Arcelor Mittal - Single-Storey Steel Buildings

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STEEL BUILDINGS IN EUROPE

Single-Storey Steel Buildings

Part 1: Architect’s Guide









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FOREWORD

This publication is part one of the design guide, Single-Storey Steel Buildings.

The 11 parts in theSingle-Storey Steel Buildingsguide are:

Part 1: Architect’s guide

Part 2: Concept design

Part 3: Actions

Part 4: Detailed design of portal frames

Part 5: Detailed design of trusses

Part 6: Detailed design of built up columns

Part 7: Fire engineering

Part 8: Building envelope

Part 9: Introduction to computer software

Part 10: Model construction specification

Part 11: Moment connections

Single-Storey Steel Buildings is one of two design guides. The second design guide isMulti-Storey Steel Buildings.

The two design guides have been produced in the framework of the European project“Facilitating the market development for sections in industrial halls and low risebuildings (SECHALO) RFS2-CT-2008-0030”.

The design guides have been prepared under the direction of Arcelor Mittal,Peiner Träger and Corus. The technical content has been prepared by CTICM and SCI,collaborating as the Steel Alliance.




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ContentsPage No

FOREWORD i

SUMMARY v 1 INTRODUCTION 1

1.1 Steel as a construction material 1 1.2 Steel in single storey buildings 7

2 ADVANTAGES OF CHOOSING A STEEL STRUCTURE 8 2.1 Low weight 8 2.2 Minimum construction dimensions 9 2.3 Speed of construction 9 2.4 Flexibility and adaptability 10 2.5 A sustainable solution 11

3 FORM OF PRIMARY STEEL STRUCTURE 12 3.1 Structure types 12 3.2 Connections between columns and beams 26

4 BUILDING ENVELOPE 28 4.1 Cladding systems 29 4.2 Secondary steelwork 30 4.3 Roofs 30

5 FIRE SAFETY 33

6 OVERHEAD CRANES 34

7 CONCLUSIONS 36

8 FURTHER READING 37




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SUMMARY

This publication presents an introduction for architects to the use of steel in singlestorey steel-framed buildings. The primary application of such buildings is for industrialuse but single storey solutions are appropriate for many other applications. The

advantages of the use of steel, in terms of low weight, minimum constructiondimensions, speed of construction, flexibility, adaptability and sustainability areexplained. The primary forms of steel structure and the methods of cladding them areintroduced. It is noted that the requirements for fire resistance are usually modest, sinceoccupants can usually escape quickly in the event of fire. The influence of providing acrane inside a single storey building, in terms of the structural design, is brieflyaddressed.




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1 INTRODUCTION

1.1 Steel as a construction material

Steel is synonymous with modern architecture. Throughout the twentiethcentury, the material has inspired architects and engineers, for it combinesstrength and efficiency with unparalleled opportunities for sculpturalexpression.

The key attribute of steel is its high strength to weightratio, which gives remarkable spanning and load carrying ability. Steel lendsitself to prefabrication. Whole structures can be created in a factoryenvironment and then constructed quickly on site. Steel buildings are highlyadaptable, in that frames can be modified and altered. Costs are low, recyclingsimple and aesthetic opportunities rich and varied. As designers, fabricators

and constructors continually advance the boundaries of steel design, bothtechnically and expressively, steel has a crucial role in modern architecture.

Steel is basically a simple alloy of iron and carbon, but its properties can beenhanced and modified by the addition of other alloying elements and by themanufacturing process. The material is then made into sections, plate, or sheet,and these simple products used to produce structures and buildingcomponents.

Standard approaches have evolved for many types of single storey structuresbut they are not constraining: departures from norms are commonplace, for

steel lends itself to creative solutions. Modern architecture is rich withsolutions that defy simple categorization, even in single storey structures. These do not have to be utilitarian. They can be formed into gentle arcs orstartling expressed structure. Although greatest economy is often achieved withregular grids and standardization, steel structures offer outstanding opportunityfor architectural expression and outstanding design opportunities. Someillustrations of the dramatic structural forms that are possible in steelconstruction are shown inFigure 1.1toFigure 1.5.




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Figure 1.1 Single storey structure with curved roof

Figure 1.2 Single storey warehouse with exposed steelwork truss




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Figure 1.3 Single storey curved and cranked steelwork for an art gallery

Figure 1.4 Modern industrial building with curved steel roof




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Figure 1.5 Roof steelwork for a transport museum

Structural steel frames generally rely on the use of hot rolled steel sections:for such sections, the material is heated and passed as a billet or blank throughheavy rollers that gradually reduce and shape the cross-section whilst at thesame time increasing the length; the final shape is generally in a standardisedrange. Typical cross section ranges are shown inFigure 1.6.




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Section IPE UPE HD HE HLHeight (mm) 80 - 750 80 - 400 260 - 400 100 - 1000 620 - 1100

Figure 1.6 Typical hot rolled prof iles

For larger spans, deep beams or other structural members can be fabricated

from hot rolled sections and plate to form geometrically complex members.Hot rolled sections can be curved after manufacture, using bendingequipment, or be converted to perforated web profiles using a variety of approaches, some of which split the beam into two in such a way that the twoparts can be welded together as a deeper beam, with its spanning ability muchincreased.

Lighter steel sections can be formed by bending thin sheet steel into C orZ profiles. Normally this is done using either a cold rolling line (for standardsections) or by using a press or folding machine (for special sections).Common structural profiles range from around 80 mm to 350 mm deep, as

shown inFigure 1.7, and are particularly suitable for roof purlins and side railsthat support cladding, for lightweight frames, and as support to internal wallsand partitions.

Wide thin sheets can be formed by cold rolling into profiled cladding for roofsand walls (see typical profiles inFigure 1.8) and into profiled floor decking.




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Sheet thickness 1,5 – 3 mm

H

H

H 175 mm 195 mm 210 mm 240 mm 260 mm

Z shape


min. 30 mm max. 100 mm

max. 350 mm

min. 80 mm

H

C shape


max. 100 mmmin. 30 mm

H

max. 350 mm

min. 80 mm

U shape

Figure 1.7 Typical cold-rolled section profiles






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2 ADVANTAGES OF CHOOSING A STEELSTRUCTURE

A very large proportion of all industrial and commercial single storey buildingsutilise a steel structure, which demonstrates the cost-effectiveness of a steelsolution. Architects and engineers use steel not only as an economical solutionbut also to achieve:

low structural weight

minimum construction dimensions

a short construction time

flexibility in use

a sustainable solution

2.1 Low weight A steel structure has a relatively low self-weight compared to masonry orconcrete structures. This advantage not only reduces the foundations requiredfor the structure, but also means that the structure is lightweight, reducingmaterial delivery to the site. The off-site prefabrication of steel construction isa significant contribution to reduced transport of materials to site and reducedsite activities, minimising construction disruption and environmental impact.

Figure 2.1 The relatively low self weight of steel structures reduces material

delivery to site




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2.2 Minimum construction dimensionsSteel enables large spans to be constructed with relatively small constructiondepths. The typical construction solution of an insulated external envelopesupported on steel secondary members is a very well-developed solution,optimised over many years, leading to a structurally efficient and cost effectivesolution.

For pitched roofs or short span flat roofs, the construction depth of the roof beams or rafters can be as low as 1/40 of the span between columns. If internalcolumns are required for multi-span structures, they may be chosen to be smallmembers, or the internal columns may be provided on every second (or everythird) frame, maximising internal space and flexibility. Steelwork supportingthe external envelope may be very slender, as shown on Figure 2.2, providingthe opportunity for maximum natural lighting.

Figure 2.2 Slender construction takes up less space and results intransparent buildings.

2.3 Speed of constructionStructural steel components are pre-fabricated off site by a steelwork

contractor; any protective coating that is required is applied at this stage. Thesite activity is primarily an assembly operation, bolting steelwork partstogether, which leads to short construction periods. The building can be madeweather tight quickly, allowing the following trades early access to commencetheir work.

Modern fabrication is achieved using numerically controlled machines, withdata from three-dimensional electronic models of the complete structure.Modern fabrication is therefore extremely accurate, and errors that needrectification on site are rare. Three-dimensional building models can be usedby other trades to ensure that their own contribution (for example, the cladding,

or the mechanical and electrical services) can be properly co-ordinated with thestructural frame before the building is constructed. All these facilitiescontribute to minimizing the period from conception to completion.




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Figure 2.3 Prefabricated components are easily and rapidly connected on site

2.4 Flexibi lity and adaptability A steel structure is both flexible and adaptable – design in steel is certainly notlimited to rectangular grids and straight members, but can accommodatedramatic architectural intent, as shown in Figure 2.4.

Figure 2.4 Dramatic, expressed steelwork

Thanks to the numeric control of modern fabrication, components may bedesigned and fabricated to almost any shape desired. In most cases, a structurewith an irregular floor plan or curved components is manufactured as easily asa rectilinear design, although there will be cost implications of the morecomplex fabrication.




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The building can also be made adaptable for future changes in use. Column-free floor space facilitates future changes in internal layout, which is likely tohappen several times in the life of a structure. The building structure can bemodified, strengthened and extended. The facility to extend the structure atsome future stage can be incorporated into the original design and construction

details. The external envelope maybe renewed, upgraded or modified. Futureowners/users with different requirements can readily adapt a steel building totheir requirements.

2.5 A sustainable solutionSteel can be recycled any number of times without loss of quality or strength.Significant quantities of recycled steel are used in the manufacture of new steelproducts and there is a commercial value in scrap steel for this reason.Figure 2.5 shows scrap material being recycled to make new steel.

Steel building components are fabricated under controlled conditions withminimal waste (off-cuts are recycled as scrap). As the site activity is mainlyassembly, there is rarely any waste on site.

Steel structures can often be dissembled, as they are primarily bolted skeletalstructures. The steel members may reused in other structures – portal framesand similar structures are frequently dismantled and used at other locations.

Figure 2.5 Modern steel making technology has the ability to recycle scrap




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3 FORM OF PRIMARY STEEL STRUCTURE

Single storey steel buildings are generally built with an external cladding

envelope, supported in many cases on relatively short span secondary steelmembers, which are in turn supported on the primary steel structure. ThisSection describes the structural possibilities that may be considered andcomments on the type of structural sections that can be used.

3.1 Structure types There are four basic structural configurations that provide a clear interior spacefor a single storey building:

Rigid framed structures (portal frames and rigid-frame trusses)

Pinned frame beam-and-column structures

Cable-supported roofs

Arched roofs

For the first three configurations, the designer has the option of providingeither a flat roof or a pitched roof.

Typical spans and span/depth ratios for the primary roof members in pinnedand rigid framed buildings are given in Table 3.1.

Table 3.1 Typical spans and structural depths for single storey structuresStructure type Roof beam depth Typical span range

Pinned frames

Simple beam span/30 to span/40 Up to approximately 20 m

Fabricated Beam span/20 to span/25 Up to approximately 30 m

Perforated web beam span/20 to span/60 Up to approximately 45 m

Truss roof (pitched) span/5 to span/10 Up to approximately 20 m

Truss roof (flat) span/15 to span/20 Up to approximately 100 m

Rigid frames

Portal frame span/60 15 m – 45 m

Truss roof (flat) span/15 to span/20 Up to approximately 100 m

3.1.1 Rigid-framed structures

Rigid frames are achieved by providing a rigid (moment resisting) connectionbetween the ends of the roof beams (or trusses) and the columns. The stiff frame that is created is much more efficient in carrying the imposed loads onthe roof than a simply supported roof member (with nominally pinnedconnections at its ends) and the frame also provides resistance against windforces on the sides of the building. Because the frames are self-supporting inthe plane of the frame, the bracing in the roof can be reduced, compared to a

structure with simply supported roof beams.

Rigid framed structures broadly fall into two categories, portal framedstructures and truss framed structures.




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Portal frames

Portal frames typically use hot-rolled I-section beams and columns for the roof rafters and supporting columns, although cold formed sections may beadequate for small span structures. Portal frames come in a variety of differentshapes and sizes, with flat and pitched roofs.

A typical configuration is shown in Figure 3.1. The roof and wall cladding issupported on purlins and side rails that span between the portal frames. Bracingis not needed between every frame but is needed in at least one bay to transferlongitudinal forces (normal to the frames) to the side walls and thus to groundlevel.

In some special design situations, the cladding can be used as the bracing – thisis known as stressed skin design. The design of the cladding and the fixings tothe supporting members will be assessed by the structural engineer. In mostcases, bracing will be provided that does not rely on the sheeting.

Figure 3.1 Typical structural configuration of a portal frame structure




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25 - 40 m

6 m

6°

6 m

25-30 m (a) Portal frame – medium span (b) Curved portal frame

8 m

8 m 9 m 8 m

3.5 m

25 m

8 m

6°

(c) Portal frame with mezzanine floor (d) Portal frame with overhead crane

25 m

6 m

6°

(e) Two bay portal

frame

10 m

8 m

3.5 m

6°6°

(f) Portal frame withintegral office

40 m

6 m

10° 3.00°

(g) Mansard portal

frame

Figure 3.2 Forms of portal frame

Portal frames typically have straight rafters, as shown in Figure 3.3. The samestructural principles can be followed to form a portal frame with a curvedrafter, as shown in Figure 3.4. In each case, the connection of the rafter to thecolumn is substantial, and usually the rafter is haunched locally to the column.

The dimensions of the haunch should be allowed for when considering theclear height requirements.




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Figure 3.3 Pitched roof portal frame

Figure 3.4 Curved roof portal frame




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Figure 3.5 Typical roof and wall bracing in portal framed structu res

In most cases, the rafter (and possibly the column) will need local restraints, asshown on Figure 3.6. In some countries, special provision must be made whenusing this form of restraint, to ensure that the purlins align with the roof bracing system. The location of these restraints will be specified by thestructural engineer.

Figure 3.6 Stabilizing the bottom flange of a roof beam

Rigid framed trusses

When flat trusses are used, both top and bottom chords can easily be connectedto the supporting columns, thus creating a rigid frame. For larger spans, roof

trusses provide an effective and economic alternative. Typical flat truss shapesare shown in Figure 3.7, and a truss roof is illustrated in Figure 3.8.




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Figure 3.7 Typical truss shapes

Figure 3.8 Rigid frame flat truss (N-type)

In some situations, the columns are also of lattice form and then the buildingconfiguration is typically as shown in Figure 3.9.

Figure 3.9 Rigid frame flat truss with lattice columns




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The lateral stability of the top chords of trusses is usually provided by thepurlins (and by one panel of bracing, as for portal frames) but where stressedskin design is permitted, it may provide the restraint without bracing, as shownin Figure 3.10.

Figure 3.10 Roof cladding acting as stressed skin in a rigid-framed truss roof

3.1.2 Pinned frame beam and column structures

In a pinned frame beam and column structure, the basic configuration is aseries of parallel beams, each supported by columns at its ends, with a pinnedor flexible connection between the beam and the column. Bracing has to be

provided in the roof to transfer horizontal forces due to wind loads to the endand side walls; the walls are braced to transfer the forces to the foundations.(Alternatively, some countries allow the roof cladding to act as a ‘stressedskin’, thus largely eliminating the need for separate bracing.) A typicalstructural configuration is shown in Figure 3.11.

Figure 3.11 Typical structural configuration for a beam and column structure




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There are numerous options for the beams:

Hot rolled sections (I-beams)

Plate girders

Fabricated beams with holes in the webs

Trusses

Hot rolled I section beams

The most common type of beam and column structure uses hot rolled steel Isections for both beams and columns. These sections are produced inaccordance with international standards and there are design tables available toallow for an easy selection of section size to suit the loading. The mostcommon section sizes are readily available from stockists and can be ordered atshort notice.

Deep sections with relatively narrow flanges are preferred for roof beams, asshown in Figure 3.12, where they primarily resist bending. Columns, whichprimarily resist compression, are usually thicker, shallower sections with widerflanges.

The span/depth ratio for the roof beams is typically 30 to 40 for spans up to20 m.

Figure 3.12 Pinned frame beam and column structure

Plate girders

Plate girders are built up beams consisting of two flange plates, welded to aweb plate to form an I-section. This type of beam offers a solution when thestandard I and H beams are not suitable. The section dimensions are chosen tosuit the design bending moments and shear forces; the beams can be profiled inelevation, as shown in Figure 3.13.

The span/depth ratio is typically 20 to 25 for spans up to 30 m.

An alternative that is sometimes used for large spans, to reduce the thickness of the web plate, is the use of a corrugated plate (profiled in plan). The span/depthratio with a profiled web plate is typically 30 to 40 for spans up to 100 m.




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Figure 3.13 Tapered plate girders

Plate girders are likely to be more expensive than hot-rolled standard sections.

Beams with web openings

Because roof beams generally carry relatively light uniformly distributed loads,beam sections that span large distances can be created by fabricating sectionswith openings in the webs. Historically, the first beam of this type was thecastellated beam, with hexagonal holes. Now beams with circular openings arecommonly used.

In both cases, the beam is fabricated from a rolled I section by cutting along theweb, to a special profile, separating the two halves and then displacing one half relative to the other and welding them back together. This is illustrated inFigure 3.14. The major advantage of this type of beam is the weight reduction:approximately 30% less than a beam with a solid web of similar depth andbending resistance.

An example of the use of beams with circular openings is shown inFigure 3.15.

Beams with web openings are less suitable for heavy concentrated loads.

The span/depth ratio is typically 30 for spans up to 50 m.

Hexagonal holes

Circular holes

Figure 3.14 Fabrication of beams with web openings




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Figure 3.15 Beams with circular web openings

Trusses

Trusses are a triangulated assembly of members. Two basic configurations areused in single storey buildings – pitched roof trusses and ‘flat’ trusses of nearuniform depth.

Pitched roof trusses

A variety of pitched roof truss forms are used in pinned frames, as illustrated in

Figure 3.16.

The trusses illustrated in Figure 3.16 are commonly fabricated from T andangle sections, and are used to create a sloped roof. The large (mostly unused)space between the trusses may be considered a disadvantage, requiring heatingand raising the overall height of the structure, but it is a cost effective solutionfor modest spans and provides space for services.

Because these trusses are used with a steeply sloping roof, the span/depth ratiois typically 5 to 10 for spans up to 20 m




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Fink or Polonçeau truss(small span)

Fink or Polonçeau truss(large span)

Belgian truss

English truss

Mansard truss

Figure 3.16 Types of pitched roof truss

Flat trusses

Flat trusses are used mainly in rigid frames (see Section 0 for a morecomprehensive description) but they are also employed in pinned frames – anexample is shown in Figure 3.17.

Figure 3.17 Flat truss in pinned frame building

Trusses typically have a greater depth than single beams or plate girders. Thedeflection of a truss is modest, and can be controlled, making trusses especiallysuitable when significant loads have to be supported from the roof structure, orwhen a flat (or nearly flat) roof is to be provided. The larger depth of thetrusses increases the dimensions of the façade, but also provides space forservices to be placed in the roof structure instead of below.

The weight of a trussed roof structure per unit area of roof in general is lessthan that of single beam girders, but the fabrication costs are higher. Trusses




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may be exposed in the completed structure, which may increase the fabricationcosts if, for example, hollow sections are used for the members.

The span/depth ratio for flat trusses is typically 15 to 20 for spans up to 100 m.

Trusses are usually planar and will generally require bracing of some form toprovide stability. As an alternative, three-dimensional trusses can be created, asshown in cross section in Figure 3.18 and illustrated in Figure 3.19. This formof truss is generally expensive to fabricate, because of the complexintersections of the internal members.

The span/depth ratio for three-dimensional trusses is typically 16 to 20 forspans over 50 m.

Triangular truss (with circular hollow sections) Triangular truss (with rectangular hollowsections)

Figure 3.18 Three dimensional triangular trusses

Figure 3.19 Three-dimensional trusses supporting a roof

3.1.3 Cable stayed roofs

In a cable-stayed structure, tensile members (wire ropes or bars) are providedto give intermediate support to members such as roof beams, thus allowing

those members to be reduced in size. The stays need to be supported bycolumns or masts and those members need to be anchored or braced with otherstays. The bracing arrangement is usually very conspicuous and the aesthetics




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of the building must be considered carefully. An example of a cable stayedbuilding structure is shown in Figure 3.20.

Figure 3.20 Cable stayed roof beams of a storage facil ity

Alternative configurations for a flat roof building are shown in Figure 3.21.

Cable stayed configurations are most economical for spans between 30 m and90 m.

As most of the structure is outside of the building, maintenance costs can behigh. Care must be taken in detailing the waterproofing where the stays passthrough the cladding.

1

2

3

1 2 3

Roof beam Bendingmoment

+ ++ +

Compressionforce

-- - +

Anchorage Tensile force ++ -- --

Figure 3.21 Comparison of the three main configurations for cable stayed structures

The arrangement of the structure has a significant effect on the internal forcesand therefore the member sizes. The building arrangement should be developedin collaboration with the structural engineer.




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3.1.4 Arches

Arches have a parabolic or circular form, as illustrated in Figure 3.22. Uniformloading is carried by compression in the arch members; modest bendingmoments are induced by non-uniform loading and point loads. Thecompression forces must be resisted by horizontal forces in the foundation of

the building – or by tie members between the foundations, as shown inFigure 3.22.

Arch members can be formed by cold bending I-section beams.

The span/depth ratio for the arch members is typically between 60 and 75 forspans up to 50 m.

An example of an arched roof building is shown in Figure 3.23.

Tie rod connecting supports

Both supports fixed

Figure 3.22 Methods of supporting arch members

Figure 3.23 Fire brigade station




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3.2 Connections between columns and beams3.2.1 Moment-resisting connections

In a portal frame structure, the connections between beams and columnstransfer bending moments, as well as shear and axial forces, and they must be

designed as rigid connections.

A rigid connection typically has a full depth end plate. The roof beam is oftenhaunched locally and the column web is stiffened in order to resist the localforces from the end of the roof beam. In general, stiffeners should be avoided if possible, as they add significant fabrication cost.

1

2

3

1 Extended end plate2 Extended end plate with stiffener3 Haunched connection with

stiffener

Figure 3.24 Rigid bolted connections between roof beams and columns

Connections between trusses and columns are usually achieved by end plateson the top and bottom chords, bolted to the face of the column. A typicalexample is illustrated in Figure 3.25.




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Figure 3.25 Truss-column connection in a rigid framed structu re

3.2.2 Nominally pinned connections

In a beam and column structure, the connections are nominally pinned and arenot assumed to transfer any moments between the connected members.Externally applied actions, such as wind forces, must be resisted by bracingsystems. The bracing system may be steel bracing, or a stiff core. For singlestorey structures, a system of steel bracing is almost universally adopted.

Pinned connections are relatively easy (and cheap) to fabricate. Typical

connections use partial depth end plates, fin plates or angle cleats; the membersare bolted together on site.

2

1

3

1 End plate connection2 Angle cleat connection3 Fin plate connection

Figure 3.26 Nominally pinned bolted connections




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4 BUILDING ENVELOPE

The steel structure of a single storey building generally comprises three

principal components: a primary construction (roof beams and columns, withbracing); secondary steelwork, such as purlins and side rails that support theroof panels and wall cladding; and the roof panels and cladding themselves.

The roof panels and cladding are generally referred to as the building envelope.

The building envelope provides a weather-tight enclosure to the building space.In most cases, it also provides thermal insulation from the exteriorenvironment. The exterior appearance is often a major consideration in thechoice of the form of the envelope. The architect must therefore choose asystem that balances the demands of sustaining actions such as wind pressureand (on flat or near-flat-roofs) imposed loads, of achieving thermal

performance that meets criteria for low energy use, and of producing anappearance that meets the client’s aspirations.

A single type of cladding system is often used for both roof and walls.

Detailing will be an important element of envelope design. Drainage systemsthat do not block or leak are essential and the integration of openings (windowsand doors) with the cladding must not compromise thermal insulation.

A striking example of using coloured profiled sheeting is shown in Figure 4.1.

Figure 4.1 Car repair workshop with steel roof and façade




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4.1 Cladding systems The principal options for cladding systems are:

Profiled steel sheeting

- Single-skin

- Double-skin, built up on site from a liner panel, insulation and an outersheet

- Composite sandwich panels, pre-fabricated off site from an inner sheet,and outer sheet and insulation.

Steel sheeting with insulation, covered by a waterproof membrane –commonly used on flat roofs.

Wooden panels/decking

Precast concrete slabs

Blockwork (for walls)

4.1.1 Profiled sheet cladding

The basic types of profiled steel sheeting system, used in roofs and walls, aresummarized in Table 4.1.

Table 4.1 Basic types of cladding system

System Insulated? Benefits

Built upsystems

yes free choice for exterior profiled sheeting

high fire resistance

good sound proofing and good sound absorption

fast construction, with simple mechanical fasteners

Compositepanels

yes fast construction

fully prefabricated

singlesheeting

no cheap and fast construction

easy to dismantle

large freedom of form

4.1.2 Precast concrete slabs

For flat roofs with significant imposed loads, cellular concrete slabs provide

both a relatively easily installed building component and a thermal insulationlayer.

Precast concrete slabs (either hollow core or sandwich panel) provide thenecessary strength where there are heavy snow loads or a heavy roof isrequired for safety reasons (e.g. resisting explosive pressures in accidentalsituations). However, precast slabs are much heavier than profiled steelcladding and the primary steel structure must be correspondingly stronger.

4.1.3 Blockwork

Blockwork construction is often used for the walls of single storey buildings,

either full height or partial height (with sheet cladding for the top of the wall). The blockwork provides insulation and robustness; it may also be chosen forappearance.




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4.2 Secondary steelwork Secondary beams are used when the spacing of the main beams or trusses is toolarge for the cladding or roof panels to span between them, or where thecladding spans parallel to the main beams, which is usually the case withpitched roofs.

For these secondary members, there is a choice between cold-formed and hot-rolled steel sections. The profiles of typical cold formed sections are shown inFigure 4.2. A cold formed section can be up to 30% lighter than a hot rolledsection.

1 2 3 4 C profile

ℓ max =10 m

140 mm <h <300 mm

profile

ℓ max =12 m

140 mm <h <300 mm

profile

ℓ max =16 m

250 mm <h <420 mm

Z profile

ℓ max =12 m

120 mm <h <400 mm

Figure 4.2 Typical cross sections of cold formed beams

Cold formed sections are manufactured from galvanized steel and this normally

provides sufficient protection against corrosion in the internal environment of the building (an exception might be, for example, in aggressive environmentssuch as cattle sheds, where ammonia is present).

Secondary members of cold-formed sections are used at relatively low spacing,typically between 1,6 m and 2,5 m. Very long secondary members can befabricated as small trusses.

4.3 Roofs

The choice between a flat roof and a pitched roof often depends on theparticular preferences in the local or national region. Some countries favour flatroofs that are able to sustain significant imposed loading, other countriesfavour pitched roofs that facilitate drainage and which are subject to only verymodest imposed loading. Clearly, the type of cladding that is appropriatedepends on those choices and circumstances.

4.3.1 Pitched roofs

The slope of a pitched roof also depends on local circumstances and custom. Aslope of at least 10% (6°) is normally provided.

Where profiled sheeting is used, the profiles run down the slope, to facilitatedrainage. Insulation must therefore be below the outer sheeting (possibly as acomposite panel). The sheeting is supported on purlins spanning between the




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roof beams and is fastened with screws or bolts. The lapped sheets do notrequire a waterproof membrane; the panels are simply lapped, the higher abovethe lower on the slope.

A typical arrangement of a pitched roof at the eaves is shown in Figure 4.3. It

is important that the drainage system is adequate for the run-off from the wholeroof.

1

1

3

2

1 Sandwich roof panel and sandwich façade panel

2 Roof slope >6 3 Hot rolled or cold formed section

Figure 4.3 Insulated sloped roof

4.3.2 Flat roofs

Where the roof is flat, it must be fully watertight against standing water and itis therefore usual to apply a waterproofing membrane on its top surface.

Where profiled steel sheeting is used, it is typically a deep profile, spanning

between the primary structural members. Insulation is then placed on top of thesheeting, fixed with bolts or screws. The waterproof membrane is then appliedon top of the insulation. An example is shown in Figure 4.4.

Where flat roofs are provided, there is a risk of ponding. Water can accumulatein the central area if the roof deflects significantly. If there is inadequatedrainage, water can also be retained by kerbs or other details around the edgeof the roof. It is vitally important to minimise the risk of ponding byprecambering the roof and providing adequate drainage.




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1

2

3

7

6

5

4

1 Insulation2 Liner panel3 Exterior profiled sheeting4 Screw

5 Insulation6 Additional metal strip7 Single roof sheeting

Figure 4.4 Insulated flat roof




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5 FIRE SAFETY

Requirements for fire safety are defined by national regulations but there are

recognised international rules for assessing the fire resistance of steelstructures. The minimum level of safety for structural fire design aims toprovide an acceptable risk associated with the safety of building occupants, firefighters and people in the proximity of the building. Levels of safety can beincreased to protect the building contents, the building superstructure, heritage,business continuity, corporate image of the occupants or owner, and theenvironmental impact.

Requirements are usually expressed in relation to:

Spread of fire: combustibility of the materials expressed in relation to timeuntil flashover. It is classified as A1 (flashover not possible) down to E

(flashover in less than 2 minutes) and F (not tested).

Smoke intensity: materials are classified from class A2 to F depending onthe smoke produced on combustion.

Fire resistance: the period of time for which a structural component canperform in a standardized fire test. The three criteria of load-bearingcapacity, integrity and insulation (commonly expressed as R, E and I) areconsidered and the rating is expressed as R30, R60 etc. where the numberrefers to the period in minutes.

In order to achieve the required fire safety level in a single storey building the

following items should be taken in account:

regulatory requirements

fire partitioning

fire spreading

escape routes

Single storey buildings often have very modest requirements for fire resistancebecause occupants can escape quickly. The main requirement is often theprevention of fire spread to adjacent properties.

To protect contents, especially in large production facilities and warehouses,partitioning may be needed or, where that is not feasible, alternative measuresmay be taken, such as the installation of a sprinkler system.




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6 OVERHEAD CRANES

Certain industrial buildings require overhead cranes – examples are printing

shops (for moving rolls of paper) and engineering shops (for moving heavyequipment and components). An example is shown in Figure 6.1.

Most overhead cranes use single or twin beams spanning across the buildingand with a hoist mounted on the beams. The crane beams are supported onrunway beams that run the length of the building. The crane serves the wholefloor by moving along the runway beams and by moving the hoist along thecrane beams (Figure 6.2).

Incorporating an overhead crane in a building always influences the design of the building structure, even when the hoisting capacity is very modest. A keydesign consideration is to limit the spread of the columns at the level of thecrane. For this reason, portal frames are not appropriate for heavy cranes aslimiting the column movement becomes uneconomic. Crane use also results inhorizontal forces from movement of the loads, so additional bracing is usuallyprovided.

A crane with a lifting capacity up to a safe working load of about 10 tons(100 kN) can usually be carried on runway beams that are supported off thecolumns that support the roof. For larger cranes, it is more economical to useseparate columns (or vertical trusses) to support the runway beams and avoidexcessive loads on the building structure.

Figure 6.1 Heavy crane in a large industr ial building




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2 1

3

7

6

4

5

13

8

910

12

11

min. 500 mm

1 Lifting2 Hoist drive3 Crane drive4 Motor drive5 Hoist

6 Crane beams7 Wheel cabinet8 Hoist9 Crane beam

10 Runway beam11 Console12 Hook13 Crane operation

Figure 6.2 Typical overhead crane with gantry and hois t




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7 CONCLUSIONS

Steel is a versatile material that allows the architect and engineer to design any

type of structure, ranging from orthodox portal frames for industrial use to stateof the art buildings with architectural features, unorthodox shapes or any otherrequirements the stakeholders might have.

Structural steel design is familiar and efficient, providing elegant cost effectivesolutions. Structural steel can be combined with other materials to achieve thedesired look, properties or functionality.

Fabrication of a steel building is carried out in a workshop, ensuring a highquality product and contributing to a low waste, sustainable solution.Standardised details and forms of construction are available which allow fasterection on site, with minimised disruption to the surroundings.

Steel has a very high resistance to weight ratio, resulting in a light, attractivesolution with minimal intrusion into the working area of the structure. Thetransportation of highly prefabricated elements reduces deliveries to site, whichis especially important in congested areas, such as city centres. The structuralefficiency of steelwork results in lower loads being transferred to thefoundations, leading to further economy.

Long span buildings can easily be designed in steel, resulting in large clearareas. This increases the functionality of the structure, offering flexibility of building use. Steel buildings are adaptable and may be easily extended, makingrefurbishment of the building a realistic solution for future use, instead of demolition.

Steel has excellent sustainability credentials. Steel buildings can easily bedismantled and reused. The steel can always be recycled without any loss of strength, minimising the amount of raw material required.

Steel’s low weight, sustainability and versatility, make steel the optimumchoice for any type of building.




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8 FURTHER READING

Best Practice in Steel Construction: Industrial Buildings, Guidance for

Architects, Designers and Constructors RFCS project deliverable for Euro-BuildAvailable from the Steel Construction Institute, UK It can be downloaded fromwww.eurobuild-in-steel.com





Part 2: Concept Design









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FOREWORD

This publication is a second part of a design guide, Single-Storey Steel Buildings.




Part 3: Actions











The design guides have been prepared under the direction of Arcelor Mittal, Peiner Träger and Corus. The technical content has been prepared by CTICM and SCI,collaborating as the Steel Alliance.






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ContentsPage No

FOREWORD iii

SUMMARY vi 1 INTRODUCTION 1

1.1 Hierarchy of design decisions 1 1.2 Architectural design 2 1.3 Choice of building type 6 1.4 Design requirements 9 1.5 Sustainability 12

2 CASE STUDIES ON SINGLE STOREY BUILDINGS 14 2.1 Manufacturing hall, Express Park, UK 14 2.2 Supermarket, Esch, Luxembourg 15 2.3 Motorway Service station, Winchester, UK 16 2.4 Airbus Industrie hanger, Toulouse, France 17 2.5 Industrial hall, Krimpen aan den Ijssel, Netherlands 17 2.6 Distribution Centre and office, Barendrecht, Netherlands 18

3 CONCEPT DESIGN OF PORTAL FRAMES 19 3.1 Pitched roof portal frame 20 3.2 Frame stability 22 3.3 Member stability 23 3.4 Preliminary Design 25 3.5 Connections 27 3.6 Other types of portal frame 29

4 CONCEPT DESIGN OF TRUSS BUILDINGS 35 4.1 Introduction 35 4.2 Truss members 36 4.3 Frame stability 38 4.4 Preliminary design 39 4.5 Rigid frame trusses 40 4.6 Connections 40

5 SIMPLE BEAM STRUCTURES 42

6 BUILT-UP COLUMNS 43

7 CLADDING 45 7.1 Single-skin trapezoidal sheeting 45 7.2 Double-skin system 45 7.3 Standing seam sheeting 47 7.4 Composite or sandwich panels 47 7.5 Fire design of walls 47

8 PRELIMINARY DESIGN OF PORTAL FRAMES 49 8.1 Introduction 49 8.2 Estimation of member sizes 49

REFERENCES 52




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SUMMARY

This publication presents information necessary to assist in the choice and use of steelstructures at the concept design stage in modern single storey buildings. The primarysector of interest is industrial buildings, but the same information may also be used in

other sectors, such as commercial, retail and leisure. The information is presented interms of the design strategy, anatomy of building design and structural systems that arerelevant to the single storey buildings. Other parts in the guide cover loading, theconcept design of portal frames, the concept design of trusses and cladding.




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1 INTRODUCTION

Single storey buildings use steel framed structures and metallic cladding of all

types. Large open spaces can be created, which are efficient, easy to maintainand are adaptable as demand changes. Single storey buildings are a “core”market for steel. However, the use of steel in this type of construction varies ineach European country.

Single storey buildings tend to be large enclosures, but may require space forother uses, such as offices, handling and transportation, overhead cranes etc.

Therefore, many factors have to be addressed in their design.

Increasingly, architectural issues and visual impact have to be addressed andmany leading architects are involved in modern single storey buildings.

This section describes the common forms of single storey buildings that maybe designed and their range of application. Regional differences may existdepending on practice, regulations and capabilities of the supply chain.

1.1 Hierarchy of design decisions The development of a design solution for a single storey building, such as alarge enclosure or industrial facility is more dependent on the activity beingperformed and future requirements for the space than other building types, suchas commercial and residential buildings. Although these building types are

primarily functional, they are commonly designed with strong architecturalinvolvement dictated by planning requirements and client ‘branding’.

The following overall design requirements should be considered in the conceptdesign stage of industrial buildings and large enclosures, depending on thebuilding form and use:

Space use, for example, specific requirements for handling of materials orcomponents in a production facility

Flexibility of space in current and future use

Speed of construction Environmental performance, including services requirements and thermal

performance

Aesthetics and visual impact

Acoustic isolation, particularly in production facilities

Access and security

Sustainability considerations

Design life and maintenance requirements, including end of life issues.

To enable the concept design to be developed, it is necessary to review theseconsiderations based on the type of single storey building. For example, therequirements for a distribution centre will be different to a manufacturing




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facility. A review of the importance of various design issues is presented in Table 1.1 for common building types.

Table 1.1 Important design factors for single storey buildings

Type of singlestorey buildings

S p a c e r e q u

i r e m e n

t s

F l e x

i b i l i t y o

f u s e

S p e e

d o

f c o n s

t r u c

t i o n

A c c e s s a n

d S e c u r i t y

S t a n

d a r d

i z a

t i o n o

f c o m p o n e n

t s

E n v

i r o n m e n

t a l p e r f o r m a n c e

A e s

t h e

t i c s a n

d v

i s u a

l i m p a c

t

A c o u s

t i c i s o

l a t i o n

D e s

i g n

l i f e ,

m a

i n t e n a n c e a n

d r e - u

s e

High baywarehouses

Manufacturing facility

Distribution centres

Retail superstores

Storage/cold storage

Office and lightmanufacturing

Processing facility

Leisure centres

Sports halls

Exhibition halls

Aircraft hangars

Legend: No tick =Not important =important =very important

1.2 Architectural designModern single storey buildings using steel are both functional in use and aredesigned to be architecturally attractive. Various examples are presented belowtogether with a brief description of the design concept. A variety of structural

solutions are possible, which are presented in Sections 2 and 3.

1.2.1 Building form

The basic structural form of a single storey building may be of various generictypes, as shown in Figure 1.1. The figure shows a conceptual cross-sectionthrough each type of building, with notes on the structural concept, and typicalforces and moments due to gravity loads.




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Simple beam

Portal frame

Truss

Portal truss

Figure 1.1 Structural concepts

The basic design concepts for each structural type are described below:

Simple roof beam, supported on columns.

The span will generally be modest, up to approximately 20 m. The roof beammay be pre-cambered. Bracing will be required in the roof and all elevations, toprovide in-plane and longitudinal stability.

Portal frame

A portal frame is a rigid frame with moment resisting connections to provide

stability in-plane. A portal frame may be single bay or multi bay as shown inFigure 1.2. The members are generally plain rolled sections, with the resistanceof the rafter enhanced locally with a haunch. In many cases, the frame willhave pinned bases.

Stability in the longitudinal direction is provided by a combination of bracingin the roof, across one or both end bays, and vertical bracing in the elevations.If vertical bracing cannot be provided in the elevations (due to industrial doors,for example) stability is often provided by a rigid frame within the elevation.

Trusses

Truss buildings generally have roof bracing and vertical bracing in eachelevation to provide stability in both orthogonal directions, as in Figure 1.4.

The trusses may take a variety of forms, with shallow or steep external roof slopes.




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A truss building may also be designed as rigid in-plane, although it is morecommon to provide bracing to stabilise the frame.

Other forms of construction

Built–up columns (two plain beams, connected to form a compound column)

are often used to support heavy loads, such as cranes. These may be used inportalised structures, but are often used with rigid bases, and with bracing toprovide in-plane stability.

External or suspended support structures may be used, as illustrated inFigure 1.6, but are relatively uncommon.

Figure 1.2 Multi bay portal frame structu re




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Figure 1.3 Use of curved cellular beams in a portal frame

Figure 1.4 Roof trusses and built-up columns




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Figure 1.5 Curved cellular beams used in a leisure centre

Figure 1.6 External structure supporting a single storey building

1.3 Choice of bui lding typePortal frames are considered to be a highly cost-effective way to provide asingle storey enclosure. Their efficiency depends on the method of analysis,

and the assumptions that are made regarding the restraint to the structuralmembers, as shown in Table 1.2. The assumptions about member stability mayvary between countries.




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Table 1.2 Effic ient portal frame design

Most Efficient Less Efficient

Analysis using elastic-plastic software Elastic analysis

Cladding considered to restrain the flange of the purlins and side rails

Purlins and side rails unrestrained

Purlins and side rails used to restrain bothflanges of the hot-rolled steelwork

The inside flange of the hot rolled steelwork isunrestrained

Nominal base stiffness utilised Nominal base stiffness ignored

The reasons for choosing simple beam structures, portal frames or trusses areshown in Table 1.3.

Table 1.3 Comparison of basic structu ral forms for single storey buildings

Simple beam Portal frame Truss

Advantages

Simple design Long span Very long spans possible

Designed to be stablein-plane

Heavy loads may be carried

Member sizes and haunchesmay be optimised forefficiency

Modest deflection

Disadvantages

Relatively short span Software required for efficientdesign

Generally more expensivefabrication

Bracing needed for in-planestability

Limited to relatively lightvertical loading, and modest

cranes to avoid excessivedeflections

Generally bracing is used forin-plane stability

No economy due to continuity

1.3.1 Cladding types

The main types of roofing and wall cladding used in single storey buildings aredescribed as follows:

Roofing

‘Built-up’ or double layer roofing spanning between secondary memberssuch as purlins.

Composite panels (also known as sandwich panels) spanning betweenpurlins.

Deep decking spanning between main frames, supporting insulation, withan external metal sheet or waterproof membrane.

Walls

Sheeting, orientated vertically and supported on side rails.

Sheeting or structural liner trays spanning horizontally between columns.

Composite or sandwich panels spanning horizontally between columns,eliminating side rails.

Metallic cassette panels supported by side rails.




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Different forms of cladding may be used together for visual effect in the samefaçade. Examples are illustrated in Figure 1.7, Figure 1.8 and Figure 1.9.Brickwork is often used as a “dado” wall below the level of the windows forimpact resistance, as shown in Figure 1.8.

Figure 1.7 Horizontal spanning sheeting

Figure 1.8 Large windows and use of composite panels with “ dado” brick wall




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Figure 1.9 Horizontal composite panels and ‘ribbon’ windows

1.4 Design requirementsDesign requirements for single-span buildings are presented as follows:

1.4.1 Actions

Permanent actions

Permanent actions are the self weight of the structure, secondary steelwork andcladding. These may be calculated from EN 1991-1-1.

Typical weights of materials used in roofing are given in Table 1.4.

If a roof only carries normal imposed roof loads (i.e. no suspended machinery

or similar) the self weight of a steel frame is typically 0,2 to 0,4 kN/m2

whenexpressed over the plan area of the roof.




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Table 1.4 Typical weights of roof ing materials

Material Weight (kN/m2 )

Steel roof sheeting (single skin) 0,07 – 0,12

Aluminium roof sheeting (single skin) 0,04

Insulation (boards, per 25 mm thickness) 0,07

Insulation (glass fibre, per 100 mm thickness) 0,01

Liner trays (0,4 mm – 0,7 mm thickness) 0,04 – 0,07

Composite panels (40 mm – 100 mm thickness) 0,1 – 0,15

Steel purlins (distributed over the roof area) 0,03

Steel decking 0,2

Three layers of felt with chippings 0,29

Slates 0,4 – 0,5

Tiling (clay or plain concrete tiles ) 0,6 – 0,8

Tiling (concrete interlocking) 0,5 – 0,8

Timber battens 0,1

Variable actions

Variable actions should be determined from the following Eurocode parts:

EN 1991-1-1 for imposed roof loadsEN 1991-1-3 for snow loadsEN 1991-1-4 for wind actions

EN 1991-1-1 recommends a uniform load of 0,4 kN/m2 for roofs not accessibleexcept for normal maintenance and repair (category H). A point load of 1,0 kNis also recommended, but this will only affect the design of the sheeting andnot the main structural elements.

EN 1991-1-3 includes several possible load cases due to snow, includinguniform snow and drifted snow, which typically occurs in valleys, behindparapets etc. There is also the possibility of exceptional snow loads.

