:

or

Verification for confined masonry given by EU 6, part 1 [9]

(1) In the verification of confined masonry members subjected to bending and/or axial

loading, the assumptions given for reinforced masonry members should be adopted. In

determining the design value of the moment of resistance of a section a rectangular stress

distribution may be assumed, based on the strength of the masonry, only. Reinforcement in

compression should also be ignored.

(2) In the verification of confined masonry members subjected to shear loading the shear

resistance of the member should be taken as the sum of the shear resistance of the masonry

and of the concrete of the confining elements. In calculating the shear resistance of the

masonry the rules for un-reinforced masonry walls subjected to shear loading should be used,

considering for l c the length of the masonry element. Reinforcement of confining elements

should not be taken into account.

Ψ = 1 – (2 e* / t)

e* = et + ea

e* = 0.6 (et + ea) + ec

Nur = Ψ * σ’mo * Bm




110

(3) In the verification of confined masonry members subjected to lateral loading, the

assumptions set out for un-reinforced and reinforced masonry walls should be used. The

contribution of the reinforcement of the confining elements should be considered.

(1)P Confined masonry members shall not exhibit flexural cracking nor deflect excessivelyunder serviceability loading conditions.

(2)P The verification of confined masonry members at the serviceability limit states shall be

based on the assumptions given for un-reinforced masonry members.

Confined masonry out-of-plane verification

Confined masonry walls are subjected to loads in the direction perpendicular to its plane.

These loads are generated by the inertia due to the gravitational loads subjected to the seismic

action. The determination of these loads can be as follow:

The ultimate bending moment originated by this loads can be computed:

Muv = qs (H²/8)

qs = 3.5 * C * q




111

5.5 “Simplified method” allowed by the Argentinean code

The code allows the use of a “simplified method” which assumes that in-plane wall

deformations are governed by shear, and the distribution of ultimate shear stresses across thewall is uniform. In the annex a calculation of a building of solid clay bricks is presented to

describe with more detail the procedure.

Figure 5.5.1 Gravity loads, INPRES-CIRSOC 103 [7]

Figure 5.5.2 Seismic coefficient of design, INPRES-CIRSOC 103 [7]

Seismic Zone Type of masonry




112

Figure 5.5.3 Determination of torsion moments and shears of each story, INPRES-CIRSOC 103 [7]

Figure 5.5.4 Determination of the elastic constants, INPRES-CIRSOC 103 [7]

Type of brick Type of mortar




113

Figure 5.5.5 Geometric characteristics, INPRES-CIRSOC 103 [7]

Figure 5.5.6 Wall stiffness, INPRES-CIRSOC 103 [7]

Considerate wall at

each floor

Contribution of transversal

walls

Flexural

deformation dfi

Shear

deformation dci

Total deformation




114

Figure 5.5.7 Total design shear for each story, INPRES-CIRSOC 103 [7]

Figure 5.5.8 Design bending moment of each wall for each story, INPRES-CIRSOC 103 [7]

Direction of

anal sis

Distribution by

relative Stiffness

Distribution by

relative Stiffness




115

Figure 5.5.9 Design normal resistance for each wall, INPRES-CIRSOC 103 [7]

Figure 5.5.10 Verification of shear strength, INPRES-CIRSOC 103 [7]

Accumulated

slab reactions

Accumulated

wall wei ht etc

T e of brick T e of mortar

It must be verified that:




116

Figure 5.5.11 Verification of gravitational loads, INPRES-CIRSOC 103 [7]

Figure 5.5.12 Reinforcement dimensions of bond-beams, INPRES-CIRSOC 103 [7]

Bond-beams

Longitudinal reinforcement

1) Minimum diameter for stirrups = 4.2mm

2) Length in critical zones: lc = 60cm

3) Minimum longitudinal reinforcement = 4ø8

Stirrups




117

Figure 5.5.13 Reinforcement dimensions of tie-columns, INPRES-CIRSOC 103 [7]

Figure 5.5.14 Verification of flexion and compression, INPRES-CIRSOC 103 [7]

Tie-columns

Longitudinal reinforcement

4) Minimum diameter for stirrups = 4.2mm

5) Length in critical zones: lc = H/5 or 2dc or 60cm

6) Minimum longitudinal reinforcement = 4ø8

Stirrups




118

5.6 Comparison between codes

Different recommendations are given by codes; the following tables summarized the principal

requirements for confined masonry. The majority of the codes belong to South America:Argentina [7], Peru [41], Colombia [39] and Mexico [40]. The EuroCode [1, 9 and 36] and

the Italian code [2] are compared as well.

Table 5.6.1 gives a comparison between the elastic and shear modulus adopted by them. Table

5.6.2 gives minimum requirements for load-bearing walls. Table 5.6.3 compares different

geometrical conditions of confined masonry walls. Table 5.6.4 gives resistance verification in

plane and Table 5.6.5 out-of-plane resistance verifications. Table 5.6.6 and 5.6.7 gives

specifications and requirements for confining elements and their reinforcement.

Table 5.6.1 Elastic properties of masonry given by the different codes

Country Name of the Code Year

Elastic molulus E Shear Modulus G

Argentina INPRES-CIRSOC 103 1983

Normas Argentinas para construcciones Dinamic loads:

sismorresistentes - Parte III: Construcciones

de mamposteria Static loads:

EU prEN 1996-1-1, Eurocode 6: 1996 E: secant value of fig. 3.2

Design of Masonry Structures, The shear modulus, G,

Part 1-1: Common rules for reinforced may be taken as 40%

and unreinforced masonry structures. of the elastic modulus, E.is the final creep coefficient

Italy Norme Tecniche per le Costruzioni in 1984

Zone Sismiche – documento di studio For different typologies For different typologies

GNDT, Dicembre 1984. of masonry units, different of masonry units, different

values of E values of G

Colombia NSR-98 1998 Concrete units:

Titulo D: Mampostería Estructural

Clay units:

Perú Normas tecnicas E.070 2006 Clay units:

Albañileria Silico-calcareas units:

Concrete units:

Mexico Gaceta Oficial del Distrito Federal, 2004 Concrete blocks:

6 de Octubre de 2004, No. 103-BIS. Normas Dinamic loads:

Técnicas complementarias para diseño y Static loads:

construcción de estructuras de mapostería Adobe:

(Tomo I). Normas Técnicas complementarias Dinamic loads:

para diseño por sismo (Tomo II). Static loads:

Elastic properties of masonry

Em 800 σmo⋅:= σmo

Em 300 σmo⋅:= σmo

Gm 0.3 E⋅:= E




119

Table 5.6.2 Minimum conditions for walls to be considered as load-bearing walls

Country Minimum length of Minimum thickness

load-bearing walls of the walls

Argentina Two constrains:

Three or more constrains: For buildings of not more than 3m

height:

EU

Italy

Colombia

Perú

Seismic zone 2 and 3:

Seismic zone 1:

Mexico

H

L2.2≤ L 1 .5m≥

H

L2.6≤ L 0.90m≥

t 17cm≥

t 13cm≥

t 100mm≥

H

t30≤

t 110mm≥

H

t25≤

tefmin mm( ) 240mm≥

hef

tef

⎛

⎝

⎞

⎠15≤

t 24cm≥

h

t15≤

H

L2.≤




120

Table 5.6.3 Geometric conditions for confined masonry given by the different codes, Decanini et al. [38]




121

Table 5.6.4 Resistance verifications in plane given by codes

C o u n t r

y

S h e a r s t r e n g t h

F l e x u r a l s t r e n g t h

C

o m p r e s s i v e s t r e n g t h

F l e x u r a l - c o m p r e s s i v e s t r e n g t h

S i m p l i f y e q u a t i o n

A r g e n t i n a

f o r f l e x u r a l - c

o m p r e s s i v e

I f t h i s e x p r e s i o n i s v e r i f y

i t v e r i f y t o f l e

x u r a l - c o m p

E U

S h e a r r e s i s t a n c e o f t h e m e m b e r s h o u l d B e n d i n g l o a d i n g , E N 1 9 9 6 - 1 - 1 f o r

A x i a l l o a d i n g , E N 1 9 9 6 - 1 - 1 f o r

B e n d i n g a n d / o r a x i a l l o a d i n g , E N 1 9 9 6 - 1 - 1 f o r

b e t a k e n a s t h e s u m o f t h e s h e a r

r e i n f o r c e d m a s o n r y m e m b e r s s h o u l d b e

r e i n f o r c e d

m a s o n r y m e m b e r s s h o u l d b e

r e i n f o r c e d m a s o n r y m e m b e r s s h o u l d b e

r e s i s t a n c e o f t h e m a s o n r y a n d o f t h e

a d o p

t e d . B a s e d o n t h e s t r e n g t h o f t h e

a d o p t e d . B

a s e d o n t h e s t r e n g t h o f t h e

a d o p t e d . B a s e d o n

t h e s t r e n g t h o f t h e

c o n c r e t e o f t h e c o n f i n i n g e l e m e n t s .

m a s o n r y , o n l y .

m a s o n r y , o

n l y

m a s o n r y , o n l y .

R u l e s f o r u n r e i n f o r c e d m a s o n r y s h o u l R e i n

f o r c e m e n t i n c o m p r e s s i o n s h o u l d

R e i n f o r c e m

e n t i n c o m p r e s s i o n s h o u l d

R e i n f o r c e m e n t i n

c o m p r e s s i o n s h o u l d

b e u s e d . R e i n f o r c e m e n t i s n o t t a k e n

a l s o

b e i g n o r e d .

a l s o b e i g n

o r e d .

a l s o b e i g n o r e d .

i n t o a c c o u n t .

T h e d e s i g n r e s i s t a n c e o f e a c h




s t r u c t u r a l e l e m e n t e v a l u a t e d w i t h



s t r u c t u r a

l e l e m e n t e v a l u a t e d w i t h

E N 1 9 9 6 - 1 - 1

E N 1 9 9 6 - 1 - 1

E N 1 9 9 6 - 1 - 1

E N 1 9 9 6 - 1 - 1

I t a l y

C o l o m b

i a

P e r ú

I f t h i s r e l a t i o n

i s v e r i f y

n o n e e d t o d o

o u t - o f - p l a n e v

e r i f i c a t i o n

M e x i c o

I n c a s e t h

e a b o v e

r e l a t i o n i s v e r i f y , w a l l s

m u s t v e r i f y f l e x i o n

I n

- p l a n e v e r i f i c a t i o n

V U R

0 . 6

τ m o

⋅

0 . 3

σ o

⋅

+

(

)

B ⋅

: =

τ m o

V U R

1 . 5

τ m o

⋅

B ⋅

≤

M

U R

0

A c

β s

⋅

L e

⋅

: =

M

U R

0

i f

N U

N U o 3

≤

⎛ ⎝

⎞ ⎠ . M U

R

M U R

0

0 . 3 N U

⋅

L ⋅

+

: =

N U

N U o 3

≤

⎛ ⎝

⎞ ⎠ . M U

R

N U o

σ m o B M ⋅

: = σ m o

i f

N U

N U o 3

>

⎛ ⎝

⎞ ⎠ . ⋅ M

U R

⋅

1 . 5 M U R

0

⋅

0 . 1 5 N U

⋅

L ⋅

+

⎝

⎠

1

N U

N U o

−

⎛ ⎜ ⎝

⎞ ⎟ ⎠

⋅

: =

N U

N U o 3

>

⎛ ⎝

⎞ ⎠ . ⋅ M

U R

⋅

H t

L

2 .

≤

N U R

Ψ σ

m o

⋅

B M ⋅

: = Ψ

Ψ

1

2 e t

⋅

⎛ ⎝

⎞ ⎠

−

: =

t

N U R

2 . 6 N

U

≥

d c

2 h b ⋅

≤

d c

5 m

≤

H L

2

≥




122

Table 5.6.5 Resistance verification out-of-plane and verification of the confining elements

C o u n t r y

O u t - o f - p l a n e v e r i f i c a t i o n

V e

r i f i c a t i o n o f

V e r i f i c a t i o n o f

V e r i f i c a t i o n o f

C r a c k c o n

t r o l

C o m p

r e s s i v e d i a g o n a l

T i e - c o l u m n

B o n d - b e a m

A r g e n t i n a

E U

C o n f i n e d m a s o n r y

m e m b e r s

s h a l l n o t e x h i b i t f l e

x u r a l

c r a c k i n g n o r d e f l e c

t

e x c e s s i v e l y u n d e r s e r v i c e a b i l i t

l o a d i n g c o n d i t i o n s .

