DCB 68 Sliding Hinge Joint

HERA Steel Design & Construction Bulletin Page 1 No. 68, June/July 2002

No. 68The author(s) of each article in this publication are noted at thebeginning of the article.

June/July 2002The procedure detailed herein has been the subject ofreview by a number of people. The effort and input of thesereviewers is greatly appreciated.

Introduction

As readers will be aware, HERA and the Universityof Auckland are engaged in a long-term researchproject aimed at developing new forms of semi-rigid joints for moment-resisting, steel framedseismic-resisting systems (MRSFs). Two jointtypes have been developed from this programmeas the preferred options for the beam to columnconnections of MRSFs. These are the FlangeBolted Joint (FBJ) and the Sliding Hinge Joint(SHJ).

The experimental and analytical phases of thisproject are now completed and the final phase(writing up and presenting of results) has begun.

The FBJ was the first joint to be developed.Design and detailing procedures for it have alreadybeen published in DCB No. 58, principally, andwith a minor corrigenda in DCB No. 62 and anextension to its original scope of application inDCB No. 64. This joint has been used in at leasttwo building developments (one in Auckland andone in Napier), which was the intention behind thedesign and detailing requirements being publishedprior to the release of the full research report [1].

With the completion of the analytical work on theSHJ, the research has now reached the stagewhere final design and detailing provisions for theSHJ can be made. This issue presents theserecommendations, covering the design anddetailing of the joint itself and the design ofmoment-resisting steel framed systemincorporating the joint. It also presents a detaileddesign example on a particular SHJ.

Also covered is a short article on a specific designissue that has arisen in recent times.

In This Issue

Page

The Sliding Hinge Joint

1

Member Compression Capacity of aSolid Section

33

References

33

The Sliding Hinge Joint: Designand Detailing Provisions andDesign Example

This article has been written by G Charles Clifton, HERAStructural Engineer, John Butterworth, Senior Lecturer at theUniversity of Auckland Department of Civil and ResourceEngineering and Tanja Miller, Undergraduate Student from theFachhochschule Weingarten on Study Leave (IndustrialPractice) at HERA.

1. Introduction and Scope of Article

1.1 Brief history of the overall project

HERA and the University of Auckland are in thefinal stages of a long-term research project aimedat developing innovative new forms of semi-rigidjoints for moment-resisting steel framed seismic-resisting systems (MRSFs).

These joints are designed and detailed to achievethe following performance characteristics:

Remain fully rigid up to the design levelserviceability limit state earthquake moment

Remain reasonably rigid above theserviceability limit state level and up to thedesign level ultimate limit state earthquakemoment

Allow inelastic rotation between beam andcolumn to occur when the design ultimate limitstate earthquake moment is exceeded

17-19 Gladding PlaceP O Box 76 134Manukau City,New Zealand

Phone: +64-9-262 2885Fax: +64-9-262 2856Email: [email protected]


Be able to withstand the inelastic rotationdemand associated with the design levelearthquake with negligible damage, such thatthe post-design earthquake building responseunder serviceability conditions is notsignificantly affected

Withstand greater levels of rotation demandwith increased damage but not failure.

Of the five joint types that have been researchedfor this project, two joint details have emerged aspreferred options for the beam to columnconnections of MRSFs. These are the FlangeBolted Joint (FBJ) and the Sliding Hinge Joint(SHJ).

These two joints are designed and detailed to meetthe performance criteria in different ways. Verybriefly:

The FBJ is designed for higher strength, lowdesign ductility demand applications. It is verysimple to fabricate and erect. It has a lowinelastic rotation damage threshold, but iscapable of withstanding high levels of inelasticrotation demand if necessary.

The SHJ is designed for lower strength, highdesign ductility demand applications. It isslightly less simple than the FBJ to fabricateand more complex to erect and is designed towithstand fully ductile levels of design inelasticrotation with minimum damage.

The FBJ development was completed in 2001.Guidance on design and detailing of the FBJ andMRSFs incorporating the FBJ has been given inDCB No. 58, with a minor corrigenda in DCB No.62. In the latter half of 2001, it became apparentfrom the numerical integration time history (NITH)analyses that the originally proposed scope ofapplication of the FBJ, which was for low ductilitydemand applications only, could be widened, andwork on this was undertaken, with the resultspublished in DCB No. 64.

Up to the end of 2001, all NITH studies wereundertaken in accordance with NZS 4203:1992 [2].However, the March 2002 version of the draftreplacement to that standard, which has beenunder development for several years, containeddetailed guidance on the selection and scaling ofearthquake records for NITH. The selection andscaling of earthquake records used up to the endof 2001 was rather ad-hoc (see details in section6.4 of HERA Report R4-88 [3] and summarydetails in section 3.4, pp. 16-17 of DCB No. 64)and so, in 2002, the opportunity has been taken touse the provisions of DR1170.4 [4] to produce arevised suite of earthquake records and scalefactors. Details of these will be summarised inDCB No. 69. The FBJ designs were thenreanalysed under this new suite of earthquake

records. Given that the records were selected andscaled in accordance with the new draft, [4], whilethe FBJ frames had been designed to the existingstandard [2], some comparative 5 and 10 storeyframes were redesigned to the new standard tosee the differences in seismic design actions, P - Deffects and subsequent member sizes. Themember sizes for a given application turned out tobe the same from both standards for each casestudied.

For the SHJ NITH studies, the frames were alldesigned and analysed to the draft provisions.Because the suite of earthquake records coverthree soil/fault conditions, the designs wereundertaken for these conditions. The threeconditions covered were:

(1) Class C shallow soil [4] with near faultaction

(2) Class C shallow soil [4] without nearfault action

(3) Class D soft soil [4] without near faultaction

Designs were undertaken for two seismic zones(Auckland, Wellington). The near fault actionoption is only applicable to Wellington.

The SHJ NITH studies were completed in June2002. With their completion, the design anddetailing provisions for the SHJ have beenfinalised and are presented herein.

Summary details of the NITH studies and theframe options will be given in DCB No. 69. Writingup of the entire project is also progressingconcurrently and is due for completion in the firstquarter of 2003 [1].

A summary paper [7] of the research into bothjoints and systems was presented at the 2001NZSEE Technical Conference.

1.2 Scope of This Article

This article presents the design and detailingprovisions for the SHJ and for MRSFs using theSHJ. The former is presented in section 3 and thelatter in section 4. This is followed with a SHJdesign example, in section 5.

However, prior to presenting these provisions, thisarticle looks briefly at the performance of SHJs insevere earthquakes, in terms of the designphilosophy, target performance requirements andbehaviour from experimental tests. These issueshave already been mentioned in DCB No. 59 andthat article will be cross-referenced as appropriate.They will also be covered in detail in the thesisreport [1] on the whole project.


Fig. 68.1Sliding Hinge Joint: Isometric and Exploded View


Fig. 68.2Layout and Notation for the

Sliding Hinge Joint

Fig. 68.3Lever Aims for Moment Capacity Determination


Before commencing with the performance inearthquakes, a description of the SHJ is in order.Fig. 68.1 shows an isometric and exploded viewof the joint. Fig. 68.2 shows an elevation with thelayout and notation, while Fig. 68.3 shows thelever arms for determination of the joint momentcapacity.

2. Performance of the Sliding Hinge Jointin Severe Earthquakes

2.1 Design philosophy and modes ofoperation

The design philosophy behind this joint has beento establish dependable behaviouralcharacteristics for the SHJ and for the MRSFsystem for the serviceability limit state conditionand for two levels of ultimate limit state conditions.These are described in section 2.3. The first levelof ULS condition is the design level ultimate limitstate earthquake, as stipulated by NZS 4203 [2] orDR 1170.4 [4] and the second is the more severemaximum considered event. All the experimentaland analytical work undertaken on the SHJ hasbeen planned and executed with this philosophyin mind.

The mode of operation of the SHJ is relativelysimple. The beam is pinned laterally at the topflange level, using nominal sized bolt holes andFBJ details. This keeps lateral movement in thefloor slab to 2-3 mm, thus minimising undesirablefloor slab participation and slab damage. Jointrotation is achieved through sliding at thebottom flange and the web bottom bolt level (seeFig. 68.1 for the location of these components andFig. 64.10, DCB No. 64, for an illustration of thismechanism).

The sliding details are shown in the isometric viewof Fig. 68.1. The sliding layers are between thebrass shims and plate (web plate, bottom bolts forbottom flange plate). The holes for the webbottom bolts in the web plate and for the bottomflange bolts in the bottom flange plate are slottedto allow this sliding to occur. The beam flange orweb and the associated cap plates all havenominal sized holes.

When the moment demand on the SHJ fromearthquake generates internal beam axial forceswhich exceed the sliding resistance availablethrough the bottom flange bolts and web bottombolts, the joint will slide, allowing beam rotation tooccur. As sliding occurs, the cap plate isanchored in position relative to the beam flange orweb by the bolts, allowing the cap plate to alsoslide relative to these surfaces. Once theimposed moment reduces, there comes a pointwhere the sliding stops and the joint becomesrigid again. This is illustrated in Fig. 59.28 of DCBNo. 59, which shows the joint rotation versus

moment from the large-scale test 4 and inFig. 68.4 herein, which shows the joint rotationversus moment from the large-scale test 3.