The value of the snow load depends on the building’s location and heightabove sea level.

EN 1991-1-4 is used to determine wind actions, which depend on altitude,

distance from the sea and the surrounding terrain.

The determination of loads is covered in detail in a separate chapter of thisguidance.

Loading due to services will vary greatly, depending on the use of the building.A typical service loading may be between 0,1 and 0,25 kN/m2 as measured onplan, depending on the use of the building. If air handling units or othersignificant equipment loading is to be supported, the service load should becalculated accurately.

1.4.2 Temperature effectsIn theory, steel frames expand and contract with changes in temperature. Often,the temperature change of the steelwork itself is much lower than any change






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steelwork on that elevation, and designing the elevation steelwork to resist theforces applied by any other parts of the structure that have collapsed.

For many building types, such as exhibition halls, fire engineering analysismay be carried to out demonstrate that active protection measures are effective

in reducing fire temperatures to a level where the structure is able to resist theapplied loads in the fire scenario without additional fire protection.

1.5 Sustainability Sustainable construction must address three goals:

• Environmental criteria

• Economic criteria

• Social criteria

These three criteria are met by construction in steel:

Environmental criteria

Steel is one of the most recovered and recycled materials. Some 84% isrecycled with no loss of strength or quality, and 10% reused. Beforedemolishing a structure, extending a building’s life is generally morebeneficial. This is facilitated by steel construction, since large column-freespaces give flexibility for change in use. Advances in the manufacturing of rawmaterials means that less water and energy is used in production, and allowsfor significant reductions in noise, particle and CO2 emissions.

Economic criteria

Steel construction brings together the various elements of a structure in anintegrated design. The materials are manufactured, fabricated and constructedusing efficient production processes. The use of material is highly optimisedand waste virtually eliminated. The structures themselves are used for allaspects of modern life, including logistics, retail, commercial, andmanufacturing, providing the infrastructure on which society depends. Steelconstruction provides low investment costs, optimum operational costs andoutstanding flexibility of building use, with high quality, functionality,aesthetics and fast construction times.

Social criteria

The high proportion of offsite fabrication in steel buildings means that workingconditions are safer, controlled and protected from the weather. A fixedlocation for employees helps to develop communities, family life and the skills.Steel releases no harmful substances into the environment, and steel buildingsprovide a robust, safe solution.

Single storey structures

The design of low-rise buildings is increasingly dependent on aspects of sustainability defined by criteria such as:

Efficient use of materials and responsible sourcing of materials

Elimination of waste in manufacturing and in construction processes




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Energy efficiency in building operation, including improved air-tightness

Measures to reduce water consumption

Improvement in indoor comfort

Overall management and planning criteria, such as public transportconnections, aesthetics or preservation of ecological value.

Steel framed buildings can be designed to satisfy all these criteria. Some of therecognised sustainability benefits of steel are:

Steel structures are robust, with a long life. Properly detailed andmaintained, steel structures can be used indefinitely

10% of structural steel sections are re-used[1]

Approximately 95% of structural steel sections are recycled

Steel products can potentially be dismantled and reused, particularlymodular components or steel frames

Steel structures are lightweight, requiring smaller foundations than othermaterials

Steel is manufactured efficiently in factory controlled processes

All waste is recycled in manufacture and no steel waste is produced on site

Construction in steel maximises the opportunity and ease of extendingbuildings and change of use

High levels of thermal insulation can be provided in the building envelope

Prefabricated construction systems are rapidly installed and are much saferin terms of the construction processes.

Different sustainability assessment measures exist in various Europeancountries[2].




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2 CASE STUDIES ON SINGLE STOREY BUILDINGS

The following case studies illustrate the use of steel in single storey buildings,such as show rooms, production facilities, supermarkets and similar buildings.

2.1 Manufacturing hall, Express Park, UK

Figure 2.1 Portal frame during construction

The portal frame shows in Figure 2.1 forms part of a new production facilityfor Homeseeker Homes, who manufacture portable homes for residential parks.

The project comprises a 150 m long production hall, an adjacent officebuilding and a separate materials storage building.

The production hall is a duo-pitch portal frame with a 35 m clear span and aheight of 9 m to the underside of the haunch. The production hall has toaccommodate four overhead gantry cranes, each with a safe working load of

5 t. Two cranes may be used in tandem, and the forces arising from thisloading case had to be carefully considered. The longitudinal surge from thecranes is accommodated by bracing in the elevations, which also provideslongitudinal stability. There are no expansion joints in the production hall –the bracing was designed to resist any loads from thermal expansion.

To control the lateral deflection at the level of the crane rail, the frames, at 6 mcentres, are rather stiffer than an equivalent structure without cranes. Thecolumns are 762 mm deep and the rafters 533 mm deep.

The gable frames are portal frames instead of a braced gable frame constructed

from columns and simply-supported rafters, to reduce the differentialdeflection between the end frame and the penultimate frame.




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The facility is relatively close to the site boundary, which meant that theboundary elevations had to have special consideration. A fire load case wasanalysed and the column bases designed to resist the overturning moment fromgrossly deformed rafters. The cladding on the “boundary” elevations was alsospecified to prevent fire spread.

The 380 t of steelwork in the project was erected in six weeks.

2.2 Supermarket, Esch, Luxembourg

Figure 2.2 Supermarket in Esch , Luxembourg using curved cellular beams

Curved 20 m span cellular beams were used to provide an exposed steelstructure in a supermarket in Esch, Luxembourg, as shown in Figure 2.2. Thebeams used HEB 450 sections that were cut and re-welded to form beams with400 mm diameter openings. The curved cellular frames were placed 7,5 mapart and the columns were also 7,5 m high and are illustrated in Figure 2.3.

The structure was designed using fire engineering principles to achieve anequivalent 90 minutes fire resistance without additional fire protection.




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Figure 2.3 Portal frame structu re using curved cellular beams

2.3 Motorway Service station, Winchester, UK Cellular beams provide an attractive solution for long span public spaces, as inthis motorway service restaurant in Winchester, UK, shown in Figure 2.4. The

600 mm deep doubly curved cellular beams spanned 18 m onto 1,2 m deepcellular primary beams that spanned 20 m between H section columns. Thecellular beams also provided for service distribution above the kitchen area.

Figure 2.4 Double curved cellular beams and primary beams




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2.4 Airbus Industrie hanger, Toulouse, France The Airbus production hall in Toulouse covers 200000 m2 of floor space and is45 m high with a span of 117 m. It consists of 8 m deep lattice trussescomposed of H sections. Compound column sections provide stability to theroof structure. The building is shown in Figure 2.5 during construction. Sliding

doors create a 117 m 32 m opening in the end of the building. Two parallelrolling cranes are installed each of 50 m span and 20 tonnes lifting capacity.

Figure 2.5 View of Airbus Industrie hanger during construction

2.5 Industrial hall, Krimpen aan den Ijssel,Netherlands

This production hall is 85 m in length, 40 m wide and 24 m high with full

height doors at the end of the building, as shown in Figure 2.6. The roof structure consists of an inclined truss. Because of the lack of bracing in the endwalls, the structure was designed to be stabilised through the columns assistedby in-plane bracing in the roof and side walls.

Figure 2.6 View of doors being lifted into place in Hollandia’s building inKrimpen aan den Ijssel




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2.6 Distribution Centre and office, Barendrecht,Netherlands

This 26000 m2 distribution centre for a major supermarket in the Netherlandscomprises a conventional steel structure for the distribution area and a two

storey high office area that is suspended above an access road, as shown inFigure 2.7. This 42 m long office building comprises a 12 m cantileversupported by a two storey high internal steel structure with diagonal bracing.

The structure uses H section beams and columns with tubular bracing.

Both the warehouse and office buildings are provided with sprinklers to reducethe risk of fire, and the steelwork has intumescent coating so that it can beexposed internally. The warehouse internal temperature is 2°C and thesteelwork of the office is thermally isolated from the warehouse part.

Figure 2.7 Distr ibut ion centre, Barendrecht, NL showing the braced cantilever office structure






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3.1 Pitched roof portal frameA single-span symmetrical portal frame (as illustrated in Figure 3.2) istypically of the following proportions:

A span between 15 m and 50 m (25 m to 35 m is the most efficient)

An eaves height (base to rafter centreline) of between 5 and 10 m (7,5 m iscommonly adopted). The eaves height is determined by the specified clearheight between the top of the floor and the underside of the haunch.

A roof pitch between 5 and 10 (6° is commonly adopted)

A frame spacing between 5 m and 8 m (the greater frame spacings beingused in longer span portal frames)

Members are I sections rather than H sections, because they must carrysignificant bending moments and provide in-plane stiffness.

Sections are generally S235 or S275. Because deflections may be critical,the use of higher strength steel is rarely justified.

Haunches are provided in the rafters at the eaves to enhance the bendingresistance of the rafter and to facilitate a bolted connection to the column.

Small haunches are provided at the apex, to facilitate the bolted connection

1

34

5

6

7

2

1 Eaves

2 Roof pitch

3 Apex

4 Rafter5 Eaves haunch

6 Apex haunch

7 Column

Figure 3.2 Single-span symmetric portal frame

The eaves haunch is typically cut from the same size rolled section as therafter, or one slightly larger, and is welded to the underside of the rafter. Thelength of the eaves haunch is generally 10% of the span. The length of the

haunch means that the hogging bending moment at the “sharp” end of thehaunch is approximately the same as the maximum sagging bending momenttowards the apex, as shown in Figure 3.3.




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3

h

h

1

2

1 Moment at the “sharp” end of the haunch

2 Maximum sagging moment3 Haunch length

Figure 3.3 Rafter bending moment and haunch length

The final frames of a portal frame are generally called gable frames. Gableframes may be identical to the internal frames, even though they experiencelighter loads. If future extension to the building is envisaged, portal frames arecommonly used as the gable frames, to reduce the impact of the structuralworks. A typical gable frame is shown in Figure 3.4.

4

3

5

1

2

1 Rafter

2 Column

3 Personnel door

4 Roller shutter door

5 Dado wall (brickwork)

Figure 3.4 Typical details of an end gable of a portal frame building




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Alternatively, gable frames can be constructed from columns and short rafters,simply supported between the columns as shown in Figure 3.5. In this case,gable bracing is required, as shown in the figure.

Figure 3.5 Gable frame (not a portal frame)

3.2 Frame stabil ity In-plane stability is provided by frame continuity. In the longitudinal direction,stability is provided by vertical bracing in the elevations. The vertical bracingmay be at both ends of the building, or in one bay only. Each frame isconnected to the vertical bracing by a hot-rolled member at eaves level.A typical bracing arrangement is shown in Figure 3.6.

1

2

2

3

1 Vertical bracing in the gable

2 Vertical bracing in the walls

3 Roof bracing

Figure 3.6 Typical bracing in a portal frame

The gable columns span between the base and the rafter, where the reaction iscarried by bracing in the plane of the roof, back to the eaves level, and to thefoundations by the vertical bracing.




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If diagonal bracing in the elevations cannot be accommodated, longitudinalstability can be provided by a rigid frame on the elevation, as shown inFigure 3.7.

1

2

1 Eaves strut

2 Rigid frame

Figure 3.7 Rigid frame alternative to vertical bracing

3.3 Member stabil ity Member stability should be checked using expressions 6.61 and 6.62 of EN 1993-1-1. For economic design, restraints to the rafter and column must beconsidered. The purlins and side rails are considered adequate to restrain theflange that they are attached to, but unless special measures are taken, thepurlins and side rails do not restrain the inside flange. Restraint to the inside

flange is commonly provided by bracing from the purlins and side rails, asshown in Figure 3.8. The bracing is usually formed of thin metal straps,designed to act in tension, or from angles designed in compression if bracing isonly possible from one side.

If the bracing shown in Figure 3.8 is not permitted by national regulations,restraint may be provided by a system of hot-rolled members.

This form of bracing will be required whenever the inside flange is incompression. This situation arises:

On the inside of the column and the inside of the rafter in the haunchregion, in the gravity load combination

Towards the apex of the rafter, in the uplift combination.




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1

2

1 Restraint to inside flange2 Purlin or side rail

Figure 3.8 Typical bracing to the inside flange

The arrangement of restraints to the inside flange is generally similar to thatshown in Figure 3.9. In some instances, it may not be possible to restrain the

inside of the column flange. In these circumstances, a larger column sectionmay have to be chosen, which is stable between the underside of the haunchand the base.

1

1

1 Restraint to inside flange of rafter and column

Figure 3.9 General arrangement of restraints to the inside flange

In all cases, the junction of the inside face of the column and the underside of the haunch, as shown in Figure 3.10, must be restrained. The restraint may beof the form shown in Figure 3.8, or may be by a hot-rolled member providedfor that purpose.

1

1 Restraint position

Figure 3.10 Restraint at the haunch / column junct ion








2 - 27

3.5 Connections3.5.1 Eaves connection

A typical eaves connection is shown in Figure 3.12. In almost all cases acompression stiffener in the column (as shown, at the bottom of the haunch)

will be required. Other stiffeners may be required to increase the bendingresistance of the column flange, adjacent to the tension bolts, and to increasethe shear resistance of the column web panel. The haunch is generallyfabricated from a similar size beam to the rafter (or larger), or fabricated fromequivalent plate. Typically, the bolts may be M24 8.8 and the end plate 25 mmthick S275.

2

1

1 Haunch

2 Compression stiffener

Figure 3.12 Typical eaves connection

3.5.2 Apex connection

A typical apex connection is shown in Figure 3.13. The apex connectionprimarily serves to increase the depth of the member to make a satisfactorybolted connection. The apex haunch is usually fabricated from the same

member as the rafter, or from equivalent plate. Typically, the bolts may beM24 8.8 and the end plate 25 mm thick S275.

Figure 3.13 Typical apex connection




2 - 28

3.5.3 Bases

A typical pinned base is shown in Figure 3.14. The base plate is generally atleast as thick as the flange of the column. Most authorities accept that evenwith four holding down bolts as shown in Figure 3.14, the base is still pinned.Alternatively, the base may have only two holding down bolts, on the axis of

the column, but this may make the erection of the steelwork more difficult.

Columns are normally located on a number of steel packs, to ensure thesteelwork is at the correct level, and the gap between the foundation and thesteelwork filled with cementicious grout. Large bases should be provided withan air hole to facilitate complete grouting.

Holding down bolts are generally embedded in the foundation, with somefreedom of lateral movement (tubes or cones) so that the steelwork can bealigned precisely. The holes in the base plate are usually 6 mm larger than thebolt diameter, to facilitate some lateral alignment.

~

5

4

3

2

1

1 Holding down bolts

2 Base plate3 Grout

4 Tube (or cone)

5 Anchor plate

Figure 3.14 Typical portal base detail

3.5.4 Bracing Connections

Forces in portal frame bracing are generally modest. Typical connections areshown in Figure 3.15. Gusset plates should be supported on two edges, if possible.




2 - 29

Figure 3.15 Typical bracing connections

3.6 Other types of portal frame The features of an orthodox portal frame were described in Sections 3.1 to 3.5.

The basic structural concept can be modified in a number of ways to produce acost effective solution, as illustrated below.

3.6.1 Portal frame with a mezzanine floor

1

1 Mezzanine

Figure 3.16 Portal frame with internal mezzanine floor

Office accommodation is often provided within a portal frame structure using amezzanine floor (as illustrated in Figure 3.17). The mezzanine floor may bepartial or full width. It can be designed to stabilise the frame. Often, theinternal floor of the office space requires fire protection.






2 - 31

3.6.3 Portal frame with overhead crane

Figure 3.19 Crane portal frame with column brackets

For cranes of relatively low capacity (up to say 20 tonnes), portal frames canbe used to support the crane beam and rail, as illustrated in Figure 3.19. Theoutward movement (spread) of the frame at the level of the crane rail is likely

to be of critical importance. Use of a horizontal tie member or fixed columnbases may be necessary to reduce this spread.

For larger cranes, a structure with a roof truss will be appropriate (seeSection 4) as the column spread is minimised. For very heavy loads, built-upcolumns are appropriate, as introduced in Section 6. Detail design guides coverboth the design of trusses[3] and the design of built-up columns[4].

3.6.4 Tied portal frame

1

2

1 Tie

2 Hangers (required for longer spans)

Figure 3.20 Tied portal frame

In a tied portal frame, as illustrated in Figure 3.20, the spread of the eaves andthe bending moments in the frame are greatly reduced. Large compressionforces will develop in the rafters, which reduce the stability of the members.Second-order software must be used for the design of tied portals.




2 - 32

3.6.5 Mansard or curved portal frames

Figure 3.21 Mansard portal frame

A mansard portal frame consists of a series of rafters and haunches, as

illustrated in Figure 3.21, which creates a pseudo-curved frame. Theconnections between the members may also have small haunches to facilitatethe bolted connections.

Curved rafter portals as illustrated in Figure 3.22 are often used forarchitectural applications. The rafter can be curved to a radius by cold bending.For spans greater than approximately 18 m, splices may be required in therafter because of limitations of transport.

Alternatively, a curved external roof must be produced by varying the lengthsof purlin brackets supported on a rafter fabricated as a series of straight

elements, as shown in Figure 3.23.

Figure 3.22 Curved beams used in a portal frame






2 - 34

12

1 Valley beam2 Rafter

Figure 3.25 Connection to valley beam






2 - 36

4.2 Truss membersUnless there are special architectural requirements, truss members are chosento produce a simple connection between the chords and the internal members.Common combinations as shown in Figure 4.2 are:

Tees used as chords, with angles used as web members. The angles may bewelded or bolted to the stem of the Tee.

Double angle members as chords, and single (or double) angles as internalmembers. The connections are made with a gusset plate welded betweenthe angles forming the chords.

Rolled sections as chords, with the web in the plane of the truss. Theinternal members are usually angle members, connected via a gusset platewelded to the chord.

Rolled sections as chords, but with the web perpendicular to the plane of

the truss. The connections to the chord members may be via gusset plateswelded to the web, although the connections will need careful detailing.A simple, effective alternative is to choose chords that have the sameoverall depth, and connect the internal members to the outside of bothflanges, generally by welding.

For heavily loaded trusses, rolled I or H sections, or channel sections maybe used as the internal members. In such a large truss, developing economicconnections will be important and both the members and internal membersshould be chosen with this in mind.

The detailed design of trusses is covered in Single-storey steel buildings.

Part5: Detailed design of trusses[3]

.






2 - 38

Figure 4.3 Truss fabricated from rolled sections

4.3 Frame stabil ity In most cases, frame stability is provided by bracing in both orthogonaldirections, and the truss is simply pinned to the supporting columns. To realisea pinned connection, one of the chord members is redundant, as shown in

Figure 4.4, and the connection of that redundant member to the column isusually allowed to slip in the direction of the axis of the chord.

1

1 Redundant member

Figure 4.4 Redundant member in a simply supported truss

In the longitudinal direction, stability is usually provided by vertical bracing.








2 - 41

Figure 4.7 Splice details

Ordinary bolts (non-preloaded) in clearance holes may give rise to some slip inthe connection. If this slip is accumulated over a large number of connections,the defection of the truss may be larger than calculated. If deflection is acritical consideration, then friction grip assemblies or welded details should beused.




2 - 42

5 SIMPLE BEAM STRUCTURES

For modest spans, (up to approximately 20 m) a simple beam and column

structure can be provided, as illustrated in Figure 5.1. The roof beam is asingle rolled section, with nominally pinned connections to the columns. Theroof beam may be straight, precambered, perforated or curved. The roof maybe horizontal, or more commonly with a modest slope to assist drainage.Ponding of water on the roof should be avoided with a slope, or precamberedbeam.

Figure 5.1 Simple beam and column frame

Frame stability for this form of structure is provided by bracing in eachorthogonal direction. The beam is designed as simply supported, and thecolumns as simple struts, with a nominal moment applied by the beamconnection. It is common to assume that the shear force from the beam isapplied 100 mm from the face of the column.




2 - 43

6 BUILT-UP COLUMNS

Heavily loaded columns, or columns in tall industrial buildings may be in the

form of built-up sections. Built-up columns often comprise HE or UPE sectionsin which battens (flat plate) or lacing (usually angles) are welded across theflanges, as shown in Figure 6.1.

Built-up columns are not used in portal frames, but are often used in buildingssupporting heavy cranes. The roof of the structure may be duo-pitch rafters, butis more commonly a truss, as illustrated in Figure 1.4.

Figure 6.1 Cross-sections of built-up columns

To support the roof above the level of the crane, a single member may projectfor several meters. This is often known as a “bayonet” column. The projectingmember may be a continuation of one of the two primary sections in thebuilt-up section, or may be a separate section located centrally to the built-upsection. Examples of built-up columns are shown in Figure 6.2. Buildings thatuse built-up columns are invariably heavily loaded, and commonly subjected tomoving loads from cranes. Such buildings are heavily braced in two orthogonal

directions.

The detailed design of built-up columns is covered in Single-storey steelbuildings. Part 6: Detailed design of built-up columns[4] of this guide.




2 - 44

Laced column Battened column Column withcrane girder

Figure 6.2 Examples of built-up columns in single storey buildings












2 - 49

8 PRELIMINARY DESIGN OF PORTALFRAMES

8.1 Introduction The following methods of determining the size of columns and rafters of single-span portal frames may be used at the preliminary design stage. Furtherdetailed calculations will be required at the final design stage. It should benoted that the method does not take account of:

Requirements for overall stability

Deflections at the Serviceability Limit State.

8.2 Estimation of member sizes The guidance for portal frames is valid in the span range between 15 to 40 m.and is presented in Table 8.1. The assumptions made in creating this table areas follows:

The roof pitch is 6.

The steel grade is S235. If design is controlled by serviceability conditions,the use of smaller sections in higher grades may not be an advantage. Whendeflections are not a concern, for example when the structure is completelyclad in metal cladding, the use of higher grades may be appropriate.

The rafter load is the total factored permanent actions (including self weight) and factored variable actions and is in the range of 8 to 16 kN/m.

Frames are spaced at 5 to 7,5 m.

The haunch length is 10% of the span of the frame.

A column is treated as restrained when torsional restraints can be providedalong its length (these columns are therefore lighter than the equivalentunrestrained columns).

A column should be considered as unrestrained when it is not possible torestrain the inside flange.

The member sizes given by the tables are suitable for rapid preliminary design.However, where strict deflection limits are specified, it may be necessary toincrease the member sizes.

In all cases, a full design must be undertaken and members verified inaccordance with EN 1993-1-1.





2

- 5 1

Table 8.1 (Continued) Single-span portal frame with 6° roof pitch

Span of frame (m)Rafter load (kN/m)

Eaves height (m)

15 20 25 30

Rafter 14

1414

6

810

IPE 330

IPE 330IPE 330

IPE 400

IPE 400IPE 400

IPE 450

IPE 450IPE 450

IPE 450

IPE 450IPE 450

Restrainedcolumn

141414

6810

IPE 360IPE 400IPE 400



IPE 600IPE 600

IPE 750 137

IPE

IPE

IPE

Unrestrainedcolumn

141414

6810


IPE 550IPE 600

IPE 750 137

IPE 600

IPE 750 137

IPE 750 173

IPE 750 137

IPE 750 173HE 800

IPE

Rafter 1616

16

68

10

IPE 330IPE 330

IPE 330

IPE 400IPE 400

IPE 400

IPE 450IPE 450

IPE 450

IPE 550IPE 550

IPE 50

Restrainedcolumn

161616

6810




IPE 750 137

IPE 750 137

IPE 750 137

IPE

IPE

Unrestrainedcolumn

161616

6810


IPE 550IPE 600

IPE 750 137

IPE 600

IPE 750 173HE 800

IPE 750 137HE 800HE 800

IPE




2 - 52

REFERENCES

1 SANSOM, M. and MEIJER, J.Life-cycle assessment (LCA) for steel construction

European commission, 2002

2 Several assessement methods are used. For example:

BREEAM in the UK

HQE in France

DNGB in Germany

BREEAM-NL, Greencalc+and BPR Gebouw in the Netherlands

Valideo in Belgium

Casa Clima in Trento Alto Adige, Italy (each region has its own approach)

LEED, used in various countries

3 Steel Buildings in EuropeSingle-storey steel buildings. Part 5: Design of trusses

4 Steel Buildings in EuropeSingle-storey steel buildings. Part 6: Design of built-up columns





Part 3: Actions






Part 3: Actions











Part 3: Actions

3 - vi

SUMMARY

This document provides guidelines for the determination of the actions on a

single-storey building according to EN 1990 and EN 1991. After a short description of

the general format for limit state design, this guide provides information on the

determination of the permanent loads, the variable actions and the combinations of actions. The determination of the snow loads and the calculation of the wind action are

described and summarized in comprehensive flowcharts. Simple worked examples on

the snow loads and the wind action are also included.















Part 3: Actions

3 - 7

3.3.4 Quasi-permanent combination

The quasi-permanent combination is normally used for long-term effects and

the appearance of the structure.

Permanent

actions

Variable

actions

E d = 1

jk,

j

G +

1

ik,i2,

i

Q

For example:

E d = G (since 2 = 0 for both the wind action and the snow load)



















Part 3: Actions

3 - 16

7 SNOW LOADS

7.1 General

This document gives guidance to determine the values of loads due to snow to be used for a typical single-storey building according to EN 1991-1-3. The

design procedure is summarized in a flowchart (Figure 7.5). A worked example

dealing with the determination of the snow loads on a single-storey building is

given in Appendix A.

The guidance does not apply to sites at altitudes above 1500 m (unless

otherwise specified).

Snow loads shall be classified as variable, fixed actions, unless otherwise

stated in EN 1991-1-3. For particular conditions like exceptional snow loads

and/or loads due to exceptional snow drifts, they may be treated as accidentalactions depending on geographical locations.

Snow loads should be classified as static actions.

Two design situations may need to be considered:

Transient/persistent situation should be used for both the undrifted and

drifted snow load arrangements for locations where exceptional snow falls

and exceptional snow drifts are unlikely to occur.

Accidental design situation should be used for geographical locations where

exceptional snow falls and/or exceptional snow drifts are likely to occur.

The National Annex may define which design situation to apply.

7.2 Methodology7.2.1 Snow load on the ground

Different climatic conditions will give rise to different design situations. The

possibilities are:

Case A: Normal case (non exceptional falls and drifts)

Case B1: Exceptional falls and no exceptional drifts

Case B2: Exceptional drift and no exceptional falls (in accordance with

EN 1991-1-3 Annex B)

Case B3: Exceptional falls and exceptional drifts (in accordance with

EN 1991-1-3 Annex B)

The National Authority may choose the case applicable to particular locations

for their own territory.

The National Annex specifies the characteristic value sk of snow load on the

ground to be used.





Part 3: Actions

3 - 18

Table 7.1 Snow load shape coefficients

Angle of pitch of roof 0° 30° 30° < < 60° 60°

1 0.8 0.8 (60 – )/30 0

2 0.8 + 0.8 /30 1.6 -

These values 1 and 2 apply when the snow is not prevented from sliding off

the roof (no snow fences or other obstructions like parapets). If obstructions

exist, the snow load shape coefficient should not be reduced below 0.8.

The snow load shape coefficient that should be used for monopitch roofs is

shown in Figure 7.1, where 1 is given in Table 7.1.

The load arrangement should be used for both the undrifted and drifted load

arrangements.

1( )

Figure 7.1 Snow load shape coefficient – Monopitch roof

The snow load shape coefficients that should be used for pitched roofs are

shown in Figure 7.2, where 1 is given in Table 7.1.

Case (i) corresponds to the undrifted load arrangement.

Cases (ii) and (iii) correspond to the drifted load arrangements.

0,5 )

)

1 2

(i)

(ii)

(iii)

(i) Undrifted load arrangement

(ii) and (iii) Drifted load arrangement

Figure 7.2 Snow load shape coefficient – Pitched roof













Part 3: Actions

3 - 24

The effects of orography may be neglected when the average slope of the

upwind terrain is less than 3°. The recommended value of co( z ) is 1,0, but the

National Annex may give the procedure to calculate the orography factor.

Annex A3 of EN 1991-1-4 gives the recommended procedure to determine co

for hills, cliffs, etc.

7. Turbulence factor k l

The recommended value is 1,0 but the National Annex may give other values.

8. Peak velocity pressure q p( z )

)(2

1)(71)(

2

mv p z v z I z q

where:

I v( z ) is the turbulence intensity which allows to take into account thecontribution from short-term fluctuations

)/ln()()(

0o

lv

z z z c

k z I for z min ≤ z ≤ z max

)()( minvv z I z I for z < z min

z max = 200 m

vm( z ) is the mean wind velocity at height z above the terrain:

vm( z ) = cr ( z ) co( z ) v b

Alternative for step 8:

For single-storey-buildings, the determination of the mean wind velocity vm( z )

is not absolutely necessary. The peak velocity pressure can be directly obtained

from the exposure factor ce( z ):

be p )()( q z c z q

where:

)()()()(

71)( 2r

2o

r o

r le z c z c

z c z ck k z c

For flat terrain (co( z ) = 1) and for turbulence factor k l = 1, the exposure factor

ce( z ) can be directly obtained from Figure 4.2 of EN 1991-1-4, as a function of

the height above terrain and a function of terrain category.

8.2.2 Wind pressure on surfaces – Wind forces

There are three types of wind forces acting on a building:

External forces F w,e (see 8.2.2.1)

Internal forces F w,i (see 8.2.2.2)

Friction forces F fr (see 8.2.2.3).











Part 3: Actions

3 - 29

Multispan roofs : Figure 7.10 and the coefficients c pe are derived from

Tables 7.3 to 7.4.

Figure 8.4 of this guide shows the zones for duopitch roofs.

b

e/10

G

F

1

e/10

F

H J I

2 4

e/4

e/4

3

b

e/10

G

F

1

e/2

F

H I

I

2

e/4

e/4

G

H

Wind on the long side(perpendicular to the ridge line)

1 Wind direction

2 Ridge line

3 Upwind face

4 Downwind face

Wind on the gable(parallel to the ridge line)

e = min(b ; 2h)

b is the crosswind dimension

Figure 8.4 Zones for duopitch roofs

8.2.5 Internal pressure coefficients

The internal pressure coefficient c pi depends on the size and distribution of the

openings in the building envelope.

When in at least two sides of the building (façades or roof) the total area of

openings in each side is more than 30 % of the area of that side, the structure

should be considered as a canopy roof and free-standing walls.

A face of a building should be regarded as dominant when the area of openings

in that face is at least twice the area of openings in the remaining faces of the

building considered.

Where an external opening would be dominant when open but is considered to

be closed in the ultimate limit state, during severe windstorms (wind used for

the design of the structure), the condition with the opening open should be

considered as an accidental design situation.

For a building with a dominant face, the internal pressure should be taken as a

fraction of the external pressure at the openings of the dominant face:

Area of the openings on the dominant face = 2 area of openings in the

remaining faces:

c pi = 0,75 c pe

Area of the openings in the dominant face = 3 area of openings in the

remaining faces:c pi = 0,90 c pe



















TitleAPPENDIX A. Worked Example: Snow load applied on a single-storey

building3 of 8

3 - 38

0,52 kN/m2

Case (i)

0,26 kN/m2

0,52 kN/m2

Case (ii)

0,52 kN/m2

0,26 kN/m2

Case (iii)

Figure A.2 Snow load arrangements on the upper roof in persistent design

situation

EN 1991-1-3

Figure 5.3

Accidental design situations – exceptional load on the ground

- Case (i): Undrifted load arrangement

1( = 8,5°) = 0,8

s = 0,8 1,30 = 1,04 kN/m2

- Case (ii): Drifted load arrangement

0,5 1(= 8,5°) = 0,4

s = 0,4 1,30 = 0,52 kN/m2

- Case (iii): Drifted load arrangement

The case (iii) is symmetrical about the case (ii) because of the

symmetry of the roof ( 1 = 2 = 8,5°)

1,04 kN/m2

Case (i)

0,52 kN/m2 1,04 kN/m2

Case (ii)

1,04 kN/m2 0,52 kN/m2

Case (iii)

Figure A.3 Snow load arrangements on the upper roof in accidental designsituation

Accidental design situations – exceptional drift:

This case is not applicable. There are no parapets or valleys.











TitleAPPENDIX A. Worked Example: Snow load applied on a single-storey

building8 of 8

3 - 43

3.5.2. Roofs where drifting occurs behind parapets at eaves

1 = Min(2 h/ sk ; 2 b2/l s ; 8)

where: l s = Min(5h ; b1 ; 15 m)

h = 3,00 m

b1 = 12,50 m

b2 = 25,00 m

sk = 0,65 kN/m2

5h = 15,00 m ; l s = 12,50 m ; 2h/ sk = 9,23 ; 2b2/l s = 4,00

1 = 4,00

And: s = 1 sk = 2,60 kN/m2

EN 1991-1-3

Annex B § B.4

3.5.3. Roofs where drifting occurs behind parapets at gable end

1 = Min(2 h/ sk ; 2 b2/l s ; 8)

where: l s = Min(5h ; b1 ; 15 m)

h = 3,00 m

b1 = 40,00 m

b2 = 25,00 m

sk = 0,65 kN/m2

5h = 15,00 m ; l s = 15,00m ; 2h/sk = 9,23 ; 2b2/ls = 5,33 1 = 5,33

And: s = 1 sk = 3,46 kN/m2

EN 1991-1-3

Annex B § B.4

0,00 kN/m2

12,50 m12,50 m

15,00 m

2,60 kN/m2 2,60 kN/m2

3,46 kN/m2

Snow behind the parapet at gable end Snow behind the parapets at eaves

Figure A.8 Exceptional snow drifted on the lower roof in the case of roofswhere drifting occurs behind parapets at eaves







3 - 46

APPENDIX B. Worked Example: Wind action ona single-storey building

1 of 11

Made by DC Date 06/2009

Calculation sheetChecked by AB Date 07/2009

1. DataThis worked example deals with the calculation of the wind action on a

single-storey building according to EN 1991-1-4. The overall dimensions of

the building are given in Figure B.1.

14 °

5 m

5 m

6 m

6 m4,8 m

6 m

16 m

16 m

60 m

Figure B.1 Geometry of the building

The doors are assumed to be shut during severe gales.

The fundamental value of the basic wind velocity is:

v b,0 = 26 m/s

2. Peak velocity pressureThe peak velocity pressure is determined according to the step-by-step

procedure given in this guide.

1. Fundamental value of the basic wind velocity

v b,0 = 26 m/s

2. Basic wind velocity

For cdir and cseason, the recommended values are:

cdir = 1,0

cseason = 1,0

Then: v b = v b,0 = 26 m/s

EN 1991-1-4§ 4.2(2)





TitleAPPENDIX B. Worked Example: Wind action on a single-storey

building3 of 11

3 - 48

8. Peak velocity pressure (alternative for a single-storey building)

q p( z ) = ce( z ) q b

where:

)()()()(

71)( 2

r

2

o

r o

r le z c z c

z c z c

k k z c

EN 1991-1-4

§ 4.5(1)

56,1706,00,1706,00,1

215,00,171)( 22

e

z c

Then: q p( z ) = 1,56 423 = 659 N/m2

q p( z ) = 0,659 kN/m2 for z = 8 m

3. Wind pressure on surfaces

3.1. External pressure coefficients c pe,10

3.1.1. Vertical walls

1. Wind on gable

h = 8 m

b = 32 m (crosswind dimension)

h < b, so z e = reference height = h = 8 m

EN 1991-1-4

7.2.2 (1)

Figure 7.4

d = 60 mh/d = 8/60 = 0,13 (h/d < 0,25)

EN 1991-1-47.2.2 (2)

Table 7.1

2h = 16 m

e = 16 m (b or 2h, whichever is smaller)

EN 1991-1-4

§ 7.2.2 (1)

Figure 7.5

e < d

e/5 = 3,2 m

4/5 e = 12,8 m

d – e = 44 m

Figure B.2 defines the external pressure coefficients c pe,10 on vertical walls for

zones A, B, C, D and E with wind on the gable.

EN 1991-1-4

§ 7.2.2(2)

Table 7.1














building9 of 11

3 - 54

5. Wind forces on surfaces F / Aref = cscd q p( z e) c pe – q p( z i) c pi

with: cscd = 1 (height < 15 m)

q p( z e) = q p( z i) = 0,66 kN/m2

The figures below show the wind forces per unit surfaces:

F / Aref = 0,66 (c pe – c pi) (in kN/m2)

EN 1991-1-4

§ 6.2(1)b

Wind

+0,33

-0,99

-0,53-0,46

-0 33

-0,92

-0,66

-0,46

Ffr = 8,32 kN

Figure B.6 Wind on gable with c pi = +0,2

Wind

+0 66

-0,66

-0,20-0,13

0

-0,59

-0,33

-0,13

Ffr = 8,32 kN

Figure B.7 Wind on gable with c pi = -0,3




building10 of 11

3 - 55

-0,73

Wind

+0 33

-0,73+ 0

-0,33(+ 0)

-0,46

-0 33

-0,92

-0,66

-0,46 -0,73+ 0

-0,66+ 0

Figure B.8 Wind on long side with c pi = +0,2

The values in brackets should be used together.

Wind

+0,66

-0,40(+0,33)

-0(+0,33)

-0,13

0

-0,59

-0,33

-0,13 -0,40(+0,33)

-0,33(+0,33)

-0,40

Figure B.9 Wind on long side with c pi = -0,3

Values in brackets should be used together.





















Part 4: Detailed Design of Portal Frames

4 - vii

SUMMARY

This publication provides guidance on the detailed design of portal frames to the

Eurocodes.

An introductory section reviews the advantages of portal frame construction andclarifies that the scope of this publication is limited to portal frames without ties

between eaves. Most of the guidance is related to single span frames, with limited

guidance for multi-span frames.

The publication provides guidance on:

The importance of second order effects in portal frames

The use of elastic and plastic analysis

Design at the Ultimate and Serviceability Limit States

Element design: cross-section resistance and member stability

Secondary structure: gable columns, bracing and eaves members.

The document includes a worked example, demonstrating the assessment of sensitivity

to second order effects, and the verification of the primary members.








4 - 2

Whilst manual design may be useful for initial sizing of members and a

thorough understanding of the design process is necessary, the use of bespoke

software is recommended.


















4 - 10

M

M

M

y

p

1

1

23

2

1 True behaviour

2 Elastic-perfectly-plastic model

3 Unloading behaviour

Figure 3.4 Moment/rotation behaviour and elastic-perfect ly-plasti c model for a Class 1 section

(4)

2

6

3

5

1

V Ed

Ed

EdH

H Ed,V (7)

1 Elastic response

2 First hinge forms

3 Second hinge forms

4 Horizontal displacement

5 True behaviour

6 Elastic/perfectly plastic model

7 Increasing vertical and (in proportion)horizontal load

Figure 3.5 Simple model of a portal frame subject to increasing vertical and horizontal loads, with failure governed by a sway mechanism






4 - 12

(a)

First hinge forms

1

(b)

Load increases – rafter approaches yield

1

(c)

Load increases, second hinge forms and a

mechanism leads to collapse

11

(d)

1 Plastic resistance moment

Figure 3.6 Elastic-perfectly-plastic method of analysis, showing state of frame as horizontal and vertical l oads are increased proport ionally a) Elastic throughout; (b) Plastic hinge at eaves;(c) Raftersapproaching p lasticity; (d) Plastic hinge in rafter

It is recognised that some redistribution of moments is possible, even with the

use of elastic design. EN 1993-1-1 § 5.4.1.4(B) allows 15% redistribution, as

discussed in Section 3.2.2, although this is uncommon in practice.

Where haunch lengths of around 15% of the span are acceptable and the lateral

loading is small, the elastic bending moment diagram will be almost the same

as the plastic collapse bending moment diagram. As illustrated in Figure 3.3,

the maximum hogging moment at the end of the haunch is similar to the

maximum sagging moment in the rafter. In such cases, an elastic analysis may

provide an equivalent solution to a plastically analysed frame.




4 - 13

3.3 First order and second order analysisFor both plastic analysis and elastic analysis of frames, the choice of first-order

or second order analysis may be governed by the in-plane flexibility of the

frame, measured by the factor cr (see Section 3.3.1). In practice, the choice

between first and second order analysis is also dependent on the availability of software. Even if a portal frame was sufficiently stiff that second order effects

were small enough to be ignored, it may be convenient still to use second order

analysis software.

When a second order analysis is required but is not available, modified first

order methods can be useful for calculations. A modified first order approach is

slightly different for elastic and plastic analysis, and is described in

Sections 3.3.2 and 3.3.3. In elastic analysis, the horizontal actions are

amplified; in plastic analysis, all actions are amplified.

3.3.1 cr factor Expression 5.2 of EN 1993-1-1 § 5.2.1(4)B gives cr as:

EdH,Ed

Edcr

h

V

H

Note 1B and Note 2B of that clause limit the application of Expression 5.2 to

roofs with shallow roof slopes and where the axial force in the rafter is not

significant. Thus:

a roof slope is considered as shallow at slopes no steeper than 26°

axial force in the rafter may be assumed to be significant if Ed

y3,0

N

Af .

A convenient way to express the limitation on the axial force is that the axial

force is not significant if:

crEd 09.0 N N

Where

N cr is the elastic critical buckling load for the complete span of the rafter

pair, i.e.2

2

cr L

EI π N

L is the developed length of the rafter pair from column to column,

taken as span/Cos θ (θ is the roof slope)

If the limits are satisfied, then Expression 5.2 may be used to calculate cr. In

most practical portal frames, the axial load in the rafter will be significant and

Expression 5.2 cannot be used.

When the axial force in the rafter is significant, Appendix B provides an

alternative, approximate method to calculate the measure of frame stability,

defined as cr,est. In many cases, this will be a conservative result. Accurate

values of cr may be obtained from software.




4 - 14

3.3.2 Modified first order, for elastic frame analysis

The ‘amplified sway moment method’ is the simplest method of allowing for

second order effects for elastic frame analysis; the principle is given in

EN 1993-1-1, § 5.2.2(5B).

A first-order linear elastic analysis is first carried out; then all horizontal loadsare increased by an amplification factor to allow for the second order effects.

The horizontal loads comprise the externally applied loads, such as the wind

load, and the equivalent horizontal forces used to allow for frame

imperfections; both are amplified.

Provided cr 3,0 the amplification factor is:

cr 11

1

If the axial load in the rafter is significant, and cr,est has been calculated inaccordance with Appendix B, the amplifier becomes:

est cr,11

1

If cr or cr,est is less than 3,0 second order software should be used.

3.3.3 Modified first order, for plastic frame analysis

Design philosophy

In the absence of elastic-plastic second order analysis software, the design

philosophy is to derive loads that are amplified to account for the effects of deformed geometry (second order effects). Application of these amplified loads

through a first-order analysis gives the bending moments, axial forces and

shear forces that include the second order effects approximately.

The amplification is calculated by a method that is sometimes known as the

Merchant-Rankine method. Because, in plastic analysis, the plastic hinges limit

the moments resisted by the frame, the amplification is performed on all the

actions that are applied to the first-order analysis (i.e. all actions and not only

the horizontal forces related to wind and imperfections).

The Merchant-Rankine method places frames into one of two categories:

Category A: Regular, symmetric and mono-pitched frames

Category B: Frames that fall outside of Category A but excluding tied

portals.

For each of these two categories of frame, a different amplification factor

should be applied to the actions. The Merchant-Rankine method has been

verified for frames that satisfy the following criteria:

1. Frames in which 8h

L

for any span

2. Frames in which 3cr








4 - 17

h

0.75h

Figure 3.9 Dummy member to model nominally rigid column base

Note that the reaction at the pinned end of the dummy member will affect the

reaction at the column base. This must be corrected by taking the base reaction

equal to the axial force in the column, which equals the sum of the reactions at

the base and the pinned end of the dummy member.

3.4.1 Pinned and rocker bases

Where a true pin or rocker is used, as illustrated in Figure 3.10, the rotational

stiffness is zero. The use of such bases is rarely justified in practice. Where

they are adopted, careful consideration needs to be given to the transfer of

shear into the foundation, and temporary stability of the column during

erection.

Figure 3.10 Examples of zero sti ffness column bases

3.4.2 Nominally rig id column bases

If a column is rigidly connected to a suitable foundation, the following

recommendations should be adopted:

Elastic global analysis:

For Ultimate Limit State calculations the stiffness of the base can be taken as

equal to the stiffness of the column.