A s s u m p t i o n s g i v e n

f o r u n -

r e i n f o r c e d m a s o n r y

i s u s e d

I t a l y

C o l o m b i a

f l e x o - c o m p r e s i o n s t r e n g t h

g i v e n o n l y b y t i e - c o l u m n s

P e r ú

C o m p r e s s i v e s t r e n g t h

i n s e r v i a b i l i t y s t a t e s h o u l d

S h e a r

b e l o w e r t h a n :

T e n s i o n

C o m p r e s s i o n

M e x i c o

q s

3 . 5 C ⋅

q ⋅

: =

q

M U V

q s

H 2 8

⋅

: = q s

M U V 2 . 6 ⋅

M U R

≤




123

Table 5.6.6 Specifications and requirements of the confining elements

C o u n t r y

D e s i g n a t i o n o f t h e

M a x i m u m

d i s t a n c e b e t w e e n

M i n i m u m s e c t i o n o f c o n f i n i n g e l e m e n t s

M i n i m u m s t r e n g t h o f t h e

c o n f i n i n g e l e m e n t s

c o n f i

n i n g e l e m e n t s

c o n c r e t e o f t h e c

o n f . E l e m

A r g e n t i n a

T i e - c o l u m n s :

c o l u m n a s d e e n c a d e n a d o


B o n d - b e a m s :

B o n d - b e a m s : t h e m i n i m u m w i d t h i s t h e

v i g a s d e e n c a d e n a d o

t h i c k n e s s o f t h e w a l l a n d t h e m i n i m u m h e i g h t

1 5 c m

E U


c r o s s - s e c t i o n a l d i m e n s i o n s o f c o n f i n i n g e l e m e n t s

v e r t i c a l c o n f i n i n g e l e m e n t s

m a y n o t b e l e s s t h a n

0 . 0 2 m ² , w i t h a m i n i m u m

d i m e n s i o n i n t h e p l a

n o f t h e w a l l o f 1 5 0 m m


h o r i z o n t a l c o n f i n i n g e l e m e n t s


v e r t i c a l c o n f i n i n g e l e m e n t s

c r o s s - s e c t i o n a l d i m e

n s i o n s o f c o n f i n i n g e l e m e n t s

m a y n o t b e l e s s t h a n

1 5 0 m m


h o r i z o n t a l c o n f i n i n g e l e m e n t s

I t a l y


M i n i m u m t r a n s v e r s a l d i m e n s i o n o f t i e - c o l u m n s

c o r d o l i v e r t i c a l i

t h e t h i c k n e s s o f t h e

w a l l a n d t h e o t h e r d i r e c t i o n


c o r d o l i o r i z z o n t a l i

B o n d - b e a m s m u s t h

a v e a t l e a s t t h e h e i g h t o f t h e

s l a b

C o l o m b i a


F o r t i e - c o l u m n s a n d

b o n d - b e a m s :

c o l u m n a s d e c o n f i n a m i e n t o


v i g a s d e c o n f i n a m i e n t o

m i n i m u m d i m e n s i o n o f t h e c o n f i n i n g

e l e m e n t s : t h e t h i c k n

e s s o f t h e w a l l : t

P e r ú

M i n i m u m s e c t i o n : t h e m a x i m u m o b t a i n e d b y t h e


f o l l o w i n g e x p r e s s i o n s :

c o l u m n a s



v i g a s s o l e r a s



M e x i c o

M a x i m u m d i s t a n c e

b e t w e e n b o n d - b e a m s

d a l a s = b o n d - b e a m s

c a s t i l l o s = t i e - c o l u m n s

M a x i m u m d i s t a n c e

b e t w e e n t i e - c o l u m n s

Q = f a c t o r d e

M i n i m u m d i

m e n s i o n o f t i e - c o l u m n

c o m p o r t a m i e n t o s i s m i c o




d c

4 m

≤

o r

d c

1 . 5 H ⋅

≤

H

D i s t a n c e b e t w e e n b o n d - b e a m s

h b

3 m

≤

s e i s m i c

z o n e

M a x i m u m

a r e a

M

a x i m u m d

i e n s i o n o f t h e p a n e l

1 7 c m

≥

1 7 c m

<

a n d 1 3 c m

≥

1

3 0 m 2

7 m

4 . 5 m

2

2 5 m 2

6 m

4 m

3

2 0 m 2

5 m

4 m

σ b k

1 1

M N m 2

≥

B c

0 . 0 2 5 V p

: =

V p

c m

2

(

)

d c

4 m

≤

o r

d c

1 . 5 H ⋅

≤

d c

3 5 t ⋅

≤

H

D i s t a n c e b e t w

e e n b o n d - b e a m s

d c

2 h b ⋅

≤ d c

5 m

≤ d c

5 m

≤

h b

4 m

≤ d c

4 m

≤

h b

4 m

≤

d c

5 m

≤

h b

4 m

≤




124

Table 5.6.7 Specifications and requirements of reinforcement for confining elements

C o

u n t r y

L o n g i t u d i n a l r e i n f o r c e m e n t

M

i n i m u m l o n g . R e i n f o r .

M i n i m u m s t i r r u p s o f t h e

O p e n i n g s

o f c o n f i n i n g e l e m e n t s

o f t h e c o n f . E l e m e n t s

c o n f i n i n g e l e m e n t s

A r g e n t i n a

S e i s m

i c z o n e 1 a n d 2 :

* a r e a o f t h e o p e n i n g ≤ 3 5 %

o f t h e t o t a l a r e a o f t h e p a n e

l

* M i n i m u m l e n g t h f r o m t h e

v e r t i c a l

S e i s m

i c z o n e 3 a n d 4 :

e d g e o f t h e v e r t i c a l c o l u m n

t o t h e

v e r t i c a l e d g e o f t h e o p e n i n g : 0 . 9 0 m

o r 2 5 % o f t h e t o t a l l e n g t h o f t h e p a n e l

E U

M a y n o t h a v e a c r o s s - s e c t i o n a l

N o t l e s s t h a n d i a m

e t e r 6 m m a n d

A l l o p e n i n g s w i t h a n a r e a h

i g h e r

a r e a l e s s t h a n 2 0 0 m m ² o r t h a n

s p a c e d n o t m o r e t h a n 3 0 0 m m

t h a n 1 . 5 m ² m u s t b e c o n f i n e

d

0 . 8 %

o f t h e c r o s s - s e c t i o n a l a r e a

o f t h e

c o n f i n i n g e l e m e n t

M a y n o t h a v e a c r o s s - s e c t i o n a l

N o t l e s s t h a n d i a m

e t e r 5 m m a n d

T h e r a t i o o f t h e l e n g t h o f t h

e w a l l , l ,

a r e a l e s s t h a n 3 0 0 m m ² o r t h a n

s p a c e d n o t m o r e t h a n 1 5 0 m m

t o t h e g r e a t e r c l e a r h e i g h t , h , o f t h e

1 % o f t h e c r o s s - s e c t i o n a l a r e a

a n d s h o u l d b e p r o

v i d e d a r o u n d t h e

o p e n i n g s a d j a c e n t t o t h e w a l l , m a y

o f t h e

c o n f i n i n g e l e m e n t

l o n g i t u d i n a l r e i n f

o r c e m e n t

n o t b e l e s s t h a n a m i n i m u m

v a l u e

l / h > 0 . 3

I t a l y

D o e s n o t g i v e a n y i n f o r m a t i o n

C o l o m b i a

T o b e c o n s i d e r e d a s a l o a d - b e a r i n g

o r 1 . 5 t i m e s t h e m

i n i m u m d i m e n s i o n w a l l s , t h e w a l l p a n e l m u s t n o t h a v e

