Missouri University of Science and Technology Missouri University of Science and Technology

Scholars' Mine Scholars' Mine

International Specialty Conference on Cold-Formed Steel Structures

(1971) - 1st International Specialty Conference on Cold-Formed Steel Structures

Aug 20th, 12:00 AM

Design and Testing Composite Open Web Steel Joists Design and Testing Composite Open Web Steel Joists

J. A. Cran

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Recommended Citation Recommended Citation Cran, J. A., "Design and Testing Composite Open Web Steel Joists" (1971). International Specialty Conference on Cold-Formed Steel Structures. 3. https://scholarsmine.mst.edu/isccss/1iccfss/1iccfss-session6/3

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DESIGN AND TESTING

COMPOSITE OPEN WEB STEEL JOISTS

BY

J. A. CRANl, P. ENG.

Introduction

Traditionally, most buildings ·have been constructed in a

utilitarian manner. The consideration of low first cost far out­

weighed the considerations of maintenance and future renovation

costs. In the past decade, rapid technological change has spurred

a dramatic shift in design emphasis. In order to avoid early

obsolescence, designers are now concentrating on building flexi­

b1li ty.

For example, owners of schools and modern office buildings now

demand the utmost in flexibility of interior design. Large open

areas; long, column-free spans; moveable partitions; the ability to

move heating, ventilating and air conditioning ducts; and ease in

accomodating new electrical and communication requirements are man­

datory. A most economical method of providing long clear spans, a

high degree of services, and built-in flexibility of space utiliza­

tion is the composite open web steel joist.

Advantages

The main advantages offered by composite open web steel joists

(OWSJ) are as follows:

1. Economical long spans. Fig. 1 compares the cost per square

foot of installed floor for various spans of composite and non­

composite OWSJ. Fig. 1 is based on the following parameters:

a) Joist depth - 32. II

b) Joist spacing - 5'0"

c) Steel floor - lis" deep, ZZ ga., 6"c-c ribs

d) Welded wire fabric - 6 X 6 10/10

e) Concrete slab - 4" total thickness

f) Stud shear connectors (composite joists only)

g) Column and girder costs are not included.

It can be seen that for spans of 37' and greater, composite open

web steel joists offer cost savings. Below 40', the savings in OWSJ

steel weight do not fully compensate for the added expense of shear

connectors. Clear spans of 40' are commonplace in modern office

towers while schools designed to an open plan require spans of

50-65'. Therefore, composite joists offer significant savings when

used in schools and office towers.

Z. Increased Natural Freguency. The natural frequency and amplitude

of long span floor systems sometimes subject a building's occupants

to objectionable vibrations. Composite joists have vibration charac­

teristics superior to those on non-composite joists.

3. Increased Stiffness. The concrete-steel interaction results in

composite joists having greater stiffness than non-composite joists.

In practical terms, composite joists can be shallower than non-composite

joists, or, for a given joist depth, live load deflections are sig­

nificantly less when composite joists are used.

Shear Connection

As indicated when economy was discussed, the prime difference

between a composite and nan-composite joist is the presence of a

*The Steel Company of Canada, Limited

Sales Engineering Manager, The Steel Company of Canada, Limited, Hamilton, Ontario, Canada

shear connection. For both types of joists, the bottom chord and

web systems are identical. However, because the presence of a shear

connection in a composite joist allows the concrete floor slab to

act with the steel joist, the top chord of a composite joist oan be

considerably 1 ighter than that of a non-composite joist.

Because the material cost savings in the top chord must be

sufficient to offset the added cost of a shear connection, it is

desirable to have an economical connection. To this end, Stelco*

conducted a series of push-out tests to evaluate various types of

shear connectors .1

Five .types of push-out spec.imens as shown in Figs. 2-4 were

prepared. Two samples of each type were tested and the laboratory

results averaged. The resulting load versus slip curves are

shown in Figs. 5 and 6 while a summary of the push-out results

is given in Table I.

It will be noted that all push-out specimens consisted of two

pieces of cold formed steel OWSJ chord sections welded together.

The relatively thin chord sections (0.110") were an important

aspect of the test program. Goble2 conducted a series of tests

on thin flange push-out specimens with welded stud shear connectors

as shown in Fig. 7. It was found that there is a limiting thick­

ness below which the failure mode shifts from stud shear to flange

pull out. For Is" ~ studs, the limiting thickness is about 0.18 in.