On rotation reversal, the joint unloads abruptly,then the moment capacity builds up in the reversedirection, as shown in Fig. 59.28 or 68.4. Theincrease in moment with increasing reverserotation occurs in two stages; one as slidingoccurs along the first interface (beam to plate) andthen with a further increase in shear capacity asthe second interface (plate to cap plate) isactivated.

The slotted hole is designed to accommodate ajoint rotation of 30 mrad (radians x 10-3)multiplied by an over rotation factor of 1.25; if theinelastic rotation demand exceeds this, the jointundergoes further inelastic behaviour throughflange plate yielding, in the same manner as forthe FBJ (see DCB Issue No. 58). The first large-scale SHJ specimen, tested to destruction in test2, still developed its design moment capacity atover 120 mrad rotation!

Under the design level ULS earthquake, inelasticrotation demand is expected to be not greaterthan the 37.5 mrad accommodated within theslotted holes. At this level of rotation demand,minimum joint degradation will occur and onlyminor slab cracking, such that no post-earthquakerepair is required.

Under the maximum considered event (MCE), theMRSF with SHJs will retain its integrity, to allowevacuation and post-earthquake assessment, butwill suffer controlled joint damage, which maynecessitate replacement of components.However, the results from the NITH studies showthat, in most instances, little or no reinstatementwould be needed after most maximum consideredevents, especially for buildings not subject to nearfault action.

In terms of the force based seismic designphilosophy of [2, 4], the design proceduresdeveloped for these semi-rigid systems utiliseeither the equivalent static or modal responsespectrum methods, in conjunction with NZS 3404[5] and, where appropriate, HERA Report R4-76[6]. The preliminary sizing / design method, inparticular, is easy and rapid to use. Theseprocedures are given in sections 3 and 4 below.

2.2 Design role of joint components

This is described in section 3.3.2 herein.

2.3 Performance characteristics

The MRSFs with SHJs have been developed todeliver the following performance characteristicsfor the three levels of earthquake described insection 2.1.


(1) For the serviceability limit stateearthquake (ie. as represented by DR1170.4 [4] Section 2.1.1, involving a returnperiod of 20 years for normal structures (asdefined by Table 3.1 of AS/NZS 1170.0[8])):

(i) The joint and system shall remaineffectively rigid, with negligibleinelastic action from any component

(ii) This condition shall apply even whenthe system has been subjected to aprior ultimate limit state design levelevent.

(2) For the design level ultimate limit stateearthquake (ie. as represented by [4, 8]involving a return period of 500 years fornormal structures):

(i) Negligible inelastic demand in thebeams

(ii) Minimal inelastic demand in thecolumns at base level (such thatfixed column bases will be readilyrepairable) and none at higher levels

(iii) The rotation demand on the joints isnot to cause the bottom flange boltsto contact the ends of the slottedholes

(iv) Column panel zone rotation demandto be 1%

(v) P - D effects to be accounted foreither through provision of suitableframe stiffness (ie. satisfyingEquation 6.1 (1) of [4]) or throughincreased strength (ie. satisfyingClause 6.5.4 of (4)).

(vi) Lateral drift not to exceed 2%

(vii) The positioner bolt may needreplacement

(viii) Minor cracking only to the concretefloor slab surrounding the frame.

(3) For the maximum consideredearthquake (ie. based on a 2000 yearreturn period event or higher):

(i) Negligible inelastic demand in thebeams, except in the vicinity of boltsto the flange and web plates

(ii) Inelastic demand in the columns tobe able to be dependably resisted(this applies especially at the base,

which is the only location likely to besubjected to appreciable inelasticdemand)

(iii) In the extreme case, joint rotationdemand may cause the bolts toimpact the ends of the slotted holes,requiring replacement of the slidingbolts and possibly bottom flangeplate replacement

(iv) Panel zones may rotate in excess of1% strain demand

(v) Lateral drift to be within sustainablelimits, including the influence of P - Deffects

(vi) The positioner bolt will needreplacement

(vii) Minor cracking only to the concretefloor slab surrounding the frame.

Application of the design procedures for the force-based method of design involves:

(a) Analysing the frame for the design levelearthquake using the Equivalent StaticMethod or the Modal Response SpectrumMethod from [2,4], and sizing the membersand connection components to meet therequired strength and stiffness criteria forthis event

(b) Following the joint design and detailingprovisions given herein (section 3) suchthat the joint can sustain the MCE rotationaldemands while delivering the performancecharacteristics of (3) above.

2.4 SHJ behaviour from experimental tests

There has been extensive experimental testingundertaken on the SHJ, involving both small-scalecomponent and large-scale assemblage tests.Some details of the large-scale tests are given onpages 26-30 of DCB No. 59 and a very briefoverview of these large-scale tests is given insection 3.3 of [7]. Details of the small-scalecomponent tests are given on pages 28, 29 ofDCB No. 64.

There has also been extensive finite elementanalysis (FEA) modelling of the sliding hinge jointsliding assemblage. This work was undertaken intwo stages; that from 2001 is summarised onpages 24-33 of DCB No. 64 and presented in fulldetail in HERA Report R4-110 [9]. That from2002 will be summarised in DCB No. 70 andpresented in [1].


Fig. 68.4Experimental and Simulated Moment-Rotation

Behaviour for Large-Scale TestsWithout Belleville Springs

Fig. 68.4 shows the moment-rotationcharacteristics from the large-scale test 3, whichinvolved the final proposed joint configurationwithout Belleville Springs to the bottom flangebolts. The moment-rotation characteristics of theSHJ are markedly different to those of any othersemi-rigid joint, because of the two stage slidingfrom the sliding components. In order toaccurately represent the joint behaviour in theNITH analyses, a mathematical model of themoment-rotation characteristics has had to bedeveloped and implemented into the computerprogram, RUAUMOKO [10] used for the NITHanalyses. This has been done; see details in [11].The simulated moment from that model generatedby the experimental rotations from test 3 is alsoshown in Fig. 68.4.

As the large-scale experimental tests could onlyinvestigate one size and layout of bolt, plate andcap plate and only at a pseudo-static rate ofloading, a series of small-scale tests on thebottom flange sliding assemblage wereundertaken during 2000/2001 on representativeconnections to determine the influence of thefollowing parameters.

bolt size range from M24 to M30

bolt layout and orientation flange plate and cap plate thickness presence/absence of Belleville Springs effect of loading rate: seismic-dynamic and

pseudo-static effect of repeated loading on assemblage,

including after a delay time of 4 weeks

One of the component experimental test results isshown in Fig. 64.16, DCB No. 64.

On the basis of these component tests, a boltdesign model has been developed to give the boltsliding shear capacity. Details of that model aregiven on pages 29, 30 of DCB No. 59. The basicmechanisms assumed for that model wereconfirmed by FEA modelling, as described in [9]and more briefly on pages 24-33 of DCB No. 64.

The completion of the experimental testingprogramme, FEA modelling and NITH studies hasallowed the design and detailing provisions for theSHJ and the MRSF systems incorporating theSHJ to be finalised. These are given in the nexttwo sections, starting with the design and detailingof the joint itself, in section 3.

SHJ Test 3, 04/08/2000, Plastic Rotation vs Moment and Simultest3 from Hysteresis Model

-800

-600

-400

-200

0

200

400

600

800

-30 -25 -20 -15 -10 -5 0 5 10 15 20 25 30

Rotation [mrad]

Mo

men

t [k

Nm

]

Test 3 experimental data Simultest3


3. Design and Detailing of the SlidingHinge Joint

3.1 General

The SHJ is intended for high ductility demand;mdesign = 4 is used. In theory it is possible to usem = 6, the maximum allowed from [2 or 4]. Thejoints ductility capacity is more than adequate forthis. However, as noted in section 2.3 (1), one ofthe performance criteria set for the joint is toremain effectively rigid, even after the joint hasbeen subjected to a design level ultimate stateearthquake.

Such an event is associated with some permanentsoftening of the joint, hence the decision to usemdesign = 4 as the ULS design ductility.

Designers should be aware of the very greatadvantage that the SHJ and the FBJ offer overconventional rigid-jointed MRSF systems. Thisadvantage is the ability to de-couple seismic andgravity requirements in the frame and connectiondesign. The approach used involves a variationon the procedure for design of multi-storey wind-resisting MRSFs in non-seismically activecountries, such that:

(i) The beams are designed to resist themaximum applied gravity loads (dead, liveloads) in a simply supported condition

(ii) The joint is sized to resist only the momentgenerated by the earthquake action, ie.Mcode, mdesign. This moment is calculated andapplied independently of the beams sectionmoment capacity.

(iii) The columns are designed to resist theoverstrength action developed by the joint,not that from the beam.

Thus the beam depth can be chosen for gravitystrength and lateral stiffness control withoutimpacting on the column design.

Details of the MRSF design are given in section 4.Coverage of the joint design itself nowcommences, first with the all-important detailingprovisions and material selection. These shouldbe read in conjunction with Fig. 68.1 for generaldetails and Fig. 68.2 for specific layout andnotation. The notation used herein is consistentwith that of the Structural Steelwork ConnectionsGuide, HERA Report R4-100 [12].

3.2 Detailing requirements and materialselection

As with all structural components designed todeliver dependable performance under severeseismic action, the detailing requirements andselection of appropriate materials are as importantto the final behaviour as the design itself.

Section 3.2 presents these requirements. It startswith limitations on the flange and web plate gradeand thickness, followed by the material selectionfor the brass shims.