o f t h e t i e - c o l u m n

o r 2 0 0 m m

a n y o p e n i n g

I n h i g h s e i s m i c z o n e s m i n i m u m

s p a c i n g 1 0 0 m m

P e r ú

M e

x i c o

o p e n i n g s m u s t b e c o n

f i n e d

i f t h e w i d h o f t h e o p e n i n g

i s m o r e t h a n 6 0 0 m

m

o r 1 / 4 o f t h e d i s t a n

c e

b e t w e e n t i e - c o l u m

n s

s

1 . 5 t ⋅

≤

o r

s

2 0 0 m m

≤

t t h i c k n e s s o f c o n f . e l e m e n t

s d i s t a n c e b e

t w e e n s t i r r u p s

A c

1

0 . 2 5 k ⋅

+

(

) V p ⋅

H o

L o

β

s ⋅

⋅

: =

k

A v

V p

1 β s

⋅

: = V p

A m i n

0 . 2 5

0 . 1 3 k ⋅

+

(

) t ⋅

1 β s

⋅

: =

k

A m i n

0 . 3

5

0 . 1

8 k ⋅

+

(

) t ⋅

1 β

s

⋅

: =

k

d s

0 . 2 0

0 . 1 k

+

(

)

s e ⋅

: =

k

c m

s e

2 0 c m

≤

n o r m a

l z o n e s

A e

0 . 5

V p

d c

β s ⋅

⋅

s e c

⋅

: =

V p

s e c

1 0 c m

≤

c r i t i c

z o n e s

s e i s m i c z o n e 1 , 2

>

4 φ 6 m m

s e i s m i c z o n e 3 , 4

>

4 φ 8 m m

0 . 0 0 7 5 B m

⋅

≥

m i n i m

u m s

e c t i o n

o f l o n g . r e i n f o r c e m e n t

B m

s e c t i o n o f t h e c o n f i n i n g

e l e m e n t




125

5.7 Conclusions and possible topics to develop

In Table 5.6.1 the names of the different codes are given, also their year, being the oldest one

the Argentinean masonry code, 1983.

Table 5.6.1 refers to the calculation of the elastic modulus and the shear modulus. All the

codes have more less the same way to determine the shear modulus Gm, as the 40% of the

elastic modulus Em, only the Argentinean code adopts the 30% of Em. One difference in the

determination of Em is that the Argentinean and the Mexican code compute this modulus as a

function of static or dynamic loads, while the others used the type of masonry units to

difference the value of this modulus.

In Table 6.6.2 geometrical relations are given in order to consider the wall panel as a load-

bearing wall. Comparing the different values given for the thickness of the walls panels it wasfound that the minimum values are given by the Mexican code, 10cm, while the maximum is

given by the EU and the Italian code of 24cm. This values depend on the country may be for

the fact that the masonry units use are not the same so the thickness can vary from one

country to the other.

Only the Argentinean and the Italian codes give limitations in the relations between the height

and the length of the walls panels, being the common values approximately of 2.5. On the

other hand the relation between the height of the wall and the thickness is very low for the EU

and the Italian code (h/t = 15) in comparison with the rest of the codes that vary between 20

and 30.

Enough recommendations are given related with minimum dimensions of the wall panels and

confining elements in Table 5.6.3. The minimum distance of confining elements (tie-columns

or bong-beams) is more less the same, between 4 and 5m. This limitation is due to the fact

that the confined masonry panel can develop a compressive diagonal needed for the resisting

mechanisms.

In Table 5.6.4 and 5.6.5 expressions to compute the resistance of the wall to shear,

compression and flexion are given, in the plane of the wall and in the out-of-plane direction as

well. Verification of the confining elements is given only by the Colombian and the Peruviancode, crack control is only analyzed in the Peruvian code. The fact that the others codes do

not give any verification for the confining elements is that the contribution of the tie-columns

and bond-beams to the lateral resistance of the masonry building is normally not taken into

account for design. Consequently specific design calculations for confining elements are not

required.

Another difference found was if the confining elements collaborate or not with the resistance

of the confined masonry wall. In the Argentinean code not contribution of the confining

elements to shear is taken into account. For the EU to compute the shear resistance of the wall

the sum of the masonry and the concrete of the confining elements is used, no collaboration of

the reinforcement is taken into account. For axial and bending resistance the reinforcement is




126

ignored as well. In the Mexican and the Argentinean code the resistance to bending is given

by the reinforcement of the confining elements.

In Table 5.6.6 and 5.6.7 the minimum sections for the confining elements are given. Only the

Argentinean and the Peruvian codes give the minimum section as function of the shear baseacting in the structure. The rest of the codes give a minimum section dimension of 200cm².

The longitudinal reinforcement is only calculated in function of the shear base by the

Colombian, Peruvian and Argentinean codes. The rest of the codes give minimum values of

longitudinal reinforcement. In the Argentinean code the amount of reinforcement depends on

the thickness of the wall and the number of stories above the analyzed one. Frequently, the

amount of reinforcement in vertical and horizontal confining elements is determined on an

empirical basis. Although the tie-columns and bond-beams do not provide frame system

contribution to the wall, adequate splicing and anchoring of re-bars is required at all joints.

In the EU6 and EU8 some differences were found. In the EU6 a percentage of 0.8% is

considered or a minimum of 200mm², while in the EU8 this percentage is 1.0% or a minimum

of 300mm². Another difference was also found in the stirrups were the EU6 has a minimum

bar diameter of 6mm each 300mm of spacing and the EU8 a minimum diameter of 5mm with

a spacing of 150mm.

Not much information was found for the openings, only limitations of section and some

recommendation codes gives when it should be necessary to confine the openings.