For Serviceability Limit State calculations the base can be treated as rigid to

determine deflections under serviceability loads.






4 - 19

A summary of the assessment of sensitivity to second order effects and the

amplification to allow for second order effects is given in Table 3.1.

Table 3.1 Second order effects: assessment and amplif ication factors

Restrictions Elastic analysis Plastic analysis

shallow slopes, andrafter axial force notsignificant

cr cr Measure of sensitivity to secondorder effects

steep slopes, andrafter axial forcesignificant

cr,est cr,est

Regular frames

cr 11

1

or

est cr,11

1

cr 11

1

or

est cr,11

1

Amplifier to allow forsecond order effects

Irregular frames, butexcluding tied portals

cr 11

1

or

est cr,11

1

cr

,

11

11or

est cr,11

1,1

Amplifier applied to: Horizontal loadsonly

All loads




4 - 20

4 SERVICEABILITY LIMIT STATE

4.1 General

The Serviceability Limit State (SLS) analysis should be performed using theSLS load cases, to ensure that the deflections are acceptable at ‘working loads’.

4.2 Selection of deflection cri teria No specific deflection limits are set in EN 1993-1-1. According to

EN 1993-1-1 § 7.2 and EN 1990, Annex A1.4, deflection limits should be

specified for each project and agreed with the client. The relevant National

Annex to EN 1993-1-1 may specify limits for application in individual

countries. Where limits are specified’ they have to be satisfied. Where limits

are not specified, Appendix A of this document presents typical limits.

If the structure contains overhead travelling cranes, the spread of the columns

at the level of the crane is likely to be an important design criterion. In many

cases, it will be necessary to provide stiffer steel sections than are necessary for

the ULS design, or to provide some fixity in the base and foundation. An

alternative is a tied portal (when second order analysis must be used) or a truss.

4.3 AnalysisThe SLS analysis is normally a first-order (elastic) analysis. The designer

should verify plastic hinges do not form at SLS, simply to validate thedeflection calculations.

4.4 Design summary The Serviceability Limit State (SLS):

Is assessed by first order analysis

Uses deflection criteria defined in the relevant National Annex or agreed

with the client.










4 - 24

Practical design addresses this interaction in several ways:

Out-of-plane stability near plastic hinges is generally addressed by the

concept of stable lengths, Lstable, Lm, Lk and Ls. These are assumed to be

independent of any interaction with in-plane stability effects (see

Section 6.4.).

Interaction between bending moment and axial load is addressed by

simultaneously satisfying Expressions 6.61 and 6.62 of EN 1993-1-1. This

is usually undertaken by considering the most onerous out-of-plane check

(from any part of the member) with the relevant in-plane check.

6.2 Buckling resistance in EN 1993-1-1The verification of buckling resistance of members is addressed by several

clauses in EN 1993-1-1. The clauses of primary interest in portal frame design

are described below.

6.3.1 Uniform members in compression. This clause covers strut buckling

resistance and the selection of buckling curves. The clause is primarily

concerned with flexural buckling, but also addresses torsional and

torsional-flexural buckling. These latter modes of failure will not govern the

IPE sections and similar cross-sections adopted for portal frames.

6.3.2 Uniform members in bending. This clause covers lateral-torsional

buckling of beams.

The distribution of bending moments along an unrestrained length of beam has

an important influence on the buckling resistance. This is accounted for by thechoice of C 1 factor when calculating M cr (See Appendix C).

6.3.3 Uniform members in bending and axial compression. This clause

addresses the interaction of axial load and moment, in-plane and out-of-plane.

The clause requires the following checks to be carried out unless full second

order analysis, including all member imperfections ( P – , torsional and lateral

imperfections), is utilised.

1

M1

Rk z,

Edz,Edz,yz

M1

Rk y,

LT

Edy,Edy,

yy

M1

Rk y

Ed

M

Δ M M

k M

Δ M M

k N

N (6.61)

1

M1

Rk z,

Edz,Edz,zz

M1

Rk y,

LT

Edy,Edy,

zy

M1

Rk z

Ed

M

Δ M M k

M

Δ M M k

N

N (6.62)

For Class 1, 2, 3 and bi-symmetric Class 4 sections, 0Ed z,Ed y, M M

It is helpful to define M1

y.Rk

y

N

as N b,y,Rd and LT M1

Rk y,

M

as M b,Rd.

M z.Ed is zero because the frame is only loaded in its plane.






4 - 26

6.3 Out-of-plane restraint

(a)

(b)

(c)

Figure 6.2 Types of restrain t to out-of-plane buck ling

Figure 6.2 shows the three basic types of restraint that can be provided to

reduce or prevent out-of-plane buckling:

(a) Lateral restraint, which prevents lateral movement of the compression

flange.

(b) Torsional restraint, which prevents rotation of a member about itslongitudinal axis.

(c) Intermediate lateral restraint to the tension flange. Such restraints are only

of limited benefit, but do modify the out-of-plane buckling mode and may

therefore allow the distance between torsional restraints to be increased.

As shown in Figure 6.3, practical details may provide more than one type of

restraint.




4 - 27

1 Stay

Figure 6.3 Example of combined lateral and tors ional restraint

Purlins attached to the top flange of the rafter and side rails attached to the

outer flange of the column provide stability to the rafter in a number of ways:

Direct lateral restraint, when the outer flange is in compression.

Intermediate lateral restraint to the tension flange between torsional

restraints, when the outer flange is in tension.

Torsional and lateral restraint to the rafter when the purlin is attached to thetension flange and used in conjunction with rafter stays to the compression

flange.

In all cases, the purlins and side rails should be tied back into a system of

bracing in the plane of the rafters (see Section 9). Generally, the assumption

that the forces are carried back to the bracing system via the roof diaphragm is

accepted in many countries, even without supporting calculations. In other

countries calculations are necessary, or the purlins can only be assumed to

provide restraint if they are aligned directly with the bracing system.

The position of the purlins and side rails will be a balance between the capacityof the purlins themselves, and the necessary spacing required to restrain the

primary steel members. The maximum spacing will usually be determined

from manufacturers’ load tables. Spacing may have to be reduced to provide

restraint to the inside flange at strategic points along the rafter or column, so it

would be common to provide purlins at reduced spacing in zones of high

bending moment, such as around the eaves haunch.

Normal practice is to locate one purlin at the ‘sharp’ end of the haunch, and

one near the apex. The intervening length is split at regular spacing – typically

about 1,6 to 1,8 m. A purlin is often located near the end plate of the rafter, and

depending on the length of the haunch, one, two or more purlins in the lengthto the ‘sharp’ end of the haunch, usually at lesser spacing than the main length

of rafter.

Additional purlins may be required to carry drifted snow – these may also be

used to provide restraint.

Side rails are usually located at positions to suit the cladding, doors and

windows. The inside of the flange at the underside of the haunch always

requires restraint – it is common to position a side rail at this level.

Purlins and side rails must be continuous in order to offer adequate restraint, asshown in Figure 6.3. A side rail that is not continuous (for example,

interrupted by industrial doors) cannot be relied upon to provide adequate

restraint.








4 - 30

Figure 6.5 Decision tree for selecting appropr iate stable length cri teria for any segment in a portal frame – Sheet 2




4 - 31

Figure 6.6 Decision tree for selection of appropriate stable length criteria in a portal f rame – Sheet 3

6.5 Design summary Before proceeding to the detailed verification of rafter and column stability,

designers should appreciate that:

Torsional and lateral restraints need to be provided at all hinge positions, as

required by § 6.3.5.2.

EN 1993-1-1 recognises four different types of stable lengths, Lstable, Lm, Lk

and Ls, adjacent to plastic hinge positions. Lateral restraints must be

provided adjacent to the hinge at no greater distance than Lstable or Lm and

torsional restraints at no greater distance than Lk or Ls, as appropriate.

In zones where there is no plastic hinge, each member must satisfy thesimplified forms of Expressions 6.61 and 6.62. These consider in-plane and

out-of-plane stability and their potential interaction.








4 - 34

plastic hinge is predicted at this position, a restraint must be located within h/2

of the hinge position, where h is the depth of the rafter. In Figure 7.2, a hinge is

predicted at point 7, and a restraint to the bottom flange has been provided. The

restraints to each flange in the haunch region are shown in Figure 7.3.

1

2

4

53

6

1. Zone A

2. Depth of haunch

3 Intermediate restraint between torsional restraints4. Torsional restraints

5. Depth of rafter

6. Restraints to flange

Figure 7.3 Restraints in the haunched region of a portal frame

It is necessary to check that the distance between torsional restraints (in

Figure 7.2 this is indicated as ‘1’ in zone A) on both sides of a plastic hinge

does not exceed Ls as given in § BB.3.2.2. In zone A, the member is tapered,

and the bending moment is not constant.

Ls is given in § BB.3.2.2 Expression BB.11 for a three flange haunch andExpression BB.12 for a two-flange haunch. In both cases, a factor C n (given in

BB.3.3.2) takes account of non-linear moment gradients by calculating relevant

parameters at the five cross-sections, as shown in Figure 7.4. The parameter c

is a taper factor, given in § BB.3.3.3(1)B. § BB.3.2.2 also demands that the

spacing of intermediate lateral restraints satisfies the requirements for Lm given

in § BB.3.2.1. In Figure 7.2, both lengths indicated ‘2’ must satisfy this check.

Expression BB.9 is used for a three flanged haunch and BB.10 for a

two-flanged haunch. A three flanged haunch would be the common situation

when the haunch is fabricated from a section cutting and welded to the

underside of the rafter.






4 - 36

countries any purlins providing restraint must be connected directly to the

bracing system.

The out-of-plane checks require the verification of the member in accordance

with Expressions 6.61 and 6.62 (see Section 6.2 of this document). Normally,

if the purlins are regularly spaced, it is sufficient to check the rafter betweenrestraints assuming the maximum bending moment and maximum axial load.

If a plastic hinge is predicted to form adjacent to the apex, it must be

restrained. In addition, the usual requirements for stability near a plastic hinge

must be satisfied:

The distance between the restraint at the plastic hinge and the next lateral

restraint must not exceed the limiting distance Lm.

The distance to the next torsional restraint each side of the hinge must not

exceed the limiting distance Lk , or Ls, with the spacing of intermediate

restraints satisfying the requirements for Lm, all as described for zone B.

Even if there is no plastic hinge adjacent to the apex, it is normal practice to

provide a torsional restraint at this point, as this will be necessary when

considering the uplift combinations of actions – the bottom flange will be in

compression.

7.3.2 Rafter and haunch stability for uplift conditions

Under uplift, most of the bottom flange of the rafter is in compression. A

typical reversal bending moment diagram is shown in Figure 7.5.

1

1

2

E

F

3

1 Torsional restraint

2 Torsional restraint to column

3 Possible additional torsional restraint required for the uplift condition.

Figure 7.5 Typical purlin and rafter stay arrangement for wind uplift

This type of bending moment diagram will generally occur under internal

pressure and wind uplift. Normally, the bending moments are smaller than the

gravity load combinations and the members will remain elastic. The stability

checks recommended below assume that plastic hinges will not occur in thisuplift condition.




4 - 37

Haunch stability in Zone E

In Zone E, (see Figure 7.5) the top flange of the haunch will be in compression

and will be restrained by the purlins.

The moments and axial forces are smaller than those in the gravity load

combination. The members should be verified using Expression 6.62 (seeSection 6.2 of this document). By inspection, it should be clear that the rafter in

this zone will be satisfactory.

Stability in Zone F

In Zone F, the purlins will not restrain the bottom flange, which is in

compression.

The rafter must be verified between torsional restraints. A torsional restraint

will generally be provided adjacent to the apex, as shown in Figure 7.5. The

rafter may be stable between this point and the virtual restraint at the point of

contraflexure. If the rafter is not stable over this length, additional torsionalrestraints may be introduced, and each length of the rafter verified.

This verification may be carried out using Expression 6.62.

The beneficial effects of the restraints to the tension flange (the top flange, in

this combination) may be accounted for using a modification factor C m, taken

from § BB.3.3.1(1)B for linear moment gradients and from § BB.3.3.2(1)B for

non-linear moment gradients. If this benefit is utilised, the spacing of the

intermediate restraints should also satisfy the requirements for Lm, found in

§ BB.3.1.1.

7.4 In-plane stabil ity In addition to the out-of-plane checks described in Section 7.3, in-plane checks

must be satisfied using Expression 6.61.

For the in-plane checks, the axial resistanceM1

Ed y

N is based on the system

length of the rafter. The buckling resistanceM1

Rk y,

LT

M

should be taken as the

least resistance from any of the zones described in Section 7.3.

7.5 Design summary

Rafters should be IPE or similar sections with Class 1 or Class 2

proportions under combined moment and axial load. Sections containing

plastic hinges must be Class 1.

Cross-sections should be checked to Section 6 of EN 1993-1-1.

Detailed checks must be carried out to ensure adequate out-of-plane

stability under both gravity and uplift conditions – see Sections 7.3.1 and

7.3.2.

In-plane stability of the rafters and interaction with out-of-plane stability

must be verified, using Expressions 6.61 and 6.62 – see Section 6.2.














4 - 43

An eaves strut may be required in the end bays, depending on the configuration

of the plan bracing (see Section 9.3.2).

1

2

1 Eaves level

2 Position of plan bracing

Figure 9.1 Single diagonal bracing for low rise frames

1

2

1 Eaves level


Figure 9.2 K bracing arrangement for taller frames

9.2.3 Bracing using angle sections or flats

Cross braced angles or flats (within a masonry cavity wall) may be used as

bracing (as shown in Figure 9.3). In this case, it is assumed that only the

diagonal members in tension are effective.

1

2

1 Eaves level


Figure 9.3 Typical cross bracing system using angles or flats as tensionmembers

9.2.4 Bracing in a sing le bay

For vertical bracing provided in a single bay, an eaves strut is required to

transmit wind forces from the roof bracing into the vertical bracing

(Figure 9.4). Further details of eaves struts are given in Section 12.2.






4 - 45

2

1 1

1 Moment-resisting frames


Figure 9.6 Individual, local sway frames

1 12 222

3

1 Moment connection

2 Pin connection

3 Eaves strut

Figure 9.7 Hybrid frame along the full length of the building

In design of both systems, it is suggested that:

The bending resistance of the portalised bay (not the main portal frame) is

checked using an elastic frame analysis

Deflection under the equivalent horizontal forces is restricted to h/1000.

The stiffness is assured by restricting serviceability deflections to a

maximum of h/360, where h is the height of the portalised bay.

In some cases, it is possible to provide conventional bracing on one elevation,

and provide moment resisting frames on the other. The effects of racking

action due to the difference in stiffness of the sides is generally negligible due

to the diaphragm action of the roof.










4 - 49

Figure 9.13 and Figure 9.14. The bracing is usually attached to cleats on the

web of the rafter, as shown in Figure 9.15. The attachment points should be as

close to the top flange as possible, allowing for the size of the member and the

connection.

Location of vertical bracing

Position of gable posts

Figure 9.13 Plan view showing both end bays braced

Position of gable posts

Location of vertical bracing

Figure 9.14 Plan view showing both end bays braced where the gable postsare closely spaced

An eaves strut may be required in the end bays, depending on the configurationof the plan bracing. In all cases, it is good practice to provide an eaves tie along

the length of the building.






4 - 51

Figure 9.17 Effect of purlin flexibility on bracing

9.5 Bracing at plastic hingesSection 6.3.5.2 of EN 1993-1-1 recommends that bracing should be provided to

both tension and compression flanges at or within 0,5h of the calculated plastic

hinges, where h is the depth of the member (see Figure 9.18).

h

2

1

0.5 h0.5 h

1. Hinge position

2. Member must be braced within these limits

Figure 9.18 Bracing at plastic hinges

EN 1993-1-1 recommends that the bracing to a plastic hinge should bedesigned assuming that the compression flange exerts a lateral load of 2,5% of

the flange force, (taken as the plastic moment resistance/depth of section)

perpendicular to the web of the member.

In addition, according to § 6.3.5.2(5)B of EN 1993-1-1, the bracing system

must be able to resist the effects of local forces Qm applied at each stabilised

member at the plastic hinge locations, where:

1005,1

Edf,

mm

N Q

where:

N f,Ed is the axial force in the compressed flange of the stabilised member at

the plastic hinge location

αm is a coefficient to recognise the statistical benefits of restraining a

group of members compared with an individual member

m

115,0m in which m is the number of members to be restrained.

Where the plastic hinge is braced by diagonals from the purlins (seeFigure 6.3), the stiffness of the ‘U-frame’ formed by the purlin and diagonals is

especially important. Where the proportions of the members, purlins or

spacings differ from previous practice, the effectiveness should be checked. In




4 - 52

the absence of other methods, the stiffness check may be based on the work of

Horne and Ajmani[4]. Thus, the support member (the purlin or sheeting rail)

should have I y,s such that:

21

22

3

y

f y,

sy,

10190 L L

L L L f

I

I

where:

f y is the yield strength of the frame member

I y,s is the second moment of area of the supporting member (purlin or

sheeting rail) about the axis parallel to the longitudinal axis of the

frame member (i.e. the purlin major axis in normal practice)

I y,f is the second moment of area of the frame member about the major

axis

L is the span of the purlin or sheeting rail

L1 and L2 are the distances either side of the plastic hinge to the eaves (or

valley) or points of contraflexure, whichever are the nearest to the

hinge (see Figure 9.18).

Hinges that form, rotate then cease, or even unload and rotate in reverse, must

be fully braced. However, hinges that occur in the collapse mechanism but

rotate only above ULS need not be considered as plastic hinges for ULS

checks. These hinges are easily identified by elastic-plastic or graphical

analysis.

Analysis cannot account for all of the section tolerances, residual stresses and

material tolerances. Care should be taken to restrain points where these effects

could affect the hinge positions, e.g. the shallow end of the haunch instead of

the top of the column. Wherever the bending moments come close to the

plastic moment capacity, the possibility of a hinge should be considered.

9.6 Design summary Bracing must be provided with adequate strength and stiffness to act in

conjunction with the purlins, side rails and eaves beams to resist horizontal

actions, including wind, to provide overall stability to the building and to provide local stability to the columns and rafters. Bracing must be provided:

To side walls, in a vertical plane; see Section 9.2

On plan at or near the roof of the building; see Section 9.3

Stays are required to stabilise inner flanges of the columns and rafters

where they are in compression and potentially unstable; see Section 9.4

At, or near, plastic hinge positions to provide torsional restraint; see

Section 9.5.




4 - 53

10 GABLES

10.1 Types of gable frame

Gable frames are typically of two forms:

An identical portal frame to the remainder of the structure. The gable

columns do not support the rafter. This form of gable is used for simplicity,

or because there is the possibility of extending the structure in the future.

A gable frame comprising gable posts and simply supported rafters. The

gable posts support the rafters. Gable frames of this form require bracing in

the plane of the gable, as shown in Figure 10.1. The advantage of this form

of gable is that the rafters and external columns are smaller than those in a

portal frame.

Figure 10.1 Gable frame from columns , beams and bracing

10.2 Gable columnsGable columns are designed as vertical beams, spanning between the base and

the rafter. At rafter level, the horizontal load from the gable column is

transferred into the roof bracing, to the eaves, and then to the ground via the

bracing in the elevations.

The gable column will be designed for pressure and suction. The maximum

suction may be when the gable is on the downwind elevation, as shown in

Figure 10.2(a), or more likely when the gable is parallel to the wind direction,

as shown in Figure 10.2(b).




4 - 54

1

2

(a)

1

2

2

(b)

1 Apex

2 Gable under suction

1 Apex

2 Gable under suction

Figure 10.2 Wind loads on gables

The internal pressure or suction contributes to the net loads on the gable. When

the net loads are equivalent to an external pressure, the outside flanges of the

gable columns are in compression, but are restrained out-of-plane by the side

rails. When the net loads are equivalent to an external suction, the inside

flanges of the gable columns are in compression. This design case may be the

most onerous of the two conditions. It may be possible to reduce the length of

the unrestrained inside flange of the gable columns by introducing column

stays from the side rails, as illustrated in Figure 6.3.

10.3 Gable raftersIf the gable is of the form shown in Figure 10.1, the gable rafters are generally

simply supported I section members. In addition to carrying the vertical loads,

the gable rafters often act as chord members in the roof bracing system and this

design case must be verified.

If a portal frame is adopted as a gable frame, it is common to adopt an identical

frame size, even though the vertical loads on the end frame are rather less.

Generally, the reduced vertical loading will mean that the rafter canaccommodate the axial force as part of the roof bracing system without needing

to increase the section size.










4 - 58

2

3

4

1

5

6

(a) For columns greater than or equal to 400 mm deep the holding down bolts may be locatedentirely within the section profile

4

2

1

5

6

3

(b) For columns less than 400 mm deep the bolts may be located outside the section profile

1 Top of concrete foundation

2 Holding down bolts in clearance holes(bolt diameter +6 mm)

3 Base plate, usually 15 mm thick

4 Bedding space ( 50 mm)

5 Location tube

6 Anchor plate

Figure 11.3 Typical nominally pinned bases






4 - 60

11.3.2 Safety in erection

It is usual to provide at least four bolts in the base plate for stability during

erection. The alternative is to provide temporary support immediately after the

erection of the column, which on most sites would be impractical and is likely

to create hazards.

11.3.3 Resistance to horizontal forces

The highest horizontal forces acting at the base of the column are generally

those that act outwards as a result of bending in the column caused by vertical

loading on the roof.

Horizontal reactions acting outwards can be resisted in a number of ways, by:

Passive earth pressure on the side of the foundation, as indicated in

Figure 11.6(a)

A tie cast into the floor slab connected to the base of the column, as shown

in Figure 11.6(b)

A tie across the full width of the frame connecting both columns beneath or

within the floor slab as illustrated in Figure 11.6(c) and (d).

By far the most popular method of resisting horizontal forces is to use passive

earth pressure. This has economic advantages in that the foundation size

required to resist uplift is usually adequate to provide adequate passive bearing

against the ground. However, the passive resistance of the surrounding ground

can be less than anticipated if the ground is not compacted correctly, and

drainage and service trenches alongside the frame can reduce the passive

resistance considerably.

As an alternative, a bar connected to the column and cast into the floor slab,

and wrapped at the end to allow vertical movement, can be relatively cheap.

This detail may lead to some local cracking of the floor slab and, where a high

specification floor slab is used, the warranty on the slab may be invalidated.

The length of the bar should be determined by the ultimate pull out resistance

required to resist the horizontal force.

A tie across the full width of the frame connected to the column at each side is

the most certain way of resisting horizontal forces. It is more expensive in

terms of materials and labour and can be damaged by site activities. A fullwidth tie will generally impede the erection of the structure, which will be

undertaken from within the footprint of the building.

11.3.4 Base plates and holding down bolts

The steelwork contractor will usually be responsible for detailing the base plate

and holding down bolts. However, it should be made clear in the contract

documentation where the responsibility lies for the design of the foundation

details, as special reinforcement spacing or details may be required.

Base plates will usually be in grade S235 or S275 steel.














4 - 66

Where the internal columns provide significant stiffness, it is uneconomic to

ignore them and a detailed analysis of the entire frame by software would be

preferable.

13.4 Snap through instabili ty

Figure 13.4 Snap through instabilit y

As shown in Figure 13.4, the reduced sway stiffness of frames with three or

more bays may lead to snap through instability of an internal bay. Suchstructures may be checked with appropriate software to ensure satisfactory

behaviour. Appendix B may be used to calculate an estimate of the sensitivity

to snap through.

13.5 Design summary

Many aspects of behaviour of multi-bay portal frames are similar to single

bay frames

Special consideration should be given to the sway stability and snap

through stability of multi-bay frames.






4 - 68




4 - 69

APPENDIX A Practical deflection limits for single-storey buildings

A.1 Horizontal deflections for portal frames

Figure A.1 Definition of horizontal deflection

Horizontal deflection limits for portal frame structures are not explicitly

covered in the structural Eurocodes. Generally, limits are set nationally, either

by regulation or by accepted industry practice.

Typical limiting values for horizontal deflection are given in Table A.1.




4 - 70

Table A.1 Typical horizontal deflection limits

Country StructureDeflection

limitsu

Comments

France Portal frames without gantry cranes Buildings with no particularrequirements regarding thedeflection.

Deflection at the top of thecolumns

H /150

Difference of deflectionbetween two consecutiveportal frames

B/150

Values are given in the FrenchNational Annex to EN 1993-1-1and should be used if nothingelse is agreed with the client.

The values of the deflectionscalculated from thecharacteristic combinationsshould be compared to theselimits.

Member suppo rting metal cladding

Post H /150

Rail B/150Other sing le-storey buildings Buildings with particularrequirements regarding thedeflection (brittle walls,appearance etc..

Deflection at the top of thecolumns

H /250

Difference of deflectionbetween two consecutiveportal frames

B/200

Germany There are no nationaldeflection limits. The limitsshould be taken frommanufacturers instructions(technical approvals) or shouldbe agreed with the client.

Spain Portal frames (without fragileelements susceptible tofailure in the envelopes,façade and roof)

H/150 Values are given in the nationaltechnical document for steelstructures] and in the TechnicalBuilding Code and should beused if nothing else is agreedwith the client.

Single-storey buildings withhorizontal roofs (withoutfragile elements susceptibleto failure in the envelopes,façade and roof)

H/300






4 - 72

w c : precamber in the unloaded structural member

w 1 : Initial part of the deflection under permanent loads of the relevant combination of actions

w 2 : Long-term part of the deflection under permanent loads, not to be considered for

single-storey steel buildings,

w 3 : Additional part of the deflection due to the variable actions of the relevant combination

of actions

w tot =w 1 +w 2 +w 3

w max : Remaining total deflection taking into account the precamber

Figure A.3 Definition of vertical deflections

Table A.3 Recommended limiting values for vertical deflections

Deflection limit sCountry Structure

W max W a

Comments

France Roofs in general L/200 L/250

Roofs frequentlycarrying personnelother than formaintenance

L/200 L/300

Roofs supportingplaster or other brittletoppings or non-flexible

parts

L/250 L/350

Values are given in the NationalAnnex to EN 1993-1-1 and shouldbe used if nothing else is agreedwith the client.

The values of the deflectionscalculated from the characteristiccombinations should be comparedto these limits.

Germany There are no national deflectionlimits. The limits should be takenfrom manufacturers’ instructions(technical approvals) or should beagreed with the client.

Roofs in general L/300(*) -

Roofs with access onlyfor maintenance

L/250(*)

Spain Values are given in the nationaltechnical document for steelstructures and in the TechnicalBuilding Code and should be used if nothing else is agreed with theclient.

(*) This values refers to w 2 +w 3 but w 2 =0 for steel structures.

Ultimate limit state: Ponding

Where the roof slope is less than 5%, additional calculations should be made to

check that collapse cannot occur due to the weight of water:

either collected in pools which may be formed due to the deflection of

structural members or roofing material

or retained by snow.

These additional checks should be based on the combinations at the Ultimate

Limit States.

Precambering of beams may reduce the likelihood of rainwater collecting in

pools, provided that rainwater outlets are appropriately located.

w c

w max

w1

w2 w3

wtot




4 - 73

APPENDIX B Calculation of cr,est

B.1 General

EN 1993-1-1 § 5.2.1 (4) B gives:

EdH,Ed

Edcr

h

V

H

However, this can only be applied when the axial load in the rafter is not

significant. Note 2B of § 5.2.1(4)B describes significant as when

Ed

y3,0

N

Af , which may be rearranged to indicate that the axial load is not

significant when cr Ed 09,0 N N

Where:

N cr is the elastic critical buckling load for the complete span of the rafter

pair, i.e.2

2

cr L

EI π N

L is the developed length of the rafter pair from column to column,

taken as span/Cos θ (θ is the roof slope).

If the axial load in the rafter exceeds this limit, the expression in EN 1993-1-1

cannot be used.

An alternative expression, accounting for the axial force in the rafter, has been

developed by J. Lim and C. King[6] and is detailed below.

For frames with pitched rafters:

cr,est = min estr,cr,ests,cr, ;

where:

cr,s,est is the estimate of cr for sway buckling mode

cr,r,est is the estimate of cr for rafter snap-through buckling mode.

This mode need only be checked when there are three or more

spans, or if the rafter is horizontal, or when the columns are not

vertical.

B.2 Factor cr,s,est

The parameters required to calculate cr,s,est for a portal frame are shown in

Figure B.1. NHF is the lateral deflection at the top of each column when

subjected to a notional lateral force H NHF. (The magnitude of the total lateral

force is arbitrary, as it is simply used to calculate the sway stiffness). Thehorizontal force applied at the top of each column should be proportional to the

vertical reaction.




4 - 74

The practical application of this recommendation is to calculate H NHF as 1/200

of the vertical reaction at the base of the column. In combinations including

wind actions, H NHF should still be calculated as 1/200 of the vertical reaction at

the base.

In calculating NHF only the notional lateral forces, H NHF, are applied to theframe. Base stiffness may be included in the analysis (as described in

Section 3.4).

L

h

H H NHFNHF

NHF NHF

3

1

Ed Ed

2

N N

1 Frame dimensions

2 ULS analysis, andN Ed in rafter

3 Sway analysis, underH NHF alone

Figure B.1 Calculation of cr

cr can then be calculated as:

NHF

cr200

h

The lowest value of cr for any column is taken for the frame as a whole.

cr,s,est can then be calculated as:

cr

maxRcr,

Ed est s,cr, 18,0

N

N

where:

maxRcr,

Ed

N

N is the maximum ratio in any rafter

Ed N is the axial force in rafter at ULS (see Figure B.1)

2r

2

Rcr, L

EI N

is the Euler load of the rafter for the full span of the rafter

pair (assumed pinned).






4 - 76

APPENDIX C Determination of M CR and N cr

C.1 M cr for uniform members

C.1.1 General expressionThe method given in C.1.1 only applies to uniform straight members for which

the cross-section is symmetric about the bending plane.

g2

2

g2

z2

t

2

z

w

2

w2

z2

1cr z C z C EI

GI kL

I

I

k

k

kL

EI C M

In the case of a portal frame, k = 1 and k w = 1. The transverse load is assumed

to be applied at the shear centre and therefore C 2 z g = 0. The expression may be

simplified to:

z2

t2

z

w

2

z2

1cr EI

GI L

I

I

L

EI C M

E is Young modulus (E = 210000 N/mm2)

G is the shear modulus (G = 81000 N/mm2)

I z is the second moment of area about the weak axis

I t is the torsional constant

I w is the warping constant L is the beam length between points of lateral restraint

C 1 depends on the shape of the bending moment diagram

C.1.2 C 1 factor

The factor C 1 may be determined from Table C.1 for a member with end

moment loading, and also for members with intermediate transverse loading.




4 - 77

Table C.1 C 1 factor

End Moment Loading C 1

M M

-1 +1

+1,00+0,75+0,50

+0,250,00–0,25–0,50–0,75–1,00

1,001,171,36

1,561,772,002,242,492,76

Intermediate Transverse Loading

0,94 1,17

2/3

1/3

0,62 2,60

0,86 1,35

0,77 1,69

C.2 M cr for members with discrete restraints to thetension flangeIt is possible to take beneficial account of restraints to the tension flange. This

may lead to a greater buckling resistance of the member.

Tension flange restraint is usually provided by elements connected to the

tension flange of the member (e.g. purlins).

The spacing between tension flange restraints must satisfy the requirements for

Lm as given in § BB.3.1.1 in EN 1993-1-1.

C.2.1 General expression

For the general case of a beam of varying depth but symmetrical about theminor axis, subject to a non-uniform moment:

cr0m2

cr M C c M for beams with a linearly varying moment diagram

or

cr0n2

cr M C c M for beams with a non-linearly varying moment diagram

where

M cr0 is the critical moment for a beam subject to uniform moment.

Expressions of M cr0 is given in C.2.2

c accounts for taper (c = 1 for uniform straight member)

The value of c is given by EN 1993-1-1 Annex BB.3.3.3 based on the




4 - 78

depth at the shallower end of the member and limited to members

where 1 ≤ hmax/hmin ≤ 3. Note that the expression for c was derived in

reference 4 for elements with 1.05, which is the common case for

haunches in portal frames

C m accounts for linear moment gradients. The value is given by the

Expression BB.13 of EN 1993-1-1 Annex BB. It is recommended that

C m ≤ 2,7

C n accounts for non-linear moment gradients. The value is given by the

Expression BB.14 of EN 1993-1-1 Annex BB. It is recommended that

C n ≤ 2,7

When using EN 1993-1-1 Annex BB.3.3.2, the following points need

clarification:

The same definition of ‘positive’ and ‘negative’ moments applies as in

BB.3.3.1: Moments that produce compression in the non-restrained flange

should be taken as positive.

This is fundamental as only positive values of R should be taken.

BB.3.3.2 assumes that the loads are applied at the shear centre.

C.2.2 Calculation of M cr0

For uniform sections, symmetric about the minor axis, restrained along the

tension flange at intervals:

t

2t

w2

2t

2z

2

cr0

2

1GI

L

EI

L

a EI

a M

but

z2

t2

z

w2

z2

cr0

π

EI

GI s

I

I

s

EI M

where:

a is the distance between the restrained longitudinal axis (e.g. the

centroid of the purlins) and the shear centre of the member. This takes

account of the fact that the effective restraint is provided slightly away

from the flange Lt is the length of the segment along the member between torsional

restraints to both flanges

s is the distance between the restraints along the restrained longitudinal

axis (e.g. the spacing of the purlins).

For tapered or haunched members, M cr0 is calculated using the section

properties of the shallow ends.

The parameters a, Lt and s are shown in Figure C.1




4 - 79

1 Shear centre of the shallowestcross-section

2 Axis where restraint is provided

3 Intermediate lateral restraints (purlins)

4 Lateral restraints to both flanges, providingtorsional restraint

5 Compression flange

Figure C.1 Arrangement of tension flange restraints

C.3 N cr for uniform members with discrete restraintsto the tension flangeIt is possible to take beneficial account of restraints to the tension flange. This

may lead to a greater buckling resistance of the member.

Tension flange restraint is usually provided by elements connected to the

tension flange of the member (e.g. purlins).

C.3.1 General expression

For Class 1, 2, and 3 cross-sections, § 6.3.1.2 of EN 1993-1-1 gives

cr

y

N

Af where

2

2

cr

π

L

EI N for flexural buckling

3

5

5

5

4

4

4

4

4

4

2

2

2

3

3

1

2

L

L

L

s s

s s

s s

a

t

t

t




4 - 80

C.3.2 N crT for uni form members with discrete restraints to the tensionflange

The elastic critical buckling force for an I section with intermediate restraints

to the tension flange is given in BB.3.3.1 as:

t2

t

w2

2t

2z

2

2crT1 GI

L

EI

L

a EI

i N

s

where:

22z

2y

2s aiii

Lt is the length of the segment along the member between torsional

restraints to both flanges

a is defined in C.1.

For tapered or haunched members, N crT is calculated using the section properties of the shallow ends.




4 - 81

APPENDIX D

Worked Example: Design of portal frame using elastic

analysis



4 - 82

APPENDIX D Worded Example: Design of portal frame using elastic analysis

1 of 44

Made by CZT Date 12/2009Calculation sheet

Checked by DGB Date 12/2009

1. Elastic analysis of a single bay portal frameThis example covers the design of a portal frame for a single-storey building,

using the elastic method of global analysis. Only gravity loads are covered in

this example. The frame uses hot rolled I sections for rafters and columns.

2. Frame geometry

5°

LC

30000

6 0 0 0

5 2 7 5

3020

Spacing of portal frames = 7,2 m

The cladding to the roof and walls is supported by purlins and side rails.

The purlins have been provisionally located at intervals of between 1500 mm

and 1800 mm as shown. The side rails are provisionally located at intervals of

no more than 2000 mm. The rafter and column verifications may require theselocations to be modified.



TitleAPPENDIX D Worked Example: Design of portal frame using elastic

analysis2 of 44

4 - 83

*

* * * *

* *

6 0 0 0

5 2 7 5

1 4 7 5

1 9 0 0

1 9 0 0

3 0 2 0

1 5 0 0 0

7 2 5

8 0 0

1 3

4 5

2992

14892

15057

302

1647

1 6 5

L C

5 °

3 0 2

1 3 4 5

1 1 9 8 0

7 3 1 3

1 7 0 0

1 7 0 0

1 7 0 0

1 7 0 0

1 7 0 0

1 7 0 0

1 7 0 0

13192

11492

9792

8092

6392

4692

torsional restraint to inside flange




analysis3 of 44

4 - 84

3. Loads

3.1. Permanent loads

G = Gself-weight + Groof

Gself-weight: self-weight of the beams

Groof : roofing with purlins Groof = 0,30 kN/m2

for an internal frame: Groof = 0,30 × 7,20 = 2,16 kN/m

EN 1991-1-1

=2,16 kN/m +self weightG

30 m

3.2. Snow loads

The characteristic value for snow loading on the roof for a specific location in

a given country at certain altitude has been calculated as:

sk = 0,618 kN/m²

for an internal frame: s = 0,618 × 7,20 = 4,45 kN/m

EN 1991-1-3

30 m

=4,45 kN/ms

3.3. Imposed load on roof Characteristic values for loading on the roof (type H: not accessible).

qk = 0,4 kN/m2

for an internal frame: qk = 0,4 × 7,20 = 2,88 kN/m

EN 1991-1-1

Table 6.10

30 m

Qk =2,88 kN/m




analysis4 of 44

4 - 85

3.4. Load combinations

For simplicity, the wind actions are not considered in this example.

Therefore, the critical design combination for choosing the member size is: G

G + Q Q

Where:

Q is the maximum of the snow load and the imposed load.

G = 1,35 (permanent actions)

Q = 1,50 (variable actions)

EN 1990

The snow loads are greater than the imposed loads on the roof, therefore

Q = 4,45 kN/m

4. Preliminary sizing Single-storey steel buildings. Part 2: Concept design [2] provides a table of

preliminary member sizes, according to the rafter load and the height to

eaves.

Rafter load = 1,35( 2,16 + self weight )+1,5 4,45 = 9,6 kN/m + self weight

Say 10 kN/m to include self weight.

The section chosen for the rafter is an IPE 450, S355

The section chosen for the column is an IPE 500, S355

5. Buckling amplification factor cr In order to evaluate the sensitivity of the frame to 2nd order effects, the

buckling amplification factor, cr , has to be calculated. This calculation

requires the deflections of the frame to be known under a given load

combination.

EN 1993-1-1

§5.2.1

An elastic analysis is performed to calculate the reactions under vertical loads

at ULS, which provides the following information:

The vertical reaction at each base: V Ed = 168 kN

The horizontal reaction at each base: H Ed

= 116 kN

The maximum axial force in the rafters: N R,Ed = 130 kN

5.1. Axial compression in the rafter

According to the code, if the axial compression in the rafter is significant then

cr is not applicable. In such situations, Appendix B of this document

recommends the use of cr,est instead.

The axial compression is significant if Ed

y3,0

N

Af

or if N Ed 0,09 N cr , which is an equivalent expression.

EN 1993-1-1

§5.2.1(4)

Note 2B

N Ed is the design axial load at ULS in the rafter, noted N R,Ed in this example.




analysis5 of 44

4 - 86

Lcr is the developed length of the rafter pair from column to column.

Lcr =o5cos

30= 30,1 m

N cr =2

cr

z2

L

EI =

3

23

42

10101,30

1033740210000

= 772 kN

0,09 N cr = 77209,0 = 69 kN

N R,Ed = 130 kN > 69 kN

Therefore the axial compression in the rafter is significant and cr from

EN 1993-1-1 is not applicable.

Following the guidance from Appendix B, frame stability is assessed based

on cr,est, in Section 5.2.

5.2. Calculation of cr,est

For a pitched roof frame: cr,est = min( cr,s,est; cr,r,est)

cr,r,est only needs to be checked for portal frames of 3 or more spans. Appendix B of

this document

When assessing frame stability, allowance can be made for the base stiffness.

In this example, a base stiffness equal to 10% of the column stiffness has been

assumed to allow for the nominally pinned bases.

To calculate cr , a notional horizontal force is applied to the frame and the

horizontal deflection of the top of the columns is determined under this load.

The notional horizontal force is:

H NHF =200

1V Ed = 168

200

1 = 0,84 kN

Appendix B of

this document

The horizontal deflection of the top of the column under this force is obtainedfrom the elastic analysis as 1,6 mm.

1,6 mm 1,6 mm

H H NHF NHF

cr,s,est is calculated as follows:




analysis6 of 44

4 - 87

cr,s,est =

NHFmaxcr R,

EdR,

200

118,0

h

N

N

=

6,16000

2001

77213018,0 = 12,5

Appendix B of this document

Thus cr,est = cr,s,est = 12,5 > 10

First order elastic analysis may be used and second order effects do not need

to be allowed for.

Section 2.2 of thisdocument

6. Frame imperfectionsThe global initial sway imperfection may be determined from

= 0 h m 0 = 1/200

h = 82,00,6

22

h

m = 87,0)1

1(5,0 m

= )2

11(5,0 = 0,87

m = 2 (number of columns)

=

3

1056,387,082,0200

1

EN 1993-1-1

§5.3.2

Initial sway imperfections may be considered in two ways:

By modeling the frame out of plumb

By applying equivalent horizontal forces (EHF).

Applying equivalent horizontal forces is the preferred option and the method

that is used in this worked example. The equivalent horizontal forces are

calculated as:

H EHF = V Ed

However sway imperfections may be disregarded where H Ed 0,15 V Ed. EN 1993-1-1

§5.3.2(4)

Table 1 shows the total reactions for the structure to determine H Ed and V Ed.

Table 1 Vertical and horizontal reactions

Left column (kN)Right column(kN)

Total reaction(kN)

0,15 VEd (kN)

H Ed V Ed H Ed V Ed H Ed V Ed

Reactions 116 168 –116 168 0 336 50

H Ed = 0 0,15 V Ed Therefore the initial sway imperfections have to be taken into account.




analysis7 of 44

4 - 88

The equivalent horizontal forces:

H EHF = V Ed,column = 1681056,3 3 = 0,60 kN

This force is applied at the top of each column, in combination with the

permanent and variable actions.

For the ULS analysis, the bases are modeled as pinned. Otherwise the basedetails and foundation would need to be designed for the resulting moment.

The following figure shows the internal forces on the frame subject to the

ULS loads including the equivalent horizontal forces.




analysis8 of 44

4 - 89

5 2 7 5

M

= 0 k N m

V N M

= 0 k N m

V N M

= 6 9 3 k N m

V N M

= 2 9 2 k N m

V N M

= 3 5 6 k N m

L CV N M

= 3 5 1 k N m

V N M

= 0 k N m

V N M

= 7 0 1 k N m

V N M

= 2 9 8 k N m

3 0 1 1 5

8 6 9

3 0 1 1

5 9 4 1

3 0 0 0 0

V N M

= 6 1 6 k N m

V N M

= 6 1 0 k N m

= 1 1 8 k N

= 1 2 7 k N

= 1 2 4 k N

= 1 5 0 k N

= 1 3 0 k N =

1 1 7 k N

= 1 6 2 k N

= 1 0 k N

= 1 1 6 k N

= 0 k

N

= 1 1 7 k N

= 8 6 k N

= 1 2 4 k N

= 1 1 7 k N

= 1 2 7 k N

= 1 5 0 k N

= 1 3 0 k N

= 1 1 6 k N

= 1 6 1 k N

E d E

d E d

= 8 7 k N

E d

E d

E d

E d E d E d

E d

E d

E d

E d

E d

E d E

d

E d

E d E

d

E d E d

E d

E d

E d

E d

E d

E d

E d

E d

E d E

d




analysis9 of 44

4 - 90

7. Summary of member verificationThe cross-section resistance and the buckling resistance are verified for eachmember. Sections 7.1and 7.2 provide a summary of the checks carried out for

each member of the frame.

7.1. Cross-section verification

The resistance of the cross-section has to be verified in accordance withSection 6.2 of EN 1993-1-1.

The cross-sectional checks carried out in this worked example are:

Shear resistance

V Ed V pl,Rd =

M0

yv 3

f A

EN 1993-1-1

§6.2.6

Compression resistance

N Ed N c,Rd =M0

y

A f

EN 1993-1-1

§6.2.4

Bending moment resistance

M Ed M pl,y,Rd =M0

yy pl,

f W

EN 1993-1-1

§6.2.5

In addition, bending and shear interaction, as well as bending and axial force

interaction must be verified.