This is followed by edge distances, bolt pitchesand gauges, then by the very important provisionof clearance between the beam face and thecolumn flange.

The dimensioning of all components is thencovered. This is followed by aspects of boltselection and installation and forming of theslotted holes.

Section 3.2 ends with surface treatmentrequirements for the ply contact surfaces.

3.2.1 Material selection for the jointcomponents

The bolts used, except for the positioner bolt, areProperty Class 8.8 Structural Bolts (HSFG bolts)to AS/NZS 1252 [13]. For calculation of bolt shearcapacity, threads are assumed to be in the shearplane. These bolts are to be supplied galvanized(this is the default surface treatment specified by[13]).

The positioner bolt is a Property Class 4.6 blackbolt to AS 1111.1 [14]. Only one positioner boltper joint is used and it has the same diameter asthe bottom flange bolts. It must be supplied blackfinish, to make it visibly different from the HSFGbolts. Black finish is the default surface treatmentfor this property class of bolt.

Grade of steel for the flange, web plates and capplates is to be 250, 300 or 350. It is important,when sizing the plates, that the use of grades 300or 350 in order to reduce the plate thickness for agiven width is clearly specified in the contractdocuments so that the grade used in design issupplied in practice. Designers can always opt foruse of grade 250 material; this is also consistentwith the approach used in R4-100 [12].

The brass shim material must be specified asUNS C2600 Hard Temper, eg. to AS 1566[15]. It is very important that the HardTemper is included in the specification, as thatdefines the hardness, yield stress and tensilestrength required and on which all the researchhas been based.

3.2.2 Limit on flange and web platethickness as a function of boltdiameter

The same relationship as is used for the FlangeBolted Joints should be used for the bottom flangeplate and web plate. This is given by equation68.1 and has been determined from thecomponent testing;


ti,max = 0.9df (68.1)

where:ti,max = maximum thickness of bottom flange,

web platedf = diameter of bolt

This translates to:

16 mm for M20 bolts 20 mm for M24 bolts 25 mm for M30 bolts 32 mm for M36 bolts

For the top flange plate, which is sized on thebasis of the actions generated by the sliding bolts(bottom flange and web bottom bolts), this limitcan be relaxed slightly in the larger bolt diameters,up to:

16 mm for M20 bolts 20 mm for M24 bolts 32 mm for M30 bolts 40 mm for M36 bolts

3.2.3 Edge distances required

For the edge distances to all the nominal sizedholes, these are 2df. This applies to the web topbolts, and the top flange bolts. The relevantdistances are shown in Fig. 68.2, namely:

aet = edge distance transverse to the line ofprincipal applied force

aep = edge distance parallel to the line ofprincipal applied force

For the slotted holes, the minimum distance fromthe end of a slotted hole to an adjacent free edge,

parallel to the line of principal applied force, shall

be df. This dimension is 'epa in Fig. 68.2.

The distance between the centreline of the lastpair of sliding bottom flange bolts and thecentreline of the positioner bolt is given by;

Sp,bfbpb = Max(2aep; 0.5Lsh + 'epa + aep)

(68.2)

where:Lsh = length of slotted hole; see equation 68.6in section 3.2.6.

3.2.4 Pitches and gauges

Spf = Sgf = Sgw = 70 mm for M20 bolts

Spf = Sgf = Sgw = 90 mm for M24, M30 bolts

Spf = Sgw = 140 mm for M36 bolts

Sgf = 140 mm (preferred) for M30, M36 bolts = 90 mm (alternative) for M30, M36 bolts,

where the beam flange width is inadequateto accommodate the sum of 140 mm plus atleast 4df.

Note that the minimum beam flangewidth required from (Sgf + 2aet,f,b) willnot allow the SHJ to be used for beamswith bf < 170 mm.

Table 68.1 gives the relevant values for eachdimension that have been used for th rangeof practical bolt diameters for the SHJ, alongwith the design sliding shear capacities,determined in accordance with equations 59.4to 59.10 of DCB No. 59.

Table 68.1Bolt Sliding Shear Design Capacities and Detailing Properties

BOLT SLIDING SHEAR DESIGN CAPACITIES AND OTHER PROPERTIES

Bolt Plate

Designation Thickness

mm

M20 12 42 51 93 20 22 50 50 70 70 70 16

M20 16 38 47 93 20 22 50 50 70 70 70 16

M20 20 36 44 93 20 22 50 50 70 70 70 16

M24 12 65 78 133 24 26 50 50 90 90 90 20

M24 16 60 73 133 24 26 50 50 90 90 90 20

M24 20 56 69 133 24 26 50 50 90 90 90 20

M24 25 52 64 133 24 26 50 50 90 90 90 20

M30 16 104 124 214 30 33 65 65 90 90 90 25

M30 20 98 118 214 30 33 65 65 90 90 90 25

M30 25 91 111 214 30 33 65 65 90 90 90 25

M30 32 83 102 214 30 33 65 65 90 90 90 25

M36 16 162 190 313 36 39 75 75 140 90 140 32

M36 20 153 182 313 36 39 75 75 140 90 140 32

M36 25 144 173 313 36 39 75 75 140 90 140 32

M36 32 132 162 313 36 39 75 75 140 90 140 32

Plate thickness limit, bottom flange & web

platesSgf mm Sp mmdf

' mm aep mm aet mm Sgw mmfVfss kN fVfss, bs kN fVfn kN df mm


3.2.5 Clearance between beam face andcolumn flange

This is the dimension fSHJ shown in Fig. 68.2.

The dimension is calculated on the basis that,when the sliding hinge joint is subject to maximumdesign negative rotation, thus causing the beambottom flange to move its closest in towardsthe column flange (see the right hand figure,Fig. 64.10 of DCB No. 64), there is still a clearlength of flange plate of 2.5tbfp available. Thisgives the following requirements for fSHJ in mm;

fSHJ 10 + 1.25 qp,desdb + 2.5 tbfp (68.3)

where:10 = gap to clear weld between column and

bottom flange plate (mm)qp,des = 30 x 10

-3 radiansdb = depth of beam (mm)tbfp = thickness of bottom flange plate (mm)

The value from equation 68.3 should be roundedup to the nearest 5 mm.

As specified in section 3.2.6 of DCB No. 58, theFBJ has a constant clearance gap, fFBJ, of 20 mm.In contrast, the SHJ has a variable clearance gapthat, in practice, varies from 50 mm to 100 mm ormore.

The 10 mm gap for the weld applies, irrespectiveof the type of weld used between the bottomflange plate and the face of the column.

3.2.6 Dimensions of the bottom flangeplate

This depends on the number of bottom flangebolts, which are determined from sections 3.6 and3.7. Once this is determined, the dimensions ofthe bottom flange plate are determined as follows:

(1) Width of bottom flange plate:

bbfp,min 4df,bfb + Sgf (68.4.1)bbfp,max 1.05bfc (68.4.2)

where:df,bfb = diameter of bottom flange boltsSgf = bolt gauge

(Fig. 68.2 and Table 68.1)

Where possible, use a flat bar to minimisefabrication cost.

(2) Thickness of bottom flange plate; tbfp

The initial estimate of thickness isdetermined from section 3.5 and confirmedfrom section 3.7. The limiting thickness asa function of bolt size from section 3.2.2must also be met.

(3) Length of bottom flange plate

See Fig. 68.2 for these terms.

Lbfb = fSHJ + Lsh(0.5nbfb 0.5) + 0.5nbfp 'epa

+ Sp,bfbpb + aep(68.5)

where:

Lsh = 2.5 qpdb + 'fd (68.6)

Sp,bfbpb = as given by equation 68.2qp = 30 x 10

-3 radians'fd = diameter of nominally sized

bolthole to NZS 3404 Clause14.3.5.2.1 (mm)

db = depth of beam (mm)

3.2.7 Dimensions of bottom flange platebrass shims

(1) For both brass shims (upper and lower):

Width = bbfp + 40 mm (68.7)

where:bbfp = width of bottom flange plate

(2) Thickness of both brass shims = 3 mm

(3) Length of upper brass shim

Lubfbs = Lbfp - fSHJ (68.8)

(4) Length of lower brass shim

Llbfbs = Lbcp (68.9)

where:Lbcp = length of flange cap plate, from

section 3.2.8

3.2.8 Dimensions of bottom flange capplate

(1) Width, bbcp

bbcp = bbfp (68.10)

(2) Thickness,

tbcp = Min (tbfp ; 20 mm) (68.11)

(3) Length

Lbcp = 2aep + 0.5(nbfp 2) (LSH + 'epa )

(68.12)

3.2.9 Dimensions of web plate

(1) Depth

This is the dimension dwp in Fig. 68.3. Theweb plate should be as deep as ispracticable for the given depth of beam,


leading to the following recommendationsfor hot rolled beams.

dwp, minimum = db 2tfb 58 (68.12.1)

dwp, maximum = db 2tfb 48 (68.12.2)

dwp, average = db 2tfb 53 (68.12.3)

where:db, tfb are the beam depth, flange thickness(mm).

The limits for dwp are given in mm.

(2) Thickness

twp = tbfp is initially used and is increasedonly if required from section 3.9. This hasnot been required in any of the designsundertaken for the NITH studies.

(3) Length

As can be seen from Fig. 68.2, the spacingof the web top bolts and the web bottombolts is controlled by different criteria. Theweb top bolts align with the top flange bolts,while the web bottom bolts align with thebottom flange bolts.