However, even when the steel thickness is only 0.110", about 75%

of the maximum strength of a a;j, stud can be developed as shown in

Fig. 7.

In light of this information, and the fact that cold formed

chords in joists spanning the economic limit (40') or greater are

at least O.liO" thick, O.liO" was chosen as the chord thickness

for the tests •

Another noteworthy detail of the push-out specimen is the

presence of two open cells in the concrete slab. These cells left

a concrete "rib" which contained the shear connectors. Therefore,

the push-out test specimens represent a building having open web

steel joists, cellular steel floor and shear connectors welded

through the steel floor to the top chord of the joist. The concrete

in a rib adds appreciably to the strength of a shear connector and

hence, when discussing shear strength, the term "connection" is

usod. The strength of a "connection" is the combined strength of

the concrete rib and the mechanical shear connector it contains.

From the load-slip curves, the maximum load sustained by a

push-out specimen and its elastic modulus is readily obtained.

These values for each of the 5 specimen types are given in Table 1.

Push-out specimen Type I consisted of two connections and

hence the specimen values are halved to obtain the maximum strength

per connection (qu) and the connection elastic modulus (K).

Specimen types 2,3, and 4 consisted also of two connections;

however, each connection contained two mechanical connectors. For

design purposes, it 1s convenient to define a connection as one

concrete rfb containing one mechanical connector. Therefore, for

these speci~~~ens, the maximum test load and the specimen elastic 188 0

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modulus is divided by four to obtain qu and "K". Since "qu" and

"K" determined in this manner are less than the values which would

be obtained if each rib in the test specimen had only one mechanical

fastener, this approach is conservative.

From Figs. 5 and 6, it can be seen that the load-slip behaviour

of the puddle weld and the shear strap is not ideal. Both types

of connection tend to reach their max·;mum load value and then undergo

a rapid drop off in load. This type of behaviour is undesirable and

these types of connectors need further development.

The load-slip behaviour of the stud, hat and washer with plug

weld, however, are fully satisfactory. These three types of connectqrs

all exhibit a good amount of ductility and are able to sustain

load through large imposed deformations.

Composite OWSJ Tests

One of the objects of this paper is to analyze the results

obtained from three full size open web steel joist tests. Each test

joist was of a different type and they are designated as the

Stelco Tower Joist, the CSICC Joist and the Hambro D-500 Joist.

Construction details for the three test joists are given in Figs. 8,

9, and 10 while the test arrangements are given in Figs. ll and 12.

Stelco Tower Composite Joist

The joist shown in Fig. 8 is currently being fabricated for

the 22 storey Ste leo Tower in Hamilton, Ontario. The avera ll

composite joist depth is 30" and the typical span is 40'. All steel

members in the joist are cold-formed from sheet steel and have a

minimum yield strength of 55,000psi after cold-forming. The chords

are cold formed hat sections while the webs are cold-formed open

seam tubes which are stitch welded.

It can be seen that a 2'-0" wide opening is provided at mid­

span for an air conditioning duct. Chord reinforcements were added

at this location.

Two types of shear connector were used on the test joist as

shown in Fig. 8 - 2" deep hat connectors (see Fig. 3) and \"0

studs. All connectors were directly welded through the l\" cellular

steel floor to the underlying steel chord which was 0.153 in. thick.

The total thickness of concrete slab was 4".

As shown in Fig. 11, two point loading was used during the

test. Therefore, the bending moment is constant over the lD' long

section between the point loads. The critical section under live

load bending moment is just outs ide the ends of the chord reinforce-

ments and is shown as 11A-A 11 on Fig. 11.

CSICC* Joist

This joist test was the first in a series of 5 now being conducted

for the CSICC by Dr. Hugh Robinson at McMaster University in Hamilton.

The joist members were double angles and bars of CSA Standards G40-12

steel (Fy- 44,000psi. minimum) and details are given in Fig. 9. The

test construction incorporated l\" cellular steel floor running

continuously over the top chord of the joist. Stud shear connectors,

3/4"0, were directly welded through the steel floor to the top

chord of the joist. The test joist spanned 50', was 36" in overall

depth, and meets all of Metropolitan Toronto's SEF** schools' design

criteria.

Two point loading, as shown in Fig. ll, was used in the laboratory

test.

Hambro D- 500 Joist***

The D-500 composite joist is int@nded primarily for apartment

buildings. Its extreme shallowness makes it ideal for this appli­

cation. As shown in Fig. 10, the overall joist depth for a 20'-4"

span is only 12'>".