EN 1993-1-1

§6.2.8§6.2.9

7.2. Buckling verification

The rafters and the columns must be verified for out-of-plane buckling between restraints and in plane buckling.

The buckling checks due to the interaction of axial force and bending moment

are carried out using Expressions 6.61 and 6.62 from EN 1993-1-1.

0,1

M1

Rk z,

Edz,Edz,

yz

M1

Rk y,

LT

Edy,Edy,

yy

M1

Rk y

Ed

M

M M k

M

M M k

N

N

0,1

M1

Rk z,

Edz,Edz,

zz

M1

Rk y,

LT

Edy,Edy,

zy

M1

Rk z

Ed

M

M M k

M

M M k

N

N

EN 1993-1-1

Expressions

(6.61) and (6.62)




analysis10 of 44

4 - 91

For orthodox single-storey portal frames, these expressions can be simplifiedas follows:

Edy, M = 0 and Edz, M = 0 for Class 1, Class 2 and Class 3 sections.

M z,Ed = 0

Therefore expressions (6.61) and (6.62) can be written as:

0,1Rd b,

Edy,

yy

Rdy, b,

Ed M

M k

N

N and 0,1

Rd b,

Edy,

zy

Rdz, b,

Ed M

M k

N

N

Expression (6.61) is used to verify in-plane buckling, and expression (6.62) isused to verify out-of-plane buckling.

COLUMN: IPE 500, S355

1 4 7 5

6 0 0 0

0 kNm

616 kNm

444 kNm

1 9 0 0

1 9 0 0

221 kNm

*V

V

=117 kN

=117 kN

N

N

=162 kN

=168 kN

Ed

Ed

Ed

Ed

Section properties:

500h mm 11600 A mm2

200b mm3

y pl, 102194W mm3

2,10w t mm4

y 1048200 I mm4 204y i mm

16f t mm 4z 102142 I mm4 1,43z i mm

21r mm 4t 103,89 I mm

4

468w h mm9

w 101249 I mm6

426d mm

7.3. Cross-section classif ication

7.3.1. The web

wt

c=

2,10

426= 41,8

EN 1993-1-1Table 5.2

(Sheet 1)




analysis11 of 44

4 - 92

d N =yw

Ed

f t

N =

3552,10

168000

= 46,4

=w

Nw

2d

d d

= 4262

4,46426

= 0,55 > 0,50

The limit for Class 1 is :113

396

=

155,013

81,0396

= 52,2

Then :wt

c= 41,8 52,2

The web is class 1.

7.3.2. The flange

f t

c= 16

9,73= 4,6

The limit for Class 1 is : 9 ε = 9 0,81 = 7,3

Then :f t

c= 4,6 8,3

The flange is Class 1

EN 1993-1-1

Table 5.2 (Sheet2)

So the section is Class 1. The verification of the member will be based on the

plastic resistance of the cross-section.

7.4. Resistance of the cross-section

7.4.1. Shear resistance

Shear area: Av = A 2bt f + (t w+2r )t f but not less than hwt w

Av = 16)2122,10(16200211600 = 6035 mm2 EN 1993-1-1§6.2.6

Conservatively = 1,0. Therefore:

Av hwt w = 2,104680,1 = 4774 mm2

Av = 6035 mm2

fromEN 1993-1-1§6.2.6(3)

V pl,Rd = M0

yv 3

f A= 310

0,1

33556035 = 1237 kN

V Ed = 117 kN < 1237 kN OK

Bending and shear interaction

When shear force and bending moment act simultaneously on a cross-section,

the shear force can be ignored if it is smaller than 50% of the plastic shear

resistance.

V Ed = 117 kN < 0,5 V pl,Rd = 0,5 1237 = 619 kN

EN 1993-1-1§6.2.8

Therefore the effect of the shear force on the moment resistance may be

neglected.




analysis12 of 44

4 - 93

7.4.2. Compression resistance

N c,Rd =M0

y

A f = 310

0,1

35511600

= 4118 kN

N Ed = 168 kN N c,Rd = 4118 kN OK

EN 1993-1-1§6.2.4

Bending and axial force interaction

When axial force and bending moment act simultaneously on a cross-section,

the axial force can be ignored provided the following two conditions are

satisfied:

N Ed 0,25 N pl,Rd and N Ed M0

yww5,0

f t h

0,25 N pl,Rd

= 0,25 4118 = 1030 kN

3

M0

yww10

0,1

3552,104685,05,0

f t h= 847 kN

168 kN < 1030 kN and 847 kN, OK

Therefore the effect of the axial force on the moment resistance may be

neglected.

EN 1993-1-1§6.2.9

Bending moment resistance EN 1993-1-1§6.2.5

M pl,y,Rd =M0

y pl

f W

=6

3

100,1

355102194

= 779 kNm

M y,Ed = 616 kNm < 779 kNm OK

7.5. Out-of-plane buckling

The out-of-plane buckling interaction is verified with expression (6.62) in

EN 1993–1–1.

0,1

Rd b,

Edy,

zy

Rdz, b,

Ed

M

M k

N

N

This expression should be verified between torsional restraints.

If the tension flange is restrained at discreet points between the torsional

restraints and the spacing between the restraints to the tension flange is small

enough, advantage may be taken of this situation.

In order to determine whether or not the spacing between restraints is small

enough, Annex BB of EN 1993-1-1 provides an expression to calculate the

maximum spacing. If the actual spacing between restraints is smaller than this

calculated value, then the methods given in Appendix C of this document may

be used to calculate the elastic critical force and the critical moment of the

section.




analysis13 of 44

4 - 94

Verification of spacing between intermediate restraints

In this case the restraint to the tension flange is provided by the siderails.

These siderails are spaced at 1900 mm.

The limiting spacing as given by Annex BB of EN 1993-1-1 is:

Lm =2

y

t

2

y pl,

21

Ed

z

235756

1

4,57

1

38

f

AI

W

C A

N

i

EN 1993-1-1Annex BB§BB.3.1.1

C 1 is a factor that accounts for the shape of the bending moment diagram. C 1

values for different shapes of bending moment diagrams can be found in

Appendix C of this document.

For a linear bending moment diagram, C 1 depends on the ratio of the

minimum and the maximum bending moments in the segment beingconsidered.

The ratios of bending moments for the middle and bottom segments of the

column (without considering the haunch) are as follows:

=444

222= 0,50 1C = 1,31

Appendix C of this document

=222

0= 0 1C = 1,77

1C = 1,31 is the most onerous case and therefore this is the case that will be

analysed.

Lm =

2

4

23

2

3

235

355

103,8911600

102194

31,1756

1

11600

10168

4,57

1

1,4338

Lm = 1584 mm

Siderail spacing is 1900 mm > 1584 mm

Therefore the normal design procedure must be adopted and advantage may

not be taken of the restraints to the tension flange.

7.5.2. Whole column (5275 mm)

Firstly the whole column is verified. If the flexural buckling, lateral torsional

buckling and interaction checks are satisfied for the length of the whole

column, no further restraints are required. Otherwise, intermediate torsional

restraints will be introduced to the column, or the column size increased.

Flexural buckling resistance about the minor axis, N b,z,Rd

b

h

200

500 2,5

t f 16 mm




analysis14 of 44

4 - 95

buckling about z-z axis:

Curve b for hot rolled I sections

z 0,34

EN 1993-1-1Table 6.2Table 6.1

1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

z =1z

cr 1

i

L=

4,76

1

1,43

5275 = 1,60

z = 2zzz 2,015,0

= 260,12,060,134,015,0 = 2,02

EN 1993-1-1§6.3.1.2

z =22

1

=22 60,102,202,2

1

= 0,307

N b,z,Rd =M1

yz

Af = 310

0,1

35511600307,0

= 1264 kN

N Ed = 168 kN < 1264 kN OK

Lateral-torsional buckling resistance, M b,Rd

The lateral-torsional buckling resistance of a member is calculated as a

reduction factor, LT, multiplied by the section modulus and the yield strengthof the section. The reduction factor is calculated as a function of the

slenderness, LT , which depends on the critical moment of the member. The

expression for the critical moment, M cr , is given below. The factor C 1

accounts for the shape of bending moment diagram of the member. Appendix

C of this document provides values of C 1 for different shapes of bending

moment diagrams. For the case of a linear bending moment diagram, C 1

depends on the ratio of the bending moments at the ends of the member, given

as .

For the total length of the column (without the haunch):

0616

0 77,11 C

Appendix C of

this document

M cr =z

2

t2

z

w

2

z2

1 EI

GI L

I

I

L

EI C

=2

42

5275

10214221000077,1

42

42

4

9

102142210000

103,89810005275

102142

101249

M cr = 909 106 Nmm





analysis15 of 44

4 - 96

The non dimensional slenderness, LT , is calculated as:

LT cr

yy

M

f W =

6

3

10909

355102194

= 0,926

EN 1993-1-1§6.3.2.2

For the calculation of the reduction factor, LT, EN 1993-1-1 provides two

methods. The general method, applicable to any section, is given in §6.3.2.2.

§6.3.2.3 provides a method that can only be used for rolled sections or

equivalent welded sections.

In this example the second method is used, i.e. §6.3.2.3.

LT = 2LTLT,0LTLT15,0

EN 1993-1-1 recommends the following values:

LT,0 0,4

0,75

The values given in the National Annex may differ. The designer should

check the National Annex of the country where the structure is to be built.

EN 1993-1-1§6.3.2.3

b

h 2,5

Curve c for hot rolled I sections

LT 0,49


LT = 2926,075,04,0926,049,015,0 = 0,950

LT =2

LT2

LTLT

1

LT =22 926,075,0950,0950,0

1

= 0,685

EN 1993-1-1§6.3.2.3

22

LT

926.0

11

= 1,17

LT = 0,685

M b,Rd =M1

yy pl,LT

f W = 6

3

100,1

355102194685,0

= 534 kNm

M b,Rd = 616 kNm 534 kNm Fails

Since the check for lateral torsional buckling resistance alone fails, the

interaction of axial force and bending moment is not carried out.

It is necessary to introduce a torsional restraint between the haunch and the

base, as shown in the following figure. The bending moment is greater at thetop of the column and therefore the restraint is placed closer to the maximum

bending moment, rather than in the middle of the column.




analysis16 of 44

4 - 97

The restraint must be at a side rail position, since bracing from the side rail to

the inner flange is used to provide the torsional restraint.

3 8 0 0

1 4 7 5

6 0 0 0

0 kNm

616 kNm

444 kNm*

*=117 kNV

N =162 kNEd

Ed

=117 kNV

N Ed

Ed=168 kN

7.5.3. Upper segment (1475 mm)

As previously, the flexural buckling and the lateral torsional buckling checks

are carried out separately before proceeding to verify the interaction between

the two.


bh

200500 2,5

t f 16 mm

buckling about z-z axis:


z 0,34


1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

z =1z

cr 1

i

L=

4,76

1

1,43

1475 = 0,448

z = 2zzz 2,015,0

= 2448,02,0448,034,015,0 = 0,643

EN 1993-1-1§6.3.1.2

z =2

z2

zz

1

=

22 448,0643,0643,0

1

= 0,906

z = 0,906




analysis17 of 44

4 - 98

N b,z,Rd =M1

yz

Af = 310

0,1

35511600906,0

= 3731 kN

N Ed = 168 kN < 3731 kN OK


As previously the factor C 1 needs to be calculated in order to determine the

critical moment of the member.

616 kNm

444 kNm

1

4 7 5

721,0616

444 16,11 C


M cr =

z

2

t2

z

w

2

z2

1

EI

GI L

I

I

L

EI C

=2

42

1475

10214221000016,1

42

42

4

9

102142210000

103,89810001475

102142

101249

M cr = 5887 106 Nmm


LT cr

yy

M

f W =

6

3

105887

355102194

= 0,364

EN 1993-1-1§6.3.2.2

For hot rolled sections


LT,0 0,4

0,75

EN 1993-1-1§6.3.2.3

As previously:


LT 0,49





analysis18 of 44

4 - 99

LT = 2364,075,04,0364,049,015,0 = 0,541

LT =2

LT

2

LTLT

1

LT =22 364,075,0541,0541,0

1

= 1,02

EN 1993-1-1§6.3.2.3

LT cannot be greater than 1.0, therefore:

LT = 1,0

M b,Rd =M1

yy pl,LT

f W = 6

3

100,1

3551021940,1

= 779 kNm

M Ed = 616 kNm < 779 kNm OK

Interaction of axial force and bending moment – out-of-plane buckling

Out-of-plane buckling due to the interaction of axial force and bending

moment is verified by satisfying the following expression:

0,1Rd b,

Edy,

zy

Rdz, b,

Ed

M

M k

N

N

EN 1993-1-1§6.3.3(4)

For z 0.4, the interaction factor, k zy is calculated as:

k zy =

zRd, b,

Ed

mLTzRd, b,

Ed

mLT 25,01,01;

25,01,01max

N N

C N N

C z

EN 1993-1-1Annex B

Table B.2

C mLT = 4,06,0

=616

444= 0,721

C mLT = 721,04,06,0 = 0,888 0,4

C mLT = 0,888

EN 1993-1-1Annex BTable B.3

k zy =

3731168

25,0888,01,01;

3731168

25,0888,0448,01,01max

k zy = max (0,996; 0,993) = 0,996

Rd b,

Edy,

zy

Rdz, b,

Ed

M

M k

N

N =

779

616996,0

3731

168 = 0,832 < 1,0 OK

7.5.4. Lower segment (3800 mm)

As previously the flexural buckling resistance and the lateral-torsional

buckling resistance are checked individually and then the interaction betweenthe two is verified by using interaction Expression 6.62.




analysis19 of 44

4 - 100


As previously:


z 0,34


1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

z =1z

cr 1

i

L=

4,76

1

1,43

3800 = 1,15

z = 2zzz 2,015,0

z = 2

15,12,015,134,015,0 = 1,32

EN 1993-1-1§6.3.1.2

z =2

z2

zz

1

=

22 15,132,132,1

1

= 0,508

N b,z,Rd =M1

yz

Af = 310

0,1

355160010,508

= 2092 kN

N Ed = 168 kN < 2092 kN OK


As previously the C 1 factor needs to be calculated in order to determine thecritical moment of the member.

444 kNm

3 8 0 0

0444

0 77,11 C





analysis20 of 44

4 - 101

M cr =z

2

t2

z

w

2

z2

1 EI

GI L

I

I

L

EI C

=2

42

380010214221000077,1

42

42

4

9

102142210000

103,89810003800

102142

101249

M cr = 1556 106 Nmm


LT cr

yy

M

f W =

6

3

101556

355102194

= 0,708

EN 1993-1-1§6.3.2.2



LT,0 0,4 and 0,75

EN 1993-1-1§6.3.2.3

As previously:


LT 0,49


LT = 2708,075,04,0708,049,015,0 = 0,763

LT =2

LT2

LTLT

1

LT =22 708,075,0763,0763,0

1

= 0,822

EN 1993-1-1§6.3.2.3

2

LT

1

=

2708,0

1= 1,99

LT = 0,822

M b,Rd =M1

yy pl,LT

f W = 6

3

100,1

355102194822,0

= 640 kNm





0,1Rd b,

Edy,

zyRdz, b,

Ed

M

M k

N

N

EN 1993-1-1§6.3.3(4)




analysis21 of 44

4 - 102

For z 0.4, the interaction factor, k zy is calculated as:

k zy =

zRd, b,

Ed

mLTzRd, b,

Ed

mLT 25,0

1,01;

25,0

1,01max

N

N

C N

N

C

z

C mLT = 4,06,0

=444

0= 0

C mLT = 4,06,0 = 04,06,0 = 0,6 > 0,4

CmLT = 0,6


k zy =

2092

168

25,06,0

1,01;

2092

168

25,06,0

15,11,01max

k zy = max (0,974; 0,977) = 0,977

EN 1993-1-1Annex B

Table B.2

Rd b,

Edy,

zy

Rdz, b,

Ed

M

M k

N

N =

640

444977,0

2092

168 = 0,758 < 1,0 OK

7.6. In-plane buckling

The in-plane buckling interaction is verified with expression (6.61) in

EN 1993-1-1.

0,1Rd b,

Edy,

yy

Rdy, b,

Ed M

M k

N

N

M

M

Ed

Ed

Ed

Ed

Ed

Ed

V

V

N

N

=0 kNm

=616 kNm

=117 kN

=162 kN

=117 kN

=168 kN

The maximum design values of either column occur on the right hand column

(considering EHF applied from left to right) and are as follows:

M Ed 616 kNm

N Ed 168 kN




analysis22 of 44

4 - 103

Firstly individual checks are carried out for flexural buckling alone and

lateral-torsional buckling alone. Then the interaction expression for in-plane

buckling is applied to verify that the combination of axial force and bending

moment does not cause excessive buckling on the columns.

7.6.1. Flexural buckling resistance about the mayor axis, N b,y,Rd

b

h

200

500 2,5

t f 16 mm

buckling about y-y axis:

Curve a for hot rolled I sections

y 0,21


The buckling length is the system length, which is the distance between nodes(i.e. the length of the column), L = 6000 mm.

1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

y =1y

cr 1

i

L=

4,76

1

204

6000 = 0,385

y = 2yyy 2,015,0

= 2385,02,0385,021,015,0 = 0,594

EN 1993-1-1§6.3.1.2

y =22

1

=22 385,0594,0594,0

1

= 0,956

EN 1993-1-1§6.3.1.2

N b,y,Rd =M1

yy

Af = 310

0,1

35511600956,0

= 3937 kN

N Ed = 168 kN < 3937 kN OK

7.6.2. Lateral-torsional buck ling resistance,M b,Rd M b,Rd is the least buckling moment resistance of those calculated previously.

M b,Rd = 640;779min

M b,Rd = 640 kNm

7.6.3. Interaction of axial force and bending moment – in-planebuckling

In-plane buckling due to the interaction of axial force and bending moment is

verified by satisfying the following expression:

0,1Rd b,

Edy,

yy

Rdy, b,

Ed M

M k

N

N




analysis23 of 44

4 - 104

For C my, the relevant braced points are the torsional restraints at the end of the

member.

The interaction factor, k yy, is calculated as follows:

k yy =

Rdy, b,

Edmy

Rdy, b,

Edymy 8,01;2,01min

N

N C

N

N C

From table B.3, C my is:

C my = 4,06,0 0,4

0

C my = 04,06,0 = 0,6

k yy =

3937

168

8,016,0;3937

168

2,0385,016,0min

= 620,0;605,0min = 0,605

Rd b,

Edy,

yy

Rdy, b,

Ed

M

M k

N

N =

640

616605,0

3937

168 = 0,625 < 1,0 OK

Validity of column section

In Section 7.4 it has been demonstrated that the cross-sectional resistance of

the section is greater than the applied forces.The out-of-plane and in-plane buckling checks have been verified in

Sections 7.5 and 7.6 for the appropriate choice of restraints along the column.

Therefore it is concluded that the IPE 500 section in S355 steel is appropriate

for use as columns in this portal frame.

Rafter: IPE 450

13451345

17001700

17001700

17001700

1700

351 kNm

354 kNm

111 kNm

298 kNm




analysis24 of 44

4 - 105

V Ed 118 kN (maximum value)

N Ed 127 kN (maximum value)

M Ed 356 kNm (maximum value)

Section properties

450h mm 9880 A mm2

190b mm3

y pl, 101702W mm3

4,9w t mm4

y 1033740 I mm4 185y i mm

6,14f t mm4

z 101676 I mm4 2,41z i mm

21r mm4

t 109,66 I mm4

8,420w h mm 9w 10791 I mm6

8,378d mm


7.7.1. The web

wt

c =

4,9

8,378= 40,3

EN 1993-1-1Table 5.2(Sheet 1)

d N =yw

Ed

f t N =

3554,9127000

= 38

=w

Nw

2d

d d =

8,3782

388,378

= 0,55 > 0,50

The limit for Class 1 is :113

396

=

155,013

81,0396

= 52,1

Then :wt

c= 40,3 < 52,1

The web is class 1.

7.7.2. The flange

f t

c =

6,14

3,69= 4,7


Then :f t

c= 4,7 < 7,3

The flange is Class 1


Therefore, the section is Class 1. The verification of the member will be basedon the plastic resistance of the cross-section.




analysis25 of 44

4 - 106

7.8. Resistance of the cross-section


Shear area : Av = A - 2bt f + (t w+2r )t f but not less than hwt w

Av = 6,14)2124,9(6,1419029880 = 5082 mm2

EN 1993-1-1§6.2.6(3)

hwt w = 4,98,4200,1 = 3956 mm2

Av = 5082 mm2

fromEN 1993-1-1§6.2.6(3)

V pl,Rd =

M0

yv 3

f A=

3100,1

33555082 = 1042 kN

V Ed = 118 kN < 1042 kN OK

EN 1993-1-1§6.2.6(3)

Bending and shear interaction


the shear force can be ignored if it is smaller than 50% of the plastic shear

resistance of the cross-section.

EN 1993-1-1

§6.2.8

V Ed = 118 kN < 0,5 V pl,Rd = 521 kN OK


neglected.


N c,Rd =M0

y

A f = 310

0,1

3559880

= 3507 kN

N Ed = 127 kN < 3507 kN OK

EN 1993-1-1§6.2.4

Bending and axial force interaction


the axial force can be ignored provided the following two conditions are

satisfied:

N Ed 0,25 N pl,Rd and N Ed M0

yww5,0

f t h

0,25 N pl,Rd = 0,25 3507 = 877 kN

And

3

M0

yww10

0,1

3554,98,4205,05,0

f t h= 702 kN

127 kN < 887 kN and 702 kN OK

EN 1993-1-1§6.2.9

Therefore the effect of the axial force on the moment resistance may be

neglected.




analysis26 of 44

4 - 107

7.8.3. Bending moment resistance EN 1993-1-1§6.2.5

M pl,y,Rd =M0

yy pl,

f W = 6

3

100,1

355101702

= 604 kNm


7.9. Out-of-plane buckling

The out-of-plane buckling interaction is verified with expression (6.62) from

EN 1993-1-1

0,1Rd b,

Edy,

Rd b,z,

Ed M

M k

N

N zy

The rafter should be verified between torsional restraints. If advantage istaken of intermediate restraints to the tension flange, the spacing of the

intermediate restraints must also be verified.

7.9.1. Mid-span region

The purlin spacing in this region is 1700 mm.

1700 mm

1

1 Mid-span region

354 kNm351 kNm

1700

356 kNm

1 1: Bending moment

Flexural buckling resistance about minor axis bending, N b,z,Rd

b

h

190

450 2,37

t f 14,6 mm

buckling about z-z axis


z 0,34





analysis27 of 44

4 - 108

1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

z =1z

cr 1

i

L= 4,76

1

2,41

1700

= 0,540

z = 2zzz 2,015,0

z = 2540,02,0540,034,015,0 = 0,704

EN 1993-1-1§6.3.1.2

z =2

z2

zz

1

=

22 540,0704,0704,0

1

= 0,865

N b,z,Rd

=M1

yz

Af = 310

0,1

3559880865,0

= 3034 kN

N Ed = 127 kN < 3034 kN OK

Lateral-torsional buckling resistance for bending, M b,Rd

In this zone, lateral-torsional buckling is checked between restraints, which

are the purlins. For equally spaced purlins, the critical length is at the point of

maximum bending moment.

In order to determine the critical moment of the rafter, the C 1 factor takes

account of the shape of the bending moment diagram.

In this case the bending moment diagram is nearly constant along the segmentin consideration, so 1,0. Therefore:

11 C ,0 Appendix C of this document

M cr =z

2

t2

z

w

2

z2

1 EI

GI L

I

I

L

EI C

=2

42

1700

1016762100000,1

42

42

4

9

101676210000

109,66810001700

101676

10791

M cr = 2733 106 Nmm


LT cr

yy pl,

M

f W =

6

3

102733

355101702

= 0,470

EN 1993-1-1§6.3.2.2

2LTLT,0LTLTLT 15,0

EN 1993-1-1§6.3.2.3




analysis28 of 44

4 - 109

LT,0 0,4 and 0,75

b

h 2,37


LT 0,49


2LT 470,075,04,0470,049,015,0 = 0,60

LT =2

LT2

LTLT

1

LT =22

470,075,060,060,0

1

= 0,961

EN 1993-1-1§6.3.2.3

2

LT

1

=

2470,0

1= 4,53

LT = 0,961

M b,Rd =M1

yy pl,LT

f W = 6

3

100,1

355101702961,0

= 581 kNm


Interaction of axial force and bending moment – out-of-plane buckling Out-of-plane buckling due to the interaction of axial force and bending


0,1Rd b,

Edy,

zy

Rdz, b,

Ed M

M k

N

N

EN 1993-1-1§6.3.3(4)

For z 0,4, the interaction factor, k zy is calculated as:

k zy =

Rdz, b,

Ed

mLTRdz, b,

Ed

mLT 25,0

1,01;

25,0

1,01max

N

N

C N

N

C

z

The bending moment is approximately linear and constant. Therefore C mLT is

taken as 1.0

EN 1993-1-1Annex B TableB.3

k zy =

3034

127

25,01

1,01;

3034

127

25,01

540,01,01max

= max (0,997; 0,994) = 0,997


Rd b,

Edy,

zy

Rdz, b,

Ed

M

M k

N

N =

581

356997,0

3034

127 = 0,653 < 1,0 OK




analysis29 of 44

4 - 110

7.9.2. End-of-span region

In this region the bottom flange is in compression and stability must be

checked between torsional restraints.

2930 mm

1 1

1 End of span region

1230 1700

298 kNm1

2

111 kNm

1 Simplified bending moment

2 Bending moment

The buckling length is taken from the torsional restraint at the sharp end of

the haunch to the ‘virtual’ restraint which is the point of contraflexure of the

bending moment diagram, i.e. where the bending moment is equal to zero. In

some countries the assumption of a virtual restraint may not be common

practice. If the practice is not allowed, the buckling length should be taken to

the next purlin (i.e the first restraint to the compression flange).

From the analysis, the buckling length to the point of contraflexure is

2930 mm.

If the tension flange is restrained at discreet points between the torsional

restraints and the spacing between the restraints to the tension flange is small


In order to determine whether or not the spacing between restraints is small

enough, Annex BB of EN 1993-1-1 provides an expression to calculate the

maximum spacing. If the actual spacing between restraints is smaller than this

calculated value, then the methods given in Appendix C of this document may be used to calculate the elastic critical force and the critical moment of the

section.

Verification of spacing between intermediate restraints

In this case, the restraint to the tension flange is provided by the purlins.

These purlins are spaced at 1700 mm.

Lm =2

y

t

2

y pl,

21

Ed

z

235756

1

4,57

1

38

f

AI

W

C A

N

i

EN 1993-1-1Annex BB§BB.3.1.1




analysis30 of 44

4 - 111

=298

111= 0,37 1C = 1,42


Lm =

2

4

23

2

3

235

355

109,669880

101702

42,1756

1

9880

10127

4,57

1

2,4138

Lm = 1669 mm

Purlin spacing is 1700 mm > 1669 mm

Therefore the normal design procedure must be adopted and advantage may

not be taken of the restraints to the tension flange.


As previously:


z 0,34


1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

z =1z

cr 1

i

L=

4,76

1

2,41

2930 = 0,931

z = 2zzz 2,015,0

z = 2931,02,0931,034,015,0 = 1,06

EN 1993-1-1§6.3.1.2

z =2

z2

zz

1

=

22 931,006,106,1

1

= 0,638

N b,z,Rd =M1

yz

Af = 310

0,1

35598800,638

= 2238 kN

N Ed = 127 kN < 2238 kN OK


As previously the C 1 factor needs to be calculated in order to determine thecritical moment of the member. For simplicity, the bending moment diagram

is considered as linear, which is slightly conservative.

=298

0= 0 1C = 1,77





analysis31 of 44

4 - 112

M cr =z

2

t2

z

w

2

z2

1 EI

GI L

I

I

L

EI C

= 2

42

293010167621000077,1

42

42

4

9

101676210000

109,66810002930

101676

10791

M cr = 1763 106 Nmm


LT cr

yy pl,

M

f W =

6

3

101763

355101702

= 0,585

EN 1993-1-1§6.3.2.2



EN 1993-1-1§6.3.2.3

LT,0 0,4 and 0,75

As previously:


LT 0,49


LT = 2585,075,04,0585,049,015,0 = 0,674

LT =2

LT2

LTLT

1

LT =22 585,075,0674,0674,0

1

= 0,894

EN 1993-1-1§6.3.2.3

2

LT

1

=

2585,0

1= 2,92

LT = 0,894

M b,Rd =M1

yy pl,LT

f W = 6

3

100,1

355101702894,0

= 540 kNm EN 1993-1-1§6.2.5(2)




0,1Rd b,

Edy,

zy

Rdz, b,

Ed M

M k

N

N

EN 1993-1-1§6.3.3(4)




analysis32 of 44

4 - 113

For z 0,4, the interaction factor, k zy, is calculated as:

k zy =

Rdz, b,

Ed

mLTRdz, b,

Ed

mLT

z

25,0

1,01;

25,0

1,01max

N

N

C N

N

C

0298

0

C mLT = 4,06,0 = 04,06,0 = 0,6


k zy =

2238

127

25,06,0

1,01;

2238

127

25,06,0

931,01,01max

= max ( 0,985; 0,983 ) = 0,985


Rd b,

Edy,

zy

Rdz, b,

Ed M

M k

N

N

=540298985,0

2238127 = 0,601 < 1,0 OK

7.10. In-plane buckling

The in-plane buckling interaction is verified with expression (6.61) in

EN 1993-1-1.

0,1Rd b,

Edy,

yy

Rdy, b,

Ed M

M k

N

N

M

M

M M

Ed

Ed

EdEd

Ed

EdEd

Ed

EdEd

=351 kNm

V

V V

N

N N

=298 kNm =701 kNm

Assumed maximum moment

=356 kNm

=118 kN

=127 kN

=150 kN

=130 kN

=10 kN

=116 kN

Maximum bending moment and axial force in the rafter, excluding the

haunch.

M Ed 356 kNm

N Ed 127 kN

The haunch is analysed in Section 8.




analysis33 of 44

4 - 114

7.10.1. Flexural buckling resistance about the major axis, N b,y,Rd

b

h

190

450 2,37

t f 14,6 mm

buckling about y-y axis:

Curve a for hot rolled I sections

0,21


The buckling length is the system length, which is the distance between the joints (i.e. the length of the rafter, including the haunch), L = 15057 mm

1 =y f

E =

355

210000 = 76,4

EN 1993-1-1§6.3.1.3

y =1y

cr 1

i

L=

4,76

1

185

15057 = 1,065

y = 2yyy 2,015,0

y = 2065,12,0065,121,015,0 = 1,158

EN 1993-1-1§6.3.1.2

y =2

y2

yy

1

=

22065,1158,1158,1

1

= 0,620

N b,y,Rd =M1

yy

Af =

310

0,1

3559880620,0

= 2175 kN

N Ed = 127 kN < 2175 kN OK

7.10.2. Lateral-torsional buckling resistance, M b,Rd

M b,Rd is the least buckling moment resistance of those calculated before.

M b,Rd = 540;581min

M b,Rd = 540 kNm

7.10.3. Interaction of axial force and bending moment – in-planebuckling

In-plane buckling due to the interaction of axial force and bending moment isverified by satisfying the following expression:

0,1Rd b,

Edy,

yy

Rdy, b,

Ed M

M k

N

N

The interaction factor, k yy, is calculated as follows:

k yy =

Rdy, b,

Edmy

Rdy, b,

Edymy 8,01;2,01min

N

N C

N

N C




analysis34 of 44

4 - 115

The expression for C my depends on the values of h and .

=351

298 = 0,849.

h =s

h

M

M =

356

351= 0,986

Therefore C my is calculated as:

C my = h05,095,0 = 986,005,095,0 1,0


k yy =

2175

1278,00,11;

2175

1272,0065,110,1min

= 047,1;05,1min = 1,047


Rd b,

Edy,

yy

Rdy, b,

Ed

M

M k

N

N =

540

356047,1

2175

127 = 0,749 < 1,0 OK

The member satisfies the in-plane buckling check.

7.11. Valid ity of rafter section

In Section 7.8 it has been demonstrated that the cross-sectional resistance of

the section is greater than the applied forces.

The out-of-plane and in-plane buckling checks have been verified inSections 7.9 and 7.10 for the appropriate choice of restraints along the rafter.

Therefore it is concluded that the IPE500 section in S355 steel is appropriate

for use as rafter in this portal frame.

8. Haunched lengthThe haunch is fabricated from a cutting of an IPE 550 section. Checks must

be carried out at end and quarter points, as indicated in the figure below.

312

45

5°

2740

IPE 450

IPE 500

7 2 5

3020

685685685685




analysis35 of 44

4 - 116

From the geometry of the haunch, the following properties can be obtainedfor each of the cross-sections 1 to 5, as shown in Table 2.

Table 2 Section propert ies of haunched member at cross-section , as per

figure aboveCross-sectionno.

Cuttingdepth(mm)

Overalldepth(mm)

Grossarea, A (mm

2)

I y

(cm4)

W el,min

(cm3)

N Ed

(kN)

M Ed

(kNm)

1 503 953 15045 200500 4055 129 661

2 378 828 13870 144031 3348 129 562

3 252 702 12686 98115 2685 128 471

4 126 576 11501 62258 2074 127 383

5 0 450 9880 33740 1500 127 298

The section properties are calculated normal to the axis of the section.

For simplicity, the section properties above have been calculated assuming aconstant web thickness of 9,4 mm and neglecting the middle flange.

The actual and the equivalent cross-sections are shown in the following figure

for cross-section No.1:

190 190

210210

11,1

9,4

9,4 953

503

450

14,6

14,6

17,2

Actual cross-section Equivalent cross-sectionFor cross-section No.1 the values of N Ed and M Ed are taken at the face of thecolumn.


8.1.1. The web

The web can be divided into two webs, and classified according to the stress

and geometry of each web. The upper section (i.e. the rafter) is called theupper web and the lower section (i.e. the cutting) is called the lower web.




analysis36 of 44

4 - 117

Upper web

By inspection the upper web will be Class 3 or better, because it is mostly intension.

Lower web

Stress in the section caused by axial load:

N = 31015045

129 = 8,57 N/mm2

Assuming an elastic stress distribution in cross-section No.1, the maximumstress available to resist bending is:

M = N

M0

y

f = 57,8

0,1

355 = 346 N/mm

2

9 5 3

4 5 0

5 0 3

4 5 1 , 4

5 0 1 , 6

31 N/mm²

346 N/mm²

The distance from the bottom flange to the elastic neutral axis is:

z = 451,4 mm

Distance from underside of middle flange to neutral axis: 51,6 mm

Bending axial stress at the top of cutting section:

= 57,84,4516,51346 = 31 N/mm2




analysis37 of 44

4 - 118

For Class 3 check, determine :

=346

31= 0,09

Considering section 1 parallel tocolumn flange, the depth of web

excluding root radius is:

cw = 242,17503 = 461,8 mm

w

w

t

c=

1,11

8,461= 41,6

190

210

11,1

9,4

503

450

14,6

14,6

17,2

461,8

51,6

E.N.A

Z =451,4 _

EN 1993-1-1Table 5.2

For 1, the limit for Class 3 is: EN 1993-1-1Table 5.2

33,067,0

42

=

09,033,067,0

81,042

= 53,1

wt

c= 41,6 < 53,1

The web is Class 3

8.1.2. The flangesTop flange

f t

c =

6,14

3,69= 4,7


Then :f t

c = 4,7 < 7,3

The top flange is Class 1


Bottom flange

f t

c =

2,17

45,75= 4,4


f t

c= 4,4 < 7,3

The bottom flange is Class 1

Therefore the overall section is Class 3.




analysis38 of 44

4 - 119

8.2. Cross-sectional resistance

IPE 450

IPE 500

312

45

5°298 kNm

383 kNm471 kNm

562 kNm

661 kNm701 kNm

7 2 5

3020


The shear area of cross-section No.1 can be conservatively estimated as:

Av = A (bt f )topfl (bt f ) botfl = 2,172106,1419015045 = 8659 mm2

V pl,Rd =

M0

yv 3

f A=

3100,1

33558659 = 1775 kN

V Ed = 147 kN < 1775 kN OK

EN 1993-1-1§6.2.6

Bending and shear interaction:


the shear force can be ignored if it is smaller than 50% of the plastic shear resistance.

V Ed = 147 kN < 0,5 V pl,Rd = 888 kN


neglected.

The same calculation must be carried out for the remaining cross-sections.

The table below summarizes the shear resistance verification for the haunchedmember:

Table 3 Shear verification for cross-sections 1 to 5

Cross-sectionno.

V Ed

(kN) Av (mm

2)

V pl,Rd

(kN)V Ed V Rd 0,5V Rd

(kN)Bending andshearinteraction

1 147 8659 1775 Yes 888 No

2 140 7484 1534 Yes 767 No

3 132 6300 1291 Yes 646 No4 125 5115 1048 Yes 524 No

5 118 5082 1042 Yes 521 No




analysis39 of 44

4 - 120


The compression resistance of cross-section No.1:

N c,Rd = M0

y

A f

=

3

100,1

35515045

= 5341 kN

N Ed = 129 kN < 5341 kN OK

EN 1993-1-1

§6.2.4

Bending and axial force interaction:


the total stress, x,Ed, must be less than the allowable stress.

x,Ed = N + M

M =y

Ed

I

z M =

4

6

10200500

6,50110661

= 165 N/mm2

x,Ed = N + M = 8,57 + 165 = 174 N/mm2

EN 1993-1-1§6.2.9.2

The maximum allowable stress is:

max =M0

y

f =

0,1

355= 355 N/mm

2

x,Ed = 174 N/mm2

< 355 N/mm2

OK

A similar calculation must be carried out for the remaining cross-sections.

The table below summarize compression resistance verification for the

haunched member:

Table 4 Compression verification for cross-sections 1 to 5

Cross-section(i)

N Ed

(kN) A (mm

2)

N c,Rd

(kN)N Ed N c.Rd Bending and

axialinteraction

1 129 15045 5341 Yes No

2 129 13870 4924 Yes No

3 128 12686 4504 Yes No

4 127 11501 4083 Yes No

5 127 9880 3507 Yes No

8.2.3. Bending moment resistance

The bending moment resistance of cross-section No.1 is:

M c,y,Rd = M el,y,Rd =M0

yminel,

f W =

63

100,1

355104055

= 1440 kNm


EN 1993-1-1§6.2.5(2)

A similar calculation must be carried out for the remaining cross-sections.

The table below summarizes bending moment resistance verification for the

haunched member.




analysis40 of 44

4 - 121

In this case, all cross-sections have been treated as Class 3, and therefore theelastic properties have been used. This is conservative. However, from

previous calculations carried out to check the rafter, it is observed thatcross-section No.1 is Class 1. It may be that other sections between

cross-sections No.1 and No.5 are plastic sections and therefore a greater moment resistance could be achieved.

Table 5 Bending verification for cross-sections 1 to 5

Cross-section(i)

M Ed

(kNm)W el,min

(mm3)

103

M el,Rd

(kNm)M Ed M el,Rd

1 661 4055 1440 Yes

2 562 3348 1189 Yes

3 471 2685 953 Yes

4 383 2074 736 Yes

5 298 1500 533 Yes

8.3. Buckling resistance

There is a torsional restraint at each end of the haunched length.

298 kNm

661 kNm

471 kNm

2740 mm

Buckling length considered

When the tension flange is restrained at discreet points between the torsionalrestraints and the spacing between the restraints to the tension flange is small


In order to determine whether or not the spacing between restraints is smallenough, Annex BB of EN 1993-1-1 provides an expression to calculate the

maximum spacing. If the actual spacing between restraints is smaller than thiscalculated value, then the methods given in Appendix C of this document may

be used to calculate the elastic critical force and the critical moment of the

section.

On the contrary, if the spacing between restraints is larger than the calculatedvalue, an equivalent T-section may be used to check the stability of the

haunch.




analysis41 of 44

4 - 122

8.3.1. Verif ication of spacing between intermediate restraints

Lm =2

y

t

2

y pl,

21

Ed

z

235756

1

4,57

1

38

f

AI

W

C A

N

i

EN 1993-1-1Annex BB§BB.3.2.1

For simplicity, the purlin at mid-span of the haunched member is assumed to

be aligned with the cross-section No. 3.

Equally, the purlin at the end of the haunched member is assumed to be

aligned with the cross-section No. 1.

=661

471= 0,71 1C = 1,2


According to the Eurocode, the ratiot

2

pl

AI

W should be taken as the maximum

value in the segment.

In this case cross-sections No.1 and 3 have been considered, as shown in

Table 6.

Table 6 t

2

pl

AI

W ratio for cross-sections No.1 and 3

Cross-section(i)

A(mm

2)

I t(mm

4)

104

W pl

(mm3)

103

t

2 pl

AI

W

1 15045 81 4888 1961

3 12686 74 3168 1069

EN 1993-1-1

Annex BB§BB.3.2.1

For simplicity, in the calculation of I t and W pl, the middle flange has been neglected.

The section properties of cross-section No.1 give the maximum ratiot

2

pl

AI

W .

Therefore Lm is calculated using the section properties of cross-section No.1.

I z = 2168 104 mm4

iz = A I z =

15045102168

4

= 38 mm

Lm =

2

4

23

2

3

235

355

108115045

104888

2,1756

1

15045

10129

4,57

1

3838

Lm = 700 mm

Purlin spacing is 1345 mm 700 mm

Therefore the design procedure taking advantage of the restraints to the

tension flange given in Section C.2 of Appendix C cannot be used.




analysis42 of 44

4 - 123

8.3.2. Verification of flexural buckling about minor axis

Maximum forces in the haunched member (at the face of the column) are:

N Ed 129 kN

M Ed 661 kNm

EN 1993-1-1 does not cover the design of tapered sections (i.e. a haunch), and

the verification in this worked example is carried out by checking the forces

of an equivalent T-section subject to compression and bending.

The equivalent T-section is taken from a section at mid-length of the

haunched member.

The equivalent T-section is made of the bottom flange and 1/3 of the

compressed part of the web area, based on §6.3.2.4 of EN 1993-1-1.

The buckling length is 2740 mm (length between the top of column and the

first restraint).

Properties of cross-section No.1:

Section area A = 15045 mm2

Elastic modulus to the compression flange W el,y = 4527 103 mm3

Properties of cross-section No.3:

Properties of the whole section

y

y

f

f

/

/

M

M

312 329

104

Elastic neutral axis (from bottom flange): z = 329 mm

Section area A = 12686 mm2

Properties of the equivalent T-section in compression:




analysis43 of 44

4 - 124

9,4

210

104

17,2

Area of T-section:

Af = 4590 mm2

Second moment of area about the

minor axis:

I f,z =1328 104 mm4

Compression in the T-section

The total equivalent compression in the T-section is calculated for

cross-section No.1 by adding the direct axial compression and the

compression due to bending.

N Ed,f = f

yel,

Edf Ed A

W

M

A

A N = 4590

104527

10661

15045

4590129

3

6

= 670 kN

Verification of buckling resistance about the minor axis

Buckling curve c is used for hot rolled sections

z 0,49

1 =y f

E =

355

210000 = 76,4

if,z =f

zf,

A

I

= 4590

101328 4= 53,8

zf, =1zf,

cr 1

i

L=

4,76

1

8,53

2740 = 0,667

z = 2zf,zf,z 2,015,0

z = 2667,02,0667,049,015,0 = 0,837

EN 1993-1-1§6.3.1.2

z =

2zf,2zz

1

=22

667,0837,0837,0

1

= 0,745EN 1993-1-1§6.3.1.2

N b,z,Rd =M0

y

z

Af

= 3100,1

3554590745,0

= 1214 kN

N Ed,f = 670 kN < 1214 kN OK




analysis44 of 44

4 - 125

9. DeflectionsThe horizontal and vertical deflections of the portal frame subject to the

characteristic load combination, as per Expression 6.14 of EN 1990 are as

follows:

20 mm 16 mm

240 mm

Appendix A of this document provides typical deflection limits used in some

European countries. These limits are only intended to be a guideline. The

requirements for a given portal frame design must be agreed with the client.





Part 5: Detailed Design of Trusses









5 - ii




5 - iii

FOREWORD

This publication is part five of the design guide, Single-Storey Steel Buildings.