Thus the length of the web plate iscontrolled by:

Lwp Max [(fSHJ + 2aep + (nwtb 1) Sg,w);( 'epa + nwbb (Lsh +

'epa )]

(68.13)

where:nwtb = number of web top bolts, from

section 3.8nwbb = number of web bottom bolts, from

section 3.6

3.2.10 Dimensions of web brass shims

(1) Depth of web inner brass shim

As shown in Fig. 68.1 and, to a lesserextent, in Fig. 68.2, the inner brass shimextends the full depth of the web plate, witha return at the top of 15 mm. This return isto allow the brass shim to hook over theweb plate during erection, thus making itself-supporting while the beam is being putinto position.

The inner web brass shim is therefore dwpclear depth with a 15 mm return to eitherthe left or right as appropriate.

(2) Depth of web outer brass shim

This is equal to the web cap plate depth.

(3) Thickness of both brass shims

This is 3 mm.

(4) Length of inner web brass shim

Liwbs = Lwp - fSHJ (68.14)

(5) Length of outer web brass shim

Lowbs = Liwbs + 30 mm (68.15)

The additional length of the outer brassshim is to allow it to be held for positioningduring erection, once the web cap plate isin place.

3.2.11 Dimensions of web cap plate

(1) Depth

dwcp = 2aet (68.16)

(2) Thickness

twcp = Min (twp ; 20mm) (68.16)

(3) Length

Lwcp = Lwp fSHJ (68.18)

3.2.12 Dimensions of top flange plate

This depends on the number of top flange bolts,which are determined from section 3.11. Oncethis is determined, the dimensions of the topflange plate are determined as follows:

(1) Width of top flange plate

btfp, min 4df,tfb + Sgf (68.19.1)

btfp, max 1.05bbfc (68.19.2)

where:df,tfb = diameter of top flange bolts


(2) Thickness of top flange plate

This is determined from section 3.11; thelimit of section 3.2.2 as a function of boltsize must also be met.

(3) Length of top flange plate

Ltfp = fSHJ + 2aep + (0.5ntfb 1) Spf (68.20)

where:ntfb = number of top flange bolts, from

section 3.12.


3.2.13 Dimensions of optional deckingsupport shim

This is shown in Fig. 68.1. Its use facilitateslaying of decking around the connection, reducingcost and enhancing constructability. It is formedfrom 3 mm thick steel plate.

(1) Width

bdss Max (btfp ; bbf) + 100 (68.21)

where:bbf = width of beam flange (mm)

This allows 50 mm overlap each side of thewider of the beam flange or the top flangeplate.

(2) Thickness = 3 mm

(3) Length

Ldss = Ltfp 20 mm (68.22)

The outer edge of the decking support shimand top flange plate coincide; the inneredge extends past the face of the beamtowards the column face, as shown in Fig.68.2, with a gap of 20 mm adjacent to thecolumn.

3.2.14 Preferred bolt sizes and boltgroupings

For an initial guesstimation of bolt sizes, use M24for beams up to 600 mm deep and M30 for beamsabove 600 mm deep.

The minimum sliding bolt group layout is:

4 bottom flange bolts (2 rows of 2 bolts) 3 web bottom bolts (3 rows)

This is the layout shown in Fig. 68.2.

When increasing the number of sliding bolts todevelop the design moment, do this as follows:

Add one row of bottom flange bolts to give 6bottom flange bolts (3 rows) and 3 web bottombolts (3 rows); then

Add one row to each bolt group (ie. increasethe sliding bolt numbers in groups of 3 at atime).

This keeps sliding bolt group proportions in linewith those experimentally tested.

Other constraints on bolt sizes and groupings are:

nwtb nwbb df,bfb = df,wbb = df,wtb df,tfb = df,bfb is preferred df,positioner bolt = df,bfb

3.2.15 Use of Belleville Springs

These are optional for the bottom flange bolts.They increase the bolt sliding shear capacity, asdescribed in DCB No. 59 pages 29, 30, throughreducing the loss of installed bolt tension due tothe interaction of moment and axial force in thebolt shank when the joint is sliding. This benefit isof principal importance for the bottom flange boltsand, throughout this project, the research hasconcentrated on the following options:

no Belleville Springs Belleville Springs to the bottom flange bolts

Fig. 68.1 shows the former option, while Fig. 68.2shows the latter.

If Belleville Springs are to be used, then they mustbe of sufficient number and strength to developclose to the bolt proof load, from NZS 3404 Table15.2.5.1, when fully compressed.

From the manufacturers load charts [16] foralloy/carbon steel springs, the followingdesignation and number of springs are required toachieve this:

For a M20 bolt, 2 No. 12-EH-168 springs For a M24 bolt, 3 No. 16-H-168 springs For a M30 bolt, 3 No. 20-H-225 springs For a M36 bolt, 3 No. 24-H-262 springs

When Belleville Springs are installed, they are tobe placed under the nut end of the bolt, betweenthe hardened washer and the face of the capplate, as shown in Fig. 68.2.

When determining the nut rotation from the snug-tight position to apply, for the given bolt length,from Table 15.2.5.2 of [5], an extra turn must beadded to allow for compression of the BellevilleSprings. This extra turn applies for all boltdiameters used (M20 to M36). A background tothis will be given in [1].

3.2.16 Allowance for manufacturingtolerances in the supported beam andinclusion of a decking support shim

As described on pages 23, 24 of DCB No. 56 andin section 3.2.7 of DCB No. 58 for the FBJ,allowance must be made for manufacturingtolerances in the beams by offsetting the positionsof the top and bottom flange plates.


In the case of the FBJ, the magnitude of the offsetwas not important to the operation of the joint,thus the final recommendations were driven onlyby constructability considerations.

However, in the case of the SHJ, it is important tominimise the offset between the bottom face ofthe beam and the bottom flange plate. In the firstlarge-scale test specimen, this offset was 3 mm,whereas in the second test specimen, it was only1 mm. The greater offset from the first testresulted in an appreciable loss of bolt tension andhence sliding shear capacity of the joint. Howeverthe effect of the offset in test 1 was exacerbatedby allowing for a minimum gap between thebottom corner of the beam and the column face ofonly 15 mm under maximum negative rotation. Inthe second test specimen, this minimum gap wasincreased to 40 mm using equation 68.3, thusreducing the pull-down effect on the bolts by afactor of 18. While this increase in clearance hasa significant effect, it is also desirable to limit themaximum net extent of mismatch likely betweenthe top surface of the bottom flange plate and thebottom surface of the beam to 2 mm. This resultsin the following recommendations formanufacturing tolerance allowances in the SHJ:

(1) The allowances are provided as an offset ofeach flange plate away from thespecified centreline position of the beam(see Fig. 68.2)

(2) The up offset for the top flange plate is asfollows:

3 mm for beam depths up to 610 mm 4 mm for beam depths above 610 mm

3 mm is added to all the above toaccommodate a decking support shim,where used.

(3) The down offset for the bottom flange plateis as follows:

2 mm for all beam depths 3 mm is added to all the above to

accommodate the flange upper brassshim, which is always required.

In practice, these tolerance allowances will lead toa gap existing between the beam flange and topflange plate in most instances; this gap is readilyclosed by the bolt tightening, for which themoment developed is at least an order ofmagnitude greater than the weak axis plasticmoment capacity of the plate.

The web plate must also be offset from thecolumn flange centreline by an amount equal tohalf the beam web thickness plus 3.5 mm. 3 mmof this is to accommodate the web inner brassshim.

Finally, note mention of the decking support shimin Fig. 68.1. This is made from 3 mm thick Grade250 or 300 plate. It extends 50 mm beyond thetop flange plate on whichever side(s) of the beamsupport(s) steel decking and provides a support tothe decking during construction. It is also detailedin item 35 of HERA Report R4-58 [17]; seeespecially item 35c therein in this regard. It is anextra component to consider in fabrication anderection but one which greatly facilitates placing ofthe decking around the connection. Note also the3 mm thick plate extensions welded onto theunderside of the top tension/compressionstiffeners in Fig. 68.1 and Fig. 68.5 for the samepurpose.

3.2.17 Bolt tightening sequence and methodof tightening

The bolts are to be positioned in the directionsshown in Fig. 68.1 and tightened from the nut end.This is particularly important to avoid clashesbetween the web and flange bolts duringinstallation.

The positioner bolt is used during erection tostabilise the bottom of the joint and to preventundue rotation.

Once the frame is aligned, the bolts should all besnug tightened, starting with the bottom flangebolts and working up.

The tightening pattern should be to NZS 3404Clause 15.2.4.1. For each group of bolts (eg. thebottom flange bolts) this means starting with thebolts closest to the column face and workingalong the row away from the column face. For theflange bolts, this may require two or more roundsof snug tightening to get all bolts snug tight,pulling the flange plate in hard against the flangeupper brass shim.

The bolts are then fully tensioned, starting againwith the bottom flange bolts and working up.Tensioning is to the part turn method of NZS 3404Clause 15.2.5.2. For bottom flange bolts whereBelleville Springs are installed, tighten by an extra turn from snug tight over that specified in Table15.2.5.2 of [5].

3.2.18 Tightening of large diameter HSFGbolts

The SHJ connections will routinely require the useof fully tensioned M30 high strength structuralbolts and occasionally the use of M36 bolts. It isimportant to ensure that, when this size isspecified, they are fully tensioned.