The test joist consisted of a cold-formed steel top chord

having 39,200 psi minimum yield strength after cold forming. The

yield strength of the two rods (5/8"~ and 3/4"~) in the bottom

chord was 45,000 psi average.

The vertical leg of the top chord has slots punched into it

which permits the use of a unique slab forming system. Roll bars,

inserted into the slots, support the plywood formwork and provide

-6-

*Canadian Steel Industries Construction Council

**Study of Educational Facilities

***Patented

lateral support to the top chord of the joist until the concrete

hardens. Prior to concrete hardening, the allowable load the top

chord may carry is dictated by local buckling. The unsupported

vertical leg, with a w/t ratio of 23, governs and results in an

allowable stress3 of 15,300 psi or an allowable force in the top

chord of 8.25 kips.

After the concrete has hardened, the top chord is firmly

embedded in the concrete and serves as the connecting structural

link between the concrete slab and the steel joist.

As shown in Fig. 12, a 3-joist assembly was loaded with con­

crete blocks to evaluate its composite behaviour.

Test Results

The test results for the Stelco Tower Joist, the CSICC test

Joist and the D-500 Joist are reported in References 4,5, and 6.

Relevant data from these references in given in Figs. 13-19.

Load-Deflection Curves

Load-Deflection curves are given in Figs. 13, 14, and 15. For

all three joists, the load is taken as the externally applied

live load and does not include the dead load of the steel joist,

steel floor and concrete slab. Fig. 11 defines "W" for The Stelco

Tower and CSICC Test Joists, while as shown in Fig. 12, the live

load on the Hambro joist was applied by the use of concrete blocks.

In Figs. 13-15, it can be seen that the Load-Deflection curve

follows an initial stra~ght line. At some load, the composite

interaction between the steel and the concrete begins to break down

and the curve starts to deviate from the straight line. The curve

continues to rise and eventually levels out as the joist undergoes

plastic deformation. All three joists exhibited this same plasticity

as evidenced by the large deflections which occured during the tests.

Both the Stelco Tower and CSICC joists were loaded to failure.

Collapse in both cases was due to a web member buckling. The web

failures were not due to any inherent weaknesses in either web

system; rather, it is the author's opinion that the large deflec­

tions near failure gave rise to high secondary stresses which

eventually combine with the compressive axial load to cause web

buckling.

189

The fact that the Load-Deflection curves levelled out prior

to failure indicates that the ultimate bending strength of the

Stelco Tower and CSICC test joists was realized prior to web

buckling. Although the Hambro joist was not loaded to the point

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of collapse, once again, the levelling out of the Load-Deflection

curve would indicate that the maxi rum 1 ive load applied during

the test is close to the ultimate live load the test assembly

could have supported.

A theoretical r.oad-Oeflection curve is also plotted on Figs.

13-15. In all cases, complete interaction between the concrete

and steel and elastic behaviour is assumed. For the Stelco Tower

and CSICC test joists, the theoretical deflections shown were

computed by the method of virtual work and considering the joists

as trusses. The concrete slab area was converted to an equivalent

steel area and this area was taken as that of the top chord. (The

area of the steel portion of the top chord was neglected for reasons

to be explained later.) The theoretical deflections shown for the 5Wl3

Hambro joist were calculated from bending theory ('3!!'4rl) with a

10% increase to account for shear deflection. Once again, the area

of the steel top chord was neglected.

In all deflection calculations, the full width of the concrete

slab was used.

A comparison of the theoretical and actual deflections are

given in Table II. No discernable pattern is evident which would

indicate that the theoretical methods applied (virtual work and

bending theory) are, strictly speaking, applicable to a composite

joist. It is felt that more accurate theoretical results can

be obtai ned if composite beam theory which accounts for s 1 i p

between the concrete and steel (partial interaction) were applied.

In any event, it appears that the normal method of calculating

joist deflections - bending theory with a 10% shear deflection

allowance- gives a reasonable estimation of deflection.

As exnlained in Reference 7, the presence of open cells causes

increased slip between the concrete and steel. Thus, when cellu-

lar steel floor is employed (as in the case of the Stelco Tower

and CSICC test joists), it is suggested that the deflections cal-

cu l a ted by bend i 09 theory with a shear a 11 owance, be further in-

creased 10-20. to account for slip.