The 10 parts in the Single-Storey Steel Buildings guide are:



Part 3: Actions









Single-Storey Steel Buildings is one of two design guides. The second design guide is

Multi-Storey Steel Buildings.

The two design guides have been produced in the framework of the European project

“Facilitating the market development for sections in industrial halls and low rise

buildings (SECHALO) RFS2-CT-2008-0030”.

The design guides have been prepared under the direction of Arcelor Mittal, Peiner Träger and Corus. The technical content has been prepared by CTICM and SCI,

collaborating as the Steel Alliance.




5 - iv




5 - v

ContentsPage No

1 INTRODUCTION 1 1.1 Definition 1

1.2 Use of trusses in single-storey buildings 1 1.3 Different shapes of trusses 4 1.4 Aspects of truss design for roof structure 7 1.5 Design of wind girders 9

2 INTRODUCTION TO DETAILED DESIGN 11 2.1 General requirements 11 2.2 Description of the worked example 12

3 GLOBAL ANALYSIS 15 3.1 General 15 3.2 Modelling 15

3.3 Modelling the worked example 16 3.4 Simplified global analysis of the worked example 18 3.5 Secondary forces 19 3.6 Effect of clearance of deflection 21 3.7 Modification of a truss for the passage of equipment 23

4 VERIFICATION OF MEMBERS 28 4.1 Verification of members under compression 28 4.2 Verification of members in tension 41

5 VERIFICATION OF CONNECTIONS 45 5.1 Characteristics of the truss post connection 45 5.2 Chord continuity 47 5.3 Connection of diagonals to chords 48

REFERENCES 51

APPENDIX A Worked Example – Design of a continuous chord connection usingsplice plate connections 53

APPENDIX B Worked example – Design of a truss node with gusset 79




5 - vi

SUMMARY

This publication provides guidance on the design of trusses for single-storey buildings.

The use of the truss form of construction allows buildings of all sizes and shapes to be

constructed. The document explains that both 2D and 3D truss forms can be used. The

2D form of truss is essentially a beam and is used to supporting a building roof,spanning up to 120 metres for large industrial buildings. The 3D form of truss can be

used to cover large areas without intermediate supports; this form is often used for large

exhibition halls. The detailed guidance in this document relates mainly to 2D truss

structures composed of rolled profiles but the principles are generally applicable to all

forms of truss structure.




5 - 1

1 INTRODUCTION

1.1 Definition

A truss is essentially a triangulated system of (usually) straight interconnectedstructural elements; it is sometimes referred to as an open web girder. The

individual elements are connected at nodes; the connections are often assumed

to be nominally pinned. The external forces applied to the system and the

reactions at the supports are generally applied at the nodes. When all the

members and applied forces are in a same plane, the system is a plane or 2D

truss.

F

1 2

1

2

1 Compression axial force2 Tension axial force

Figure 1.1 Members under axial forces in a simple truss

The principal force in each element is axial tension or compression. When the

connections at the nodes are stiff, secondary bending is introduced; this effect

is discussed below.

1.2 Use of trusses in single-storey buildingsIn a typical single-storey industrial building, trusses are very widely used to

serve two main functions:

To carry the roof load:

- Gravity loads (self-weight, roofing and equipment, either on the roof or

hung to the structure, snow loads)

- Actions due to the wind (including uplift due to negative pressure). To provide horizontal stability:

- Wind girders at roof level, or at intermediate levels if required

- Vertical bracing in the side walls and/or in the gables.

Two types of general arrangement of the structure of a typical single-storey

building are shown in Figure 1.2 and in Figure 1.3.

In the first case (Figure 1.2), the lateral stability of the structure is provided by

a series of portal trusses: the connections between the truss and the columns

provide resistance to a global bending moment. Loads are applied to the portalstructure by purlins and side rails.




5 - 2

For the longitudinal stability of the structure, a transverse roof wind girder,

together with bracing in the side walls, is used. In this arrangement the forces

due to longitudinal wind loads are transferred from the gables to the side walls

and then to the foundations.

Lateral stability provided by portal trusses

Longitudinal stability provided by transverse wind girder and vertical cross bracings (blue)

No longitudinal wind girder

Figure 1.2 Portal frame a arrangement

In the second case, as shown in Figure 1.3, each vertical truss and the two

columns on which it spans constitute a simple beam structure: the connection

between the truss and a column does not resist the global bending moment, and

the two column bases are pinned. Transverse restraint is necessary at the top

level of the simple structure; it is achieved by means of a longitudinal wind

girder carries the transverse forces due to wind on the side walls to the braced

gable walls.




5 - 3

Vertical trusses are simply supported by columns

Lateral stability provided by longitudinal wind girder and vertical bracings in the gables (blue)

Longitudinal stability provided by transverse wind girder and vertical bracings (green)

Figure 1.3 Beam and column arrangement

A further arrangement is shown in Figure 1.4.The roof structure is arranged

with main trusses spanning from column to column, and secondary trusses

spanning from main truss to main truss.




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A A

L

On this plan view, main trusses aredrawn in blue: their span L is the longside of the column mesh.

The secondary trusses have a shorterspan A (distance between maintrusses).

This arrangement is currently used for“saw tooth roofs”, as shown on thevertical section:

Main beams are trusses withparallel chords

Secondary beams (green) have atriangular shape.

in red, members supporting the northoriented windows

Figure 1.4 General arrangement 3

1.3 Different shapes of trussesA large range is available for the general shapes of the trusses. Some of the

commonly used shapes are shown in Table 1.1.




5 - 5

Table 1.1 Main types of trusses

Pratt truss: In a Pratt truss, diagonal membersare in tension for gravity loads. Thistype of truss is used where gravityloads are predominant

In a truss as shown, diagonalmembers are in tension for upliftloads. This type of truss is usedwhere uplift loads are predominant,such as open buildings.

L o n g s p a n s : r a n g e f r o m 2 0

t o 1 0 0 m

Warren t russ: In this type of truss, diagonalmembers are alternatively intension and in compression

This type of truss is also used forthe horizontal truss of gantry/cranegirders (see Figure 1.5)

There are two different types of X truss :

if the diagonal members are designedto resist compression, the X truss isthe superposition of two Warrentrusses.

if the resistance of the diagonalmembers in compression is ignored,the behaviour is the same as a Pratttruss.

This shape of truss is more commonlyused for wind girders, where the diagonalmembers are very long.

It is possible to add secondary members inorder to : create intermediate loading points

limit the buckling length of members incompression (without influencing theglobal structural behaviour).

A l l t h e s e

t y p e s o f t r u s s e s c a n b e u s e d e i t h e r i n p o r t a l t r u s s s t r u c t u r e s ( s e e f i g u r e 1 . 2 )

o r i n s i m p l e t r u s s s t r u c t u r e s ( s e e f i g u r e 1 . 3

) .

For any of the forms shown above, it ispossible to provide either a single or adouble slope to the upper chord of a roof supporting truss

This example shows a duo-pitch truss

Single slope upper chord for thesetriangular trusses, part of a “saw toothroof”North oriented windows

S i m p l y s u p p o r t e d ,

s m a l l e r s p a n s

R a n g e f r o m 1

0 t o 1 5 m

Fink truss: This type of truss is more commonly usedfor the roof of houses.




5 - 6

The horizontal truss is positioned at thelevel of the upper flange of the gantrygirder in order to resist the horizontalforces applied by the wheels on the rail(braking of the crane trolley, crabbing)

1

32

1 Crane girder2 Crane rail3 Horizontal bracing (V truss)

Figure 1.5 Horizontal bracing for a crane girder

Figure 1.6 and Figure 1.7 illustrate some of the trusses described in Table 1.1.

Figure 1.6 N-truss – 100 m span




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Figure 1.7 N-truss (also with N-truss purlins)

1.4 Aspects of truss design for roof structure1.4.1 Truss or I-beam

For the same steel weight, it is possible to get better performance in terms of

resistance and stiffness with a truss than an I-beam. This difference is more

sensitive for long spans and/or heavy loads.

The full use of this advantage is achievable if the height of the truss is not

limited by criteria other than the structural efficiency (a limit on total height of

the building, for example).

However, fabrication of a truss is generally more time consuming than for an

I-beam, even considering that the modernisation of fabrication equipment

allows the optimisation of fabrication times.

The balance between minimum weight and minimum cost depends on many

conditions: the equipment of the workshop, the local cost of manufacturing; the

steel unit cost, etc. Trusses generally give an economic solution for spans over 20 or 25 m.

An advantage of the truss design for roofs is that ducts and pipes that are

required for operation of the buildings services can be installed through the

truss web.

1.4.2 General geometry

In order to get a good structural performance, the ratio of span to truss depth

should be chosen in the range 10 to 15.

The architectural design of the building determines its external geometry and

governs the slope(s) given to the top chord of the truss.




5 - 8

The intended use of the internal space can lead either to the choice of a

horizontal bottom chord (e.g. where conveyors must be hung under the chord),

or to an inclined internal chord, to allow maximum space to be freed up (see

the final example in Table 1.1).

To get an efficient layout of the truss members between the chords, thefollowing is advisable:

The inclination of the diagonal members in relation to the chords should be

between 35° and 55°

Point loads should only be applied at nodes

The orientation of the diagonal members should be such that the longest

members are subject to tension (the shorter ones being subject to

compression).

1.4.3 Section of the members

Many solutions are available. The main criteria are:

Sections should be symmetrical for bending out of the vertical plane of the

truss

For members in compression, the buckling resistance in the vertical plane

of the truss should be similar to that out of the plane.

A very popular solution, especially for industrial buildings, is to use sections

composed of two angles bolted on vertical gusset plates and intermediately

battened, for both chords and internal members. It is a very simple and efficient

solution.

For large member forces, it is a good solution to use:

Chords having IPE, HEA or HEB sections, or a section made up of two

channels (UPE)

Diagonals formed from two battened angles.

The web of the IPE / HEA / HEB chord section is oriented either vertically or

horizontally. As it is easier to increase the resistance to in-plane buckling of the

chords (by adding secondary diagonal members) than to increase their to out-

of-plane resistance, it is more efficient to have the web horizontal, for chords incompression. On the other hand, it is easier to connect purlins to the top chord

if it has a vertical web.

It could be a good solution to have the top chord with a vertical web, and the

bottom chord with a horizontal web.

Another range of solutions is given by the use of hollow sections, for chords

and/or for internals.

1.4.4 Types of connections

For all the types of member sections, it is possible to design either boltedconnections or welded connections. Generally, bolted connections are preferred

on site. Where bolted connections are used with bolts loaded perpendicular to

their shank, it is necessary to evaluate the consequences of slack in




5 - 9

connections. In order to reduce these consequences (typically, the increase of

the deflections), solutions are available such as use of pre-stressed bolts, or

limiting the hole size.

1.4.5 Lateral stabili ty

It is necessary to design the chords in compression against the out-of-plane buckling. For simply supported trusses, the upper chord is in compression for

gravity loading, and the bottom chord is in compression for uplift loading. For

portal trusses, each chord is partly in compression and partly in tension.

Lateral restraint of the upper chord is generally given by the purlins and the

transverse roof wind girder.

For the restraint of the bottom chord, additional bracing may be necessary, as

shown in Figure 1.8. Such bracing allows the buckling length of the bottom

chord to be limited out of the plane of the truss to the distance between points

laterally restrained: they serve to transfer the restraint forces to the level of thetop chord, the level at which the general roof bracing is provided. This type of

bracing is also used when a horizontal load is applied to the bottom chord (for

example, forces due to braking from a suspended conveyor).

A A

A A A

A

Truss

AA

Cross bracing between trusses

Thick black dots: twoconsecutive trusses

Blue The purlin whichcompletes the bracing inthe upper region

Green The longitudinalelement which closes thebracing in the lowerregion

Red Vertical roof bracing

Figure 1.8 Lateral bracing

The roof purlins often serve as part of the bracing at the top chord. Introduction

of longitudinal members at the lower chord allows the trusses to be stabilised

by the same vertical bracing.

It is possible to create a horizontal wind girder at the level of the bottom

chords, with longitudinal elements to stabilize all the trusses.

1.5 Design of wind girders

1.5.1 Transverse wind girder

In general, the form of a transverse wind girder is as follows (see Figure 1.2):

The wind girder is arranged as an X truss, parallel to the roof plane.




5 - 10

The chords of the wind girder are the upper chords of two adjacent vertical

trusses. This means that the axial forces in these members due to loading on

the vertical truss and those due to loads on the wind girder loading must be

added together (for an appropriate combination of actions).

The posts of the wind girder are generally the roof purlins. This means that

the purlins are subject to a compression, in addition to the bending due to

the roof loading.

It is also possible, for large spans of the wind girder, to have separate posts

(generally tubular section) that do not act as purlins.

The diagonal members are connected in the plane of the posts. If the posts

are the purlins, the diagonal members are connected at the bottom level of

the purlins. In a large X truss, diagonals are only considered in tension and

it is possible to use single angles or cables.

It is convenient to arrange a transverse wind girder at each end of the building, but it is then important to be careful about the effects of thermal expansion

which can cause significant forces if longitudinal elements are attached

between the two bracing systems, especially for buildings which are longer

than about 60 m.

In order to release the expansion of the longitudinal elements, the transverse

wind girder can be placed in the centre of the building, but then it is necessary

to ensure that wind loads are transmitted from the gables to the central

wind-bracing.

Transverse wind girders are sometimes placed in the second and penultimate

spans of the roof because, if the roof purlins are used as the wind girder posts,

these spans are less subject to bending by roof loads.

The purlins which serve as wind girder posts and are subject to compression

must sometimes be reinforced:

To reinforce IPE purlins: use welded angles or channels (UPE)

To reinforce cold formed purlins: increase of the thickness in the relevant

span, or, if that is not sufficient, double the purlin sections (with fitting for

the Zed, back to back for the Sigma).

1.5.2 Longitudinal wind girder

It is necessary to provide a longitudinal wind girder (between braced gable

ends) in buildings where the roof trusses are not “portalized”.

The general arrangement is similar to that described for a transverse wind

girder:

X truss

The chords are two lines of purlins in small buildings, or additional

elements (usually tubular sections)

The posts are the upper chords of the consecutive stabilized roof trusses.




5 - 11

2 INTRODUCTION TO DETAILED DESIGN

The detailed design of trusses is illustrated in the following Sections by

reference to a ‘worked example’. This Section summarizes the generalrequirements and introduces the example. The topics covered in subsequent

Sections are:

Section 3: Global analysis

Section 4: Verification of members

Section 5: Verification of connections

Fully detailed calculations for verification of a gusset plate connection and a

chord splice are given in Appendices A and B.

2.1 General requirementsThe parameters to be taken into account in design are:

Aesthetics

Geometry (span length, height, rise, etc)

Actions.

The following requirements have to be considered:

Regulatory requirements

Contractual requirements with regard to standards

Specific contractual requirements.

The resulting outcome of a design is the set of execution documents for the

structure.

The nature of regulatory requirements varies from one country to another.

Their purpose is usually to protect people. They exist in particular in the area

of seismic behaviour, and for the behaviour of buildings during a fire (see

Single-Storey Steel Buildings. Fire engineering Guide1 ).

The requirements in standards concern the determination of actions to be

considered, the methods of analysis to be used, and the criteria for verification

with respect to resistance and stiffness.

There is no limit to the number of specific requirements which may be imposed

for any particular building but these mainly concern construction geometry;

they influence determination of actions, in particular climatic actions.

Obligations and interface arrangements for detailed design might include:

Banning the use of tubes for the bottom chord of trusses to which theindustry client wishes to hang equipment

Obligation to use tubes for truss chords for reasons of appearance




5 - 12

Use of the roof to stabilise certain structural elements.

The flowchart below illustrates the main steps in the design of a structural

element.

DATA

CHOICE OF GLOBAL

ANALYSIS

MEMBER RESISTANCE

VERIFICATION

CONNECTIONSRESISTANCE

VERIFICATION EC3-1-8

EC3-1-1

Contractual data Geometrical data

Incidence of neighbouring construction

Obligations or restrictionsin choice of sections

Nature and position of perm anent l oads

Nature and position of imposed loads

Stabilising role of envelope

Regulatory data and Standards Climatic loads

Seismic loads Exploitation loads

…

SLSVERIFICATION

CRITERIA

CHAPTER 3

CHAPTER 4

CHAPTER 5

EC1

EC8

Figure 2.1 Flowchart for the design of a structural element

2.2 Descript ion of the worked example

The worked example that is the subject of subsequent Sections is a large spantruss supporting the roof of an industrial building, by means of purlins in the

form of trusses. This example is directly transposed from a real construction

and has been simplified in order to clarify the overview.




5 - 13

1

2

1 Main truss2 Purlin truss

Note: the horizontal bracing is not displayed in this diagram but it is designed in such a way thatthe purlins provide efficient lateral restraints to the main trusses.

Figure 2.2 Worked example - General layout of the roof

The roof is a symmetrical pitched roof; the slope on each side is 3%.

Each main truss has a span of 45,60 m and is simply supported at the tops of

the columns (there is no moment transmission between the truss and the

column).

General transverse stability of the building is provided by fixity of the columns

at ground level; longitudinal stability is provided by a system of roof bracings

and braced bays in the walls.

1

2 5

6

4

3

7

1

2

4

1 Upper chord IPE 330 with horizontal web2 Lower chord IPE 330 with horizontal web3 Post - Single angle L100x100x104 Top of the column (IPE 450)

5 Diagonals - Double angle6 Secondary truss members7 Sketch of the cross-section

Figure 2.3 Worked example – View of truss

The truss is illustrated in Figure 2.3. The truss chords are parallel and are made

up of IPE 330 profiles with the webs horizontal. The diagonals are made of

twinned angles: two 120 120 12 angles for diagonals in tension under gravity loads (in blue in the diagram above), two 150 150 15 angles for




5 - 14

diagonals in compression under gravity loads (in red in the diagram above); the

posts are single angles 100 100 10.

Note that, in the central panels, secondary diagonals and posts are present.

They would generally be installed with one or other of the following

objectives:

To permit application of a point load between main nodes, without causing

further bending in the upper chord

To reduce buckling, in the plane of the truss of central members of the

upper chord.

In this example, the secondary trusses reduce the buckling length.

The pairs of angles which make up the section of a diagonal are joined by

battens, to ensure combined action with respect to buckling between the truss

nodes. To be efficient, battens must therefore prevent local slip of one angle inrelation to the other. See Section 4.1.3 for more information.

Each chord is fabricated in two pieces (see Figure 3.6). The diagonals and

posts are bolted at their two ends to vertical gusset plates, which are themselves

welded to the horizontal webs of the IPE 330 chords. Detailed diagrams of this

type of connection are given in Appendix A and in Sections 5.2 and 5.3.

The columns on which the truss is supported are IPE 450, for which the web is

perpendicular to the plane of the truss beam.

In order to illustrate all of the topics here, the truss beam in the workedexample is designed for two situations: a gravity load case and an uplift load

case. The loads correspond to the combination of actions, determined

according to EN 1990 for verification with respect to the ultimate limit state

(ULS).

91 kN136 kN

182 kN182 kN 182 kN

136 kN91 kN

ULS combination n°1: Gravity loading(without self-weight)

43,50 kN 65,25 kN 87 kN87 kN 87 kN 65,25 kN 43,50 kN

ULS combination n°2: Uplift loading

Figure 2.4 Worked example – Load Combinations




5 - 15

3 GLOBAL ANALYSIS

3.1 General

Section 1.1 describes the general behaviour of a truss. In reality, structuresdeviate from this theoretical behaviour and their global analysis involves

consideration of the deviations. In particular, the deviations include the

occurrence of bending in the members, in addition to the axial forces. These

bending moments, known as “secondary moments”, can cause significant

additional stresses in the members which make up the truss.

The deviations in design take various forms:

All the members which make up the structure are not usually articulated at

their original node and their end node. Truss chords, in particular, are

usually fabricated in one length only, over several truss purlins: thecontinuous chord members are then connected rigidly to their original and

end nodes. Rotation of the nodes, resulting from general deformation of the

truss beam then causes bending moments in the rigidly connected members;

the more rigid the chord members, the bigger the moments (see

Section 3.4).

The members are not always strictly aligned on their original and end

nodes. Bending moments which result from a misalignment of axes

increase in proportion to the size of the eccentricity and the stiffness of the

members. This phenomenon is illustrated in Section 3.6.

Loads are not always strictly applied to the nodes and, if care is not taken tointroduce secondary members to triangulate the point of application of the

loads between nodes, this results in bending moments.

3.2 Modelling Several questions arise in respect of the modelling of a truss.

It is always convenient to work on restricted models. For example, for a

standard building, it is common and usually justified to work with 2D models

(portal, wind girder, vertical bracing) rather than a unique and global 3D

model. A truss can even be modelled without its supporting columns when it is

articulated to the columns.

Nonetheless, it is important to note that:

If separate models are used, it may be necessary, in order to verify the

resistance of certain elements, to combine the results of several analyses;

example: the upper chord of a truss also serves as chord of the wind girder.

If a global 3D model is used, “parasitic” bending can be observed, which

often only creates an illusory precision of the structural behaviour process.

That is why 2D models are generally preferable.

In the worked example, where the truss is simply supported on the columns,

the design model chosen is that of the truss only.




5 - 16

Once the scope of the model has been decided and adapted according to use to

be made of the results, it is important to consider the nature of the internal

connections. In current modelling of member structures, the selection is made

between “a pin-jointed member at a node” and a “member rigidly connected to

a node”; the possibility offered by EN 1993 to model connections as semi-rigid

is rarely used for truss structures.

For trusses, the model is commonly represented as either:

Continuous chords (and therefore chord members rigidly connected at

both ends)

Truss members (diagonals and verticals) pin jointed to the chords.

3.3 Modelling the worked exampleIn the worked example, the truss diagonals are pin jointed to the chords,

although the connections are carried out using high strength bolts suitable for

preloading with controlled tightening. This provides a rigid connection without

slack between the diagonal and the connection gusset plates. The connection

can be considered as pinned due to the fact that the vertical gusset plates are

welded in the middle of the horizontal, not very stiff, IPE 330 web.

The modelling is shown in Figure 3.1, with the numbering of the members.

Left part

Right part

Figure 3.1 Computer model

It is important for the model to be representative of the eccentricities which

exist in the real structure. They can have a significant effect, as illustrated in

Section 3.6.1.

It is also important that modelling of the loads is representative of the real

situation. In particular, the fact of applying to the truss nodes loads which, in

reality, are applied between nodes, risks leading to neglect of the bending with

quite significant outcomes.

The main results of the analysis are given in Figure 3.2 for the left part of the

truss.




5 - 17

ULS Load combination n°1 (Gravity loading) – Axial force (N) in kN

ULS Load combination n°1 (Gravity loading) – Bending moment (M) in kNm

ULS Load combination n°2 (Uplift loading) – Axial force (N) in kN

ULS Load combination n°2 (Uplift loading) – Bending moment (M) in kNm

Figure 3.2 Worked example – Axial forces and bending moments

It is interesting to note the form of the moment diagrams in the member:

In the chords and the diagonals, the self weight generates a bending

moment with a parabolic shape

In the chords, continuous modelling (members rigidly connected at bothends) leads to moments at the nodes.




5 - 18

3.4 Simplified global analysis of the worked exampleA triangulated beam, with a constant depth, can be equated to an I-beam. This

equivalence is possible and provides a good approximation, for example, for a

truss with parallel chords.

The global shear force V global and the global bending moment M global in the

equivalent beam vary very little along a panel and can be equated with the

mean values in the panel. Therefore the axial load can be assessed using the

following expressions (see Figure 3.3 for the notations):

N ch = ± M global/h in the chords

N d = ±V global/cos θ in a diagonal

hθ

Figure 3.3 Truss with parallel chords - Notation

An estimate can also be made for the deflection of the truss beam by

calculating that for an equivalent beam, for the same loading. In order to do

this, the classic approach is to use elementary beam theory, giving theequivalent beam a second moment of area equal to:

22

1

ch, i

i

id A I

where:

Ach,i is the section area of the chord i

d i is the distance from the centroid of both chords to the centroid of the

chord i.

In order to take into account global shear deformations, not dealt with in

elementary formulae, a reduced modulus of elasticity is used. Global shear

deformations are not, in fact, negligible in the case of trusses, since they result

from a variation in length of the diagonals and posts. The value of the reduced

modulus of elasticity clearly varies depending on the geometry of the truss, the

section of the members, etc. For a truss beam with “well proportioned” parallel

chords, the reduced modulus of elasticity is about 160000 N/mm2 (instead of

210000 N/mm2).




5 - 19

4 0 0 0

101 kN 158 202 202 202 158 101

7100 7200 8500 8600 7100 7100

Truss (combination n°1), including self-weight

461 (616)

303 (405)

101 135

-101 (-135)

-303 (-405)

-461 (-616)

562

-562

Diagram of the global shear force V (kN) In parentheses: values of N d =V /cos

3273(818)

5455

(1364) 6320

(1580)

5455

(1364)

3273

(818)

Diagram of the global bending moment M (kNm)In parentheses: values of N ch =M /h

Figure 3.4 Worked example – Approximate calculation

The values of the axial forces in the chords obtained by the simplified

approach, M global/h, are shown in Figure 3.4. The values are very close to the

values obtained using structural analysis software (see Figure 3.2), for the

sections close to the applied loads. The small difference comes from the slope

(3%) of the chords of the truss in the worked example, not taken into account

in the hand calculation.

The values of the axial forces in the diagonals obtained by the simplified

approach, V global/cos θ , are also very close to the values obtained using

software.

3.5 Secondary forces

3.5.1 Influence of chord rig idi ty

Chord members in trusses which are used in construction are rarely pinned at

the nodes and are more often rigidly connected; this means that members

connected to the same node have to keep their respective angles. During

deformation of the structure under load, the ends of the members all rotate at

the same angle around the node. In these conditions, bending loads (bending

moments and shear forces) called secondary forces are added to the axial loads

in the members calculated assuming the nodes are pinned (primary forces).




5 - 20

It is routine in design to use continuous chord members and to pin the truss

members.

In fact, transforming pinned connections into rigid nodes hardly leads to any

modification to the axial forces in the members, because the shear transmitted

by the members has little influence on the equilibrium equation of nodal forcesand, on the other hand, bending of the member due to secondary bending

moments only causes a slight variation in the distance between the ends of this

member compared to the difference in length due to axial force.

Nevertheless, it is essential that the triangulated structures be designed properly

so that the members are adequately arranged to withstand bending stresses, but

not too slender so as to avoid buckling. Note that the greater the stiffness of the

chords (which are usually continuous), compared to the global stiffness of the

truss beam, the bigger the moments developed in the chords. For instance, for a

wind girder in a roof, the stiffness of the chords is relatively small and the

secondary moments remain small as well.

For a stocky truss, i.e. when the flexural stiffness of the individual chords is not

significantly lower than the global stiffness of the truss, it can be necessary to

take into account the secondary moments. Then the members and the

connections must be designed accordingly.

This phenomenon can be illustrated in the worked example by arranging the

IPE 330 sections as ‘standing up’ chord members, instead of being flat in the

initial design (Figure 3.5). The chords therefore bend in the vertical plane of

the truss member, mobilising their strong inertia. The calculation results

demonstrate well a significant increase in the secondary moments.

Figure 3.5 Options for the orientation of the chords

In the upper chord in a standing up IPE 300 section near the half-span, the bending moment under gravity loads (ULS) is 28,5 kNm, compared to

2,7 kNm for the flat IPE 330 section.

Similarly, in the lower chord, the bending moment is 23,4 kNm, compared to

1,7 kNm.

The multiplier of the bending moments is 11 for the upper chord, and 14 for the

lower chord. This is comparable with the ratio of the inertia in an IPE 330

section (about 15).

3.5.2 Assumption of rigid connections

In another evaluation of the effect of member stiffness on the value of the

secondary moments, the truss in the example was recalculated by making all




5 - 21

the internal connections rigid (diagonal and verticals fixed on their original end

nodes). The comparison is summarized in Table 3.1, where it can be seen that

the end moments are in the same range as the moments resulting from the self-

weight of the diagonals.

Table 3.1 Effect of rigid connection instead of pinned Horizontal web Vertical web

End moment in a diagonal in tension(Double angles 120 x12)

1,03 1,17

End moment in a diagonal in compression

(Double angles 150 15)

1,30 2,35

Moment resulting from the self-weight (for comparison) 1,36 1,36

Assumption of bi-hinged diagonals Acceptable Acceptable

Note: the bending moments are given in kNm.

3.6 Effect of clearance of deflectionWhen the connections between elements which make up a truss beam are

bolted connections, with bolts in shear (category A in EN 1993-1-8[2]), the

clearance introduced into these connections can have a significant effect on

displacement of the nodes.

In order to facilitate erection, the bolts are in fact inserted in holes which are

larger than the bolts themselves. For standard bolt sizes, holes more than 2 mm

bigger than the bolt are usually made (usually referred to as a 2mm clearance).

In order for a connection with clearance to transmit to the node the loadrequired by the attached member, the bolt must come into contact with one or

other of the connected parts: this is called often referred to as ‘taking up slack’.

For a connected tension member, this slack can be assimilated as an additional

extension that is added to the elastic elongation of the member in tension.

Likewise, for a connected compression member, the slack is assimilated as a

reduction in length that is added to the elastic shortening of the compressed

member.

The total slack in the many different connections of a truss structure can lead to

a significant increase in displacements, which can have various and more or

less serious consequences. Amongst these, note:

In most of the cases, the visual effect is the worst consequence.

Increased deflection can lead to a reduction of free height under the bottom

chord, which might prevent or upset the anticipated usage. For example, the

additional deflection of a truss holding doors suspended in a gable of an

aeroplane hangar could prevent the passage of the aeroplane.

Increase in the deflection can result in reduction in the slope of the

supported roof and even, if the nominal slope were small, to a slope

inversion; a risk of water accumulation is therefore associated with an

inversion in pitch. If the truss structure is not a statically determinate system, this may lead to

unexpected internal forces.




5 - 22

It is therefore essential, where truss structures are concerned, to control the

effect of connection slack on the displacements. In order to do this, it is often

necessary:

either to limit slack in category A connections: drilling at +1 mm, even

+0,5 mm and using shear bolts on a smooth bolt shank (to limit the increase

in slack by deformation) of the threads and pieces; or

to use ‘fit bolts’; or

to use preloaded bolts (category C connections); or

to use welded connections instead of bolted connections.

In cases where loading in the members does not result in reversal of axial

force, it is possible to calculate a value for the effect of slack in all the

connections. The following calculation illustrates this phenomenon for the

worked example.

Each of the chords, upper and lower, has a continuous connection with bolted

splice plates around the mid-span. In addition, the diagonals are connected by

bolting on gusset plates welded to the chords. Holes are 2 mm larger than the

bolt diameter.

Figure 3.6 Worked example – Position of the chord connections using splice plates

In a spliced connection of a chord, the effect of slack on the deflection can be

evaluated by assuming that the bolts are initially centred on their holes. If the

diameter of the holes is d + 2 mm (where d is the bolt diameter), a chord in

tension is extended by 4 mm, as shown in Figure 3.7.

1 mm 1 mmd 1 mm 1 mmd

g

g +4 mm

Figure 3.7 The effect of slack under load

In order for a diagonal to be loaded, 2 mm has to be recovered at each end: the

length of a diagonal in tension is increased by 4 mm; a diagonal under

compression is reduced by a further 4 mm.




5 - 23

The deflection of a truss due to the slack can be evaluated by considering the

effect of a unit load applied at mid span, using the Bertrand Fontviolant

equation.

-0,5 0,66 -0,68 0,66 -0,68 0,71 -0,75 0,17 -0,75 0,72 -0,68 0,66 -0,68 0,66 -0,5

2,85

Figure 3.8 Worked example – Axial forces ( N 1,i ) under unit load

The deflection is given by:

bi

i i

iii

ES

l F N v

1

1,

Where:

N 1,i is the axial force produced in the member i by a unit force applied at

the point where the deflection is required

il is the length of member i

iS is the section area of the member i

b is the number of members with bolted connection(s).

i

ii

ES l F is the variation in length of member i due to the slack recovery

= ±4 mm according to whether the chord is in compression or tension.

Then:

v = 4 × (2,31 + 2,85 + 0,5 + 0,66 + 0,68 + 0,66 + 0,68 + 0,71 + 0,75 +…

+ 0,17 + 0,75 + 0,72 + 0,68 + 0,66 + 0,68 + 0,66 + 0,5)

v = 58,4 mm

This is a significant additional deflection, compared with the deflection due to

the ULS combination (127 mm).

3.7 Modification of a truss for the passage of equipment It frequently occurs that it is necessary to modify the form of a truss in order to

allow equipment to pass (a large section duct for example).

Several solutions can be provided (Figure 3.9):

Either to increase the passage area available by an eccentricity in the

connection of one of the chords (case 1)

Or “break” the straight form of a diagonal, by triangulating the breakage

point (case 2).




5 - 24

Case 1 Case 2

Figure 3.9 Passage of a duct – Local modification of the truss

In case 1, the secondary moments which result from the introduction of an

eccentricity increase in relation to the size of the eccentricity. If there is a

choice, it is always preferable to introduce an eccentricity in the least stressed

chords.

In case 2, care must be taken with several phenomena:

The axial force can increase significantly in certain chords situated in the

immediate proximity of the modified panel (as a result of modification tothe position of the members).

“Secondary” moments appear as a result of the lack of stiffness in a broken

diagonal compared with a straight diagonal, even if the breakage point is

triangulated.

The breakage point must of course be triangulated in the plane of the truss;

it must also be restrained out-of-plane (where three members meet) if the

broken diagonal is in compression.

These two phenomena (case 1 and case 2) are illustrated using the worked

example.

3.7.1 Introduction of an eccentric ity axis in a diagonal (case 1)

The truss panel through which the passage of equipment is required is the

second panel from the support on the right. Figure 3.10 shows a part of the

truss, with the eccentricity of a diagonal.

300 mm

Figure 3.10 Passage of a duct – Eccentr ici ty of a diagonal

Changes in axial forces in the modified area are represented on the Figure 3.11.




5 - 25

Axial force (kN) Bending moment (kNm)Initial structure

Modified structure

Figure 3.11 Effects of the eccentricity of diagonal under ULS gravity loading

The 300 mm eccentricity makes the triangulation imperfect.

The main consequence of this arrangement is a significant increase in the

bending moments in the lower chord that receives the eccentric diagonal. A

74,15 kNm moment is calculated in the second chord member from the right

hand support, a 62,72 kNm moment in the first chord member, much higher

than in the initial structure without eccentricity.

The elastic moment resistance of an IPE 330 horizontal section is:

69,2 0,355 = 24,57 kNm

The bending capacity is therefore greatly exceeded, apart from any other

interactions. Reinforcement of the lower chord member will therefore be

required in order to support the axis eccentricity introduced.

3.7.2 “ Broken” diagonal (example 2)

The panel of the penetration equipment is the same as in 3.6.1. Figure 3.12 is a

diagram of the diagonal “breakage”.




5 - 26

Figure 3.12 Passage of a duct – Broken diagonal

Development of stress in the modified area is represented on the section

diagrams in Figure 3.13.

Axial force (kN) Bending moment (kNm)

Initial structure

Axial force (kN) Bending moment (kNm)

Modified structure

Figure 3.13 Effects of a broken diagonal under ULS gravity loading

The effects of modification on the calculated stresses are mainly:

A noticeable increase is observed in the axial force in the second lower

chord member from the right hand support (in the panel with the broken

diagonal): the tension calculated increases from 818 to 1350 kN.

A significant increase is also observed in the compression force in the

broken diagonal compared with the rectilinear diagonal of the initial

structure: compression increases from 624 to 1090 kN.

As far as the additional triangulation member is concerned, this supports a

normal compression force of 755 kN.

In the lower chord, as well as an increase in the normal tension force, an

increase in “secondary” moments is also observed on the three right panels




5 - 27

The modification to the structure (broken diagonal) therefore has a significant

effect on the size of the members.




5 - 28

4 VERIFICATION OF MEMBERS

As seen in the preceding section, which dealt with the global analysis, the

members are mainly subjected to axial forces.

It was also observed that, in many cases, members are also subject to stress by

bending moments, i.e. secondary moments.

4.1 Verification of members under compressionThe resistance of a member to compression is evaluated by taking into account

the different modes of instability:

Local buckling of the section is controlled using section classification, and

when necessary, effective section properties (class 4)

Buckling of the member is controlled by applying a reduction coefficient in

the calculation of resistance.

For a compression member, several buckling modes must be considered. In

most truss members, only flexural buckling of the compressed members in the

plane of the truss structure and out of the plane of the truss structure need be

evaluated.

For each buckling mode, the buckling resistance is obtained from

EN 1993-1-1[3] by applying a reduction to the resistance of the cross-section.

This reduction factor is obtained from the slenderness of the member, whichdepends on the elastic critical force.

For the diagonals and the verticals stressed in uniform compression. the elastic

critical force is determined from the buckling length of the member in

accordance with EN 1993-1-1, 6.3.1.3. The following can be observed,

according to Annex BB §BB.1 of EN 1993-1-1:

For buckling in the plane of the truss beam: the buckling length is taken

equal to 90% of the system length (distance between nodes), when the truss

member is connected at each end with at least two bolts, or by welding

(EN 1993-1-1 §BB.1.1(4)).(An exception is made by Annex BB for angle truss members, for which a

different evaluation is given; it is not specified in this annex if the particular

rule also concerns members made up to two pairs of angles: by way of

simplification, it is recommended that a buckling length of 0,9 times the

length of the axis be retained.)

For buckling out of plane of the truss beam, the buckling length is taken

equal to the system length.

For buckling in the plane of the truss of the chord members in uniform

compression, the buckling length may be taken as 90% of its system length(distance between nodes).




5 - 29

For buckling out of plane of the truss, it can be more difficult to determine the

elastic critical force for the following reasons:

There is not necessarily a lateral support at each node of the truss

The lateral support points are not necessarily effectively rigid.

When there is no lateral support at each node along the chord, the segment

located between support points is subject to variable compression between

bays. In these circumstances:

A conservative approach would be to use the normal compression force at

its maximum value and to take the buckling length as the distance between

supports but this can lead to an under-estimate of the chord resistance.

Refined methods can be adopted by investigating an equivalent buckling

length under constant compression.

In the worked example, where the truss supports a roof, with purlins at thelevel of the upper chord of the truss:

All the purlins connected to a roof bracing can be considered as lateral rigid

support points.

Intermediate purlins can also be considered as a rigid point of support.

Insofar as a diaphragm role has been attributed to the roof (class 2

construction according to EN 1993-1-3).

With regard to the lower chord, these lateral support points are provided by

additional vertical bracing elements between trusses (see the braces under

the truss purlins in Figure 2.2).

Another point to note, which is very common, concerning determination of the

compression resistance, is the case of pairs of members. It is quite common, as

was stated, to make up members from a truss structure using two angles, or two

channels (UPE).

In order to ensure that such built-up members will behave as sole members in

the flexural buckling mode, the two components are connected by small battens

(Figure 4.1). Since the role of these members is to prevent relative slip of one

component compared with the other, they must be connected without slack.

The gap between the angles, and the thickness of the battens, should be thesame as the thickness of the gusset to which the built-up member is connected.




5 - 30

11

2

A

A

A-A

1 Batten2 Gusset

Figure 4.1 Members composed of two angles

The maximum spacing of the connections between members is limited by

EN 1993-1-1 to 15 times the minimum radius of gyration of the isolated

component. Otherwise a more complex verification needs to be carried out, bytaking into account the shear stiffness of the composed member. This limitation

is very restrictive. By way of example, in order to link two 50 × 50 × 5 angles

by respecting the spacing limit, it would be necessary to provide a batten every

15 cm.

In order to illustrate the different principles stated above, justifying

calculations are developed in the following sections for the different types of

compressed members in the worked example truss structure. The results are

taken from the basic worked example:

IPE 330 chords with horizontal web

Web members are assumed to be hinged at both ends

Chords are assumed to be continuous.

4.1.1 Upper chord in compression

The verifications set out below, concern the upper chord member adjacent to

mid span (element B107 in Figure 3.1), in which the normal compression force

calculated under gravity ULS loads is greatest and equal to:

N Ed = −1477 kN

The checks take into account the coincident bending moments.

Note that the verification should also be carried out on the first member from

the mid span, which is not restrained by the secondary truss: axial force of

lesser compression, but with increased buckling length in the plane of the truss.

Since the calculation is identical, it is not set out formally below. If this

verification indicated a lack of resistance, the reinforcement solution would of

course consist of extending the installation of the secondary truss.

The shear force and the bending moments are given in Figure 4.2.




5 - 31

2,86 kNm

-1,05 kNm

2,151

Bending moment M Ed

-1,82 kN

Shear force V Ed

Figure 4.2 Bending moment and shear force in the upper chord

Cross-section properties

For an IPE 330 with horizontal web (steel grade S355)

A = 62,6 cm2

I y = 11770 cm4

I z = 788 cm4

W el,z = 98,5 cm3

Class of the cross-section

The material parameter is:

= 0,81

As simplification, the cross-section can be classified in uniform compression,

even if it is subjected to combined axial force and bending moment.

The compressed flanges are classified as outstand flanges (EN 1993-1-1 Table

5.2, Sheet 2):

29,791,55,11

25,58

t

c

The flange is Class 1.

The web is classified as an internal compressed part (EN 1993-1-1 Table 5.2,

Sheet 1):

02,34421,365,7

271

t

c

The web is Class 4.

Effective properties of the cross-section

The effective area Aeff is calculated for pure compression.

The flanges are Class 1, so fully effective.




5 - 32

The effective width of the web is evaluated according to EN 1993-1-5 (Table

4.1):

41 k

673,0782,0481,04,28

5,7

271

4,28 σ

p

k

t b

mm5,1245,0

mm249271919,0919,0)3(055,0

673,0782,0481,04,28

5,7

271

4,2841

eff 2e1e

eff 2 p

p

σ

pσ

bbb

b

k

t

b

k

beff = 0,919 × 271 = 249 mm

be1 = be2 = 0,5 beff = 124,5 mm

The effective area of the section is:

Aeff = 6260 – (271 – 249) × 7,5 = 6095 mm2

The effective elastic modulus about the weak axis (W eff,z) is calculated for pure

bending.

In simple bending in the plane of the truss, about the weak axis, the flanges are

inevitably Class 1, whilst the web is not stressed. Then the section is fully

effective:

W eff,z = W el,z = 98,5 cm3

Resistance of cross-section

In compression (EN 1993-1-1 §6.2.4):

0,1

355,06095

M0

yeff Rdc,

f A N = 2164 kN

1683,02164

1477

Rdc,

Ed N

N OK

In bending in the plane of the truss (EN 1993-1-1 §6.2.5):

kNm97,340,1

355,05,98

M0

yzeff,Rdz,

f W M

1082,097,34

86,2

Rdz,

Ed M

M OK

In shear (EN 1993-1-1 §6.2.6):

Av,y = 2×160×11,5 = 3680 mm2




5 - 33

kN7540,1

3

355,03680

3

M0

yyv,

Rd pl,

f A

V

1002,0754

82,1

Rd pl,

Ed

V

V

OK

Since V Ed/V pl,Rd is less than 0,5, there is no influence of the shear force on the

resistance of the cross-section under bending moment and axial force.

M-N interaction (EN 1993-1-1 §6.2.93):

The M-N interaction is taken into account using the following criterion:

0,683 + 0,082 = 0,765 < 1 OK

Buckl ing resistance of member Buckling resistance in the plane of the truss, i.e. about the weak axis of the

cross-section (EN 1993-1-1 § 6.3.1)

The buckling length of the upper chord member is equal to 90% of the system

length (EN 1993-1-1 §B.B.1.1):

Lcr,z = 0,9 × 2151 = 1936 mm

The elastic critical force is:

kN43576,193

78821000ππ

2

2

2z

z2

zcr,

l

EI

N

The slenderness is given by:

705,04357

355,06095

,

eff

z cr

y z

N

f A

The buckling curve to use is curve b (EN 1993-1-1 Table 6.2), and the

imperfection factor is:

= 0,34

8344,0))2,0(1(5,02

zz Φ

781,0705,08344,08344,0

11

222z

2zz

ΦΦ

z

The design buckling resistance is then:

kN16900,1

355,06095781,0

M1

yeff zRdz, b,

f A N

N Ed / N b,z,Rd = 1477/1690 = 0,874 OK




5 - 34

Buckling resistance out of the plane of the truss, i.e. about the strong axis

of the cross-section (EN 1993-1-1 § 6.3.1)

The lateral supports of the upper chord are composed of truss purlins at

8504 mm intervals.