Summarizing, enough recommendations are given for limitation of spacing of confining

elements, where they should be positioned, and minimum dimensions. More investigations in

resistance verification should be done, with emphasis in which parts of the confined masonry

walls collaborates in the different solicitations. Also clear equation for computing this last

topic must be developed and for reinforcement as well. More specifications about minimum

section of openings must be analyzed, the same as crack control, out-of-plane verification and

control of the compressive diagonal, especially for the cases of hollow concrete masonry

units.



127



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128

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earthquake of December 26, 2003, Iran,” Institute of Industrial Science, Volume 3, Number 4.

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masonry structures through shaking table test,” Proceedings of 13th World Conference on Earthquake Engineering , Vancouver, B.C., Canada. Paper no 2130.

[28] Tomazevic, M., Klemenc, I., Petrovic, L., Lutman, M. [1996] “Seismic behaviour of confined

masonry buildings. Part one: Shaking-table tests of model buildings M1 and M2 - Test

results,” A Report to the Ministry of Science and Technology of Republic of Slovenia. Grant

no. J2-5208-1502ZAG/PI-95/04, Ljubljana: National buildings and civil engineering institute.

[29] San Bartolomé, A., Quiun, D. and Torrealva, D. [1992] “Seismic behaviour of a three-story

half scale confined masonry structure,” Proceedings of 10th World Conference on Earthquake Engineering, Madrid, Spain.

[30] Scaletti, H., Chariarse, V., Cuadras, C., Cuadros G. and Tsugawa T. [1992] “Pseudos dynamictests of confined masonry buildings,” Proceedings of 10th World Conference on Earthquake

Engineering, Madrid, Spain.

[31] Alcocer, S.M., Zepeda, J.A. “Behavior of multi-perforated clay bricks walls under earthquake

clay brick walls,”

[32] Decanini, L.D., Payer, A., Serrano, C., Terzariol, R. [1985] “Investigación experimental sobre

el comportamiento sismoresistente de prototipos a escala natural de muros de mamposteria

encadenada,” XXIII Jornadas Sudamericanas de Ingenieria Estructural.

[33] Irimies, M.T. [2000] “Confined Masonry walls. The influence of the tie-column vertical

reinforcement ratio on the seismic behaviour,” Proceedings of 12th World Conference on Earthquake Engineering .

[34] Marinilli, A., Castilla, E. [2004] “Experimental evaluation of confined masonry walls with

several confining columns,” Proceedings of 13th World Conference on Earthquake Engineering , Vancouver, B.C., Canada. Paper no 2129.

[35] Ishibashi, K., Meli, R., Alcocer, S.M., Leon, F. and Sanchez, T.A. [1992] “Experimental study

on earthquake-resistant design of confined masonry structures,” Proceedings of 10th WorldConference on Earthquake Engineering, Madrid, Spain.

[36] Eurocode 8 [1995] “Design provisions for earthquake resistance of structures. Part 1-2:

General rules- General rules for buildings,” ENV 1998-1-2: 1995 (CEN, Brussels, 1995)

[37] “Construction under seismic condition in the Balkan region,” Vol. 3: Design and constructionof stone and brick masonry buildings (UNIDO/UNDP, Vienna, 1984)

[38] Decanini, L., Liberatore, L., De Sortis, A., Benedetti, S. [2006] “Esame e raffronto delle

prescrizioni di diverse normative, nazionali e internazionali, sugli edifici a struttura mista

muratura-c.a.,” Rete dei Laboratori Universitari di Ingegneria Sismica (RELUIS), Progettoesecutivo 2005-2008. Progetto di ricerca N.1

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construcción de estructuras de mampostería,” Gobierno del Distrito Federal de México.

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Annex

A1

2. ANNEX

Seismic Verification of a confined masonry building following the seismic

regulation of the Argentinean Code INPRES-CIRSOC 103

Example for a masonry building, Figure 1 with the following conditions:

Seismic zone: I

Solid clay bricks, type B

Mortar joint with normal resistance

Seismic coefficient, C : (3.1.4.2. INPRES-CIRSOC 103)

Cnm 0.1:=

γd 1:=

C Cnm γd⋅:=

C 0.1= Seismic coefficient

Cn

Normalized seismic coefficient for masonry constructions, depends of the masonry unit

and the seismic zone, Table 1 of INPRES-CIRSOC 103

γd Safety factor, Part I of the INPRES-CIRSOC 103 for concrete structures

Minimum conditions for load-bearing walls: (7.4.2. INPRES-CIRSOC 103)

The minimum thickness for a load-bearing wall is 17cm except for the following cases:

b) For seismic zone 1 and 2:

It is allowed to consider like a load-bearing wall of 13cm of thickness if:

Is a wall M1 or M2

Be a construction type B or C of the part I of the INPERES-CIRSOC code

Maximum number of story 1

Maximum height 3m

In this case it can be consider like load-bearing walls those which have a thickness of 13cm,

because they present the minimum conditions described above



Annex

A2

Figure annex I Dimensions of the building



Annex

A3

Minimum length for load-bearing walls: (7.4.3. INPRES-CIRSOC 103)

a) Two horizontal constrains

H

L 2.2≤ or L 1.5≥ For confined masonry

Like H= 2.6m: L 1.2≥

Concluding: to be a load-bearing wall, if it has two horizontal

constrains L 1.5≥

b) Three or more constrains

H

L2.6≤ or L 0.9≥ For confined masonry

Like H= 2.6m: L 1.0≥

Concluding: to be a load-bearing wall, if it has two horizontal

constrains L 1.0≥

With this conditions listed above the following walls, Figure 2, that will act like load-bearing

walls were chose.