Measured St.!'_~

l·leasures strains for the Stelco Tower and CSICC test joists

are shown in figs. 16-19. In Figs. 16 and 18, strain blocks are

shown for two conditions a) when the live load uwll on the

composite joist results in a bending moment aporoximately equal

to the working bending moment, and b) the last recorded strains

prior to joist failure. In Figs. 17 and 19, Load -Strain curves

are oresented for the top chord of each joist.

The yield strain for the bottom chords of the Stelco Tower

and CSJCC joists are 1893 and 1776 in./in. x 10-6 respectively.

From Fiqs. 16 a) and 18 a) it can be seen that the measured strains

when the joists were under working bending moment were well below

the yield strains of the bottom chords. The loads associated K

with working moment were about 13.64 for the Stelco Tower Joist

and l0.95K for the CSICC joist. From Figs. 13 and 14, it can be seen

that under working conditions, both joists are on the straight line

portion of the Load - Deflection curve and are behaving elastically.

From Figs. 16 b) and 18 b), it can be seen that prior to

joist collapse, the bottom chord strain of both the Stelco Tower

and CSICC test joists was well beyond the yield strain of the

steel. The presence of these yield strains is a further indication

that the ultimate bending strength of both joists was achieved

prior to joist collapse.

In Figs. 17 and 19, it can be seen that the dead load of the

steel floor and concrete slab, as well as shrinkage, places com­

pressive strains of -394 and -484 in./in. x 10-6 in the top chords

of the Stelco Tower and CSICC test joists respectively. As the

live load "W" was applied to the test joists, contrary to what

might be expected, the top chord compressive strain diminished. I~

fact, the Stelco Tower top chord went into tension and the CSICC

top chord approached zero stress. This behaviour indicates that the

top chord is actually working in conjunction with the bottom

chord to resist the compressive force in the concrete slab of the

composite joist. In practice, the short lever a1"11! between the

top·chord and the concrete slab, as well as the relatively low

level of strain in the top chord, means that once the concrete

slab has hardened, the top steel chord of a composite joist can

be neglected.

Ultimate Strength Calculation

The ultimate bending strength calculations for the Stelco Tower,

CSICC and Hambro composite joists are given in Figs. 20, 21, and 22.

Basically, the calculations hinge around the strength of the connec-

tion between the concrete and steel. The compressive force in the

concrete cannot exceed the crushing strength of the concrete, the

tensile force in the steel cannot exceed the yield strength of the

steel, and neither the compressive force nor the tensile force can

exceed the ultimate strength of the concrete-steel connection.

For example, let's consider the Stelco Tower joist. Referring to

Figs. 8 and 20, the maximum tensile force the bottom chord of the

joist can develop is 130K. From Table I, the ultimate strength

of a hat connection is 4.75K, and the ultimate strength of a 1/2"

diameter stud connection is 5.5K. On one side of the joist there

were 24 hats between the point of zero and maximum moment, and on

the other side there were 21 studs. Therefore, the ultimate strength

of the concrete-steel connection is ll4K on the side containing the

hats and ll5K on the side containing the studs. Consequently, the

force in the steel cannot reach its yield force but is 1 imited to

the strength of the steel-concrete connection- ll4K. The force in

the concrete is equal and opposite to that in the steel. The couple

created by the steel force and concrete force is, of course, the

theoretical ultimate strength of the joist, in this case, 3276 in. K.

A similar procedure was used for the CSICC and Hambro joists. In

the case of the CSICC joist, the ultimate strength of a 3/4" diameter

stud was taken as 14.3K from Reference 8. For the Hambro joist, the

steel-concrete connection was assumed fully effective because the

the top chord is securely embedded in the concrete slab and the

concrete slab does not contain open cells.

From Figs. 20, 21 and 22, it can be seen that the maximum error between the theoretical ultimate strength calculated in the above manner and the ultimate strength for the two joists, which were loaded to failure, was 1-l/2%, and for the Hambro joist, which was not loaded to failure, was 7-1/2%. It should be noted that the theoretical moment in all cases is less than the maximum applied moment.

Composite OWSJ Design

The author's design procedure for composite open web steel

joists is given in Appendix "A" for the Stelco Tower joist.

As explained under the section on"Measured Strains", the top

chord can be neglected under live load conditions. Therefore, 192

the top chord is selected to satisfy only construction loads. In

the case of the Stelco Tower, it was decided to apply the Construction

Safety Act9 and a safety factor of 2 was used.