This task is beyond the scope of a standardimpact wrench. Suitable equipment is readily


available; details are given on page 24 of DCBissue No. 56.

3.2.19 Forming of the slotted holes

The slotted holes in the bottom flange plate andweb plate (see Figs. 68.1 68.3 for their location)can be formed by machine flame cutting or waterjet cutting to the required dimensions.

However, they can also be formed by drilling anominally sized hole at each end of the slottedhole, then gas cutting across the top and bottomof this pair of drilled holes to form the slotted hole.This gas cut surface need be no smoother thanthat from good practice hand gas cutting, providedthat the rounded ends of the slotted hole are ofdrilled surface smoothness. If this method isadopted, then the width of slotted hole, asmeasured between the adjacent gas cut surfaces,

must lie between 'fd and 2) (d'f + mm. This has

been the approach used in all the SHJexperimental tests undertaken.

3.2.20 Surface treatment of the ply contactsurfaces

The sliding surfaces are between steel and brass.

The steel surfaces must be clean and free of anysurface coatings, loose scale, loose rust, visiblegrease or oil marks.

The brass surfaces must be clean and free ofsurface coatings, visible grease or oil marks.

Because of these restrictions on surface conditionof the sliding surfaces, the SHJ is principallyintended for application in corrosion category C1to ISO 9223 [18] (very low rate of corrosion,typically found inside heated or air conditionedbuildings with clean atmospheres).

The contact surfaces for the bottom flange boltsand web bottom bolts must be as specified abovefor SHJs in corrosion categories C2 to C5 of [18].Non-contact surfaces can be protected with anappropriate surface treatment; the edges of thecontact surfaces should be sealed against wateringress. The positioner bolt will need to bepainted in these applications.

3.3 Design concepts for the sliding hingejoint

3.3.1 Development of moment and shearcapacity

The design moment capacity of the SHJ, fMSHJ, isdetermined as follows:

(1) Calculate the design sliding shear capacity,SfVfss, of the bottom flange plate bolt groupand the web bottom bolt group.

(2) Take moments of each of the sliding shearcapacities from (1) about the top of steelbeam. The lever arms are shown in Fig.68.3. The sum of these moments = fMSHJ.

The design vertical shear (seismic plus gravity) iscarried by the web top bolts; thus the designshear capacity, fVSHJ, of the SHJ = the designshear capacity of the web top bolt group. Formost applications, only one row of web top boltswill be required to carry the design vertical shear,however, if this is large, two rows of web top boltsmay be needed.

In all the representative designs undertaken in thisproject, only one row has been needed.

The joint is sized to develop the following designmoment and shear capacities:

fMSHJ *designM (68.23)

fVSHJ VGQU + VEmdesign (68.24)

fVSHJ VGQmax (68.25)

where:*designM = design moment for the SHJ from

the most critical of earthquake orwind; see section 3.4.

VGQU = design shear force from loadcombination G + Qu (dead andlive load for use in conjunctionwith earthquake).

VEmdesign = design shear force derived fromout-of-balance design seismicmoments acting on the clearbeam length.

VGQmax = design shear force for full factoredloading, eg. 1.2G + 1.6Q from [2].

3.3.2 Design role of joint components

Refer to Fig. 68.1 in conjunction with this section.The design roles of the SHJ components are asfollows:

The top flange bolts act as the anchor pointfor joint rotation, pinning the beam top cornerin place relative to the column

The web top bolts resist the applied verticalshear force. They are subject to only smallmovement in the longitudinal direction due to


their proximity to the pinning action of the topflange bolts.

The web bottom bolts and bottom flange boltsdevelop the sliding shear resistance

The cap plates provide the support to the boltend remote from the beam of the sliding boltgroups

The brass shims facilitate smooth slidingbetween the steel surfaces at a near constantlevel of shear friction, which is essential to themaintenance of stable and sufficient bolttension when the joint is sliding

The Belleville Springs, which are optionaladditions to the bottom flange bolts, assistthese bolts to retain bolt tension under sliding.This sustains the bolt sliding shear capacity,Vss, at a higher level than is the case withoutthe springs and retains joint stiffness in thepost-sliding regime of behaviour.

The positioner bolt is a black finish class 4.6bolt that connects between the beam flangeand bottom flange plate only, through nominalsized holes in each ply. It has the samediameter as the rest of the bolts (which are allgalvanised finish property class 8.8 structuralbolts). The positioner bolt has three veryimportant roles, namely:

(i) It acts as a stability bolt for erectionpurposes, making the joint rigid forerection by developing momentresistance in conjunction with the topflange bolts

(ii) It functions as a locater bolt for thesliding bolts, ensuring that they arelocated in the middle of the slottedholes in the erected joint

(iii) It provides a rapid visual indicator asto whether the joint has gone into thesliding mode following a severeearthquake; if this happens and thejoint inelastic rotation exceeds around10 mrad, the positioner bolt shearsthrough and the lower half drops out.

3.3.3 Sequence of design actions

The full SHJ design procedure involves thefollowing 14 steps:

Step 1 : Determine design moments and shears

Step 2 : Determine sliding bolt group layouts

Step 3 : Determine initial bottom flange platewidth and thickness and initial web platethickness

Step 4 : Determine bolt size and numbers formoment adequacy, then finalise bottomflange plate width and thickness

Step 5 : Design web top bolts for vertical shearresistance

Step 6 : Design web plate

Step 7 : Design top flange bolts and plate

Step 8 : Check on reduced tension capacity ofthe beam at the bolted connection

Step 9 : Design welds between plates andcolumn

Step 10 : Dimension flange and web plates

Step 11 : Design, detail positioner bolt and shims

Step 12 : Design tension/compression stiffeners

Step 13 : Calculate joint overstrength capacity

Step 14 : Design joint panel zone

The full SHJ design procedure, starting withdetermination of joint design moment and designshear, is given in sections 3.4 to 3.21.

3.4 Calculation of the design moment anddesign shear

3.4.1 Design earthquake moment

As has been mentioned in section 3.1, the jointitself is sized to resist the code-derivedearthquake moment alone, ignoring joint momentsinduced by gravity only, with the beam designedto resist the full factored gravity load (ie. 1.2G +1.6Q to NZS 4203 [2]) as a simply supportedbeam.

For the SHJ, the design earthquake moment,*

designEM m , is determined from [2 or 4, 5 and 6] for

low-rise and medium-rise MRSFs. The jointdesign earthquake moment is given in sections4.2 and 4.3 herein.

3.4.2 Design shear force

This is given by the largest of equations 68.24 and68.25.

The seismic component of shear, * designEV m , is

given by:

) - (

3

cb

*designE*

designE dL

MV mm = (68.26)


where:3 1.4 x 1.1 x 2

1.4 = overstrength factor on joint1.1 = allowance for fMSHJ / M*2 = moment pattern factor (equal

and opposite end moments)(Lb dc) = clear length of beam

3.4.3 Design wind moment

The SHJ has been developed as a semi-rigid jointfor seismic-resisting systems. However, it mustalso perform satisfactorally under wind loading.

In designs for New Zealand application, inaccordance with NZS 4203 [2] or its proposedreplacement [4, 19], it is possible that ultimatelimit state wind design may govern some joints inbuildings over around 10 storeys high. This willbe especially the case for designs to the draftLoadings Standards [4] which are located in thelowest seismic regions.

Also, because the levels of wind loadingassociated with the serviceability and ultimate limitstates are closer (see eg. Table 5.4.2 of [2]) thanfor earthquake, it is possible that either wind limitstate may govern some SHJs in buildings as lowas 10 storeys high.

For this reason, brief guidance on SHJ design foreach wind limit state is given below.

3.4.3.1 Wind ultimate limit state

The joint design for the wind ultimate limit state

moment, *WULSM , uses the principles andprocedures as given in sections 3.5 to 3.16. Insaying this, it is conservative to apply the relevantoverstrength factors as ductility demand is not

anticipated under *WULSM .

It follows, in checking for the wind ultimate limit

state, that if *EM >*WULSM , then the earthquake

condition governs design for the ultimate limitstate.

3.4.3.2 Wind serviceability limit state

The SHJ must remain rigid at the windserviceability limit state. This is easily checked asfollows:

Step 1: Calculate the design wind serviceability

limit state moment, *WSLSM .

Step 2: Determine the moment associated withrigid action of the joint from equation 68.27.

fMSHJ,WSLS = 0.75fMSHJ,Em (68.27)

where:0.75 = (0.7/0.8) x 0.85

0.7 = strength reduction factor for tensionfriction action

0.8 = strength reduction factor for bolt slidingshear capacity determination

0.85 = kh for short slotted holes, from NZS 3404Clause 9.3.3.1.

In practice, the length of slotted hole will typicallybe such as to classify it as a long slotted hole.However the cap plate provides much morerobust confinement than an oversized washer,thus the value of kh for short slotted holes ratherthan for long slotted holes is used.

Step 3: Check if equation 68.28 is satisfied

fMSHJ,WSLS *WSLSM (68.28)

If it is, the joint design is satisfactory.

If it isnt, then add an extra set of sliding bolts inaccordance with section 3.2.14 and recheck. Thiswill affect the overall joint design and overstrengthaction and require reconsideration of the joint andsystem design for earthquake.