Loads:

1) Slabs

γc 240:= tn/m3 Concrete density

γco 180:= tn/m3 Cover density

ts 0.1:= cm Slab thickness

tc 0.15:= cm Cover thickness

A 4.20 7.35⋅( ) 3.80 10.40⋅( )+:=

A 70.39= m 2 Total area of the slab

Total weigh of the slab W

WL ts γc⋅( ) tc γco⋅( )+A

1000⋅:=

WL 35.899= tn



Annex

A4

Figure annex II Load-bearing walls



Annex

A5

2) Walls: approximation to obtain the total weighs of the walls

x-x direction:

Walls of 15cm thickness:q15 24:= kg/cm

2 Weigh of the 15cm walls per unit

area

hw 2.60:= m Height of the walls

Lx15 2.15 3.35+ 2.25+( ):= Total length in the x-x direction

Lx15 7.75= m

wxm15 Lx15hw⋅( )q15

1000⋅:=

wxm15 4.836= tn Total weigh of the wall of 15cm thickness in the x-x direction

Walls of 30cm thickness:q30 480:= kg/cm2

hw 2.6= m Height of the walls

Lx30 2.15 3.35+ 3.40+ 3.10+( ):= Total length in the x-x direction

Lx30 12= m

wxm30 Lx30hw⋅( )

q30

1000⋅:=

wxm30 14.976= tn Total weigh of the wall of 30cm thickness in the x-x direction

y-y direction:

Walls of 15cm thickness:

Ly15 17.30( ):= Total length in the x-x direction

Ly15 17.3= m

wym15 Ly15 hw⋅( ) q15

1000⋅:=

wym15 10.795= tn Total weigh of the wall of 15cm thickness in the x-x direction



Annex

A6

Walls of 30cm thickness:

Ly30 9.5( ):= Total length in the x-x direction

Ly30 9.5= m

wym30 Ly30 hw⋅( )q30

1000⋅:=

wym30 11.856= tn Total weigh of the wall of 30cm thickness in the x-x direction

3) For this example no water tank is supposed

Total weigh

Total weigh in the x-x direction:

Wx wxm15 wxm30+ W+:=

Wx 55.711= tn

Total weigh in the y-y direction:

Wy wym15 wym30+ W+:=

Wy 58.55= tn

Base Shear: (3.1.4.1. INPRES-CIRSOC 103)

x-x direction: y-y direction:

V0x Wx C⋅:= V0y Wy C⋅:=

V0x 5.571= tn V0y 5.855= tn



Annex

A7

Determination of the stiffness of the walls panels: (4.2. INPRES-CIRSOC 103)

Em 10:= tn/cm2 Elastic modulus of the masonry hw 2.6= m

t30 0.3:= m P 1:= tn Applied loadThickness

t15 0.1:= m

f P hw

3⋅

30 Em⋅ J⋅0.288P⋅

hw

Em Am⋅⋅+

⎛

⎝ :=

Am

Stiffness R R P

f :=

f

J td

3

12⋅:= t Being t: thickness of the wall; d: length of the wall and J: inertia of the wall

Am t d⋅:= t Area of the wall

Walls

x-x direction y-y direction

Mx1 R x1 1.66:= tn/cm My1 R y1 4.7:= tn/cm My5 R y5 5.32:= tn/cm

Mx R x2 2.85:= tn/cm My2 R y2 3.4:= tn/cm My6 R y6 5.98:= tn/cm

Mx R x3 2.50:= tn/cm My3 R y3 0.8:= tn/cm My7 R y7 0.96:= tn/cm

My4 R y4 5.8:= tn/cm My8 R y8 5.32:= tn/cm

Determination of the mass centre:

It is determined by the sum up of all the walls area multiplied by the distance of each one to an

arbitrary axis divided after by the total area

cΣAm dc⋅

At

:=ΣAm

Where: Am is the area of the wall; dc is the distance between the centre of

the wall and an arbitrary axis and At is the total area of the walls


cx 5.9:= m cy 4.3:= m



Annex

A8

Determination of the stiffness centre:

It is determined by the sum up of all the walls area multiplied by the distance of each one to an

arbitrary axis divided after by the total area

R ΣR w dc⋅

R t

:=ΣR w

Where: Rw is the stiffness of the wall; dc is the distance between the

centre of the wall and an arbitrary axis and Rt is the total stiffness of the

wallsx-x direction y-y direction

R x 5.51:= m R y 4.3:= m

Distribution of the shear base: (3.1.4.3. INPRES-CIRSOC 103)

x-x directionR xt R x1 R x2+ R x+:=

R xt 7.01= tn/cm

Vx1 R x1

V0x

R xt

⋅:= Vx2 R x2

V0x

R xt

⋅:= Vx1 1.319= tn Vx2 2.265= tn

Vx3 R x3

V0x

R xt

⋅:= Vx3 1.987= tn

y-y direction

R yt R y1 R y2+ R y3+ R y4+ R y5+ R y6+ R y7+ R y8+:=

R yt 32.52= tn/cm

Vy1 R y1

V0y

R yt

⋅:= Vy5 R y5

V0y

R yt

⋅:= Vy1 0.857= tn Vy5 0.958= tn

Vy2 R y2

V0y

R yt

⋅:= Vy6 R y6

V0y

R yt

⋅:= V

y2

0.619= tn V

y6

1.077= tn

Vy3 R y3

V0y

R yt

⋅:= Vy7 R y7

V0y

R yt

⋅:= Vy3 0.155= tn Vy7 0.173= tn

Vy4 R y4

V0y

R yt

⋅:= Vy8 R y8

V0y

R yt

⋅:= Vy4 1.059= tn Vy8 0.958= tn



Annex

A9

Torsional moments of the structure: (3.1.5. INPRES-CIRSOC 103)

Static eccentricity:

ex cx R −:= ey cy R y−:=ex 0.42= m ey 10 10 3−×= m

Torsional moments:

x-x direction

ly 9.4:= m Maximum dimension measured in the building plane perpendicular to the

seismic actionMtx1 2 ey⋅ 0.1 ly⋅+( ) V0⋅:=

Mtx1 5.348= tnmMtx1 Mtx2>

Mtx2 ey 0.1 ly⋅−( ) V0⋅:=

Mtx2 5.181−= tnm

y-y direction

lx 8.0:= m Maximum dimension measured in the building plane perpendicular to the

seismic action

Mty1 2 ex⋅ 0.1 lx⋅+( ) V0y⋅:= Mty1 9.602= tnm

Mty1 Mty2> Mty2 ex 0.1 lx⋅−( ) V0y⋅:=

Mty2 2.225−= tnm

Distribution of the torsional moment between the walls:

Ft

R i di⋅

ΣR i di2

⋅

⎛

⎝

Mt⋅:=R i

Ri: stiffness of each wall; di: distance between the walls and the

stiffness centre


Mx1 Mxt1 1.1:= tn My1 Myt1 0.6−:= tn My5 Myt5 0.3−:= tn

Mx Mxt2 0.2:= tn My2 Myt2 0.4−:= tn My6 Myt6 0.2:= tn

Mx Mxt3 0.1−:= tn My3 Myt3 0.0−:= tn My7 Myt7 0.0:= tn

My4 Myt4 1.0−:= tn My8 Myt8 0.2:= tn



Annex

A10

Total loads applied to the walls: Shear base + torsional moment

The torsional effect is only consider in the case that it add force to the shear base due to

translation only

x-x direction

Mx1 Vx1 Mxt1+:= Mx1 2.509= tn Total shear base acting in wall Mx1

Mx2 Vx2 Mxt2+:= Mx2 2.465= tn Total shear base acting in wall Mx2

Mx3 Vx:= Mx3 1.987= tn Total shear base acting in wall Mx3

y-y direction

My1 Vy1:= My1 0.857= tn Total shear base acting in wall My1





My6 Vy6 Myt6+:= My6 1.347= tn Total shear base acting in wall Mx6





Annex

A11

Resistance verification: (10. INPRES-CIRSOC 103)