The normal compression force is almost constant between lateral supports(see 3.2).

There is therefore no need to use a method which allows for non-uniform force.


kN33734,850

1177021000ππ

2

2

2y

y2

ycr,

l

EI N

The slenderness is given as:

8009,03373

355,06095

ycr,

yeff y

N

f A

The buckling curve is curve a (EN 1993-1-1 Table 6.2), and the imperfection

factor is:

= 0,21

8838,0))2,0(1(5,02

yy yΦ

7952,08009,08838,08838,0

11222

y2

yy

y

And so the compression resistance is therefore:

kN17200,1

355,060957952,0

M1

yeff yRdy, b,

f A N

N Ed / N b,y,Rd = 1477/1720 = 0,859 OK

M-N interaction (EN 1993-1-1 §6.3.3):

There is no effect of lateral torsional buckling to consider for a member in

bending about its weak axis (no bending about the strong axis). The criteria

are:

1// M1yzeff,

Edz,yz

M1yeff y

Ed f W

M k

f A

N (Eq. 6.61 in EN 1993-1-1)

1// M1yzeff,

Edz,zz

M1yeff z

Ed f W

M k

f A

N (Eq. 6.62 in EN 1993-1-1)

Using resistances already calculated, these criteria can also be written as:




5 - 35

1Rdz,

Edz,yz

Rdy, b,

Ed M

M k

N

N

1Rdz,

Edz,zz

Rdz, b,

Ed M

M k

N

N

The interaction factors k yz and k zz are calculated according to Annex A of

EN 1993-1-1, for a Class 4 section:

zcr,

Ed

y

mzyz

1 N

N C k

where:

zcr,

Edmz )33,0(36,021,079,0

N N C

367,086,2

05,1

C mz = 0,628

8624,0

3373

14777952,01

3373

14771

1

1

ycr,

Ed

ycr,

Ed

y

N

N

N

N

y

819,0

4357

14771

8624,0628,0yz

k

First interaction criterion (eq. 6.61)

1926,097,34

86,2819,0

1720

1477 OK

zcr,

Ed

zmzzz

1 N

N C k

where:

C mz = 0,628

899,0

4357

1477781,01

4357

14771

1

1

zcr,

Ed

zcr,

Ed

N

N

N

N

z

z




5 - 36

Then, the factor k zz can be calculated:

854,0

4357

14771

899,0628,0

zz k

Second interaction criterion (eq. 6.62)

1944,097,34

86,2854,0

1690

1477 OK

Note on secondary trusses

The presence of secondary trusses in the central part of the truss (see

diagram 2.3) permitted the reduction by half of the buckling length of the upper

chord in the plane of the truss.

The secondary truss is sized in order to support a buckling restraint load whosevalue depends on the compression force in the supported chord and on its

slenderness ratio (see EN 1993-3-1 on subject of design of pylons in annex

H4).

4.1.2 Lower chord in compression

With respect to the complete design of the structure, it is also of course

essential to check the lower chord, subject to the lower compression force, but

without support from a secondary truss.

Verification of the lower chord in compression is similar to that described for

the upper chord in compression, in 4.1.1.

Lateral restraint of the lower chord is provided at each purlin (Figure 2.2).

The only specific point which would be interesting to develop is an analysis of

the buckling out of plane of the truss.

Buckling of the lower chord is to be considered similarly to that of the upper

chord, for a length equal to the distance between truss panels, thanks to the

presence of sub-panel braces (See Figure 2.3).

The difference is that the axial force in the lower chord varies along the buckling length, in two panels, whereas the force was constant along the

buckling length for the upper chord.

It should also be noted here that, for the chord member with the greatest

bending moment, the variation in axial force is very small; in a real design, the

small reduction in buckling length due to variation of normal axial force can

safely be ignored.




5 - 37

545 kN 470 kN

Axial force N Ed

Figure 4.3 Axial force in the lower chord

4.1.3 Diagonal in compression

The diagonal, whose resistance is calculated here, by way of example, is the

second diagonal from the right support (element B40 in Figure 3.1), under ULS

gravity loading.

The compression force is:

N Ed = −624,4 kN

Initially, as in common practice, the bending moment due to the self weight of the member is ignored.

The effect of this moment will be evaluated later.

Cross-section properties of a single angle

For a 150 × 150 × 15 L

A = 43 cm2

z G = yG = 4,25 cm

I y = I z = 898,1 cm4

I v = 369 cm4

For a pair of angles

Section area:

A = 2 × 43 = 86 cm2

Second moment of area out of plane of the truss (the section is assumed to

be homogeneous), assuming the gap between the angles is 10 mm:

I y = 2 × 898,1 + 2 × 43 × (4,25+1,0/2)2 = 3737 cm4.

Second moment of area in the plane of the truss:

I z = 2 × 898,1 = 1796 cm4

Class of section in uniform compression

Material parameter for f y = 355 N/mm2: = 0,81

For an angle (EN 1993-1-1 Table 5.2 (Sheet 3)):

31,95,11 10152

1502

2

15,1215 1015

150

t

bh

t

h




5 - 38

The section is a Class 4 and it is therefore not fully effective in uniform

compression. The effective area of the cross-section should be calculated with

reference to EN 1993-1-5. Such a calculation leads to a fully effective area:

Aeff = A = 86 cm2

Resistance of the cross-section

The resistance of the section in uniform compression is therefore given by:

kN30530,1

355,08600

M0

yRdc,

Af N

Buckl ing resistance of member

Buckling resistance in the plane of the truss

The buckling length is equal to:

0,9 × 5,464 = 4,918 m


kN15398,491

179621000ππ

2

2

2y

z2

zcr,

l

EI N

The slenderness is given by:

408,11539

355,08600

zcr,

y

z

N

Af

The buckling curve is curve b (EN 1993-1-1 Table 6.2), and the imperfection

factor is:

34,0

697,1))2,0(1(5,02

zz z Φ

378,0408,1697,1697,1

11

2222

z z z

z

And the buckling resistance is then:

kN11540,1

355,08600378,0

M1

yzRdz, b,

Af N

Buckling resistance out of plane of the truss

The buckling length is equal to the system length: Lcr,y = 5,464m.

The critical axial force is:

kN25945,546

373721000ππ

2

2

2y

y2

ycr,

l

EI N






5 - 40

For this type of buckling the elastic critical force is:

kN836810956

10369210000ππ 3

2

42

2v

v2

vcr,

l

EI N

The slenderness for a single angle is:

427,08368000

3554300

,

vcr

yv

N

Af

The buckling curve to use is curve b and the imperfection factor is: = 0,34

630,0))2,0(34,01(5,02 vvvΦ

915,0

427,0630,0630,0

11

222v

2vv

v

ΦΦ

Conservatively, the resistance to the compression may be evaluated calculating

the reduction factor as the product of that for the whole member and that for an

individual angle between battens:

= Min( y ; z) × v = 0,378 × 0,915 = 0,346

The design buckling resistance of the diagonal is:

kN1056100,1

3558600346,0 3

M1

yRd b,

Af N

0,1591,01056

4,624

Rd b,

Ed N

N

The compression resistance is adequate.

Local verification of the section to the right of the gusset plateconnection

This verification carried out in Appendix B

Effect of bending moment due to self weight of the diagonal

The bending moment is:

M y,Ed = 2,20 kNm (see 3.2 above).

The elastic modulus of the cross-section for bending in the plane of the truss is:

W el,z = 167 cm3.

Interaction criteria are given in EN 1993-1-1 §6.3.3:

1// M1yzel,

Edz,yz

M1yy

Ed f W

M k

Af

N

1// 1,

,

1

M y z el

Ed z

zz

M y z

Ed

f W

M k

f A

N




5 - 41

where:

The k yz factor is:

zcr,

Ed

ymzyz

1 N

N C k

863,0

2594

4,624544,0915,01

2594

4,6241

1

1

ycr,

Edyv

ycr,

Ed

y

N

N

N

N

012,11539

4,62403,0103,01

zcr,

Edmz

N

N C

47,1

1539

4,6241

863,0012,1yz

k

The k zz factor is:

zcr,

Ed

zmz

1 N

N C k zz

691,0

1539

4,624378,0915,01

1539

4,6241

1

1

zcr,

Edzv

zcr,

Ed

z

N

N

N

N

18,1

1539

4,6241

691,0012,1zz

k

From which:

1465,00,1/355167000

1020,247,1

0,1/3558600544,0915,0

624400 6

1635,00,1/355167000

1020,218,1

0,1/3558600378,0915,0

624400 6

When the bending moment due to self weight of the diagonal is taken into

account, the resistance criterion increases from 0,591 to 0,635: that is an

increase of 7%.

4.2 Verification of members in tension

A particular feature when checking the resistance of tension members is theexistence of criteria which bring into play the net section of the member. This

is explored for the worked example.




5 - 42

4.2.1 Lower chord in tension (flat IPE 330)

The lower chord in tension is verified for calculated forces near the mid-span.

Given the results shown in 3.2 above:

N Ed = 1582 kN

M Ed = 1,69 kNm

The tension resistance of the section is determined by two conditions, one in a

“gross” section and the other in a “net” section :

Gross section

A = 6260 mm2

kN2222

0,1

355,06260

M0

yRd pl,

x Af N

Net section

2net mm4661)5,7223()5,11244(6260 A

kN171125,1

51,046619,09,0

M0

unetRdu,

f A N

Tension resistance is given by:

kN1711),min( Rdu,Rd pl,Rdt, N N N

In simple bending, in the truss plane (EN 1993-1-1 (6.2.5)), class 1 of the

section allows the plastic modulus to be mobilised:

32

pl cm2,1474

1615,12

W

kNm3,520,1

355,02,147

M0

y plRd pl,

f W M

The verification is:

03,03,52

69,1

93,01711

1582

Rd

Ed

Rdt,

Ed

M

M

N

N

N-M Interaction: 0,93 + 0,03 = 0,96 < 1

4.2.2 Diagonal in tension (double angles L120 120 12)

Checking is done for the diagonal at the left hand support, under gravity loads.

Given the results shown in 3.2 above:

N Ed = 616,3 kN




5 - 43

M Ed =1,36 kNm

Tension resistance

The tension resistance of the section is determined by two conditions, on ingross section and the other in net section:

Gross section

kN19560,1

355,05510

M0

yRdpl,

x Af N

Net section (See arrangements described in Annex 2)

2net mm4886)12262(5510 A

For angles connected by a single leg, EN 1993-1-8 gives an additional

requirement for the effect of eccentricity of the tension force in the angle(distance between the neutral axis and the gauge marking) on the forces(appearance of secondary moments).

This method involves the application of an ultimate resistance reduction factorfor the angle (EN 1993-1-8 Clause 3.10.3(2))

M2

unet3Rdu,

γ

f A β N

The reduction factor β 3 depends on the distance between axes p1.

For, p1=2,5d 0 =65 mm: 3=0,5 (EN 1993-1-8 Table 3.8)

N.B.: The reduction factors β are only provided for a simple angle; the methodis conservative for a “double angle”. It is recommended that, within theconnection, the behaviour of the two simple diagonals is considered withrespect to these local phenomena.

kN99725,1

51,048865,05,0

M0

unetRdu,

f A N

Then:kN997),min( Rdu,Rdpl,Rdt, N N N

Bending resistance

In simple bending in the truss plane (EN 1993-1-1 (6.2.5)):

3el cm46,85W

kNm3,300,1

355,046,85

M0

yelRdel,

f W M

Verification:




5 - 44

05,03,30

36,1

162,0997

3,616

Rd

Ed

Rdt,

Ed

M

M

N

N

And the M-N interaction criterion is: 0,62 + 0,05 = 0,67 < 1




5 - 45

5 VERIFICATION OF CONNECTIONS

5.1 Characterist ics of the truss post connection

5.1.1 General

It is essential to connect the truss and post according to the assumptions in the

modelling.

In particular, the choice between a fixed connection and a pinned connection

must be respected. The difference between these two types of connection is

that the pinned connection allows a rotation independent deflection of the truss

and the post. The outcome in terms of loading is that the hinge does not

transmit any bending moment from the truss to the post, whereas a fixed

connection does.

The rotation at the support of a truss is manifested by a differential horizontal

displacement between the original node of the upper chord and the original

node of the lower chord.

In order to permit global rotation, it is therefore necessary to allow the

horizontal displacement of the end of one of the chords in relation to the post:

usually, the displacement of the chord which does not receive the diagonal on

support is released.

A

Figure 5.1 Elongated hole on the bottom chord of the truss

With such an arrangement, the axial force is zero in the lower chord in the first panel. The lower chord of the first truss node could therefore be stopped short

(A in the diagram); nevertheless it is preferable to lengthen the lower chord and

to connect it to the post in order to provide lateral stability of the truss at the

level of the lower chord.

An application of this type of hinge action in the worked example is given in

5.1.2 below.

By contrast, in order to carry out a rigid truss-column connection, it is

necessary to make a connection without slack from each of the chords of the

truss to the column.




5 - 46

5.1.2 Convergence of the axes at the truss-column connection

Another question to be asked when carrying out the connection of a truss on a

post is that of convergence of the axes of the connected members and of its

effect on the modelling. The choices are illustrated in Figure 5.2.

Convergence of the axescolumn/chord/diagonal:solution to avoid

Axis convergence of the axes chord/diagonal at the internalface of the column: recommended solutio n

1

1 : Rigid links

Figure 5.2 Rigid truss-column connection

In the first example, the actual physical connection and the model are not

consistent: there is a risk of causing significant secondary moments in the

diagonal and the chord. In the second example, the consistency is much

greater; the eccentric moment is clearly supported by the post, which has a

higher bending resistance than the chord or the diagonal, particularly when the

truss is hinged at the post.

Note that this not the case in the worked example in which the posts have their

web perpendicular to the plane of the truss: the convergence of the three axes

happens then without causing secondary moments.

5.1.3 Worked example: detailing a pinned joint

The Figure 5.3 represents horizontal displacements of the lower and upper nodes of the two support sections, for cases of ULS gravity load combinations

and for cases of ULS uplift load combinations. We can observe that, when the

structure is symmetric or symmetrically loaded, each load case produces equal

global rotations in the two support sections.




5 - 47

35,6 mm 8,6 mm

44,2 mm

(44,2 – 8,6 =35,6 mm) Gravity loading

12,2 mm 3,1 mm

15,2 mm

(15,3 – 3,1 =12,2 mm)

Uplift loading

Figure 5.3 Rotations at truss supports

In order for the global rotations at the supports to be free (assumption for truss

with pinned connections to the column), the elongated holes introduced into the

column on lower chord connection must allow a 35,6 mm movement towards

the outside and 12,2 mm towards the inside. It is of course prudent to allow for a certain safety margin on the sizing of the elongated holes (say 50 mm), and to

check after erection that, under self weight, the freedom of movement remains

adequate in both directions.

5.2 Chord continuity It is often necessary to deliver large span truss beams to site in several sections;

it is therefore necessary to provide continuous chord joints between these

sections. Generally, the preferred method is to make such connections on site

by bolting rather than by welding.

The design of these bolted connections depends on the type of chord section to

be connected. However, we can distinguish between two types of such

connections:

Those in which the bolts are mainly loaded in tension : these use end plates

Those in which bolts are loaded perpendicular to their shank: these use

splice plates.

When the chords are made of a single profile/section in I or H, either of the

connections can be used.

When the chords are made of two double angle or channel sections, splice

connections are generally used.




5 - 48

When the chords are made of hollow sections end plate connections are

generally used (use of hollow sections is outside the scope of this guide).

Continuity using end plate connections

Continuity using splice plate connections

Figure 5.4 Chord continuit y

The splice plate connection shown Figure 5.4 has double cover splice plates on

the web and flanges (giving two interfaces for shear forces). If the force in the

splice is low, single external spliced plates can be used, although double plates

are normally used on the web, to preserve symmetry in the transmission of the

axial force.

The resistance of the splice connections of truss chords must be verified under

dominant load with secondary bending moment in the truss plane, according to

EN 1993-1-8, by adapting the components method developed for beam-post

connections. Software is freely available for this verification (see theSteelBizFrance.com website developed by CTICM). Verification of this type

of connection, for the worked example, is given in Appendix A.

As well as verifying the resistance, it is essential to ensure the stiffness of the

continuous chord connections. Generally, when the resistance of a beam-beam

connection using end plates is selected, it can be considered as rigid.

Spliced plate connections are only effectively rigid when the slack is controlled

(see Section 3.6 for evaluation example of the effect of slack in the bolted

connections of the truss in the worked example). For splice connections, it is

therefore recommended that one of the following options is selected:

Use preloaded bolts with controlled tightening, allowing transmission of

loads by friction (non-slip)

Use fit bolts, preferably loaded on the shank in order to avoid slip under

load by distortion of the thread of the connected pieces.

5.3 Connection of diagonals to chordsConnection of diagonals and posts to chords can be made in different ways,

according to the type of sections to be connected.

When the chords are made of double members (two angles or two UPE

sections), common practice is to insert gusset plates between the two




5 - 49

component members of the chord. The gussets are, therefore, either welded or

bolted on the chords. The diagonals and posts are connected to the gussets,

usually by bolting.

When the chords are made of IPE or HEA/HEB sections, the most common

connection method is also to use a welded gusset plate on the chord. The gusset plate is attached to the flange when the section is upright (vertical web), and to

the web when the section is flat (horizontal web).

(a) Bolted gusset in the space between doubleangle chords, truss members in bolteddouble angles onto gusset

(b) Welded gusset on HEA chord flange,double angle truss members bolted togusset

(c) Gusset welded to web of flat IPE chord

Figure 5.5 Truss connections on chord

When the chord sections are flat, it is also common to use IPE or HEA truss

members with the same depth as the chords and to connect them by double

gussets, one on each flange. An alternative solution is to design a welded

connection without gussets, as shown in Figure 5.6.




5 - 50

1

2

1

2

1

3

4

5

1 Truss members

2 Chord

3 Fillet weld

4 Half-V fillet weld

5 K-fillet weld

Figure 5.6 Welded connection between truss members and chord

When the chords are hollow sections (outside the scope of this guide), the

connection using a gusset welded on the chord is also used. Direct welding of

the diagonals and posts to the chords is also used; this requires profiling for

connections to circular section chords.

In the gusset connections described above, verification of the resistance of the

bolted or welded connection clearly defined in EN 1993-1-8. However,

verification of the resistance of the gusset plate is not. Verification of a gusset

plate connection for the worked example is given in Appendix B.

Special attention must be given to checking of the gussets, particularly those

which have a large non stiffened part: many truss problems have been caused

local buckling of the gusset plate. For example, in the connections in

Figure 5.5(c), if the height of the flat chord web is insufficient for the angles

making up the truss members to be connected near the web, the unstiffened

part of the gusset and its stability must be examined carefully.

Although hollow section trusses are not the subject of the present guide, note

that EN 1993-1-8 devotes a Section to the design of welded connections of

hollow sections.

In the connections to the chords, slip must also be controlled (as indicated for

continuous chords), in order to control displacements of the structural

components, and, as a result, the distribution of forces if the structure is

hyperstatic.




5 - 51

REFERENCES

1 Single-Storey Steel Buildings. Part 7: Fire engineering.

2 EN 1993-1-8:2005 Eurocode 3: Design of steel structures. Part 1.8 Design of

joints.3 EN 1993-1-1: 2005, Eurocode 3: Design of steel structures. Part 1.1 General rules

and rules for buildings.




5 - 52




5 - 53

APPENDIX AWorked Example – Design of a continuous chord

connection using splice plate connections



5 - 54

Appendix A Worked Example: Design of acontinuous chord connection using splice plate connections

1 of 24

Made by PM Date 02/2010

Calculation sheet Checked by IR Date 02/2010

1. Splice joint using bolted cover platesThis calculation sheet refers to the splice plate connection located on the

Figure A.1. This connection has double spliced plates on the web and single

external spliced plate on the flanges (see Figure A.2).

1

1 Splice plate connection studied

Figure A.1 Location of the splice plate connections

2

3

2

31

3

1 Longitudinal axis2 Lower chords to assembly (IPE 330)3 Splice plate connection

Figure A.2 Chord continuit y by splice plate connections

The resistance of this connection must be verified under tension axial force

with secondary moment in the plane of the truss.

Four bolted cover plates must be verified (See Figure A.3)

It is also essential to ensure the stiffness of the continuous chord connection.

A slip resistant connection is required.



TitleAPPENDIX A Worked Example: Design of a continuous chord

connection using splice plate connections2 of 24

5 - 55

213

Z

Y

X

Y

1 cover plates of web chord2 cover plate of flange 1 (on the right-hand side)

3 cover plate of flange 2 (on the left-hand side)

Figure A.3 Cover plates

The global coordinates system is such as:

The XOZ plane is that of the truss plane

The XOY plane is that of the web chord

2. Basic dataThe sizes of the cover-plates and the positioning of holes are shown on the

Figure A.4.





5 - 56

35

70

70

140

70

70

35

35

70

70

35

100

40 95 95 40

1411,5

7 / 7,5 / 7

165 165

5

50

50

30

30

Figure A.4 Sizes (in mm) and posi tioning

Material data (except bolts)

The I-profile and the cover-plates are grade S355 to EN 10025-2.

Steel grade S355

Yield strength f y = 355 N/mm2

Ultimate tensile strength f u = 510 N/mm2

EN 1993-1-1

Table 3.1

I Beam data

Depth h = 330 mm

Flange width b = 160 mm

Web thickness t w = 7,5 mm

Flange thickness t f = 11,5 mm

Radius of root fillet r = 18 mm

Cross-section area A = 62,61 cm2

Second moment of area I y = 788,1 cm4

Plastic modulus W pl,y = 153,7 cm3





5 - 57

Bolted connections data

Category of bolted connections Category C

Bolt Class Class 10.9

Yield strength f yb = 900 N/mm2

Ultimate tensile strength f ub = 1000 N/mm2

For flanges cover plates

Nominal bolt diameter d f = 22 mm

Hole diameter d 0,f = 24 mm

For web cover plates

Nominal bolt diameter d w = 18 mm

Hole diameter d 0,w = 20 mm

EN 1993-1-8Table 3.1

Partial Factors (Recommended values)

Structural steel M0 = 1,00


Bolts M2 = 1,25

Bolts M3 = 1,25

EN 1993-1-1

6.1 NOTE 2B

EN 1993-1-8

2.2 NOTE

Internal forces

For the direction of the internal forces see Figure A.5

M Ed = 1,71 kNm (about y-y axis)

V Ed = 1,7 kN

N Ed = 1567,4 kN (tension force)

Note: the bending moment and the shear force can be ignored. For all that in

some phases we take them into account so as to show the concept of the

calculation in the presence in such internal forces.





5 - 58

Y

X

Y

Z

V Ed

N Ed M Ed

M Ed

Figure A.5 Internal forces and moment

3. Classification of cross-section chord

For the classification of the cross-section, it’s necessary to know thedistribution of the normal stresses.

EN 1993-1-1Table 5.2

Sheet 2 of 3

For the web we consider a uniform stress equal to:

A

N Edw = -250,34 N/mm2

For the flanges we have:

iyy

EdEdi

v I

M

A

N

Where vi is the position of the considered fibre.

And for the upper part ( Z > 0) of the flange:

2f 1 /bv and r t v 2w2

1 = 180,91 N/mm2, 2 = 245,62 N/mm2

And for the inner part ( Z < 0) of the flange:

2f 1 /bv and r t v 2w2

1

= 319,78 N/mm2,2

= 255,06 N/mm2

In view of these results, the cross-section being all over in tension is

considered of class 1.





5 - 59

4. Global checking of the cross-section chord

4.1. Effect of the shear force EN 1993-1-1

6.2.10

Determination of Rd pl,

Ed

V V

With: wv t h A A w = 3959 mm2

M0

yv

Rd pl,

3

f AV = 811,3 kN EN 1993-1-1

6.2.6(2)

From whereRd pl,

Ed

V

V = 0,002< 0,5

So, no reduction due to the shear force needs to be taken into account.

EN 1993-1-1

6.2.10 (2)

4.2. Combination M + N – Effect of the axial force EN 1993-1-1

6.2.9.1

kN4,8174,1567M0

yww

Ed

f t h N

Allowance has to be made for the effect of the axial force.

EN 1993-1-1

6.2.9.1 (5)

4.3. Combination M + N – Consideration of fastener holes

Ax ial force

Under tension axial force, the fastener holes should be considered.

Category C connection the design tension resistance is:

M0

ynet

Rdnet,Rdt,

f A N N

EN 1993-1-1

6.2.3(4)

For the net cross-section, we consider 7 holes for fastener (2 by flange and 3

for the web).

The net area is: net A = 4707 mm2

Therefore: Rdnet, N = 1671 kN

Bending moment

With f f t b A and f f 0,f netf, 2 t d A A

For each flange in tension, one checks:

kN2,6534739,0

M0

yf

M2

unetf,

f A f A

So, the holes for fasteners in the flange should be considered.

EN 1993-1-1

6.2.5 (4)





5 - 60

With ww0,f f 0,net 34 t d t d A A

For the full tension area, one checks:

kN7,22224,17289,0

M0

y

M2

unet

Af f A

So, the holes for fasteners in the web should be considered.

EN 1993-1-1

6.2.5 (5)

Design resistance for bending

With for a IPE 330: y pl W , = 153,7 cm3

d z = 50 mm = distance from centre of holes of flange to z-z axis

zf f 0,holesy, pl, 4 d t d W = 55,2 cm3

The design plastic moment resistance of the net section is:

M0

yholesy, pl,y pl,

Rd pl,

f W W M

= 34,967 kNm EN 1993-1-1

6.2.5(2)

4.4. Combination M + N – Verification

The following criterion should be verified:

Rd N,Ed M M EN 1993-1-1

6.2.9.1(1)

With:

Rdnet,

Ed

N

N n = 0,938

5,0;/)2(min At b Aa f = 0,412

EN 1993-1-1

6.2.9.1(3)

We obtain :

2

Rd pl,Rd N,1

1a

an M M = 6,99 kNm

M Ed = 1,71 < M N,Rd = 6,99 kNm OK

5. Distr ibution of the internal forces EN 1993-1-8

2.5

Note that the web is in the horizontal plane.

5.1. Axial force

The axial force is distributed between the web and the flanges. This

distribution is based on the ratio of the gross cross-section of the web and the

flanges. The fillets are appointed to the flange.

So, with: wf w )2( t t h A 2302,5 mm2

2/)( wf A A A 3958,5 mm2 (per flange)

Then: A A N N /wEdw N, = 576,4 kN

2/w N,Edf N, N N N = 495,5 kN





5 - 61

5.2. Shear force

The shear force is fully transferred by the flanges.

So: 2/Edf V, V V (per flange)

5.3. Bending moment

The bending moment about the weak axis is fully transferred by the flanges:

f M, M 0,855 kNm for each flanges

6. Internal forces in each connected parts

6.1. Connection of the webs

The cover plate of webs (and its bolts) is only subjected to an axial force:

N N,w = 576,4 kN

6.2. Connection of the flanges

Each of cover plates of flanges (and its bolts) is subjected to:

An axial force N N,f = 495,49 kN,

A shear force V V,f = 0,85 kN

A bending moment M M,f = 0,855 kNm

The moment due to the eccentricity of the shear force against the centroid of

the joint (see Figure A.6):

Vf V,f V, eV M

With: eV= 140 mm M V,f = 0,119 kNm

V V,f G

ev

M V,f

Figure A.6 Moment due to the eccentricit y of the shear force





5 - 62

6.3. Summary of the internal forces and moments

In the web: N w = 576,42 kN

In one flange: N f = 495,49 kN

V f = 0,85 kN

M f = 0,97 kN

7. Verification of the web connectionThe connection of the webs is a double lap joint.

The web component will be verified and by symmetry only one plate

component.

7.1. Design details EN 1993-1-8Table 3.3

It is assumed that the structure is not exposed to the weather or other

corrosive influences.

The design details are verified for the web component and for the plate

component in the tables below

Table A.1 Connection of the webs – Web component – Design details

Distance or spacing Min. value Design value Max. value

e1 24 47,5

e2 24 1)

p1 44 70 105

p2 48 95 105

1)Not applicable because of the proximity of the flange

Table A.2 Connection of the webs – Plate component – Design details


e1 24 35

e2 24 40

p1 44 70 98

p2 48 95 98

7.2. Design shear force F V,Ed for each bolt

6

wwEd,V,

N F = 96,07 kN for the component web

EN 1993-1-8

3.12 (3)

6

2/w pEd,V,

N F = 48,03 kN for each component plate

7.3. Design slip resistance F S,Rd

By considering: Bolts in normal holes 0,1s k





5 - 63

Class friction surfaces = Class A 5,0

And with: ws, A 192 mm2 tensile stress area of the bolt

ws,ubc p, 7,0 A f F 134,4 kN pretension force

n number of the friction surfaces

2wn relatively to the web component

1 pn relatively to the plate component

Then: c p,

M3

wswRd,s, F

nk F

107,52 kN

c p,

M3

ps

pRd,s, F nk

F

53,76 kN

EN 1993-1-8

3.9.1 (1)

7.4. Design bearing resistance F b,Rd for each bolt

Table 3.4 of EN 1993-1-8 gives the expressions of the design bearing

resistance. In these expressions, the coefficients b and 1k depend on the

orientation of the loading, the position compared with the ends of the

component and also the position of the other bolts.

EN 1993-1-8

Table 3.4

The general expression for the design bearing resistance is:

M2

u b1Rd b,

t d f k F

EN 1993-1-8

Table 3.4

According to Table 3.4 of the Eurocode 1993-1-8, the coefficients b and k 1

are determined from:

For end bolts

0,1;;3

minu

ub

0

1end b,

f

f

d

e

5,2;7,18,2;7,14,1min0

2

0

2end1,

d

e

d

pk

For inner bolts

0,1;;4

1

3min u

ub

0

1

b,inner f

f

d

p

5,2;7,14,1min0

21,inner

d

pk

Web component

Figure A.7 shows how it is processed for the determination of the coefficients

b and 1k .





5 - 64

N w

F V,Ed,w

k 1 k 1

b,inner

k 1,end

b,inner

k 1,inner

b,inner

k 1,end

b,end

k 1,end

b,end

k 1,inner

b,end

k 1,end

b4 b5 b6

b1 b2 b3

Figure A.7 Connection of the webs – Web component – Determination of type of bolts

The determination of coefficients k 1 is carried out perpendicularly to the

direction of load transfer. But two directions are conceivable for this

perpendicular and it is difficult for some bolts (b1, b4, b3, and b6) to determine

if they are end or inner bolts.

In these cases we consider the minimum value of k 1,inner and k 1,end. And by

noticing that end1,end1,1,inner ;min k k k , these bolts are considered as end

bolts.

In addition, for the web component, it is reminded that the edge distance e2 is

not applicable because of the proximity of the flange. So, the expressions of

k 1,inner and k 1,end are identical.

As the design shear force is identical for each bolt and furthermore:

k 1,inner = k 1,end = 2,50

So only one row of bolts is considered, for example the bolts b1 and b4.

Then, for the bolt b1:

79,0end b1, b, b1 b,

kN01,109wRd, b1, b, F

And for the bolt b4:

92,0inner b4, b, b4 b,

kN23,126wRd, b4, b, F





5 - 65

Therefore, in the end for the web component,

kN01,109wRd, b, F

Plate component Compared with the web component, for the plate it can be noticed that the

bolts b1, b2, b3 become inner bolts and the bolts b4, b5, b6 become end bolts

(see Figure A.8).

Then, for the bolt b1:

92,0inner b1, b, b1 b,

kN81,117 pRd, b1, b, F

And for the bolt b4:

58,0end b4, b, b4 b,

kN97,74 pRd, b4, b, F

In the end, for the plate component, it should retained:

kN97,74 pRd, b, F

Figure A.8 Connection of the webs – Plate component – Determination of type of bolts

k 1 k 1

b,end

k 1,end

b,end

k 1,inner

b,end

k 1,end

b,inner

k 1,end

b,inner

k 1,inner

b,inner

k 1,end

F V,Ed,w

b4 b5 b6

b1 b2 b3

N w/2





5 - 66

7.5. Checking bol ts

7.5.1. With regard to the web component

Individual checking

Design bearing resistance kN01,10907,96 wRd, b,wEd,V, F F

Design slip resistance kN52,10707,96 wRd,s,wEd,V, F F

EN 1993-1-8Table 3.2

Group of fasteners

The shear resistance per shear plane Rdv, F is taken as:

M2

ubvRdv,

A f F

EN 1993-1-8

Table 3.4

By considering that the shear plane does not pass through the threaded portion

of the bolt in normal holes:

v = 0,6

A = 254,47 mm2 (gross cross-section of the bolt)

Then: Rdv, F = 122,15 kN

Since wRd, b,Rdv, F F for only three bolts as a result the design of our group

of fasteners:

kN06,65401,1096min wRd, bi, b, biwRd, b,r, F n F g

EN 1993-1-8

3.7

Then: kN06,65442,576 wRd, b,r,w g F N

7.5.2. With regard to the plate component

Individual checking

Design bearing resistance kN97,7403,48 pRd, b, pEd,V, F F

Design slip resistance kN76,5303,48 pRd,s, pEd,V, F F

EN 1993-1-8

Table 3.2

Group of fasteners

The shear resistance per shear plane Error! Objects cannot be created from editing

field codes. is equal to:

Rdv, F = 122,15 kN

Since Error! Objects cannot be created from editing field codes. for each of the bolts

as a result the design of our group of fasteners:

kN34,57897,74381,11731

Rdh, bi, b,Rdh, b,gr, bin

F F

EN 1993-1-8

3.7

Then: kN34,57821,2282/ Rd b,r,w g F N





5 - 67

7.6. Design of net cross-section

For a connection in tension, the design plastic resistance of the net cross-

section at bolt holes should be verified:

b

1

Rdnet,EdV,

n N F

where n b is the number of bolts located in the considered net cross-section.

EN 1993-1-8Table 3.2

7.6.1. Web component

The net cross-section is taken as 2ww0,wnetw, mm5,18523 t d A A

The design resistance is: kN64,657M0

ynetw,

Rdnet,w,

f A N

Then: kN21,28807,96364,6573

1

wEd,V,Rdnet,w, F N

7.6.2. Plate component

The net cross-section is taken as 2 pw0, pnet p, mm14703 t d A A

The design resistance is: kN85,521M0

ynet p,

Rdnet, p,

f A N

Then: kN10,14403,48385,521

3

1

wEd,V,Rdnet,w, F N

7.7. Design for block tearing

The Figure A.9 shows the block tearing for the web and for the plate. EN 1993-1-8

3.10.2 (1)


The bolt group is subjected to concentric loading.

And with: 2w02nt mm1125)22( t d p A

2w011nv mm5,1312)5,1(2 t d pe A

EN 1993-1-8

3.10.2 (2)

Then: kN01,728Rdeff,1, V

576,42kN01,728 wRdeff,1, N V


Two block tearing are defined. For the both, the shear area is the same, so the

case giving the minimum area subjected in tension is considered. The bolt

group is subjected to concentric loading.

EN 1993-1-8

3.10.2 (2)

And with:2

p02nt mm420)2( t d e A 2

p011nv mm1050)5,1(2 t d pe A







5 - 69

8.1. Design details EN 1993-1-8

Table 3.3

It is assumed that the truss is not exposed to weather or other corrosive

influences.

The design details should be verified in both directions of loading. By taking

into consideration the limits specified in Table 3.3 of EN 1993-1-8, the

following requirement have to be fulfilled:

021 2,1;min d ee

021 2,2;min d p p

mm200;14min;max 21 t p p

The tables below check the design details for each component.

Table A.3 Connection of the flanges – Plate component – Design detailsDistance or spacing Min. value Design value Max. value

21 ee ;min 28,8 30

21 p p ;min 52,8 70

21 p p ;max 100 161

Table A.4 Connection of the flanges – Plate component – Design details


21 ee ;min 28,8 30

21 p p ;min 52,8 70

21 p p ;max 100 196

8.2. Design shear force F V,Ed for each bolt

With regard to the flange component

The components of the design shear force are calculated in the basis vh ,

(see Figure A.10). The group of bolts is subjected to a axial force f N , a shear

force f V and a bending moment f M (see 6.2)

The axial force f N generates a horizontal shear force:

kN58,826

f h bi, N,

N F for each bolt

The shear force f V generates a vertical shear force:

kN14,06

f v bi,V,

V F for each bolt





5 - 70

The moment f M is divided out the bolts according to the distance ir between

the centre of bolts bi and the centre of gravity of the group of bolts

6

1

2i

if biM,

r

r M F

This shear force biM, F resolved in the basis vh , gives:

6

1

2i

if h bi,M,

r

v M F a horizontal component for the bolt bi.

6

1

2i

if v' bi,M,

r

h M F a vertical component for the bolt bi.

With ih and iv coordinates of centre of bolt bi.

In the end, for each bolt:

h bi,M,h bi, N,Edh, bi,V, F F F Horizontal design shear force

v bi,M,v bi,V,Edv, bi,V, F F F Vertical design shear force

2,,,

2,,,Ed bi,V, Ed vbiV Ed hbiV

F F F Resulting design shear force

The Figure A.10 shows the distribution of the internal forces.

V f

G

M f N f

F V,bi,v

F N,bi,h F M,bi

h

v

b1 b2 b3

b6 b5 b4

Figure A.10 Distribution of the internal forces for the flange component.

The Figure A.11 shows the directions of the resulting force and its

components.





5 - 71

F V,v,Ed

F V,h,Ed F V,Ed

h

v

Figure A.11 Directions of the design shear force

Table A.5 sums up the determination of the design shear forces.

The vertical component of the load can be neglected. We will confine to the

horizontal direction for the design bearing resistance checking.

In addition, if we had not considered the shear force EdV and the moment

Ed M , the unique horizontal design shear force would be:

h bi, N,Edh, bi,V, F F = -82,58 kN

That is a difference of 2%

So the value of 84,02 kN can be retained (= maximum value obtained for

Ed bi,V, F ) for the design shear force: kN02,84EdV, F .





5 - 72

Table A.5 Connection of the flanges – Flange component – Design shear

forces in kN in the basis vh , .

Bolt b1 b2 b3 b4 b5 b6

ih -70 0 70 -70 0 70

iv 50 50 50 -50 -50 -50

ir 86,02 50 86,02 86,02 50 76,02

biM, F 2,42 1,41 2,42 2,42 1,41 2,42

h bi,M, F 1,41 1,41 1,41 -1,41 -1,41 -1,41

v bi,M, F 1,97 0 -1,97 1,97 0 -1,97

h bi, N, F -82,58 -82,58 -82,58 -82,58 -82,58 -82,58

v bi,V, F 0,14 0,14 0,14 0,14 0,14 0,14

Ed bi,V, F 81,20 81,17 81,20 84,02 83,99 84,01

Edh, bi,V, F -81,17 -81,17 -81,77 -83,99 -83,99 -83,99

Edv, bi,V, F 2,11 0,14 -1,83 2,11 0,14 -1,83

With regard to the plate component

The connection of the flanges is a single lap joint so the design shear forces

for each bolt with regard to the plate component are directly deduced from the

previous results.

The value of 84,02 kN can be retained.

8.3. Design slip resistance F S,Rd

By considering: Bolts in normal holes 0,1s k

Class friction surfaces = Class A 5,0

And with: f s, A 303 mm2 tensile stress area of the bolt

f s,ubc p, 7,0 A f F 212,1 kN pretension force

n number of the friction surfaces

Single lap joint 1n for each component

Then: c p,

M3

s pRd,s,f Rd,s, F

nk F F

84,54 kN

EN 1993-1-8

3.9.1

8.4. Design bearing resistance F b,Rd for each bol t EN 1993-1-8

Table 3.4

We confine to the horizontal direction for the determination of the design

bearing resistance (see 8.2).





5 - 73

Flange component

Figure A.12 shows for each bolt how the factors b and 1k are determined.

k 1 k 1

k 1k 1

b,end

k 1,end

b,inner

k 1,end

b,inner

k 1,end

b,end

k 1,end

b,inner

k 1,end

b,inner

k 1,end

F V,h,Ed

b1 b2 b3

b4 b5 b6

Figure A.12 Connection of the flanges – Flange component – Determination of type of bolts

For all the bolts: k 1,end = 1,80.

For the bolt b1 and b4: 94,0end b,

kN19,174f Rd, b, F

For the other bolts: 72,0 b,inner

kN19,134f Rd, b, F

In the end for the flange component, the minimum value is retained:

kN19,134f Rd, b, F

Plate component

For all the bolts, k 1,end = 1,80.

For the bolt b3 and b6: 49,0end b,

kN32,90 pRd, b, F





5 - 74

For the other bolts: 72,0b,inner

kN19,134pRd,b, F

In the end for the plate component, the minimum value is retained:kN32,90pRd,b, F

8.5. Verification of the bolts

8.5.1. With regard to the flange component

Individual checking

Design bearing resistance kN19,13402,84 wRd,b,wEd,V, FF

Design slip resistance kN54,8402,84 wRd,s,wEd,V, FF

EN 1993-1-8 Table 3.2

Group of fasteners

The design shear resistance per shear plane Rdv,F is taken as:

M2

ubvRdv,

Af F

EN 1993-1-8 Table 3.4

By considering that the shear plane does not pass through the threaded portionof the bolt in normal holes:

v =0,6

A=380,13 mm2 (gross cross-section of the bolt)

Then: Rdv,F =182,46 kN

Since wRd,b,Rdv, FF for all the bolts, the design resistance of our group of

fasteners is equal to:

kN15,88519,134419,1742bi

1

f Rd,bi,b,wRd,b,r, n

g FF EN 1993-1-83.7

Then: kN15,88549,495 f Rd,b,r,f gFN

8.5.2. With regard to the plate component

Individual checking

Design bearing resistance: kN32,9002,84 pRd,b,pEd,V, FF

Design slip resistance: kN54,8402,84 pRd,s,pEd,V, FF EN 1993-1-8

Table 3.4

Group of fasteners

The shear resistance per shear plane Rdv,F is equal to:

Rdv,F =182,46 kN





5 - 75

Since wRd,b,Rdv, FF for all the bolts, the design of our group of fasteners is

equal to:

kN40,71719,134432,902

bi

1pRd,bi,b,pRd,b,r,

n

g FF

EN 1993-1-8

3.7

Then: kN40,71749,495 pRd,b,r,f p gFNN

8.6. Design of net cross-section

For a connection in tension, the design plastic resistance of the net cross-section at bolt holes should be verified:

b

1

Rdnet,EdV,

n

NF

Wherenb is the number of bolts located in the considered net cross-section.

EN 1993-1-8 Table 3.2

8.6.1. Flange component

The net section area is: 2f f 0,netf, mm25,14272 tdAA f

And: kN67,506M0

ynetf,Rdnet,f,

f AN

Then: kN04,16802,84267,5062

1

f Ed,V,Rdnet,f, FN


The net cross-section is taken as 2pw0,pnetp, mm15682 tdAA

From where kN64,556M0

ynetp,

Rdnet,p,

f AN

Then: kN0416802842645562

1

pEd,V,Rdnet,p, ,,, FN

Note: The global cross-section of the beam has been verified accountingfor the holes for fastener and the combination of the internal forces(see 4).

The net cross-section of the plate component should also be verifiedunder this combination of internal forces.

Assuming a uniform distribution of the load in the section, it isproposed that:

y22

max 3 f

Where:vI

MAN

tnep,

p

netp,

p andnetp,

p

AV





5 - 76

Assuming a uniform distribution of the shear stresses, this leads to a

conservative situation.

With 2net p, mm1568 A

4holes p,gross p,net p, cm643062317187477 ,,, I I I

Then: 2 N/mm316 and 2 N/mm3125,

Finally: 2y

2 N/mm355 N/mm31341 f ,max

8.7. Design for block tearing EN 1993-1-8

3.10.2


The bolt group is subjected to a concentric loading N f and an eccentric

loading V f but considering the presence of the web we only consider the casewith a concentric loading.