In practice, for designs to either the currentStandard [2] or the new draft [4 and 19], it is likelythat, where wind action governs, it will be the ULSrather than SLS that is critical. This is because

the ratio of ( *WSLSM /*WULSM ) will typically be less

than 0.75.

Having determined the design moment and shear,the joint design proceeds as follows. For thisprocedure, the ULS design moment is designated

*designM , which covers the critical ULS moment

being from either earthquake or from wind, asappropriate.

3.5 Determine bottom flange plate width andinitial thickness

3.5.1 Bottom flange plate width

See section 3.2.6 (1) for the limits. Select a platewidth, bbfp, within these limits.

3.5.2 First estimate of bottom flange platethickness

The bolt sliding shear capacity is a function of theplate thickness, hence the joint moment capacityis also a function of the plate thickness. Thismeans it is desirable to obtain a rapid estimate ofbottom flange plate thickness as soon as the jointdesign moment is known. This is determined fromthe following two equations.

b

*design*

designt,

1.2

d

MN = (68.29)


bfpy,'f bfp

*designt,

bfp )2 ( 0.9

fdb

Nt

- (68.30)

where:0.9 = strength reduction factor

d'f = bolt hole diameter for nominal sizedhole, from NZS 3404 Clause 14.3.5.2.1.

fy,bfp = bottom flange plate yield stress

3.5.3 Check plate thickness limit in relationto bolt size

Check that the limit of section 3.2.2 is satisfied; ifit isnt, then a larger bolt diameter is needed forthe given plate thickness.

3.5.4 Apply this estimate of thickness tothe web plate

3.6 Determine sliding bolt size and numbersfor moment adequacy

3.6.1 Start with the following

Bolt size, numbers and layout from section3.2.14.

3.6.2 Calculate moment capacity of joint

(1) Joints with no Belleville Springs

fMSHJ = nbfb fVfss db + nwbb fVfss ewb(68.31)

(2) Joints with Belleville Springs in bottomflange

fMSHJ = nbfb fVfss,bs db + nwbb fVfss ewb(68.32)

where:nbfb = no. of bottom flange bolts

= 4 for initial trial, from section3.2.14

nwbb = no. of web bottom bolts= 3 for initial trial, from section

3.2.14

fVfss = design sliding shear capacity, noBS

= get from Table 68.1 for bolt sizeand plate thickness Bolt size for initial trial from

section 3.2.14 Plate thickness from section

3.5.2

ewb = db tfb 26.5 aet (mm)

(68.33)

fVfss,bs = design sliding shear capacity,with BS

= get from Table 68.1 for boltsize and plate thickness

Fig. 68.3 shows the lever arms for the momentcapacity determination. The value of 26.5 used inequation 68.33 comes from the average web platedepth, from equation 68.12.3.

3.6.3 Check moment adequacy

This is given by:

fMSHJ *designM (68.34)

where:*designM is from section 3.4; typically section 3.4.1.

3.6.4 Review bolt numbers and size

If equation 68.34 is easily satisfied, reducebolt size to M20 and recalculate; this givesfMSHJ, minimum for the given beam size.

If equation 68.34 is not satisfied;

Either increase the bolt numbers in accordancewith section 3.2.14 and recalculate; or

Increase the bolt size and recalculate; or Increase the bolt numbers and bolt size and

recalculate.

3.7 Design of bottom flange plate

There are four cases to consider, three of whichrequire calculation and the fourth of which is dealtwith by detailing. These are:

(i) Suppression of net tension yield prior to thebolts developing their sliding shearcapacity; see section 3.7.1

(ii) Suppression of net tension fracture whilejoint is in active sliding mode; see section3.7.2

(iii) Suppression of compression yielding whilejoint is in active sliding mode; see section3.7.3

(iv) Suppression of premature bolt shearfracture when end of slotted hole isreached; this is covered by compliance withthe bottom flange plate thickness to boltdiameter ratio given by equation 68.1.

3.7.1 Net tension yield

* bfpty,N = 1.15 nbfb fVfss,bfp (68.35)


bfpbfpy,'fbfpbfpty, )2 - ( 0.9 tfdbN =f (68.36)

where:nbfp = no. of bottom flange bolts from

section 3.6.41.15 = 0.9 / 0.8 = difference in f between bolt

and plate(b,t,fy)bfp = from section 3.6

'fd = function of df,bfb from NZS 3404 Clause

14.3.5.2.1 (see also Table 68.1)fVfss,bfp = fVfss or fVfss,bs as appropriate, from

Table 68.1.

fNty,bfp * bfpty,N is required (68.37)

3.7.2 Check for net tension failure

This is determined from the design actiondeveloped under the design level of rotation. Theideal capacity of the plate is used to resist thisaction, therefore the ideal capacity factor isincorporated into the design action determination,thus:

0.9 spbfpfss,

bfb*

bfpu, CV

nNf

f= (68.38)

where:nbfb = no of bottom flange bolts, from section

3.6Csp = 1.45 when no springs are used

= 1.55 when Belleville Springs are used0.9 = ideal capacity factor

bfpbfpu,'fbfpbfptu, )2 - ( 0.77 tfdbN =f (68.39)

where:fu,bfp = ultimate tensile strength of bottom flange

plate

fNtu,bfp * bfpu,N is required (68.40)

3.7.3 Compression capacity

First the slenderness ratio of the bottom flangeplate must be checked

Le,bfp = 0.7 (fSHJ + 1.25qpdb) (68.41)

250

0.29

bfpy,

bfp

bfpe,bfpn,

=l

f

t

L(68.42)

Check if ln,bfp 25. If it is, proceed to the nextequation. If it isnt, then av for input into equation68.43 needs to be re-evaluated from Table 6.3.3of NZS 3404 [5] for the value of ln,bfp fromequation 68.42.

bfpy,bfpbfpbfpcu, 0.85 ftbN =f (68.43)

where:0.85 = 0.9 x 0.9420.942 = av from Table 6.3.3 of NZS 3404 for

an = 25 and ab = 0.5

*bfpu,bfpcu, NN f is required (68.44)

where:*

bfpu,N is given by equation 68.38

Use the resulting tbfp for the web plate.

3.8 Design of web top bolts

These are designed to resist the applied verticalshear, in bearing, with threads included in theshear plane.

3.8.1 Vertical design shear force

) ; (Max *GQmax*GQu

*designE

*wv VVVV += m (68.45)

where:*

designEV m = as given by equation 68.26*GQmax

*GQu V , V = as given by the tributary area

vertical loading for theappropriate factored maximum(dead + live) loads

3.8.2 Determine the number of web topbolts required

wtbfn,

*wv

wtb V

Vn

f (68.46)

where:fVfn,wtb = design capacity, threads included,

same bolt diameter as for webbottom bolts. (See eg. [20] for thisinformation).

If nwtb < nwbb, where nwbb has been determinedfrom section 3.6, then add additional web top boltssuch that nwtb = nwbb. The additional web top boltsare then used to resist the forces developed bythe sliding groups of bolts, in conjunction with theflange top bolts, in section 3.11.

3.9 Design of web plate

The web plate thickness, twp, has been set equalto be bottom flange plate thickness, from step3.7.3. The web plates capability to resist thevertical shear and horizontal tension actions nowneeds to be determined.


Vertical shear will be resisted over the full depth ofplate less the width involved in resisting horizontalactions from the web bottom bolts. Horizontaltension/compression is developed by the slidingresistance of the web bottom bolts. This isresisted by the strip of web plate under the webcap plate for commencement of yield and by1.5 x bwcp for tension fracture under overstrengthaction.

3.9.1 Calculate design vertical shearcapacity of plate

vwpwpy,wcpwp

vwpwpy,wcpwpwpvn,

) - 0.27(

) - 0.83( x 0.6 x 0.6 x 0.9

a=

a=f

tfdd

tfddV

(68.47)where:

82 wpy,

wp

wcpwpv 250

) - (

if 1.0 f

t

dd=a (68.48)

av = 1.0 otherwise; see Clause 5.11.5.1 ofNZS 3404.

The second 0.6 is to account for moment / shearinteraction.

dwcp = as given by section 3.2.11(1)

3.9.2 Check vertical shear adequacy ofplate

*wvwpvn, VV f (68.49)

where:*wvV = design vertical shear force from equation

68.45.

3.9.3 Check for net tension yield

This is checked under the design sliding shear, forthe width of web plate under the cap plate only.

fsswbb*

wpty, 1.15 VnN f= (68.50)

wpy,wp'fwcpwpty, ) - 0.9( ftddN =f (68.51)

*wpty,wpty, NN f is required (68.52)

3.9.4 Check for net tension failure

This is checked for the overstrength sliding actionassociated with reaching the end of the slottedhole, with this action being resisted by a depth ofweb plate = 1.5 x depth of cap plate.

0.9 spwbfss,

wbb*

wptu, CV

nNf

f= (68.53)

wpu,wp'fwcpwptu, ) - 0.77(1.5 ftddN f (68.54)

*wptu,wptu, NN f is required (68.55)

3.9.5 Sizing of web plate

This can now be done; see section 3.2.9.

3.10 Sizing of cap plates and brass shims

3.10.1 Bottom flange cap plate

See section 3.2.8 for determining the width,thickness and length of bottom flange cap plate,using the values determined above.

3.10.2 Bottom flange upper and lower brassshims

See section 3.2.7.