Shear strength: (10.2.1.1. INPRES-CIRSOC 103)

τmo 0.2:= MN/m2 Shear strength of the masonry Table 10 - INPRES-CIRSOC 103

Q 740:= Kg/m

σoQ

Am

:=Am

Compressive stress due to the 85% of the vertical loads

x-x direction

σox1Q 2.3⋅

230 15⋅:=

σox1 0.493= Kg/cm2 Bmx1 230 15⋅:= Bmx2 247.515⋅:=

Bmx3 Bmx:= σox2

Q 2.475⋅

247.515⋅:=

σox2 0.493= Kg/cm2 σox3 σox2:= σox σox1:=

σox 0.493= kg/cm2

Shear strength:

Vur 0.6 τmo⋅ 0.3 σo⋅+( ) B⋅:= σo

Vur M≥ The shear strength (Vur) of each wall must be

higher than the acting force MVurx1 0.6 τmo⋅ 10⋅ 0.3 σox⋅+( )Bmx1

1000

⋅:=

Vurx1 5.686= tn > Mx1 2.509= tn VERIF

Vurx2 0.6 τmo⋅ 10⋅ 0.3 σox⋅+( )Bmx2

1000⋅:=

Vurx2 6.118= tn > Mx2 2.465= tn VERIF

y-y direction

It can only be verify the wall with the minimum length; if that one verifies the rest alsoWall with less length My3



Annex

A12

σoy3Q 1.75⋅

175 15⋅:=

σox3 0.493= kg/cm2 Bmy3 175 15⋅:=

Shear strength:

Vury3 0.6 τmo⋅ 10⋅ 0.3 σoy3⋅+( )Bmy3

1000⋅:=

Vury3 4.326= tn > My3 0.155= tn VERIF

Compressive strength: (10.4.4.1. INPRES-CIRSOC 103)

Ψ 0.25:= Reduction factor due to eccentricity and slenderness, determined by theequation given in 10.4.4.2

σmo 1.5:= MN/m2 Compressive strength of masonry determined by 6.1.1

Nurx1 Ψ σmo⋅ 10⋅Bmx1

1000⋅:=

Nurx1 12.938= tn > Nx1 0.74 2.30⋅ 0.24 2.6⋅ 2.3⋅+:=

Nx1 3.137= tn VERIF


1000

⋅:=

Nurx2 13.922= tn > Nx2 0.74 2.475⋅ 0.24 2.6⋅ 2.47⋅+:=

Nx2 3.376= tn VERIF


1000⋅:=

Nurx3 13.922= tn > Nx3 Nx:=

Nx3 3.376= tn VERIF

Nury3 Ψ σmo⋅ 10⋅Bmy3

1000

⋅:=

Nury3 9.844= tn > Ny3 0.74 1.75⋅ 0.24 2.6⋅ 1.75⋅+:=

Ny3 2.387= tn VERIF

All walls verify to compressive strength



Annex

A13

Flexion-compressive strength: (10.2.2.2. INPRES-CIRSOC 103)

In part c) of the INPRES-CIRSOC 103 the verification to flexion with compression of the

wall is not necessary for those of 1 or 2 seismic zone and high less than 9m, that verify

the following condition:Ht

L2.5≤ Where Ht is the height of the wall and L the length of

the wall

In this case all the walls verify this condition so the verification of flexion with compression

in not necessary

Out-of-plane verification: (10.5. INPRES-CIRSOC 103)

q 240:= Kg/m2 Weigh of the 15cm thickness wall per unit of area

qs 3.5 C⋅ q⋅:=

qs 84= Kg/m2 Seismic load per unit length acting in the perpendicular direction of it

plane

Out-of-plane bending moment:

Mu qshw

2

8⋅:=

Mu 70.98= kgm/m

Nu 740 4.375⋅ 0.15⋅ 4.3750.15⋅ 240⋅+:=

Nu 643.125= kg

em

Mu

Nu

:= emd

6<

t15 100⋅

62.5=

em 0.11= cm

em 0.11= <d

6

Due to this relation some part is under

compression and the rest tension

B 15 437.⋅:= W437.5

615

2⋅:=

B 6.563 103

×= W 1.641 104

×=

σ p

Nu

B

⎛

⎝

Mu

W

⎛

⎝ +:=

σ p 0.102=

σn

Nu

B

⎛

⎝

Mu

W

⎛

⎝

⎞

⎠−:=

σn 0.094=

at15 100⋅

2

⎛ ⎝

0.11 100⋅−:=

a 3.5−=



Annex

A14

σmax Nu2

3 a⋅ 100⋅⋅:=

σmax 1.225= <24.6

2.69.462= VERIF

Tie-columns and bond-beams dimensions: (9.6. INPRES-CIRSOC 103)

Tie-columns

Minimum gross section area of tie-columns

V p Mx1 100⋅:= Maximum shear applied to the walls

V p 2.509 103

×=

Bm 0.025V p⋅:= cm2 Bm 62.731= cm2

Dimension of tie-columns (9.7.1. INPRES-CIRSOC 103)

Btc 15 15⋅:=

Btc 225= cm2

Bond-beams

Dimension of bond-beams (9.7.2. INPRES-CIRSOC 103)

B bb 15 15⋅:=

B bb 225= cm2

Bond-beams at the base

B bt 15 10⋅:=

B bt 150= cm2 Bond-beams at the top

Tie-columns and bond-beams reinforcement: (9.9. INPRES-CIRSOC 103)

k 0:= Number of floors above the study one

Ho hw 100⋅:= Height measured between bond-beams

βs 420:= kg/cm2 Yield stress of the steel



Annex

A15

Minimum reinforcement given by the INPRES-CIRSOC 103 - 9.10

Amin 0.25 0.13 k ⋅+( ) t15⋅ 100⋅1

βs

⎛

⎝

⎞

⎠⋅ 100⋅:= Amin 0.893= cm2 4φ6m

Longitudinal reinforcement: (9.11. INPRES-CIRSOC 103)