The Figure A.13 shows the block tearing for the flange component

AntAnv

Nf

Figure A.13 Connection of the flanges – Block tearing for flangecomponent

With: 2f 02nt mm414)5,0(2 t d e A

2f 011nv mm5,3392)5,22(2 t d pe A


And: kN49,95424,826 wRdeff,1, N V


The bolt group is subjected to a concentric loading N p and an eccentric

loading V p.

The Figure A.14 shows the block tearing for the plate component





5 - 77

For the cases with a concentric loading, only the case giving the minimum

area in tension is considered:

With : 2 p0202nt mm504)5,0(2);(min t d ed p A

2 p011nv mm3220)5,22(2 t d pe A


And: kN49,49560,865 f Rdeff,1, N V

Ant

Anv

Anv

Ant Anv

Ant

AnvV p

N p N p

1

3

2

1 First block tearing with concentric loading2 Second block tearing with concentric loading3 Block tearing with eccentric loading

Figure A.14 Connection of the flanges - Block tearing for plate component

For the case with an eccentric loading, with:

2 p011nt mm1610)5,22( t d pe A

2 p0nv mm1316)5,122( t d pe A


And: kN85,017,598 pRdeff,2, V V

So we have just verified successively the bolt group according to the two

loadings. An additional requirement based on an interactive expression should

be fulfilled:

0,1

;min 3,,2,2,,1,1,,1,

block Rd eff

p

bloc Rd eff block Rd eff

p

V

V

V V

N

Then: 0,157,017,598

85,0

60,865

49,495

OK




5 - 78




5 - 79

APPENDIX BWorked example – Design of a truss node with gusset



5 - 80

Appendix B Worked Example: Design of a t russnode with gusset

1 of 44

Made by CZT Date 12/2009

Calculation sheet Checked by DGB Date 12/2009

The truss includes several types of joints: splice joints by bolted cover plates,

T joints and KT joints. This Appendix gives the detailed design of a KT jointlocated on the upper chord, as shown in Figure B.1.

7100 7200 8500 8600 7100 7100

4000

91 kN 136 182 182 136 136 91 kN

1

1 KT joint

Figure B.1 Location of the KT joint

The values of the internal forces in the truss members (see Table B.1) result

from a gravity load case. This load case corresponds to a ULS combination of

actions, determined according to EN 1990.

Table B.1 KT joint – Internal forces in the truss members

Member N (kN) V (kN) M (kNm)

Diagonal 35 -609,4 -1,27 0

Diagonal 24 406,9 1,03 0

Post 36 2,6 0 0

Chord 101 -413,8 1,25 -0,46

102101

24

3635

136 kN

Chord 102 -1084 1,26 -0,09

1. General presentation of KT joint The KT joint studied consists of the following connections: the gusset to web

chord welded connection and the angles to gusset bolted connection (see

Figure B.2 and Figure B.3). Both connections should be verified according to

the rules from EN 1993-1-1 and EN 1993-1-8.

The gusset to web chord welded connection is a plate welded perpendicular to

the web of the chord by two fillets welds (See Figure B.7).

The angles to gusset bolted connection is composed of two back-to-back

double-angle diagonal members (See Figure B.4) and a single angle post

member (See Figure B.5).

There are three shear connections to be designed as Category C.



Title Appendix B Worked Example: Design of a truss node with gusset 2 of 44

5 - 81

136 kN

1 2 3

1 Chord (IPE 330)2 Gusset plate3 Axes of the web members

Figure B.2 General presentation of KT join t

1

2

3

4

5 6

A

A

BB

1 Web of the chord (IPE 330)

2 Gusset plate 58026015

3 Angles L150150154 Angle L100100105 Fillet weld6 Axes of the web members

Figure B.3 KT join t




5 - 82

Figure B.4 KT Join t – Section AA Figure B.5 KT Joint- Section BB

2. Gusset plate to web chord welded connection

This connection is a welded plate perpendicular to the web of the chord, see

Figure B.6. The two fillet welds are identical. The design of the gusset plateand its weld to the chord takes into account the axial forces in all three angle

members connected to it.

O

α3 α1

Y

Z

260320

30

260

Og

N1,Ed

N2,EdN3,Ed

Figure B.6 Gusset plate to web chord welded connection

The longitudinal axes of all three angle members intersect on the chord axis at

the point O in the web.




5 - 83

The gusset plane is not positioned symmetrically about the normal OY to the

web plane (see Figure B.6 and Figure B.7). The moment resulting from the

eccentricity eZ should be taken into account.

The moment resulting from the eccentricity eY

= t w/2 can be neglected.

Y

Z O

eZ=30

eY=7,5/2Og

Y

X O

Og

tw=7,5

tg=15

Figure B.7 Gusset plate to web chord – Details

The basic assumption is that gusset plate transfers axial forces acting in its

plane and in the direction of the member axes.

2.1. Data

Global coordinates system (see Figure B.6 and Figure B.7)The YOZ Plane is that of the gusset plate

The XOZ Plane is that of the chord web

Geometric data

Gusset plate thickness t g = 15 mm

Web thickness t w = 7,5 mm

Angle between gusset and web a = 90°

Number of fillet welds na = 2

Effective throat thickness a = Value to be defined

Length of welds Lw = 560 mm

Material data

Steel grade: S355

Yield strength: f y = 355 N/mm2

Ultimate tensile strength: f u = 510 N/mm2 EN 1993-1-1

Table 3.1

Note: The specified yield strength and ultimate tensile strength of the filler

metal are required to be at least equivalent to those specified for the parentmaterial.

EN 1993-1-8

4.2(2)




5 - 84

Partial Factor

Resistance of weld: M2 = 1,25 (recommended value)EN 1993-1-8

Table 2.1 NOTE

Internal forces in the truss members (see Figure B.6)

All axial forces are applied in the gusset plate XOZ plane:

Tension axial force at an angle to normal OY of 1 = 42°:

N 1,Ed = 406,9 kN

Tension axial force on the normal OY so 2 = 0°

N 2,Ed = 2,6 kN

Compression axial force at an angle to normal OY of 3 = -41,3°

N 3,Ed = -609,4 kN

2.2. Stresses in the gusset cross-section in front of welds

The approach is based on a linear-elastic analysis that leads to a safe

estimation of the resistance of the welded joint.

EN 1993-1-8

2.4(2)

2.2.1. Design forces in the gusset plate at the chord web face

The effects of the small eccentricity eY from the chord axis will be neglected.

The gusset plate section is verified for the following forces:

N g,Ed Axial force at an eccentricity of eZ = 30 mm to the centreline of the

gusset plateV g,Ed shear force

With:

3

1i

iiEdg, )cos( N N

3

1i

iiEdg, )sin( N V

and Edg, M , the moment resulting from the eccentricity, Edg,ZEdg, N e M

Then: N g,Ed = -152,83 kNV g,Ed = 674,47 kN

M g,Ed = 4,585 kNm

Note: the high axial force component N g,Ed is due to the local point load at the

joint and the self weight of the truss.

2.2.2. Normal stress

Assuming a uniform distribution of the load in the section, the normal stress

is:

v I M

A N

g

Edg,

g

Edg,maxg,




5 - 85

Where: Ag is the cross-section area

I g is the second moment of cross-section

v is the position of the end fibre

With: 58015wgg Lt A = 8700 mm2

12

3wg

g

Lt I = 243,89.106 mm4

v = 290 mm

Then: maxg, = -23,02 N/mm2

2.2.3. Shear stress

The shear mean stress is:

g

Edg,

g A

V

Then: g = 77,53 N/mm2

One usually checks the combination of axial and shear stresses in the gusset

plate section using the Von Mises criterion.

2.3. Design resistance of the fillet weld

The design resistance of a fillet weld should be determined using either the

directional method or the simplified method.

EN 1993-1-8

4.5.3.1(1)

The directional method is based on the comparison between the design tensile

strength and the applied stress in the most severely loaded part of the weld

throat. The applied stress, being determined from a Von Mises formulation,

accounts for the influence on the weld strength of the inclination of the

resultant force per unit length to the weld axis and plane.

The simplified method is based on the design shear strength of the weld to

which is compared directly to an applied weld throat shear stress obtained by

dividing the resultant force per unit of length b the weld throat size. The

simplified method is always safe compared to the directional method.

Here, the directional method is applied. EN 1993-1-8

4.5.3.2

2.3.1. Directional method

Note: a uniform distribution of stress is assumed in the throat section of the

weld.

EN 1993-1-8

4.5.3.2(4)

With: the normal stress to the throat plane

the shear stress (in the plane of throat) perpendicular to the

axis of the weld

the shear stress (in the plane of throat) parallel to the axis of the weld




5 - 86

Note: the normal stress in the weld needs not to be considered. EN 1993-1-8

4.5.3.2(5)

On the throat section of the weld, the force per unit length are:

a = )2/sin( a

a

gmaxg,

n

e= -122,08 N/mm.mm

a = )2/cos( a

a

gmaxg,

n

e= -122,08 N/mm.mm

a =a

gg

n

e = 581,44 N/mm.mm

The design resistance of the fillet weld will be sufficient if the following

conditions are both fulfilled:

w = [ 2+3 ( 2+ 2) ]0,5 ≤ f u / ( w M2)

≤ 0,9 f u / M2

EN 1993-1-8

4.5.3.2(6)

Where: w is the correlation factor for fillet weld

w = 0,8

EN 1993-1-8

Table 4.1

These conditions can be rewritten in the following forms:

(a w) / a ≤ f u / ( w M2)

(a ) / a ≤ 0,9 f u / M2

From these conditions, a minimum value for the effective throat thickness isderived.

a1,min = a w / [ f u / ( w M2)] = 2,03 mm

a2,min = a / (0,9 . f u / M2) = 0,33 mm

amin = max(a1,min ; a2,min) = 2,03 mm

The following requirements must be satisfied:

a 3 mm

l eff max(30 mm ; 6 a) with l eff = Lw – 2 a

EN 1993-1-8

4.5.2(2)

4.5.2(1)

An effective throat thickness of 4 mm is then sufficient.




5 - 87

3. Angles to gusset bolted connectionThree shear connections are designed as Category C. These connections are

shown in Figure B.8.

320 260

260

16

41.3° 42°

15

N1

N2N3

Figure B.8 Angles to gusset bolted connections

This connection is composed of two back-to-back double-angle diagonal

members (N1 and N3) and a single angle post member (N2).

The internal forces in the truss members are:

N 1,Ed = 406,9 kN tension axial force

N 2,Ed = 2,6 kN tension axial force N 3,Ed = -609,4 kN compression axial force

3.1. Basic Data

Material data (except bolts)

Steel grade S355

Yield strength f y = 355 N/mm2

Ultimate tensile strength f u = 510 N/mm2

EN 1993-1-1

Table 3.1

Gusset plate

Thickness t g = 15 mm

Length Lg = 580 mm

Width H g = 260 mm

Angle members

N1 two equal-leg angles L15015015

N2 one equal-leg angle L10010010

N3 two equal-leg angles L15015015




5 - 88

Bolted connections data

Category of bolted connections Category C

Bolt Class Class 10.9

Yield strength f yb = 900 N/mm2

Ultimate tensile strength f ub = 1000 N/mm2

Nominal bolt diameter d = 24 mm

Hole diameter d 0 = 26 mm

EN 1993-1-8Table 3.1

Partial Factors (Recommended values)




Bolts M2 = 1,25

Bolts M3 = 1,25

EN 1993-1-1

6.1 NOTE 2B

EN 1993-1-8

2.2 NOTE

3.2. Global checking of gross cross-sections of thegusset plate

The gross cross-sections of the gusset plates to check are located on the

Figure B.9.

Note: The gross cross-sections of the angles are verified afterward.

320 260

260

2

1

N 3,Ed

N 2,Ed

N 1,Ed

1 = 42°3 = 41.3°

Figure B.9 Location of the gross cross-sections of the gusset plate

Checking of gross cross-section 1

With Ag1 cross-sectional area 1 A g 1 = H g t g = 3900 mm2




5 - 89

Shear resistance

2Ed2,1Ed1,Edg1, cos;cosmax N N V = 457,82 kN

3M0yg1Rd pl,g1, f AV = 799,34 kN

Rd pl,g1,Edg1, V V OK

Ax ial force resistance

3

1i

iEdi,Edg1, )sin( N N = 674,47 kN

M0yg1Rd pl,g1, f A N = 1384,50 kN

Rd pl,g1,Edg1, N N OK

Checking of gross cross-section 2

With Ag2 cross-sectional area 2 Ag2 = Lg t g = 8700 mm2

Shear resistance

3

1i

iEdi,Edg2, )sin( N V = 674,47 kN

3M0yg2Rd pl,g2, f AV = 1783,15 kN

Rd pl,g2,Edg2,

V V OK

Ax ial force resistance

3

1i

iEdi,Edg2, )cos( N N = 152,83 kN

M0yg2Rd pl,g2, f A N = 3088,5 kN

Rd pl,g2,Edg2, N N OK

3.3. Connect ion N3 – Back-to-back double-anglediagonal member N3 to gusset bolted connection

The shear connection in compression is designed as Category C.

The sizes of the components and the positioning of the holes are shown on the

Figure B.10 and Figure B.11.




5 - 90

172

124

76

90

57

99

141

33

60

57

67,5

65

67 65

65

35

C

C

G

Figure B.10 Connection N3 – Sizes (in mm) and posit ioning

60 3357

42.5

15

1

1 Angles neutral axis

Figure B.11 Connection N3 – Section CC

3.3.1. Connect ion N3 – Design forces

With: N 3,Ed Axial compression force at an eccentricity of e N 3 to the

centre of gravity of the joint

M 3,N,Ed Bending moment resulting from the eccentricity, M 3,N,Ed =

e N3 N 3,Ed.

For the gusset:

N 3,g,Ed = 609,4 kN

e N3 = 44,5 mm

M 3,g,Ed = e N3 N 3,g,Ed = 27,12 kNm




5 - 91

For each angle:

N 3,a,Ed = 304,7 kN

M 3,a,Ed = 13,56 kNm

3.3.2. Connection N3 – Checking of angle

Resistance of gross cross-section

Longitudinal stress

Assuming a uniform distribution of the load in the section, the longitudinal

stress is:

v I

M

A

N

a3,

Eda,3,

a3,

Eda,3,

i

Where: A3,a is the section area of the angle

A3,a = 4302 mm2

I 3,a is the second moment of area of angle

I 3,a = 8,981.106 mm4

v position of considered end fibre (see Figure B.12)

v1 = 87 mm

v2 = 63 mm

Then the normal stresses are:

1 = 202,18 N/mm2 (compression)

2 = -24,29 N/mm2 (tension)

2

1

e N3

N 3,a,Ed

Com ressionTension

M 3,a,Ed = e N3 N 3,a,Ed

υ2 υ1

Figure B.12 Stresses in the angle N3




5 - 92

Class of section

20,121510 t h

36,95,11102 t hb

class 4

14,81/10/1093,7 t c

class 2

Class of angle = class 4

EN 1993-1-1

Table 5.2

Sheet 3 of 3

Table 5.2

Sheet 2 of 3

Combination M + N

Criterion to satisfy:M0

y

eff a,3,

Eda,3,

eff a,3,

Eda,3,

Edx,

f

W

M

A

N

with: A3,a,eff effective area of cross-section

leg2eff,a,3,leg1eff,a,3,eff a,3, A A A

where A3,a,eff,leg1 effective area relative to the “free” leg

A3,a,eff,leg2 effective area relative to the “connected” leg

EN 1993-1-1

6.2.9.3

determination of the effective area of cross-section A3,a,eff,leg1

11 = 1,0

buckling factor k = 0,43

p = 0,660 = 1 no reduction

EN 1993-1-5

Table 4.2

EN 1993-1-54.4 (2)


12 = -0,120



EN 1993-1-5

Table 4.2

EN 1993-1-5

4.4 (2)

Verification

a3,eff a,3, A A (no reduction)

35518,202);max(M0

y

21Edx,

f

N/mm2

criterion satisfied

Resistance of net cross-section

From 6.2.5 (5) of EN 1993-1-1, the fastener holes in tension zone need not be

allowed for, provided that the following limit is satisfied for the complete

tension zone:

M0

yt

M2

unett, 9,0

f A f A

EN 1993-1-1

6.2.5 (5)




5 - 93

Here, the holes are in the tension zone (see Figure B.12).

Accounting for a3,eff a,3, A A , the following criterion should be fulfilled:

M0

ya3,

Rdc,a,3,Eda,3,

f A

N N

With 2a3, mm4302 A :

kN2,15277,304 Rdc,a,3,Eda,3, N N

Buckling resistance

A compression member should be verified against buckling.

This condition has been verified in the section dealt with the verification of

the members (see § 4 of this document).

3.3.3. Connection N3 – Checking of gusset plate


For the determination of the gross cross-section of gusset plate, a diffusion of

45° of the axial force N g,Ed is assumed (see Figure B.13).

286,5

45°

45°

112

Figure B.13 Connection N3 – Diffus ion by 45° of the axial force

The following criteria must be satisfied:

M0

y

g3,

Edg,3,

g3,

Edg,3,

Edx,/

f

v I

M

A

N

with: 2gg3, mm5,42975,286 t A

43g3, mm2939570612/5,286 g t I

mm2/325v

Then: 2

M0

yEdx, N/mm35572,29192,14980,141

f




5 - 94

Buckling resistance

The gusset is made similar to an embedded column of characteristics:

Area 2,3 mm5,4297 g A

Height hc = 112 mm (see Figure B.13)

Second moment of area I c,zz = 80578 mm2

We should satisfy:

M1

yg3,

Rd b,g,3,Edg,3,

f A N N

Where is the reduction factor for the relevant buckling curve

EN 1993-1-1

6.3.1.1

With a buckling length of 2hc, the slenderness is given by:

c2

yc2c4

EI

f Ah

= 0,677

The buckling curve to use is curve c and the imperfection is:

= 0,49

2)2,015,0 = 0,846

22

1

= 0,739

Table 6.1

EN 1993-1-1

6.3.1.2

Then: kN11274609 Rd b,g,3,Edg,3, N N ,

3.3.4. Connection N3 – Checking of bolts with regard to thegusset component

Design shear force F V,Ed for each bolt

Due to the orientation of the axial force N 3,Ed, the load on each bolt is not

parallel to the edge of gusset. Also, the components of the design shear load

will be performed in a suitable basis.

EN 1993-1-8

Table 3.43)

In first the components are calculated in the basis vh , located at the

centre of gravity of the joint and oriented in agreement with the principal

directions of the fasteners which are also the principal directions of the angles

(See Figure B.14).

Then a change of basis is performed from the initial vh , to the basis

vh , (see Figure B.15).

In the basis vh , the normal force N 3,g,Ed causes a horizontal shear load for

each bolt bi:

5

Edg,3,

h bi, N,

N F = 101,57 kN




5 - 95

The moment due to eccentricity is divided out according to the distance ir between the centre of bolts bi and the centre of gravity of the joint:

5

1

2i

iEda,1,

biM,r

r M F

F M,b6,h’ F M,b6,v’

F M,b6

F N,b6

N 3,g,Ed

M 3,g,Ed

G

h’ v’

b4

b5

b6

b2

b3

b1

Figure B.14 Connection N3 – Gusset component – Locations

F V,b1,Ed

G h

v

F V,b1,h,Ed

F V,b1,v,Ed

b3

b2

b1b4

b5

b6

Figure B.15 Connection N3 – Gusset component – Loadings




5 - 96

This shear load F M,bi is resolved in the basis vh , :

5

1

2

i

iEda,1,

h bi,M,

r

v M F horizontal component

5

1

2i

iEda,1,

v' bi,M,

r

h M F vertical component

With ih and iv coordinates of centre of bolt bi.

And we obtain (see Table B.2):

h bi,M,h bi, N,Ed,h bi,V, F F F Horizontal shear force,

v bi,M,Ed,v bi,V, F F Transverse shear force,

2

Ed,v bi,V,

2

Ed,h bi,V,Ed bi,V, F F F Resulting shear force

Table B.2 Connection N3 – Gusset component – Design shear forces in kN

in the basis vh , .


ih 81,25 16,25 -48,75 48,75 -16,25 -81,25

i

v -30 -30 -30 30 30 30

ir 86,61 34,12 57,24 57,24 34,12 86,61

biM, F -98,34 -38,74 -64,99 -64,99 -38,74 -98,34

h bi,M, F 34,06 34,06 34,06 -34,06 -34,06 -34,06

v bi,M, F 92,25 18,45 -55,35 55,35 -18,45 -92,25

bi N, F 101,57 101,57 101,57 101,57 101,57 101,57

Ed bi,V, F 164,03 136,88 146,49 87,30 69,98 114,31

Ed,h bi,V, F 135,63 135,63 135,63 67,50 67,50 67,50

Ed,v bi,V, F 92,25 18,45 -55,35 55,35 -18,45 -92,25

The change of basis is performed with:

)cos()sin( 3Ed,v bi,V,3Ed,h bi,V,Edh, bi,V, F F F

)sin()cos( 3Ed,v bi,V,3Ed,h bi,V,Edv, bi,V, F F F

Where 3 = 41,3° (See Figure B.6)

Table B.3gives the results.




5 - 97

Table B.3 Connection N3 – Gusset component – Design shear loads in kN in

the vh , reference system


Ed bi,V, F 164,03 136,88 146,49 87,30 69,98 114,31

Edh, bi,V, F -20,21 -75,65 -131,10 -2,97 -58,41 -113,86

Edv, bi,V, F 162,78 114,07 65,36 87,25 38,54 -10,17

Design details

The structure is not exposed to the weather or other corrosive influences.

We have to verify the design details in the two directions of the components

of loading. By considering the limits specified in Table 3.3 of EN 1993-1-8,

we have to satisfy the following checks:

EN 1993-1-8

3.5 (1) and

Table 3.3

021 2,1;min d ee

021 2,2;min d p p or 021 2,1;min d p p if 04,2 d L

mm200;14min;max 21 t p p

EN 1993-1-8

Table 3.35)

For e1 and e2 observe the minimum end and edge distances according to the

directions Gh and Gv. And For p1 and p2 consider the spacing according to the

directions Gh’ and Gv’ .

The design details are verified in the table below.

Table B.4 Connection N3 – Gusset component – Design detailsDistance or spacing Minimum value Design value Maximum value

21 ee ;min 31,2 57

21 p p ;min 31,2 60

21 p p ;max 65 200

Design bearing resistance F b,Rd for each bolt

Table 3.4 of EN 1993-1-8 gives the expressions for the determination of the

design bearing resistance. These expressions bring into play two coefficients

b and 1k .

EN 1993-1-8

Table 3.4

For each bolt the value of these coefficients depend on the orientation of its

loading, its location compared with the ends of the gusset but also with the

location of the other bolts.

So we are considering successively the horizontal loading (loads in the

direction Gh) and the vertical loading (loads in the direction Gv).




5 - 98

Horizontal loading

The horizontal loading coming from the results of Table 3 is shown on the

Figure B.16.

On this figure we indicate for each bolt how we are processing for thedetermination of its coefficients b and 1k . So, we can specify for each bolt:

the end and edge distances (e1 and e2) and the spacing ( p1, p2 and L) to

consider

the type; end or inner, or end and inner

b3

b2

b1

b4

b5

b6

k 1 k 1k 1k 1

Figure B.16 Connection N3 – Gusset component – Horizontal loading

The general expression for the design bearing resistance is:

M2

u b1Rd b,

t d f k F

EN 1993-1-8

Table 3.4

According to Table 3.4 of the Eurocode 1993-1-8, the coefficients b and k 1

are determined from:

For end bolts

0,1;;

3

min

u

ub

0

1end b,

f

f

d

e

5,2;7,18,2;7,14,1min0

2

0

2end1,

d

e

d

pk

For inner bolts

0,1;;4

1

3min

u

ub

0

1 b,inner

f

f

d

p

5,2;7,14,1min0

21,inner

d

pk

Table B.6 gives the value of the horizontal component of the design bearingresistances F b,bi,h,Rd.




5 - 99

Table B.5 Connection N3 – Gusset component – Horizontal component of the design bearing resistances in kN


e1

e2 172 124 76 90

p1 1)

68,24 68,24 68,24 68,24 68,24 68,24

p2 65 652)

652)

652)

652)

65

b,inner b,inner b,inner b,inner b,inner b,inner

b 0,62 0,62 0,62 0,62 0,62 0,62

min1,k 3) min1,k

3) min1,k

3) 1,inner k 1,inner k min1,k

3)

1k 1,80 1,80 1,80 1,80 1,80 1,80

Rd hbib F ,,,

165,19 165,19 165,19 165,19 165,19 165,19

1)the distance L have been retained

2) L;65min

3) end1,;inner 1,minmin,1 k k k

Vertical loading

The vertical loading coming from the results of Table 3 is shown on the

Figure B.17

b3

b2

b1

b4

b5

b6

k 1

k 1

k 1

Figure B.17 Connection N3 – Gusset component – Vertical loading

Table B.6 gives the value of the vertical component of the design bearing

resistances F b,bi,v,Rd.




5 - 100

Table B.6 Connection N3 – Gusset component – Vertical component of thedesign bearing resistances in kN


e1 90

e2 141 99 57

p1 65 651)

651)

651)

651)

p2 2)

68,24

68,24

68,24

68,24

68,24

68,24

b,inner b,inner b,inner b,inner b,inner end b,

b 0,58 0,58 0,58 0,58 0,58 1,00

1,inner k 1,inner k 1,inner k min1,k 3)min1,k 3)

min1,k 3)

1k 1,97 1,97 1,97 1,97 1,97 1,97

Rd vbib F ,,, 169,16 169,16 169,16 169,16 169,16 289,98

1) L;65min



Design slip resistance F s,Rd

With: As = 353 mm2 tensile stress area of the bolt

subC p, 7,0 A f F = 247,1 kN pretension force

n = 2 number of the friction surfaces relatively to the gusset

EN 1993-1-8

3.9

EN 1993-1-8

3.9.1 (2)

And by considering:

Bolts in normal holes k s = 1,0

Class of friction surfaces = Class A = 0,5

EN 1993-1-8

Table 3.6

Table 3.7

Then: C p,

M3

sRdS, F

nk F

= 197,68 kN

EN 1993-1-8

3.9.1 (1)

Checking bolts – Individual checking

The criteria to satisfy are:

In relation to the design slip resistance

RdS,Ed bi,V, F F

EN 1993-1-8Table 3.2

In relation to the design bearing resistance

Rdh, bi, b,Edh, bi,V, F F

Rdv, bi, b,Edv, bi,V, F F

EN 1993-1-8

Table 3.2 and

Table 3.43)

Note: an additional check based on an interactive expression is proposed:

1

2

Rdv, bi, b,

Edv, bi,V,

2

Rdh, bi, b,

Edh, bi,V,

F

F

F

F




5 - 101

Each bolt has to be verified. The highest values of resistance do not necessary

correspond with the bolt the most loaded.

Table B.7 summarizes only the checks for the bolt b1.

Table B.7 Connection N3 – Gusset component – Checking bolt b1

Design values Resistance values

Ed b1,V, F 164,03 197,68 RdS, F

Edh, b1,V, F 20,21 165,19 Rdh, b1, b, F

Edv, b1,V, F 162,78 169,16 Rdv, b1, b, F

2

Rdv, b1, b,

Edv, b1,V,

2

Rdh, b1, b,

Edh, b1,V,

F

F

F

F 0,94 1

Checking bolts – Group of fasteners

From the Eurocode, the design resistance of a group of fasteners may be taken

as:

bin

F F 1

Rd bi, b,Rd b,gr, if for each bolt bi we have Rd bi, b,Rdv, F F

else Rd bi, b, biRd b,r, min F n F g

EN 1993-1-8

3.7

Where Rd v F , , the shear resistance per shear plane, is taken as:

M2

ubvRdv,

A f F

By considering that the shear plane passes through the threaded portion of the

bolt in normal holes:

v = 0,5

A = As= 353 mm2 (tensile stress area)

Then: Rdv, F = 141,12 kN

Finally for the design resistance we obtain:

Rdh, b,r, g F = 991,17 kN for the horizontal components

Rdv, b,r, g F = 1014,94 kN for the vertical components

And we verify that:

21,402)sin( 3,,3 Ed g N < kN17,991Rdh, b,r, g F

82,457)cos( 3,,3 Ed g N < kN94,1014Rdh, b,r, g F




5 - 102

3.3.5. Connection N3 – Checking bolts with regard to the anglecomponent

Determination of the design ultimate shear load F V,Ed for each bolts

Table B.8 gives the results of the design ultimate shear load F V,bi,Ed and itscomponents F V,bi,h,Ed and F V,bi,v,Ed (See Figure B.18).

These results are deduced from the results obtained for the gusset in the basis

vh , .

F V,b6,Ed

N 3,a,Ed

M 3,a,Ed

G

hv

b4

b5

b6

b2

b3

b1

F V,b6,v,Ed

F V,b6,h,Ed

Figure B.18 Connection N3 – Angle component – Loading

Table B.8 Connection N3 – Angle component – Design shear loads in kN


Ed bi,V, F 82,01 68,44 73,24 43,65 34,99 57,16

Edh, bi,V, F -67,81 -67,81 -67,81 -33,75 -33,75 -33,75

Edv, bi,V, F -46,13 -9,23 27,68 -27,68 9,23 46,13

Design details


Table B.9 Connection N3 – Angle component – Design details

Distance or spacing Minimum value Design value Maximum value

21;min ee 31,2 33

21;min p p 31,2 60

21;max p p 65 200

Determination of the design bearing resistance F b,Rd for each bolts

Horizontal loading The horizontal loading coming from the results of Table B.8 is shown on the

Figure B.19




5 - 103

b4

b5

b6

b2

b3

b1

k 1

k 1

k 1

b

b

Figure B.19 Connection N3 – Angle component – Horizontal loadings

Table B.10 gives the value of the horizontal component of the design bearing

resistances F b,bi,h,Rd.

Table B.10 Connection N3 – Angle component – Horizontal component of thedesign bearing resistances in kN


e1

e2 33 33 33

p1 65 65 65 65 65 65 p2

1) 68,24 68,24 68,24 68,24 68,24 68,24

b,inner b,inner b,inner b,inner b,inner b,inner

b 0,58 0,58 0,58 0,58 0,58 0,58

1,inner k 1,inner k 1,inner k min1,k 2) min1,k

2) min1,k

2)

1k 1,97 1,97 1,97 1,85 1,85 1,85

Rd hbib F ,,, 169,16 169,16 169,16 158,84 158,84 158,84


2)

end1,;inner 1,minmin,1 k k k

Vertical loading

The vertical loading coming from the results of Table B.8 is shown on the

Figure B.20




5 - 104

b4

b5

b6

b2

b3

b1

b

k 1

k 1

Figure B.20 Connection N3 – Angle component – Vertical loading



Table B.11 Connection N3 – Angle component – Vertical component of thedesign bearing resistances in kN


e1 33 33

e2 35 67,5

p11)

68,24

68,24

68,24

68,24

68,24

68,24

p2 65 65 65 65 65 65

b,inner b,inner b,inner b,inner end b, end b,

b 0,62 0,62 0,62 0,62 0,42 0,42

min1,k 2)1,inner k 1,inner k min1,k 2)

1,inner k 1,inner k

1k 1,80 1,80 1,80 1,80 1,80 1,80

Rd vbib F ,,, 165,19 165,19 165,19 165,19 111,85 111,85


2) end1,inner 1,min,1 ;min k k k

Determination of the design sl ip resistance F s,Rd

For the angle component, the number of the friction surfaces is equal to 1.

So with n = 1 we obtain:

C p,

M3

sRdS, F

nk F

= 98,84 kN

EN 1993-1-8

3.9

EN 1993-1-8

3.9.1 (2)


Each bolt has to be verified.

Table B.12 summarizes only the checks for the bolt b1.




5 - 105



Ed b1,V, F 82,01 98,84 RdS, F

Edh, b1,V, F 67,81 169,16 Rdh, b1, b, F

Edv, b1,V, F 46,13 165,19 Rdv, b1, b, F

2

Rdv, b1, b,

Edv, b1,V,

2

Rdh, b1, b,

Edh, b1,V,

F

F

F

F 0,24 1

Checking bolts - Group of fasteners

For the angle we can consider only the horizontal component. In this case:

Rdh, b,r, g F = 991,17 kN

And we verify that:

70,304,,3 Ed a N < kN03,953Rdh, b,r, g F

3.3.6. Connection N3 – Design of net cross-section

For a connection in tension, the design plastic resistance of the net cross-

section at bolt holes should be verified only for a connection in tension.

EN 1993-1-8

3.4.1 (1) c)

3.3.7. Connection N3 – Design of block tearing

Given that this connection is in compression it is not necessary to execute thedesign for block tearing.

3.4. Connect ion N1 – Back-to-back double-anglediagonal member N1 to gusset bolted connection

We have a shear connection in tension to be designed as Category C.

The sizes of the components of this connection and the positioning of the

holes are shown on the Figure B.21. The section DD is identical to the section

CC of the connection N3 (See Figure B.11).




5 - 106

35

65

65

33

60

57

54

D

D

G

76124

80

67,5

Figure B.21 Connection N1 – Sizes (in mm) and posit ioning

3.4.1. Connect ion N1 – Design forces

With: N 1,Ed the normal tension force at an eccentricity of e N 1, to the

centre of gravity of the joint

M 1,N,Ed the moment resulting from the eccentricity, M 1,N,Ed = e N1

N 1,Ed.

We have for the gusset:

N 1,g,Ed = 406,9 kN

e N1 = 44,5 mm

M 1,g,Ed = e N1 N 1,g,Ed = 18,11 kNm

And for each angle:

N 1,a,Ed = 203,45 kN

M 1,a,Ed = 9,05 kNm

3.4.2. Connection N1 – Checking of angle

Resistance of gross cross-section

Longitudinal stress

Assuming an uniform distribution of the load on the section, the longitudinal

stress is:

v I

M

A

N

a1,

Eda,1,

a1,

Eda,1,

i




5 - 107

Where: A1,a cross-sectional area of angle

I 1,a second moment of cross-section of angle

v position of considered end fibre

With: A1,a = 4302 mm2

I 1,a = 8,981.106 mm4

v1 = 87 mm and v2 = 63 mm (see Figure B.22)

We obtain (with compression positive):

1 = -134,99 N/mm2

2 = 16,22 N/mm2

Class of section

20,121510 t h

36,95,11102 t hb

class 4

14,81/10/1093,7 t c

class 2

Class of angle = class 4

EN 1993-1-1

Table 5.2

Sheet 3 of 3

Table 5.2

Sheet 2 of 3

1

2

e N1

N 1,a,Ed

Compression

Traction

M 1,a,Ed = e N1 N 1,a,Ed

G

Figure B.22 Stresses in the angle N1




5 - 108

Combination M + N

Criterion to satisfy:M0

y

eff a,1,

Eda,1,

eff a,1,

Eda,1,

Edx,

f

W

M

A

N

with: A1,a,eff effective area of cross-section

leg2eff,a,1,leg1eff,a,1,eff a,1, A A A

where A1,a,eff,leg1 effective area relative to the “free” leg

A1,a,eff,leg2 effective area relative to the “connected” leg

EN 1993-1-1

6.2.9.3


No reduction because “free” leg in traction


12 = -0,120



EN 1993-1-5

Table 4.2

EN 1993-1-5

4.4 (2)

Verification

a1,eff a,1, A A (no reduction)

35599,134);max(M0

y

21Edx,

f

criterion satisfied

Resistance of net cross-section

We should satisfy:

M0

yneta,1,

Rdnet,a,1,Eda,1,

f A N N EN 1993-1-1

6.2.3. (1) and (4)

The net cross-sections considered are shown on the Figure B.23

1 1

2

2

2

Figure B.23 Net cross-sections of angle N1




5 - 109

With: 22,,11,,1neta,1, mm3588)3588;3912min();min( net anet a A A A

we satisfy:

kN52,131745,203 Rdnet,a,1,Eda,1, N N

3.4.3. Checking of gusset


For the determination of the gross cross-section of gusset, we use an approach

based on a diffusion of 45° of the internal force N g,Ed (see Figure B.24).

45°

45°

195

Figure B.24 Connection N1 – Diffus ion by 45° of the internal force

The following criteria must be satisfied:

M0

y

g1,

Edg,1,

g1,

Edg,1,

Edx,/

f

v I

M

A

N

with: 2gg1, mm2925195 t A

43

g3, mm926859412/195 g t I

mm2/195v

We obtain: 2

M0

y

Edx, N/mm35562,32951,19011,139

f

3.4.4. Connection N1 – Checking of bolts with regard to thegusset component


Due to the orientation of the normal force N 1,Ed, the load on each bolt is not

parallel to the edge of gusset. By consequent the components of the designshear load parallel and normal to the end will be performed.

EN 1993-1-8

Table 3.43)




5 - 110

The calculation of the components is performed in the same way as for

connection N3 (see 3.3.4). We calculate the components in the basis vh ,(see Figure B.25).) then in the basis vh , (see Figure B.26).

N 1,g,Ed

M 1,g,Ed

b1

b2

b3

b4

F N,b2

F M,b2

F M,b2,h’

G

F M,b2,v’

h’

v’

Figure B.25 Connection N1 – Gusset component – Locations

Table B.13 gives the calculations and the results of the design ultimate shear

load F V,bi,Ed and its two components F V,bi,h’,Ed and F V,bi,v’,Ed for each bolt bi in

the vh , reference system.



Bolt b1 b2 b3 b4

ih -16,25 48,75 -48,75 16,25

iv -30 -30 30 30

ir 34,12 57,24 57,24 34,12

biM, F 69,56 116,70 116,70 69,56

h bi,M, F 61,16 61,16 -61,16 -61,16

v bi,M, F -33,13 99,39 -99,39 33,13

bi N, F 101,73 101,73 101,73 101,73

Ed bi,V, F 166,22 190,82 107,35 52,37

Ed,h bi,V, F 162,89 162,89 40,56 40,56

Ed,v bi,V, F -33,13 99,39 -99,39 33,13




5 - 111

b1

b2

b3

b4

h

v

G

F V,b3,Ed

F V,b2,Ed

F V,b1,Ed

F V,b4,Ed

Figure B.26 Connection N1 – Gusset component – Loadings

The change of basis is performed with:

)sin()cos( 3Ed,v bi,V,3Ed,h bi,V,Edh, bi,V, F F F

)cos()sin( 1Ed,v bi,V,1Ed,h bi,V,Edv, bi,V, F F F

Where 1 = 42° (See Figure B.6)

Table B.14 gives the results.



Bolt b1 b2 b3 b4

Ed bi,V, F 166,22 190,82 107,35 52,37

Edh, bi,V, F 84,37 182,86 -46,72 51,76

Edv, bi,V, F -143,22 -54,54 -96,65 -7,97

Design details


For e1 and e2 we observe the minimums end and edge distances according to

the appropriate direction (Gh or Gv). For p1 and p2 we consider the spacing

according to the principal direction of the joint (Gh’ or Gv’ ).

Table B.15 Connection N1 – Gusset component – Design details


21 ;min ee 31,2 54

21 ;min p p 31,2 60

21 ;max p p 65 200




5 - 112


Horizontal loading

The horizontal loading coming from the results of Table B.14 is shown on the

Figure B.27

b1

b2

b3

b4

b

b

b

k 1k 1

Figure B.27 Connection N1 – Gusset component – Horizontal loading



Table B.16 Connection N1 – Gusset component – Horizontal component of the design bearing resistances in kN

Bolt b1 b2 b3 b4

e1 80 54

e2 124 76

p1 651)

65

p2 651)

651)

651)

651)

b,inner end b, b,inner end b, b

0,58 1,00 0,58 0,69

min1,k 3) min1,k

3) 1,inner k 1,inner k

1k 1,80 1,80 1,80 1,80

Rdh, bi, b, F 154,22 264,38 154,22 183,04

1) L;65min





5 - 113

Vertical loading

The vertical loading coming from the results of Table B.14 is shown on the

Figure B.28.

b1

b2

b3

b4

b b

k 1

k 1

k 1

Figure B.28 Connection N1 – Gusset component – Vertical loading



Table B.17 Connection N1 – Gusset component – Vertical component of the

design bearing resistances in kN

Bolt b1 b2 b3 b4

e1 124 76

e2 80 98 54

p1 651)

651)

p2

651)

65

65

651)

end b, end b, b,inner b,inner b

1,00 0,97 0,58 0,58

1,inner k min1,k 2)min1,k 2)

min1,k 2)

1k

1,80 1,80 1,80 1,80

Rd vbib F ,,, 264,38 257,60 154,22 154,22

1) L;65min

2) end1,1,inner min,1 ;min k k k




5 - 114


With n = 2, the number of the friction surfaces relatively to the gusset, we

obtain:

C p,

M3

sRdS, F nk F

= 197,68 kN

EN 1993-1-8

3.9

EN 1993-1-8

3.9.1 (1)


Each bolt has to be verified.

Table B.18 and Table B.19 summarize only the checks for the bolt b1 and b2.



Ed b1,V, F 166,22 197,68 RdS, F

Edh, b1,V, F 84,37 154,22 Rdh, b1, b, F

Edv, b1,V, F 143,22 264,38 Rdv, b1, b, F

2

Rdv, b1, b,

Edv, b1,V,

2

Rdh, b1, b,

Edh, b1,V,

F

F

F

F 0,59 1



Ed b1,V, F 190,82 197,68 RdS, F

Edh, b1,V, F 182,86 264,38 Rdh, b1, b, F

Edv, b1,V, F 54,54 257,60 Rdv, b1, b, F

2

Rdv, b1, b,

Edv, b1,V,

2

Rdh, b1, b,

Edh, b1,V,

F

F

F

F 0,52 1

Checking bolts – Group of fastenersBy considering that the shear plane passes through the threaded portion of the

bolt in normal holes:

v = 0,5

A = As= 353 mm2 (tensile stress area)

We obtain:

Rdv, F = 141,12 kN




5 - 115

And for the design resistance:

Rdh, b,r, g F = 616,90 kN for the horizontal components

Rdv, b,r, g F = 616,90 kN for the vertical components

And we verify that:

27,272)sin( 1,,1 Ed g N < kN90,616Rdh, b,r, g F

39,302)cos( 1,,1 Ed g N < kN90,616Rdh, b,r, g F

3.4.5. Connection N1 – Checking bolts with regard to the anglecomponent


Table B.20 gives the results of the design ultimate shear load F V,bi,Ed and its

components F V,bi,h,Ed and F V,bi,v,Ed (See Figure B.29).

These results are deduced from the results obtained for the gusset in the basis

vh , .

N 1,a,Ed

M 1,a,Ed

b1

b2

b3

b4

h v

G

F V,b1,Ed

F V,b2,Ed

F V,b3,Ed

F V,b4,Ed

Figure B.29 Connection N1 – Angle component – Loading

Table B.20 Connection N1 – Angle component – Design shear loads in kN

Bolt b1 b2 b3 b4

Ed bi,V, F 83,11 95,41 53,67 26,19

Edh, bi,V, F 81,44 81,44 20,28 20,28

Edv, bi,V, F 16,57 -49,70 49,70 -16,57

Design details





5 - 116

Table B.21 Connection N1 – Angle component – Horizontal loading – Designdetails


21 ;min ee 31,2 33

21 ;min p p 57,2 60 200

21 ;max p p 65 200


Horizontal loading

The horizontal loading coming from the results of Table B.20 is shown on the

Figure B.30

b1

b2

b3

b4

b

b

k 1

k 1

Figure B.30 Connection N1 – Angle component – Horizontal loadings






5 - 117

Table B.22 Connection N1 – Angle component – Horizontal component of thedesign bearing resistances in kN

Bolt b1 b2 b3 b4

e1 67,5 35

e2 33 33

p1 65 65

p2 1)

68,24 68,24 68,24 68,24

end b, b,inner end b, b,inner b

0,87 0,58 0,45 0,58

1,inner k 1,inner k min1,k 2) min1,k

2)

1k

1,97 1,97 1,85 1,85

Rdh, bi, b, F 250,95 169,16 122,18 158,84

1)

the distance L have been retained2) end1,inner 1,min,1 ;min k k k

Vertical loading

The vertical loading coming from the results of Table 20 is shown on the

Figure B.31

b1

b2

b3

b4

k 1

k 1

Figure B.31 Connection N1 – Angle component – Vertical loading






5 - 118

Table B.23 Connection N1 – Angle component – Vertical component of thedesign bearing resistances in kN

Bolt b1 b2 b3 b4

e1 33

e2 67,5 35

p11)

68,24

68,24

68,24

p2

65 65 65 65

b,inner b,inner end b, b,inner b

0,62 0,62 0,42 0,62

min1,k 2)1,inner k min1,k 2)

1,inner k 1k

1,80 1,80 1,80 1,80

Rdh, bi, b, F 165,19 165,19 111,85 165,19

1)

the distance L have been retained2)

end1,;inner 1,minmin,1 k k k


For the angle component, the number of the friction surfaces is equal to 1.