The selection of a bottom chord and the number of shear connectors

is dictated by the combination of live load and dead load. The

resulting live plus dead load moment is increased by a load factor

for ultimate strength design purposes. In this case, a load factor

of 1.7 as outlined in the Plastic Design section of CSA Standard Sl61°

was used. The theoretical ultimate moment is calculated and since

it exceeds the factored live and dead load moment, the bottom chord

and number of shear connectors selected are satisfactory.

The design procedure for the joist webs is identical to that used

I for any open web steel joist. Live load plus dead load results in the

maximum web forces. In this case, CSA Standard Sl36ll was followed

for the web design.

When calculating the live load defliktion, the procedure outlined

in the section "Load-Deflection Curves" is followed. The full width

of the concrete slab is used and the top chord of the steel joist is

ignored when calculating the moment of inertia. The deflection

calculated from bending theory 1s then increased by lOS to account

for shear deflection. In this case, a 201 allowance was also made

for the effects of slip.

The joist 1s cambered 1.25 in. to account for dead load deflec­

tion and one-quarter of the anticipated live load deflection.

TABLE II - DEFLECTION COMPARISONS

JOIST l" _.L ,. (Bending Error1 Error 1

(Actual) (Virtual Work) ~Theory x 1.10) (Virtual (Bending 7 ,;:. Work) Theory x 1.10)

S TELCO TOWER 73.4 x 1o·3 59.5 X 10"3 55.5 X 10"3 -19% -24%

CSICC 60.0 X 10"3 63,5 X 10·3 53,0 X l0"3 + 6% -12%

? + 2% HAMBRO D-500 0.44 X 10":

0,45 X 10"2 -

o"/w psf -(Exterior) o"/w psf

1 . Error : Theoret i ca 1 - Ac tua 1 Actual

2;!)

-ZDV I I

I I I

I I /,

I I/ y

I 5o ~I ,..... lh u: /1.~ vi

!>: I ,l ......... IOO ~~

0 1/''li

< . It <;) ...J I' I.U

5o / :::. ::,

/ ~

/ ____ __l ____ ....~. ___ _j__ ____ J. __ ___J

0.5 /.0 /.S 'Z·P '2..5 3.0

b - MID- SPAN PEFU.c.TION ( 1111.)

FIG. 15 LoA!> Pf.~'LE.cT!~.'-1_ -~y-~v_r; ..=-.·.::..:.-~::.··:·.··~ .. ,;,.~:;~.-.;...;...:

DSoo JOIST

193

i I

., . : : rr.: f•t 'i"" .

. :.-,-.. : fl I r

···------- --

~.

H+I-I+H+i · . i -:t

! so r- v

t-loo -t-~~o .

S'ltfAIN .(11J.j11.f. •w;

t.o" --=--"--~·1 '

•. ..,. . 1-.·-:--:-.,.-,-:-:-•• --___,--.:'~~1 - -'~· '""''-··--...:·r·:..··-·n ~,~~-·:..·.:..· ·~·-.:.:..=.•.:::•·..;..~·' -·-~-----.:;·.;,::.. ··-.r-'1 .J.Jr.:~C'""'

a~• ~ '.·. :.:-~f-.·: -.i: I ·.:~.l~~~-1 -1 ---! ., ! . i---L-

I - -I - I ____ J._- .

... ---~-

: --1 -~~----;---I -1 -1 I ! I :

t* 1 " l ' I · =~;gt.:~ ... ~~:i.--'-1~::;: ::;t;::I/.IJJ,..;~;..._u...,_-;-1 ---l .. -~: ... ; __ ; -- i ~

- f I , - i ...... L. ... : I I

__ M_,J- 0 .: OOo-oOO

~-~­. 1--

' ' ,--~ ---1 ---- -

,.,....,... __ : __ .J...­___ 1_

=-~:,, , 1

'".' ', r- r-~ 1 .. ),, ,' I '\i . . . . '!'•, ~ t;;t; • • . ~ . 'I -1--. _, . ~ -t--,ii;-fi;-"T'+ -' . 1 9f _.__ -·I____ I TA '-±1-- 1-- --1--= ..,-:: -=:.=-k=-=------: ' "-"--- -'-+--+--1 I---·-··-!I __ _I I -~------ I -- --+--1---j--. --- --~ -+-+---------1 .