Spacing between longitudinal reinforcement: 20cm

Anchorage 60* 6mm a 36:= cm

αe 1:= l a:=

Splice le αe l⋅:= le 36= cm

Mx1

Tie-columns

Lx1 23:= cm Yield stress of the steel

Acx1 1 0.25 k ⋅+( ) Mx1⋅Ho

Lx1

⎛

⎝ ⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acx1 0.675= cm2

Minimum reinforcement adopted:

Bond-beams

A bx1

Mx1 1000⋅

βs

:= A bx1 0.597= cm2


Stirrups (9.12. INPRES-CIRSOC 103)

Critic zones

Spacing se 7.:= cm

Section of the reinforcement Ae

Ae 0.5Mx1 se⋅ 1000⋅

15 βs⋅⋅:=

Ae 0.149= cm2

Normal zone

sen 15c:= ds 0.2 15⋅:=

ds 3= mm → φ 4:= m

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A16

Mx

Tie-columns

Lx2 247.:= cm

Acx2 1 0.25 k ⋅+( ) Mx2⋅Ho

Lx2

⎛

⎝ ⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acx2 0.617= cm2


Bond-beams

A bx2

Mx2 1000⋅

βs

:= A bx2 0.587= cm2



Critic zones

Spacing se 7.5= cm


Ae2 0.5

Mx2 se⋅ 1000⋅

15 βs⋅⋅:= Ae2 0.147= cm2

Normal zone

sen 0.15m= ds 3=

ds 3= mm → φ 4= m

Mx

Tie-columns

Lx3 247.:= cm

Acx3 1 0.25 k ⋅+( ) Mx3⋅Ho

Lx3

⎛

⎝ ⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acx3 0.497= cm2


Bond-beams

A bx3

Mx3 1000⋅

βs

:= A bx3 0.473= cm2

Minimum reinforcement adopted: 4φ6mm 1.13cm2=φ6mm

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm

2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A17


Critic zones

Spacing se 7.5= cm


Ae3 0.5Mx3 se⋅ 1000⋅

15 βs⋅⋅:=

Ae3 0.118= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

My1

Tie-columns

Ly1 38:= cm

Acy1 1 0.25 k ⋅+( ) My1⋅Ho

Lx3

⎛

⎝ ⋅

1

βs

⎛

⎝

⎞

⎠⋅ 100⋅:=

Acy1 0.214= cm2


Bond-beams

A by1

My1 1000⋅

βs

:= A by1 0.204= cm2



Critic zonesSpacing se 7.5= cm


Ay1 0.5My1 se⋅ 1000⋅

15 βs⋅⋅:=

Ay1 0.051= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A18

My2

Tie-columns

Ly2 322.:= cm

Acy2 1 0.25 k ⋅+( ) My2⋅Ho

Ly2

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy2 0.119= cm2


Bond-beams

A by2

My2 1000⋅

βs

:= A by2 0.147= cm2



Critic zones

Spacing se 7.5= cm


A

y2

0.5My2 se⋅ 1000⋅

15 βs⋅⋅:=

Ay1 0.051= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

My3

Tie-columns

Ly3 17:= cm

Acy3 1 0.25 k ⋅+( ) My3⋅Ho

Ly3

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy3 0.055= cm2


Bond-beams

A by3

My3 1000⋅

βs

:= A by3 0.037= cm2

Minimum reinforcement adopted: 4φ6mm 1.13cm2=φ6mm

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A19


Critic zones

Spacing se 7.5= cm


Ay3 0.5My3 se⋅ 1000⋅

15 βs⋅⋅:=

Ay3 9.217 103−

×=

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

My4

Tie-columns

Ly4 437.:= cm

Acy4 1 0.25 k ⋅+( ) My4⋅Ho

Ly4

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy4 0.15= cm2


Bond-beams

A by4

My4

1000⋅

βs

:= A by4 0.252= cm2



Critic zones

Spacing se 7.5= cm


Ay4 0.5My4 se⋅ 1000⋅

15 βs⋅⋅:=

Ay4 0.063= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

cm2

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A20

My5

Tie-columns

Ly5 412.:= cm

Acy5 1 0.25 k ⋅+( ) My5⋅Ho

Ly5

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy5 0.144= cm2


Bond-beams

A by5

My5 1000⋅

βs

:= A by5 0.228= cm2



Critic zones

Spacing se 7.5= cm


Ay5 0.5My5 se⋅ 1000⋅

15 βs⋅⋅:=

Ay5 0.057= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

My6

Tie-columns

Ly6 437.:= cm

Acy6 1 0.25 k ⋅+( ) My6⋅Ho

Ly6

⎛

⎝

⎞

⎠

⋅1

βs

⎛

⎝

⎞⋅ 100⋅:=

Acy6 0.191= cm2


Bond-beams

A by6

My6 1000⋅

βs

:= A by6 0.321= cm2

Minimum reinforcement adopted: 4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A21


Critic zones

Spacing se 7.5= cm


Ay6 0.5My6 se⋅ 1000⋅

15 βs⋅⋅:=

Ay6 0.08= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

My7

Tie-columns

Ly7 17:= cm

Acy7 1 0.25 k ⋅+( ) My7⋅Ho

Ly7

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy7 0.075= cm2


Bond-beams

A by7

My7

1000⋅

βs

:= A by7 0.051= cm2



Critic zones

Spacing sec 10:= cm

Section of the reinforcement A

e

Ay7 0.5My7 sec⋅ 1000⋅

15 βs⋅⋅:=

Ay7 0.017= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A22

My8

Tie-columns

Ly8 412.:= cm

Acy8 1 0.25 k ⋅+( ) My8⋅Ho

Ly8

⎛

⎝

⎞

⎠⋅

1

βs

⎛

⎝

⎞⋅ 100⋅:= Acy8 0.18= cm2


Bond-beams

A by8

My8 1000⋅

βs

:= A by8 0.285= cm2



Critic zones

Spacing se 7.5= cm


Ay8 0.5My8 se⋅ 1000⋅

15 βs⋅⋅:=

Ay8 0.071= cm2

Normal zone

sen 0.15m= ds 3= ds 3= mm → φ 4= m

4φ6mm 1.13cm2

=φ6mm

4φ6mm 1.13cm2

=φ6mm



Annex

A23

Figure annex III Reinforcement of the confined masonry panel that results from the calculation



Annex

A24

Figure annex IV Reinforcement intersection of bond-beams



Annex

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Dissertation2007-Asinari