3.10.3 Web cap plate

See section 3.2.11.

3.10.4 Web inner and outer brass shims

See section 3.2.10.

3.11 Design of top flange bolts and plate

The top flange plate anchors the beam laterallyand operates as a hinge about which the beamcan slide. It is designed to resist the combinedshear developed by the web bottom bolts andbottom flange bolts, using bolts of the samediameter. The shear from these is the greater ofthe overstrength sliding shear or the threadsexcluded design shear capacity. The latter willalways govern and is therefore the only checkneeded. Because of this it is simply a matter ofmatching bolt numbers, incorporating any web topbolt unused capacity to resist the lateral force.

3.11.1 Number of bolts required

Using the same bolt diameter as for the webbottom bolts and the bottom flange bolts.

nftb required = ( )( )calcwtb, used,wtbbfbwbbr

- - 1

nnnnk

+

(68.56)where:kr = reduction factor for bolts in a line

from NZS 3404 Table 9.3.2.1.

nwtb,calc = no. of web top bolts required fromequation 68.46, section 3.8.2

nwtb,used = no. of web top bolts used fromsection 3.8.2

If the length of the joint, as measured from the firstto the last bolt, exceeds 15df, then kr < 1.0.


3.11.2 Determine the top flange plate widthrequired

See section 3.2.12(1) for the limits on btfp. Selecta plate width within these limits.

3.11.3 Determine required thickness tosuppress tension yielding

This is sized so that the plate can develop thesliding shear capacity of the bottom flange andweb bottom bolts, without tension yielding.

tfpy,'ftfp

fswbbfsbfbtensiontfp,

)2 - (9.0

) ( 1.15

fdb

VnVnt

f+f (68.57)

where:fVfs = fVfss or fVfss,bs as requiredfy,tfp = yield stress of top flange plate

'fd = diameter of bolt hole to NZS 3404

Clause 14.3.5.2.1


3.11.4 Undertake a slenderness ratio checkon the top flange plate, if no concreteslab is present

If there is a concrete slab in contact with the topsurface of this plate, which will be the typical case,no slenderness check is needed.

If there isnt a concrete slab, then:

Le,tfp = 0.7 (fSHJ + aep,tf,b) (68.58)

250

0.29

bfpy,

tfp

tfpe,tfpn,

=l

f

t

L(68.59)

where:fSHJ is determined from equation 68.3aep is given in Table 68.1 for the given bolt size.

Check if ln,tfp 25, when no concrete slab ispresent. If it isnt and no slab is present, then avfor input into equation 68.61 needs to be re-evaluated from Table 6.3.3 of NZS 3404 [5].

3.11.5 Check top flange plate and boltadequacy for the ULS condition

Calculate

tfpu,tfp'ftfptfpt, )2 - 0.77( ftdbN =f (68.60)

tfpy,tfptfptfpc, 0.85 ftbN =f (68.61)

Check

tfbfn,tfp tfpc,tfpt, 0.85 and VnNN fff (68.62)

If equation 68.62 is not satisfied, add an extra pairof top flange bolts and recheck.

3.12 Check on the reduced tension capacityof the beam at the bolted connectionregion

The purpose of this check is to suppress yieldingof the beam cross-section through the loaded endof the beam under moment-induced tensionduring the sliding phase of the joint. Such yieldingwould cause unwanted loss of bolt tension andhence sliding shear moment capacity.

3.12.1 Calculate the design tension action,*tbN , on the tension half of the beam,

from equation 68.63

tbsx,

SHJ*tb

1.15 x 0.5 N

MM

Nf

f= (68.63)

where:fMSHJ = the joint design moment capacity from

section 3.6fMsx,b = the design section moment capacity for

the beam size chosen; eg. from [20]Nt = the nominal section gross yielding

capacity, determined from NZS 3404Equation 7.2.1

3.12.2 Calculate the design tension capacityof the beam from the lesser of

ubnbtb 0.39 fAN =f (68.64.1)

yfbgtb 0.45 fAN =f (68.64.2)

where:Anb = net area of the beam cross section,

calculated in accordance with Clause9.1.10 of [5]

fub = tensile strength of the beamfyfb = yield stress of the beam flange

3.12.3 Check that the following is satisfied

*tbtb NN >f (68.65)

If this equation is not satisfied, use a larger beamsize so that it is satisfied. Do not use beamreinforcing plates with the SHJ.

In practice, if

ff

bsx,

SHJ

MM

0.76, the beam end

capacity is likely to be adequate. For preliminary

design, one can use )M /M( bsx,*design f 0.76/1.15

0.66 as a target value for beam selection.


3.13 Welds required between column flangeand bottom flange plate

The bottom flange plate has been sized todependably resist the maximum force expectedunder the maximum design rotation, inaccordance with section 3.7.2. This carries thejoint-overstrength factor, which means that theweld need only be designed to develop the designtension capacity of the flange plate, not theoverstrength tension capacity.

3.13.1 Design action on bottom flange plateweld

min

*bfptw,*

bfpw,2

b

Nv = (68.66)

where:

N* bfptw, = fNtu,bfp from equation 68.39bmin = lesser of (bbfp; bfc)bbfp = width of bottom flange platebfc = width of column flange

3.13.2 Select fillet weld size such that:

*bfpw,w vv f (68.67)

where:fvw = design capacity of category SP fillet

weld from [5]

Values of fvw are listed in [20]

This is the fillet weld size required on each side ofthe flange plate to column flange.

3.13.3 From consideration of weldingeconomics and clearance requirements,determine if the fillet weld size from3.13.2 will be used or if a completepenetration butt weld is required.

If tw > 12 - 15 mm, use a complete penetrationbutt weld (CPBW). For most fabricators engagedin multi-storey construction, the changeover pointto a CPBW will be tw > 15 mm.

3.14 Welds required between column flangeand top flange plate

A similar situation applies to that for the bottomflange plate, namely:

3.14.1 Design action on top flange plateweld

2

min

*tfptw,*

tfptw, b

Nv = (68.68)

where:

*twv = fNt,tfp from equation 68.60bmin = lesser of (btfp ; bfc)btfp = width of top flange plate

3.14.2 Design of welds

Determine fillet weld size required as for thebottom flange plate, see section 3.13.2. If tw > 15mm use a CPBW.

3.15 Welds required between column flangeand web plate

These welds are subject to two very different setsof conditions. The first is combined moment andvertical shear generated by the web top bolts andresisted by the clear depth of web plate for shearand the full depth for moment. The second ismoment-induced axial tension generated by theweb bottom bolts at the end of their slidingregime, taken over a thickness of web plate equalto 1.5 x the thickness of the web cap plate. Thetwo cases are considered separately and thedesign action is the maximum from the two cases,but not required to be greater than the designtension capacity of the plate. All this involves:

3.15.1 Calculate actions on weld fromvertical shear

) - (2

wcpwp

*wv*

vwp,wv, ddV

v = (68.69)

3

2wp

y*wv*

hwp,wv,d

eVv = (68.70)

( ) ( ) 5.02* hwv,2* vwv,* wpwv,

+= vvv (68.71)

where

V *wv = design vertical shear force, fromequation 68.45

dwp = average depth of web plate, fromequation 68.12.3

dwcp = depth of web cap plate, from section3.10.3

ey = fSHJ + aep,wt,b + ((nwtb 1)/2)Sg,wt(68.72)

3.15.2 Calculate actions on weld from axialtension generated by web bottombolts

3

wcp

wptu,*wpwh, d

Nv

f= (68.73)

where:fNtu,wp= capacity given by equation 68.54


3.15.3 Calculate design actions on weldbetween column flange and web plate

( ) 10 x 2

9.0 ; ; Max Min

3wpy,wp*

wpwh,*

wpwv,*

wpw,

=

ftvvv

(68.74)

3.15.4 Design weld

Select fillet weld size such that

*wpw,wpw, vv f (68.75)

where:fvw = design capacity of a category SP fillet

weld, eg. from [20]

The size is used on each side full length of theweb plate to column flange.

If tw,required 15 mm use a CPBW.

3.16 Selection and location of the positionerbolt

The role of the positioner bolt is described insection 3.3.2; its grade and appearance in section3.2.1.

This bolt connects between the bottom flangeplate and beam flange only. It is placed as shownin Fig. 68.2; the distance from the centreline ofthis bolt to the centreline of the adjacent row ofbottom flange bolts is given by equation 68.2.

This bolt is intended to be snug tightened only, butcan be fully tensioned to hold the bottom of thejoint in place during erection, if desired.

3.17 Tension/compression stiffenerrequirements

These are determined using NZS 3404 Clause12.9.5.3.1, modified as described below, inconjunction with section 3.2, page 13, DCB IssueNo. 50.

(1) Provide tension/compression stiffenerspositioned opposite the flange plates, sothat top of steel is the same for eachelement.

(2) Use NZS 3404 Equations 12.9.5.3(3) and12.9.5.3(4) to determine the area ofstiffener required for each design action,with the following modification:

The tension/compression stiffenerdesign is based on the bottom flangeplate dimensions for both the top andthe bottom pair of stiffeners. This maymean that the top pair of stiffeners areslightly thinner than the top flange plate.

Replace all terms related to the beamflange with the same term for the bottomflange plate, ie: Abfp replaces Afb; fy,bfpreplaces fyb; tbfp replaces tfb; bbfp replacesbfb; twf relates to the weld calculated from3.13.3 above.