I ~-~ c-+-+--t--- -r-- ··1-_t--+--_-___ - __ -_ ·----~+--+-+--!--- -1---1-----

1--~-+---+---+l_~l·_+_---r--r--11:---1 -~ . _j___ ~f-' '"- r. . fi -I I I I· ; -r- -- ,_,. __

I. i •

!-~ i - I

l ~

I_ .. , ___ ----------'---:·-----!~-·- _,, _______ -· 1

--:J ;-,.q;··-fj-.. -~-.6-,-,.-1-3.~/=---I.&J-.~.,.-: -,-;.«.....;_:. -+-1/J._x ___ u, __ -78-.. -_-+-.3-~!JO-H 1 .:._; I) = A.s: 1;-~i.2.8'J~ s~;2.; l;,/1(· __ -t--~ --·-! ~-- . - - )i 0 .: ·· .......... m•e,. ------- .. -· '---~ .. 1. ~-- --~ ·--· --1-- ... _J1:_=-f/'7M(3S:U.-~.,I , rc ~ llfv 1111+ "''l+•+.75t~ /If.~ ·--~--:::..1/I-~""J,oo ::__-. 1/~ : --R _-6h fv 4--j/.3..,~~./I.S~I"~-~~.' . _ · _ .. .. ... n < .(I' ~: - ?' - - "~ · -· --~--'- · . -r--~ .. 5~75 .. tl/.1<. . fC -111., 1~,4 .S 2/IIS.S IS.~. _____ , __ if/.1(. 32 • ;.·.--~-: 176.1': ---~- '1--....,±,...,...-.....,,,~)=----,.......,..,

p I( - - . . - - ---+ ' ---.:.. -- ; -- -~ ~,;_____,. .... ,,0, . ,, :: 1/.f ., ------ «. -"'- fc.. : ··- /JI.,.L ____ j_'l!.().'/2111· --.-~:----i.-

L. _ 11o.J.. (.oo____ Ni&,'i> .. ll'ljtS•11S'.cl. .... • _ II~.., _f.,,/w.1

iiC ~.1' ~ 11u .. r..-;,. ,._: t.73o --- ---j __ , ___ 1-- __ , ___ __j ___ ·Hr." -+-.,.--s-~~o ,,./c..

-moi/1-l<..- 1 _ ~, ~ ~~f __ :~~~1~~t--r=-- -~ .. ~~~·!- ---"-. .... l ... - I 1 ,4t).,,.,,~ -~ • 5'~~0 : /,{)/

:==::.

. .::; .. ·- ~~/

//4- .. ~4

ZQ.71 ill .S'.S: 2 K-S /.

111'.;/vre .. ~ = lf,._TifU~. 127'

:-----!- u'. __ --:__,~,7lf I r~~.

-· -------~r::/::b::.::z::~:===l1~~==5~rc=t~co~Tt.~tl~~~t=R.~.J~e~ts~r~- J IL-___ :__~·:::r.::"/"=· .=2='====/'t=!W= .. =c .s=~=c=c==_J=~~:s~-~-' --· -------- --------------------,

i.s" s(M.J.18 +-.tJ.306S)=-us;J.

IJ = rl.; fy = 2.2t:'>~..f$:~-=iQ/.$'

Fill. 22 Jct!.T

Co!IC!Mslons

C-slte open ""b steel joists offer Nny odvontages In building

construction. Further, they ore highly eco,...ltll for spons over

40 ft. In length.

The results of three full scole c-slte joist tests IODUld Indi­

cate that suc:h Joists can rudlly be designed by the ultiNte strength

•tllod. A design procedure ...,Joying this Mtllod Ills been given.

~

Stud Connectors and Metal Deck", October 1967, McMaster University,

Hamil ton, Ontario.

9. Construction Safety Act 1961-62 and Ontario Regulation 269-69,

Province of Ontario, Queen's Park, Ontario.

10. CSA Standard 516-1969, "Steel Structures for Buildings",

Canadian Standards Association, Rexdale 603, Ontario.

11. CSA Standard 5136-1963, "Design of light Gauge Steel

Structural Members", Canadian Standards Association,

Rexdale 603, Ontario.

APPENDIX A

DESIGN - STELCO TOWER JOIST

Span • 40'0 Fy • 55 K.S.I. fc' • 3.0 K.S.I. 5'c-c

6max. L.L." 1"

D.L. • 45 psf(Concrete, steel floor, OWSJ)

D.L. • 10 psf(Ceiling, mech. & elec.)