More simply, use equation 50.2 fromsection 3.2(2) of DCB Issue No. 50 withthe same substitutions as stated above.

(3) Design and detail the tension/compressionstiffeners to section 3.2 of DCB issue No.50 (with the above modification to section3.2(2) of that issue)

3.18 Joint overstrength moment, oSHJM

This is determined as follows:

f

ff= SHJoms

oSHJ

MM (68.76)

where:fMSHJ = joint design moment from section 3.6f = 0.8foms = 1.4 for the SHJ with or without Belleville

Springs

This overstrength factor has been derived fromthe experimental testing, using the methodologyas will be described in [1].

3.19 Joint panel zone requirements

3.19.1 Design shear force on panel zone

The panel zone design moment for input into NZS3404 Equation 12.9.5.2(1) is the joint overstrengthmoment given by equation 68.76. However,compared to the layout of a rigid welded joint, thetop and bottom flange plates are more widelyspaced apart (see Fig. 68.2) which reduces theunbalanced shear force on the connection.

These two aspects are incorporated into equation68.77, which gives the design shear force on thepanel zone of a SHJ.

( ) ( ) COLRbfpb

oSHJ

Lbfpb

oSHJ*

SHJP, -

V

tdM

tdM

V

++

+=

(68.77)where:The subscripts L and R refer to the left and righthand beams at the connection.

oSHJM = as given by equation 68.76.

For preliminary design and for most final designs,VCOL can be accounted for as described in NZS3404 Commentary Equation C12.9.5(1).


3.19.2 The design shear capacity of thepanel zone, fVc , is calculated to NZS3404 Eq 12.9.5.3.(5).

3.19.3 The panel zone has adequatecapacity when

*f pzc VV (68.78)

Doubler plates, if needed, should be designed inaccordance with sections 4 and 6 of DCB IssueNo. 57, pages 23-25 therein, which, althoughwritten for FBJ connections, actually covers bothFBJ and SHJ connections.

With the SHJ, doubler plates are not typicallygoing to be necessary when only one beamframes into the column, but will often be requiredwhen two beams frame into the column.

3.20 Connections at column bases

3.20.1 Options available and impact onbuilding performance

The most commonly used column baseconnection type for a MRSF is a fixed baseconnection. This has the advantage of reducinglateral deflection in the superstructure. Asmentioned in section 2.3, with fixed base columnsthe inelastic demand on the joints under thedesign severe seismic event is within theperformance criteria specified for the columns insections (2) and (3) therein. With pinned basecolumns, these limits are slightly exceeded insome types of earthquake record, principallythose exhibiting positive near fault directionalmotion.

A third option is a ring-spring type detail at thecolumn base. This is mentioned in [21], with apicture of such a joint shown as Fig. 15 of [21].

When subjected to a design level severe seismicevent, it is anticipated that minor damage to theyielding regions of columns adjacent to thecolumn bases would occur in columns with fixedbase connections. For columns with pinned baseconnections, minor damage would be expectedwithin the baseplate detail. In each case, minimalor no repair would be anticipated to be necessaryfrom this level of event.

The ring-spring base would be dependablyundamaged by this level of event.

Brief guidance on each type of column baseconnection is now given.

13.20.2 Fixed bases

The design actions for fixed bases are given insection 4.2.1, pages 22,23 of DCB Issue No. 50and are directly applicable to these semi-rigidsystems. The actions are based on mdesign = 4.

The effects of the slight foundation flexibilityshould be accounted for; in lieu of a more detailedanalysis, use the rotational stiffness given by NZS3404 Clause 4.8.3.4.1(b).

Design and detailing concepts for moment-resisting column baseplate connections are givenon pages 11-20 of DCB Issue No. 56.

The advice in both articles is written to utilise, asmuch as possible, the standard details andprovisions in HERA Report R4-100 [12].

3.20.3 Pinned bases

Design actions and detailing requirements aregiven in section 4.2.2, page 23 of DCB Issue No.50. The advice therein is also written for use inconjunction with [12]. Note that, for analysis, apinned connection should be assigned a realisticrotational stiffness. This can be obtained fromClause 4.8.3.4.1 (a) of [5].

3.20.4 Ring spring bases

Fig. 15 of [21] shows a ring spring test setupwhich would also be applicable to a column baseapplication.

The ring spring joint is well suited to application atthe column base of a MRSF with SHJs or FBJs.This is because it combines the benefit of thepinned base, in protecting the column frominelastic action at its base, with the ability togenerate a rapid increase in moment capacity withincreasing rotation demand. The joint also hasgood self-centering capability, which will assist inreturning the building to its pre-earthquakeposition at the end of the strong ground motionshaking.

Design of the ring spring joint for this system isrelatively straightforward. It is referenced fromsection 6.2 of [21] and will be described in [1];further details are not given herein. Contact theHERA Structural Engineer for more information.

3.21 Guidance on practical aspects of slidinghinge joint design

The flange plates should be made as wide aspossible, within the limits of sections 3.2.6 (1)and 3.2.12(1)

The top flange plate will typically be the samethickness or the next thickness up from thebottom flange plate and the web plate

The maximum number of bottom flange boltsshould be 8, in order to keep the requiredbottom flange plate thickness within themaximum thickness allowed for the given boltdiameter. If the design from section 3.6indicates that nbfb = 10 is required, look at


increasing the bolt size. With nbfp = 8, nwbb =nwtb = 4 will result, with ntfb = 12, typically.

Refer to Table 68.2, section 4.2, step 7 fortypical values of fMSHJ/fMsx,b that haveresulted from the many representative framesdesigned as part of this project.

4. Design of Moment-Resisting SteelFrames Incorporating Sliding HingeJoint Connections

4.1 General and scope of guidance given

Section 4 presents guidance on the design of theMRSF system that incorporates the SHJ. Thisguidance is very similar to that for MRSFs withFBJs and follows the same format as that given inDCB No. 58 for the FBJ systems.

Section 4.2 covers preliminary design, whilesection 4.3 covers final design.

The design procedures presented herein arebased, in format and content (wherever possible),around the procedures incorporating capacitydesign presented in sections 5 and 6 of HERAReport R4-76 [6] for preliminary and final design,respectively, of category 1 or 2 MRSFs with rigidbeam to column connections.

For such systems, strength and stiffness cannotbe de-coupled, so the columns must be designedto resist the beam section overstrength actions (orthe upper limit seismic actions Emax).

In contrast, for the semi-rigid systemsincorporating SHJs, strength and stiffness areconsidered separately and the columns aredesigned to develop only the overstrengthmoment from the joint. This requires somemodifications to the R4-76 [7] procedures, but is aconsiderable simplification from the designersview point.

Given that this guidance is being written at thetime of transition from NZS 4203:1992 [2] to thenew Loadings Standard [4], wherever practicablethe requirements of both documents arereferenced.

4.2 Procedure for MRSF preliminary design

The preliminary design procedure presentedbelow is based around that given in R4-76 section5.2 for preliminary design of category 1 and 2MRSFs. It is presented in the same step by stepformat as section 5.2 of [7] and with the sameheadings.

Step 1 Establish preliminary frame layouts

Formulate the preliminary frame layout or layoutsin terms of the beam and column spacings and

the number of bays (more than one scheme maybe required).

The SHJ has been developed for perimeter frameapplication and the guidance given in step 2.2 andstep 6 herein for the member sizes to meet framestiffness requirements is formulated on that basis.(Perimeter frames and internal frames are asdefined in NZS 3404[5]).

Step 2 Estimate beam sizes required

This estimate should be made at the first levelabove the seismic base level, at the level ofuppermost principal seismic mass level and atselected intermediate levels.

Guidance on the number of intermediate levels toconsider is given on page 5.3 of [6]. For buildingsup to 4 storeys in height, do the check at everylevel. For buildings up to 8 storeys in height,check levels 1, 3, 5 and 8. For buildings between8 and 12 storeys in height, check levels 1, 3, 5, 8,11 and 12. For buildings above 12 storeys inheight, check 1, 3, 6 then every 4th level.However, the SHJ is probably not the most cost-effective system to use on buildings above thisheight, because of the limited ductility demandrequired compared with what the SHJ can deliver.For such high-rise buildings, the FBJ offers apotentially more cost-effective solution, especiallyin low to medium seismic zones.

Step 2.1 To carry gravity loads

Use the approach given in step 2.1, section 5.2 of[6], except use the denominator value of 8 inequation 5.1 of [6] instead of 10. Thiscorresponds to a simply supported condition,which is required for design in accordance withsection 3.1(i) herein.

Use the lightest category 3 section fromNZS 3404 [6] within a particular designation toresist the design moment, such that M* fMs. (Inthe 1992 edition of NZS 3404, this category wasdesignated 3A, which is still used in HERA reportR4-76 [6]. This point is picked up in the summarynotes Tips on Seismic Design of Steel Structureswhich are included in all post-July 2000 copies of[6]).

Step 2.2 To provide suitable frame lateralstiffness

(1) For perimeter frame MRSFs, select beamdepths from the target span to depth ratiosgiven by equations 68.79.1 to 68.79.3

(1.1) For the lower half of the structure (up to0.5H)

(L/d*) = (9 or 11)10 > (11 or 13)5(68.79.1)


(1.2) For the three-quarter height (0

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DCB 68 Sliding Hinge Joint