L.L. • !!§. psf (Incl. partitions)

TOTAL 140 psf

Construction - 25 psf L.L. • 45 psf O.L. • 70 psf

f10.l • ~10_ X 5 X 40) 40x 12 • - 1000 8

M • L. L.

TOTAl M

660 in.K.

1020 in. K.

1680 in. K.

Canst. • 840 in. K.

4~ 2f _______ __J o. 37 , I ..,• .• • • • .: , 'l:l

J. -- • .i " • ' ; rf. 1.492 ) - -----=t = ~.40)Top Chord ! • o.26) I

~I ~

.L Pc

t-~

I I

--+· Py( -=.:c- • J A • 2.06 ) r x-x 1. 02) sxx 1.14)Bottom

Chord 1.00

Construction

Top Chord Axial Force = 840 • 34.0K fa= 34.0 • 22.8 K.S.I. 2'm ~

Panel Spacing • 24" . ·. M panels : 0

Sl36

0.9 (24) - 54 --o.4lf'""

Fa • QF (1-0. 0035 _ISh_) r

Q "1 .0

_ISh_ • 0 ry-y

Construction Safety Act S.F. • 2.0 .·. F • 55 • 27.5 K.S.I. 2.0

Fa • x 27.5 (1 - 0.0035 x 54) • 22.3 K.S.I. O'K

D.l. ' L .L.

Bottom Chord & Connectors

M D.L. + L.L. • 1680 in. K.

lli L.F. • 1.70 Plastic Design

• • M 1680 X 1.7 = 2860 in.K • failure reqd.

Py" 55. x 2.06 • 113K

nqu hat • 114K nqu stud •115K (see Fig. 20)

Pc • 113K

a • 113 • 0.74" 0.85 X J X 60

30 - 1.00 - a/2 • 28.63 in.

M • 28.63 x 113" 3240 in.K. D'K u

0.55~~~ ll 91K

t 19" 12" 24" -.

13.70K ~ c

·..v.

13.1

R • 0.70 x 19.585 • 13.70K

0.70 00-0.70 (19 + 12)

"'2'XT2.

P • 17.4K A • 0.69 (See Fig. 8)

fa· 17.4 • 25.2 K.S.I. ~

Fa • o.6 x 55 • 33.0 K.S.I. O'K

£.:.!!__

P • 14 .OK A • 0.61 (See Fig. 8}

fa • 14.0 • 23.0 K.S.I. Q.bT

KL • 24.9 • 43 ;:- 0-:-ID"

ill§_

d

~,

"'·

_l

Fa • QF (1-.0035 _ISh_) • 1.0 x 55 • (1-0.0035 (43))• 28.3 K.S.I. O'K r 1.65

etc.

I L.L.

~lection w • 85 psf LL

I I • 1440 in. Comp.

°Centre • [ 5(85 x 5 x 40 ) 403 x 123 ll 384 X 30 X 1 06 X l 440

1.10 X 1.20

• 0.76•1.0 O'K

D.L. Deflection W • 55 psf

DL

I Steel Joist • 526

0 5(55x5x4Ql403 xl23 Centre • 1 384 x 30 x 1 0 x 526

1.10

LL

• 1.12 in.

Camber • D.L. + '< L.L. • 1.12 + '< (0.76)

• 1.31 in. say 1.25 in.

1116

l.

) . I

CSICC JOIST UNDER TEST AT MCMASTER UNIVERSITY, HAMILTON

2.

CSICC JOIST AFTER FAILURE. NOTE BUCKLING OF THE DOUBLE ANGLE COMPRESSION WEB.

3.

STELCO TOWER JOIST OUR! NG TEST AT MCMASTER UNIVERSITY, HAMIL TON.

4.

STELCO TOWER JOIST AFTER FAILURE. NOTE LARGE DEFLECTION CAPACITY AND BUCKLED COMPRESSION WEB MEMBERS .

197

5.

OTTAWA, ONTARIO, APARTMENT BUILDING INCORPORATING HAMBRO D- 500 COMPOSITE JOISTS. SPECIAL DETAILS HAVE BEEN DEVELOPED FOR THE BALCONY FLOORS.

6.

HAMBRO JOIST SHOWING COLD FORMED TOP CHORD CONTAINING "ROLL BAR" SLOTS.

7.

1

APARTMENT FLOOR DURING CONSTRUCTION SHOWING HAMBRO JOIST, TEMPORARY "ROLL BARS" AND PL YWOOO FORMING.

I