So with n = 1 we obtain:

C p,

M3

sRdS, F

nk F

= 98,84 kN

EN 1993-1-8

3.9

EN 1993-1-8

3.9.1 (2)

Checking bolts – Individual checking Each bolt has to be verified. Table B.24 summarizes only the checks for the

bolt b2.

Table B.24 Connection N1 – Angle component – Checking bolt b2


Ed b1,V, F 95,41 98,84 RdS, F

Edh, b1,V, F 81,44 169,16 Rdh, b1, b, F

Edv, b1,V, F 49,70 165,19 Rdv, b1, b, F

2

Rdv, b1, b,

Edv, b1,V,

2

Rdh, b1, b,

Edh, b1,V,

F

F

F

F 0,32 1

Checking bolts – Group of fasteners

For the angle we can consider only the horizontal component:

Rdh, b,r, g F = 488,73 kN

And we verify that:

45,203,,1 Ed a N < kN73,488Rdh, b,r, g F




5 - 119

3.4.6. Connection N1 – Design of net cross-section

Gusset component

For a connection in tension, the design of the net cross-sections have to be

verified.Verify on the net cross-section marked 1 on the Figure B.32. For this section,

we have to satisfy:

M0

ynet1

b

Edg,1,

b

f A

n

N n

t

EN 1993-1-8

3.4.1 (1) c) and

Table 3.2

Where 2 b n number of bolts relative to the cross-section

4 bt n total number of the connection

With 1net A 2194 mm2

We satisfy: kN7784,203M0

ynet1

b

Edg,1,

b

f A

n

N n

t

Angle component

We have been already verified the net cross-section (see 3.4.2).

Moreover these checking have been realised with N Ed in loco n b F V,Ed.

3.4.7. Connection N1 – Design for block tearing

Gusset component

EN 1993-1-8

3.10.2

The Figure B.32 shows the block tearing for the gusset.

N 1,g,Ed

1

1

1

Ant

Anv

Anv

Anv

Anv

Figure B.32 Connection N1 – Block tearing for gusset

Our bolt group is subjected to eccentric loading and we have to satisfy:

Rdeff,2,Edg,1, V N

EN 1993-1-8

3.10.2 (3)




5 - 120

WhereM0

nv

M2

ntuRdeff,2,

3

15,0

A f A f V

y

With Ant = 633,6 mm2

Anv = 3533,1 mm2

We satisfy:

kN4,8539,406 Rdeff,2,Edg,1, V N

Angle component

The Figure B.33 shows the block tearing for the gusset.

N 1,a,Ed

Anv

Anv

Ant

Ant

Figure B.33 Connection N1 – Block tearing for angle

Our bolt group is subjected to eccentric loading and we have to satisfy:

Rdeff,2,Eda,1, V N

EN 1993-1-8

3.10.2 (3)

With Ant = 933,6 mm2

Anv = 1402,5 mm2

We satisfy:

kN91,40745,203 Rdeff,2,Edg,1, V N

3.5. Connect ion N2 – Single angle post member N2 togusset bolted connection

We have a shear connection in tension to be designed as Category C.

Given that the loading is low, the checking of this connection is not carry out.

Otherwise the procedure stays the same with in addition the following point.




5 - 121

We are dealing with a single angle in tension by a single row of bolts in one

leg. During the checking of the net cross-section of this angle, the design

ultimate resistance should be determined as follows:

M2

unet2Rdu,

f A N

With 4,02 ( 01 5,265 d p )

EN 1993-1-8

3.10.3 (2)

and

Table 3.8

3.6. Influences of the eccentrici ty and other parameters

We consider only the bolts with regard to the gusset component.

3.6.1. Connection N3 – Moment due to eccentr ici ty

The effects of the eccentricity depend of the locations of the bolts

comparatively with the neutral axis but also to each other.

Lets the moment due to the eccentricity equal to 0. In this case and whatever

the bolt we obtain in the basis vh , :

kN57,101Ed b,V, F (value without moment due to eccentricity)

kN03,67Edh, b,V, F (value without moment due to eccentricity)

kN30,76Edv, b,V, F (value without moment due to eccentricity)

Values to compare at the results obtained for the bolt b1:

kN03,164Ed b,V, F (value with moment due to eccentricity)

kN21,20Edh, b,V, F (value with moment due to eccentricity)

kN78,162Edv, b,V, F (value with moment due to eccentricity)

3.6.2. Connection N3 – Influence of number of bolts and spacing p1

Reduce the number of bolts from 6 to 5 by suppression of bolt marked b6 (see

Figure B.14). This modification modifies the location of the centre of gravityof the bolt group. Even if the moment due to eccentricity decrease, the design

shear loads per bolt increase. And two bolts (b1 and b3) do not again satisfy to

the criteria relative to the design bearing resistances (see tables below).




5 - 122

Table B.25 Connection N3 – Gusset component – Bolt b1 – Reduction of total number of bolts


Total number of

bolts6 5

Ed b1,V, F 164,03 189,76 197,68 RdS, F

Edh, b1,V, F 20,21 28,43 165,19 Rdh, b1, b, F

Edv, b1,V, F 162,78 187,62 169,16 Rdv, b1, b, F

Table B.26 Connection N3 – Gusset component – Bolt b3 – Reduction of total number of bolts


Total number of

bolts 6 5

Ed b1,V, F 146,49 189,76 197,68 RdS, F

Edh, b1,V, F 131,10 182,40 165,19 Rdh, b1, b, F

Edv, b1,V, F 65,36 52,36 169,16 Rdv, b1, b, F

At this stage, increase the value of the spacing p1 from 65 to 75 mm. So all

the bolts satisfy the criteria. Look for example the results for bolt b1.

Table B.27 Connection N3 – Gusset component – Bolt b1

– Increasing of spacing p1 to 75 mm


Ed b1,V, F 180,06 197,68 RdS, F

Edh, b1,V, F 28,74 225,70 Rdh, b1, b, F

Edv, b1,V, F 177,75 220,50 Rdv, b1, b, F

3.6.3. Connection N1 – Influence of number of bolts

Reduce the number of bolts from 4 to 3 by suppression of bolt marked b3 (seeFigure B.25). The moment due to eccentricity decrease whereas the design

shear loads per bolt increase. And two bolts (b1 and b2) do not again satisfy to

the criteria relative to the design bearing resistances (see tables below).







Part 6: Detailed Design of

Built-up Columns






Part 6: Detailed Design of

Built-up Columns



6 - ii



Part 6: Detailed Design of Built-up Columns

6 - iii

FOREWORD

This publication is part six of the design guide, Single-Storey Steel Buildings.




Part 3: Actions



Part 6: Detailed design of built-up columns












6 - iv




6 - v

ContentsPage No

FOREWORD iii

SUMMARY vi

1 INTRODUCTION 1

2 TYPES OF BUILT-UP MEMBERS AND THEIR APPLICATION 2 2.1 General 2 2.2 Laced built-up columns 5 2.3 Battened built-up columns 7

3 DETAILED CALCULATIONS 9 3.1 General 9 3.2 Design methodology for laced built-up columns 9 3.3 Design methodology for battened built-up columns 14

3.4 Buckling length 17

REFERENCES 19

APPENDIX A Worked Example: Design of a laced built-up column 21




6 - vi

SUMMARY

This guide covers the structural arrangements and the calculations for built-up columns

fabricated from hot rolled sections.

The calculations refer to the European Standard EN 1993-1-1, with complementaryinformation where necessary.

The design procedures of EN 1993-1-1 are presented to verify a built-up column with

lacing or battening using simplified equations and formulas.

A worked example is given in Appendix A.




6 - 1

1 INTRODUCTION

Built-up columns are used in steel construction when the column buckling

lengths are large and the compression forces are relatively low. This guidecovers two types of built-up columns:

Built-up columns with lacing

Built-up columns with battens.

This document includes an overview of common details for such members. It

describes the design method according to EN 1993-1-1[1] for the determination

of the internal forces and the buckling resistance of each member (chords,

diagonals, etc) of built-up columns made of hot rolled profiles.

It should be noted that due to the shear deformation, battened built-up columns

are more flexible than solid columns with the same inertia; this must be taken

into account in the design.

In order to derive the axial resistance of a steel built-up column, the following

must be addressed:

Analysis of the built-up column to determine the internal forces by taking

into account an equivalent initial imperfection and the second order effects

Verification of the chords and bracing members (diagonals and battens)

Verification of the connections.

A fully worked example of a built-up column with an N-shape arrangement of

lacings is given in Appendix A, which illustrates the design principles.




6 - 2

2 TYPES OF BUILT-UP MEMBERS AND THEIR APPLICATION

2.1 General In general, built-up columns are used in industrial buildings, either as posts for

cladding when their buckling length is very long, or as columns supporting a

crane girder.

When used as a post for cladding with pinned ends, the column is designed to

support the horizontal forces, mainly due to wind. Hence the bending moment

in such a built-up column is predominant compared to the compression force.

Figure 2.1 Post for cladding with pinned ends

A typical built-up column that supports a crane girder is shown in Figure 2.2.

They usually have a fixed base and a pinned end at the top, and are designed to

resist:

The compression forces that result either from the frame or from the crane

rail

The horizontal forces that result from the effects of the crane applied on the

internal chord and the wind loads applied to the external one.

In this case, the compression forces are predominant compared to the bendingmoment.




6 - 3

1 Crane girder

Figure 2.2 Built-up column supporting a crane girder

The built-up columns are composed of two parallel chords interconnected by

lacings or battens – see Figure 2.1. In general, the truss system concentrates

material at the structurally most efficient locations for force transfer.

In an industrial building and for a given height, built up columns theoretically

have the least steel weight of any steel framing system.

Any hot rolled section can be used for the chords and the web members of

built-up columns. However, channels or I-sections are most commonly used as

chords. Their combination with angles presents a convenient technical solution

for built-up columns with lacing or battens. Flat bars are also used in built-up

column as battens.

This guide covers two types of built-up columns with pinned ends that are

assumed to be laterally supported:

Laced columns

Battened columns.

1 NEd = 900 kN

MEd = 450 kNm




6 - 4

Laced column Battened column

Figure 2.3 Built-up columns

The difference between these two types of built-up columns comes from the

mode of connection of the web members (lacings and battens) to the chords.

The first type contains diagonals (and possibly struts) designed with pinned

ends. The second type involves battens with fixed ends to the chords andfunctioning as a rectangular panel.

The inertia of the built-up column increases with the distance between the

chord axes. The increase in stiffness is counterbalanced by the weight and cost

increase of the connection between members.

Built-up columns provide relatively light structures with a large inertia. Indeed,

the position of the chords, far from the centroid of the built-up section, is very

beneficial in producing a great inertia. These members are generally intended

for tall structures for which the horizontal displacements are limited to low

values (e.g. columns supporting crane girders).

The axial resistance of built-up columns is largely affected by the shear

deformations. The initial bow imperfection is significantly amplified because

of the shear strains.

It is possible to study the behaviour of built-up columns using a simple elastic

model.




6 - 5

2.2 Laced bui lt-up columns2.2.1 General

There is a large number of laced column configurations that may be

considered. However, the N-shape and the V-shape arrangements of lacings are

commonly used.

Figure 2.4 Built-up column with lacings in an industrial building

The selection of either channels or I-sections for chord members provides

different advantages. I-sections are more structurally efficient and therefore are

potentially shallower than channels. For built-up columns with a large

compressive axial force (for example, columns supporting cranes), I or

H sections will be more appropriate than channels. Channels may be adequate

in order to provide two flat sides.

Tee sections cut from European Column sections are also used for the chord

members. The web of the Tee sections should be sufficiently deep to permit

easy welding of the bracing members.

The angle web members of the laced column allow use of gusset-less welded

connections, which minimises fabrication costs. Other member types require

either gussets or more complex welding.

The centroidal axes of the compression and tension web members are not

necessarily required to meet at the same point on the chord axes. In fact, laced

columns with an eccentricity at the joints can be as efficient as those without

eccentricity. The chord-web joint can be separated without an increase in steelweight. Although eccentric joints require that local moments be designed for,

there are several advantages in doing so. Eccentric joints provide additional




6 - 6

space for welding, hence reducing fabrication complexity. In addition, the

reduced length of the compression chord provides enhanced buckling and

bending resistance which partly compensates for the additional moments

generated by the joint eccentricity. For single angles, it is recommended that

joint eccentricity is minimised.

2.2.2 Various lacing geometries

The N-shape arrangement of lacings, as shown in Figure 2.5(a), can be

considered as the most efficient truss configuration, for typical frames in

industrial buildings. The web of the N-shape arrangement comprises diagonals

and posts that meet at the same point on the chord axes.

This arrangement reduces the length of the compression chords and diagonals.

It is usually used in frames with a significant uniform compressive force.

The V-shape arrangement of lacings increases the length of the compression

chords and diagonals and provides a reduction of buckling resistance of themembers. This arrangement is used in frames with a low compressive force.

The X-shape configurations are not generally used in buildings because of the

cost and the complexity of fabrication.

(a) N-Shape (b) V-shape (c) X-shape

Figure 2.5 Different shape arrangements of lacing




6 - 7

2.2.3 Construction details

Single lacing systems on opposite faces of the built-up member with two

parallel laced planes should be corresponding systems as shown in

Figure 2.6(a) (EN 1993-1-1 § 6.4.2.2(1)).

When the single lacing systems on opposite faces of a built-up member withtwo parallel laced planes are mutually opposed in direction, as shown in

Figure 2.6(b), the resulting torsional effects in the member should be taken into

account. The chords must be designed for the additional eccentricity caused by

the transverse bending effect, which can have a significant influence on the

member size.

Tie panels should be provided at the ends of lacing systems, at points where the

lacing is interrupted and at joints with other members.

1 2 2 1

11

2 2

A B

Lacing on face A Lacing on face B

(a) Corresponding lacing system(Recommended system)

1 2 2 1

11

2 2

A B

Lacing on face A Lacing on face B

(b) Mutually opposed lacing system(Not recommended)

Figure 2.6 Single lacing system on opposite faces of a built-up member withtwo parallel laced planes

2.3 Battened bui lt-up columnsBattened built-up columns are not appropriate for frames in industrial

buildings. They are sometimes used as isolated frame members in specific

conditions, where the horizontal forces are not significant.

Channels or I-sections are mostly used as chords and flat bars are used as

battens. The battens must have fixed ends on the chords.




6 - 8

Battened built-up columns are composed of two parallel planes of battens

which are connected to the flanges of the chords. The position of the battens

should be the same for both planes. Battens should be provided at each end of

the built-up member.

Battens should also be provided at intermediate points where loads are applied,and at points of lateral restraint.

a) Chords made of channels

b) Chords made of I sections

Figure 2.7 Battened compression members with two types of chords




6 - 9

3 DETAILED CALCULATIONS

3.1 General

The design methodology described hereafter can be applied to verify theresistance of the various components of a built-up member with pinned ends,

for the most critical ULS combination. The design axial force, NEd, and the

design bending moment, MEd, about the strong axis of the built-up member are

assumed to have been determined from analysis in accordance with

EN 1993-1-1[1].

This methodology is applicable to built-up columns where the lacing or

battening consists of equal modules with parallel chords. The minimum

number of modules in a member is three.

The methodology is summarized in the flowchart in Figure 3.2 for laced built-up columns, and in Figure 3.4 for battened built-up columns. It is

illustrated by the worked example given in Appendix A.

3.2 Design methodology for laced bui lt-up columns3.2.1 Step 1: Maximum compression axial force in the chords

Effective second moment of area

The effective second moment of area is calculated using the following

expression (EN 1993-1-1 § 6.4.2.1(4)):

ch

2

0eff 5,0 AhI

where:

h0 is the distance between the centroids of chords.

Ach is the cross-sectional area of one chord.

Shear stif fness

For the stability verification of a laced built-up column, the elastic elongations

of the diagonals and the posts must be considered in order to derive the shear

stiffness Sv. Formulae for the shear stiffness Sv are given in Table 3.1 for different arrangements of lacing.

Initial bow imperfection

The built-up column is considered as a column with an initial bow imperfection

of e0, as shown in Figure 3.1:

e0 =L/500

where:

L is the length of the built-up member




6 - 10

Table 3.1 Shear sti ffness S v of built-up columns

N-shape V-shape K-shape X-shape

Ad

Av

h0

a

d Ad

Ad

h0

a

d

Ad

Av

h0

a

d

Ad

Av

h0

a

d

3d

30d3

30d

1d A

h Ad

ahnEASV

3

20d

2d

ahnEAS V 3

20d

d

ahnEAS V 3

20d2

d

ahnEAS V

n is the number of planes of lacing

Ad is the section area of a diagonal

Av is the section area of a post

d is the length of the diagonal

Figure 3.1 Initial bow imperfect ion

Maximum axial compression force in the chords

Verifications should be performed for chords using the design forces Nch,Ed

resulting from the applied compression force NEd and the bending moment MEd

at mid-height of the built-up column.

For a member with two identical chords, the design force Nch,Ed is determined

from the following expression (EN 1993-1-1 § 6.4):

Nch,Ed =eff

ch0EdEd

22 I

AhMN

N Ed

e0 = L/500

L/2

L/2




6 - 11

where:

MEd is the maximum bending moment at mid-height of the built-up column

including the equivalent imperfection e0 and the second order effects:

MEd =

v

Ed

cr

Ed

I

Ed0Ed

1SN

NN

MeN

Ncr is the effective critical force of the built-up column:

2eff

cr

²π

L

EIN

NEd is the design compression axial force applied to the built-up column.

IEdM is the design value of the maximum moment at mid-height of the

built-up column without second order effects.

3.2.2 Step 2: In-plane buckling resistance of the chord

Classification of the cross-section of the chord

The cross-section of the chord must be classified according to EN 1993-1-1

Table 5.2.

Buckl ing resistance of a chord about z-z axis

The resistance of the chord has to be verified for flexural buckling in the plane

of the built-up member, i.e. about the weak axis of the cross-section of the

chord (z-z axis). The buckling verification is given by (EN 1993-1-1 § 6.4.2):

1Rdz, b,

Edch, N

N

where:

N b,z,Rd is the design buckling resistance of the chord about the weak axis of

the cross-section, calculated according to EN 1993-1-1 § 6.3.1.

Information on the buckling length Lch of the chord is given in Section 3.4 of

this guide.

3.2.3 Step 3: Out-of-plane buckling resistance of the chordsOut-of-plane buckling of the member, i.e. buckling about the strong axis of the

cross-section of the chords (y-y axis), has to be considered. The buckling

verification is given by:

1Rdy, b,

Edch, N

N

where:

N b,y,Rd is the design buckling resistance of the chord about the strong axis

of the cross-section, calculated according to EN 1993-1-1 § 6.3.1.

The buckling length depends on the support conditions of the built-up member

for out-of-plane buckling. At the ends of the member, the supports are




6 - 12

generally considered as pinned. However intermediate lateral restraints may be

provided.

3.2.4 Step 4: Maximum shear force

The verification of the web members of a built-up column with pinned ends is

performed for the end panel by taking into account the shear force as described below.

For a built-up member subject to a compressive axial force only, the expression

for the shear force is:

L

MV Ed

Ed

where:

MEd is the bending moment as calculated in Step 2, with: 0I

Ed M

For a built-up member subject to a uniformly distributed load, the expression

for the shear force is:

L

MV Ed

Ed 4

where:

MEd is the maximum bending moment due to the distributed load.

Built-up columns are often subjected to a combination of a compressive axial

force NEd and a uniformly distributed load. Thus the coefficient varies between

π/L and 4/L. As a simplification, the shear force may be calculated by linear interpolation:

Ed

EdEd

EdEd )4(4

1M

MNe

Ne

LV

Io

o

where:

MEd is the maximum bending moment as calculated in Step 2. The bending

moment I

EdM is the maximum moment due to the distributed load.

3.2.5 Step 5: Buckling resistance of the web members in compression

Maximum compressive axial force

The maximum axial force NEd in the web members adjacent to the ends is

derived from the shear force VEd.

Classifi cation of the web members in compression

The cross-section of the web member is classified according to EN 1993-1-1

Table 5.2.

Buckling resistance

The buckling verification of the web members should be performed for

buckling about the weak axis of the cross-section, using the following criterion:




6 - 13

1Rd b,

Edch, N

N

where, N b,Rd is the design buckling resistance of the web member about the

weak axis of the cross-section, calculated according to EN 1993-1-1 § 6.3.1.

Information about the buckling length of web members is given in Section 3.4.

3.2.6 Step 6: Resistance of the web members in tension

The resistance of the cross-section of the web members should be verified

according to EN 1993-1-1 § 6.2.3 for the tensile axial force which is derived

from the maximum shear force VEd as described in Step 3.

3.2.7 Step 7: Resistance of the diagonal-to-chord connections

The resistance of the connections between the web members and the chords has

to be verified according to EN 1993-1-8[2]. This verification depends on the

details of the connection: bolted connection or welded connection. This

verification should be performed using the internal forces calculated in the

previous steps.

The worked example in Appendix A includes the detailed verification of a

welded connection.

3.2.8 Flowchart

Step 2 : In-plane buckling resistanceof the chords

Effective second moment of area I eff

LoadsULS load combination

Maximum compression force in the chord N ch

Section propertiesof the chords

Section propertiesof the web members

Global dimensionsOf the built-up member

Start

End

Shear stiffness Sv

Initial bow imperfection e0

Step 3: Out-of-plane buckling resistance

of the chords

Step 4: Maximum shear force V Ed

Step 5 : Buckling resistance of the web membersin compression

Step 7 : Design of the web members-to-chordconnections

Step 6 : Resistance of the web membersIn tension

Step 1: Maximum compression axial forcein the chords

EN 1993-1-1 §6.4.1(6)

EN 1993-1-1 §6.4.1(1)

EN 1993-1-1 Figure 6.9

EN 1993-1-1 6.4.2.1(4)

EN 1993-1-1 §6.4.2.1(2)and §6.3.1

EN 1993-1-1 §6.3.1

EN 1993-1-1 §6.4.1(7)

EN 1993-1-1 §6.3.1

EN 1993-1-1 §6.2.3

EN 1993-1-8

Figure 3.2 Flowchart of the design methodology for laced built-up columns




6 - 14

3.3 Design methodology for battened bui lt-upcolumns

3.3.1 Step 1: Maximum compressive axial force in the chords

Effective second moment of area

The effective second moment of area is calculated using the following

expression (EN 1993-1-1 § 6.4.3.1(3)):

chch

2

0eff 25,0 IAhI

where:

h0 is the distance between the centroids of chords

Ach is the cross-sectional area of one chord

Ich is the in-plane second moment of area of one chord

is the efficiency factor according to Table 3.2.

Table 3.2 Effi ciency factor (EN 1993-1-1 Table 6.8)

Criterion Efficiency factor

≥ 150 0

75 < < 150 2 – /75

≤ 75 1,0

where:0i

L

ch

10

2 A

I i chch

20 25,0 I AhI t

Shear stif fness

For the stability verification of a battened built-up column, the elastic

deformations of the battens and the chords must be considered in order to

derive the shear stiffness Sv using the following expression (EN 1993-1-1

§ 6.4.3.1(2)):

²

²π2

21²

24 ch

0

b

ch

ch

a

EI

a

h

nI

Ia

EISv

But Sv should not be taken greater than²

²π2 ch

a

EI

where:

a is the distance between the battens

n is the number of planes of battens

I b is the in-plane second moment of area of one batten.




6 - 15

V Ed a/2

a/2

h0

a/2

V Ed a/2

V Ed a/4 V Ed a/4

Bending moment diagram

V Ed a/h0

a/2

h0

a/2

V Ed/2

V Ed/2 V Ed/2

V Ed/2

V Ed a/h0

Shear forces

Figure 3.3 Bending moments and shear forces in a panel of a battened built-up column

Initial bow imperfection

The initial bow imperfection e0 is:

e0 =L/500

where:

L is the length of the built-up member

Maximum axial compressive force in the chords

The maximum axial compression Nch,Ed in the chords is calculated from the

expression given in 3.2.1.

3.3.2 Step 2: In-plane buckling resistance of a chord

Classification of the cross-section of the chord

The cross-section of the chord is classified according to EN 1993-1-1

Table 5.2.

Buckl ing resistance of a chord about z-z axis

The resistance of the chord has to be verified for bending and axialcompression and for buckling in the plane of the built-up member, i.e. about

the weak axis of the cross-section of the chord (z-z axis), according to




6 - 16

EN 1993-1-1 § 6.3.3. Depending on the geometry of the battened built-up

member, the verifications should be performed for different segments of the

chord:

For an end panel with the maximum shear force and thus the maximum

local bending moment

For a panel located at mid-height where the compression axial force may be

maximum in the chord.

3.3.3 Step 3: Out-of-plane buckling resistance of the chords

The out-of-plane buckling resistance is verified using the following criterion:

1Rdy, b,

Edch, N

N

where:N b,y,Rd is the design buckling resistance of the chord about the strong axis

of the cross-section, calculated according to EN 1993-1-1 § 6.3.1.

The buckling length depends on the support conditions of the built-up member

for out-of-plane buckling. At the ends of the member, the supports are

generally considered as pinned. However intermediate lateral restraints may be

provided.

3.3.4 Step 4: Shear force

The shear force VEd is calculated from the maximum bending moment as for a

laced built-up member, according to §3.2.4 of this guide.

3.3.5 Step 5: Resistance of the battens

As shown in Figure 3.3, the battens should be designed to resist the shear force:

0

Edh

aV

And the bending moment:

2

EdEd

aVM

The cross-section classification should be determined according to

EN 1993-1-1 Table 5.2, for pure bending. The section resistance should be

verified using the appropriate criteria given EN 1993-1-1 § 6.2.

3.3.6 Step 5: Resistance of the batten-to-chord connections

The resistance of the connections between the battens and the chords has to be

verified according to EN 1993-1-8. This verification depends on the details of

the connection: bolted connection or welded connection. This verification is

performed using the internal forces calculated in the previous steps.




6 - 17

3.3.7 Flowchart

Step 2 : In-plane buckling resistanceof the chords (M-N interaction)

Effective second moment of area I eff

LoadsULS load combination

Maximum compression force in the chord N ch

Section propertiesof the chords

Section propertiesof the battens

Global dimensionsOf the built-up member

Start

End

Shear stiffness Sv

Initial bow imperfection e0

Step 3: Out-of-plane buckling resistanceof the chords

Step 4: Maximum shear force V Ed

Step 5 : Section resistance of the battens

Step 6 : Design of the batten-to-chord connections

Step 1: Maximum compression axial forcein the chords

EN 1993-1-1 §6.4.1(6)

EN 1993-1-1 §6.4.1(1)

EN 1993-1-1 §6.4.3.1(2)

EN 1993-1-1 §6.4.3.1(3)

EN 1993-1-1 §6.3.3

EN 1993-1-1 §6.3.1

EN 1993-1-1 §6.4.1(7)

EN 1993-1-1 §6.2

EN 1993-1-8

Figure 3.4 Flowchart of the design methodology for battened built-up

columns

3.4 Buckling length3.4.1 Laced compression members

Chords

According to EN 1993-1-1 Annex BB, the buckling length Lcr of a rolled I or H

section chord member of built-up columns is taken as 0,9L for in-plane

buckling and 1,0L for out-of-plane buckling. These values may be reduced if it

is justified through detailed analysis.

L is the distance in a given plane between two adjacent points at which a

member is braced against displacement in this plane, or between one such point

and the end of the member.

Web members

Angles are mostly used as web members.

Provided that the chords supply appropriate end restraint to web members in

compression made of angles and the end connections supply appropriate fixity

(at least 2 bolts if bolted), the buckling length Lcr for in-plane buckling is takenas 0,9L, where L is the system length between joints.




6 - 18

When only one bolt is used for end connections of angle web members, the

eccentricity should be taken into account and the buckling length Lcr is taken

equal to the system length L.

The effective slenderness ratio eff of angle web members is given in

EN 1993-1-1 § BB.1.2 as follows:

7,035,0eff

where:

is the non-dimensional slenderness defined in EN 1993-1-1 § 6.3.

For sections other than angles, the web members may be designed for in-plane

buckling using a buckling length smaller than the system length and using the

non-dimensional slenderness as defined in EN 1993-1-1 § 6.3, provided that

the chords provide appropriate end restraint and the end connections provide

appropriate fixity (at least 2 bolts if bolted). In practice, the buckling length Lcr of a rolled profile is equal to the distance between joints for in-plane buckling

and for out-of-plane buckling.

3.4.2 Battened compression members

For simplicity, any potential restraint at the ends of the columns is neglected

and the buckling length of the chords may be taken as the system length.




6 - 19

REFERENCES

1 EN 1993-1-1:2005 Eurocode 3 Design of Steel structures. General rules and rules for

buildings

2 EN 1993-1-8:2005 Eurocode 3 Design of Steel structures. Design of joints




6 - 20




6 - 21

APPENDIX A

Worked Example: Design of a laced built-up

column



6 - 22

APPENDIX A. Worked Example: Design of alaced built -up column

1 of 12

Made by DC Date 02/2009Calculation sheet

Checked by AB Date 03/2009

1. IntroductionThis worked example deals with the verification of a typical built-up column

under compressive axial force and bending moment. The calculations are

carried out according to EN 1993-1-1. No National Annex is considered and

the recommended values of EN 1993-1-1 are used in the calculations.

The calculations are performed according to the design methodology given in

Section 3.2 of this guide.

2. Description

The geometry of the built-up column is described in Figure A.1 and inFigure A.2. For the most unfavourable ULS combination of actions, an axial

force and a bending moment about the strong axis of the compound section

are applied at the top of the column.

1 Lateral restraints

Figure A.1 Design model

The built-up column is restrained against out-of-plane buckling at both ends

and at mid-height.



Title APPENDIX A. Worked Example: Design of a laced built-up column 2 of 12

6 - 23

1 Chords HEA 200

2 Posts Angles 90 9

3 Diagonals Angles 80 8

y y

z

z

Figure A.2 Geometry of the built-up column

Section properties

Note that the y-y axis and the z-z axis refer to the strong axis and the weak

axis respectively, of the cross-section of each component.Chords: HEA 220 – S355

ch = 64,3 cm2

iy = 9,17 cm iz = 5,51 cm

Diagonals: Equal angles L 90 × 90 × 9 – S355

Ad = 15,52 cm2

iy = iz = 2,73 cm iu = 3,44 cm iv = 1,75 cm

Posts: Equal angles L 80 × 80 × 8 – S355

Av = 12,27 cm2

iy = iz = 2,43 cm iu = 3,06 cm iv = 1,56 cm




6 - 24

3. Step 1: Maximum compressive axial forcein the chords

3.1. Effective second moment of area

The effective second moment of area of the built-up section about the strong

axis is calculated using the following expression:

Ieff = 0,5 h02

Ach

where:

Ach is the section area of a chord

h0 is the distance between the centroids of the chords

EN 1993-1-1

§ 6.4.2.1

The value of the effective second moment of area is:

Ieff = 0,5 × 802

× 64,3 = 205800 cm4

3.2. Shear sti ffness

For N-shaped arrangement of lacings, the expression of shear stiffness is:

3

v

3

0d3

2

0dv

1dA

hAd

ahnEAS

where:

d = 2222

0 25,18,0 ah = 1,48 m

EN 1993-1-1

Figure 6.9

n is the number of planes of lacings (n= 2)

Ad is the section area of the diagonals

Av is the section area of the posts.

Therefore:

3

3

3

3

2

v 10

14801227800155211480

800125015522100002

S

Sv = 134100 kN

3.3. Initial bow imperfection

The initial bow imperfection is taken equal to:

e0 = L/500 = 10000/500 = 20 mm

EN 1993-1-1

§ 6.4.1(1)




6 - 25

3.4. Maximum axial compressive force in the chords

The maximum compressive axial force in the chords, Nch,Ed, is determined at

mid height of the built-up column as follows:

Nch,Ed =eff

ch0EdEd

22 IAhMN EN 1993-1-1

§ 6.4.1(6)

where:

MEd =

v

Ed

cr

Ed

I

Ed0Ed

1S

N

N

N

MeN

Ncr is the effective critical axial force of the built up member:

kN426501010000

10205800210000²²

² 32

4

eff cr

LEIN

The maximum bending moment, including the bow imperfection and the

second order effects is:

MEd = kNm4,481

134100

900

42650

9001

45002,0900

In the most compressed chord, the axial force is:

Nch,Ed = kN1052102058002

1034,648,04,4812

900 8

4

4. Step 2: In-plane buckling resistance of thechord

4.1. Classif ication of the cross-section of the chord

= 0,81 for steel grade S355

Flange slenderness: c/tf = 88,5 / 11 = 8,05 < 10 = 8,10 Class 2

Web slenderness: c/tw = 152 / 7 = 21,7 < 33 = 26,73 Class 1

Therefore the cross-section is Class 2 for pure compression.

4.2. Buckling resistance of a chord

The buckling resistance of the most compressed chord is verifed according to

EN 1993-1-1 § 6.3.1 for buckling about the weak axis of the cross-section,

i.e. about the z-z axis.

The buckling length of a hot-rolled H-section member can be taken equal to

0,9 a for in-plane buckling, where a is the system length between two nodesof the built-up column.




6 - 26

Buckling length of chords:

Lcr,z = 0,9 a= 0,9 × 1,25 = 1,125 mEN 1993-1-1

BB.1.1(2)B

The slenderness is:

z

zcr,

zi

L

where

iz is the radius of gyration of the gross cross-section, about the weak

axis.

therefore: 42,201,55

1125z

9,93y

1 f E With: = 0,81 for steel grade S355

06,7681,09,931

The non-dimensional slenderness is:

268,006,76

42,20

1

zz

Buckling curve c for buckling about the weak axis, since:

Steel grade S355

h/b< 1,2

tf < 100 mm

The imperfection factor is: z = 0,49

EN 1993-1-1

Table 6.2

The reduction factor z

can be calculated from the following expressions:

553,0268,02,0268,049,015,02,015,022

zzzz

965,0268,0553,0553,0

11222

z2

zz

z

EN 1993-1-1

§ 6.3.1.2(1)

The design buckling resistance is equal to:

kN2203100,1

3556430965,0 3

1M

ychz

Rdz, b,

f AN

The resistance criterion is:

1477,02203

1052

Rdz, b,

Edch, N

NOK




6 - 27

5. Step 3: Out-of-plane buckling resistance of the chords

The built-up column is pinned at both ends and is laterally supported at mid-

height. Therefore the buckling length for buckling about the strong axis of thechords is taken equal to:

Lcr,y = L/2 =10000/2 = 5000 mm

The slenderness is:

y

ycr,

yi

L

where

iy is the radius of gyration of the gross cross-section, about the strong

axis.

Therefore:

53,547,91

5000

y

ycr,

y i

L

06,769,931

The non-dimensional slenderness is:

717,006,76

53,54

1

yy

Buckling curve b for buckling about the strong axis, since:

Steel grade S355

h/b< 1,2

tf < 100 mm

The imperfection factor is: y = 0,34

The reduction factor y

can be calculated from the following expressions:

845,0717,02,0717,034,015,02,015,0 22

yyyy

774,0717,0845,0845,0

11

222

y2

yy

y

EN 1993-1-1

§ 6.3.1.2(1)

The design buckling resistance is equal to:

kN1767100,1

3556430774,0 3

1M

ychy

Rdy, b,

f AN


1595,01767

1052

Rdy, b,

Edch, NN OK




6 - 28

6. Step 4: Maximum shear forceThe maximum compressive axial force is obtained in the diagonals of the end

panels of the built-up column. It depends on the shear force in this panel. The

shear force can be assessed by the following expression:

II

IM

MNe

Ne

LV Ed

EdEdo

EdoEd )4(4

1

where:

L = 10 m

e0 = 0,02 m

NEd = 900 kN

I

Ed

M = 450 kNm

II

EdM = 482 kNm

Therefore:

VEd =

45090002,0

90002,0)4(4

10

1 482 = 191,2 kN

7. Step 5: Buckling resistance of the web

members in compressive7.1. Diagonals

7.1.1. Maximum compression axial force

The expression of the compression axial force Nd,Ed in a diagonal is derived

from the shear force as follows:

0

EdEdEdd,

cos

nh

dV

n

VN

where:

h0 = 800 mm

d = 1480 mm

n is the number of plans of lacings: n= 2

then:

kN86,1768002

14802,191Edd,

N




6 - 29

7.1.2. Classification of a diagonal in compression

h/t = 90 / 9 = 10 < 15 = 12,15

(b+h) / (2t) = (90+90) / (2 × 9) = 10 > 11,5 = 9,31 Class 4

Although the cross-section is Class 4, according to EN 1993-1-1 Table 5.2

Sheet 3, the calculation of the effective section area leads to no reduction. The

section area is therefore fully effective and the calculation is the same as for a

Class 3 Section.

EN 1993-1-1

Table 5.2

Sheet 3

7.1.3. Buckling resistance of a diagonal

The non dimensional slenderness can be calculated according to EN 1993-1-1

§ BB.1.2 in so far as the diagonals are welded at both ends and the chords are

stiff enough to ensure that the ends are clamped.

Slenderness about the weakest axis:

57,845,17

1480

v

v

i

d

Non dimensional slenderness

112,181,09,93

57,84

9,93v

Effective non dimensional slenderness

128,1112,17,035,07,035,0 vveff,

EN 1993-1-1

§ BB.1.2

Buckling curve b is used for the determination of the reduction factor:

v = 0,34

Therefore:

294,1128,12,0128,134,015,02,015,0 22

veff,veff,v

EN 1993-1-1§ 6.3.1

519,0128,1294,1294,1

11

222

veff,2

vv

v

The design buckling resistance of a compression member is equal to:

kN9,285100,1

3551552519,0 3

1M

ydv

Rdd,- b

f AN


162,09,285

8,1761

Rdd,- b

Edd, N

NOK




6 - 30

7.2. Posts

7.2.1. Maximum compressive axial force

The maximum compressive axial force is:

Nh,Ed = VEd = 191,2 kN

7.2.2. Classification of the cross-section

h/t= 80 / 8 = 10 < 15 = 12,15

(b+h) / (2t) = (80+80) / (2 × 8) = 10 > 11,5 = 9,31 Class 4

Although the cross-section is Class 4, according to EN 1993-1-1 Table 5.2

Sheet 3, the calculation of the effective section area leads to no reduction. The

section area is therefore fully effective and the calculation is the same as for a

Class 3 section.

EN 1993-1-1Table 5.2Sheet 3

7.2.3. Buckling resistance

The buckling length is equal to:

Lcr = h0 = 800 mm

Slenderness about the weakest axis:

28,516,15

800

v

yh,

v i

L

Non dimensional slenderness:

674,081,09,93

28,51

9,93

vv

Effective non dimensional slenderness:

822,0674,07,035,07,035,0 vveff,

EN 1993-1-1

§ BB.1.2

The buckling curve b is used for the determination of the reduction factor:

= 0,34

Therefore:

943,0²822,02,0822,034,015,02,015,02

veff,veff, v

712,0822,0943,0943,0

11

222

veff,2

vv

v

The design buckling resistance of a compression member is equal to:

kN310100,1

3551227712,0 3

1M

yhv

Rd b,

f AN




6 - 31


162,0310

2,191

Rd b,

Edh, N

NOK

8. Step 6: Resistance of the web members intension

It is necessary to verify the resistance of the diagonals in tension, even if this

situation is generally less critical than compression.

The verification of these members includes the verification of the resistance

of the cross-section and the verification of the resistance of the net section for

bolted connections.

Maximum design value of the tensile axial force:

Nt,Ed = 176,8 kN


0,1 Rdt,

Edt, N

N

EN 1993-1-1§6.2.3

The design tension resistance Nt,Rd is taken as the design plastic resistance of

the gross cross-section:

kN551100,1

3551552 3

M

yd

Rd pl,Rdt,

0

f A

NN


0,132,00,551

8,176

Rdt,

Ed N

NOK




6 - 32

9. Step 7: Resistance of the diagonal-to-chord welded connection

The diagonals (L90 90 9) are welded to the chord (HEA 220) by fillet

welds, see Figure A.3.

L90x90x9

26

64

3

150

HEA 220

N Ed

Figure A.3 Welded connection of a diagonal to the chord

Throat thickness: a = 3 mm

Effective longitudinal length of the fillet weld: leff-L = 150 mm

Effective transverse length of the fillet weld: leff-T = 90 mmAxial force in the diagonal: Nd,Ed = 176,8 kN

The design resistance of a fillet weld is determined using the simplified

method given in EN 1993-1-8 § 4.5.3.3.

At every point along the length of the fillet weld, the resultant of all the forces

per unit length transmitted by the weld should satisfy the following criterion:

Rdw,Edw, FF

where:

Fw,Ed is the design value of the force per unit length

Fw,Rd is the design weld resistance per unit length

The design resistance is independent of the orientation of the weld throat

plane and it is determined from:

Fw,Rd = f vw,d a

where:

f vw,d is the design shear strength of the weld

2M

udvw,

3/

w

f f

EN1993-1-8

§ 4.5.3.3




6 - 33

f u is the ultimate tensile strength of the weaker part:

f u = 510 N/mm2

w is the appropriate correlation factor:

w = 0,9 for steel grade S355

M2 = 1,25

EN 1993-1-1

Table 3.1

EN1993-1-8

Table 4.1

therefore:

N/mm3,453

901502

176800

N/mm2,78557,261

N/mm7,26125,19,0

3/5103/

eff

Edd,

Edw,

dvw,Rdw,

2

2Mw

udvw,

l

NF

af F

f f

Therefore:

Fw,Ed = 453,3 N/mm2 < Fw,Rd =785,2 N/mm2 OK

The minimum throat thickness amin = 3 mm is acceptable.

To prevent corrosion, the diagonal may be welded all around in one pass

(a= 3 mm).

To account for eccentricity a 5 mm (2 passes) throat fillet weld is

recommended on the unconnected leg side, as shown in Figure A.4.

a = 5 mm

a = 3 mm

Figure A.4 Throat thickness of the weld fillets





Part 7: Fire Engineering









7 - ii




7 - iii

FOREWORD

This publication is the seventh part of the design guide, Single-Storey Steel Buildings.




Part 3: Actions



















7 - iv




7 - v

ContentsPage No

FOREWORD iii

SUMMARY vi

1 INTRODUCTION 1

2 FIRE RISKS IN SINGLE-STOREY BUILDINGS 2 2.1 Fire safety objectives 2 2.2 Fire risk analysis 2 2.3 Main requirements of current fire regulations 3

3 PRACTICAL FIRE ENGINEERING OPTIONS IN THE EUROCODES 6 3.1 Current design approaches 6 3.2 Fire analysis 7 3.3 Heat transfer analysis 8

3.4 Structural analysis 8

4 GUIDANCE ON APPROPRIATE FIRE ENGINEERING SOLUTIONS 10 4.1 Field of application of different design methods 10 4.2 Choice of optimum design approach 11

5 DIRECT USE OF SIMPLE ENGINEERING OPTIONS FOR USE BY NONSPECIALISTS 12 5.1 Fire models 12 5.2 Thermal Models 16 5.3 Structural Models 21 5.4 Specific design rules for single-storey buildings 31

5.5 Simplified design methods 33 5.6 Design recommendations 37

6 GUIDANCE ON THE USE OF MORE ADVANCED SOLUTIONS 47 6.1 Fire models 47 6.2 Thermal Models 50 6.3 Structural models 51

REFERENCES 56

APPENDIX A German fire safety procedure for single-storey industrial andcommercial buildings 57




7 - vi

SUMMARY

This document provides guidance for the fire design of single-storey steel building

structures. It contains detailed information to allow engineers and designers to be more

familiar with the current design approaches and calculation models, which can be

applied not only to meet the prescriptive requirements but also to develop the performance-based fire safety design. The design methods introduced in the guide,

ranging from simple design rules to more sophisticated calculation models, are derived

from EN 1993-1-2 and 1994-1-2. They cover both steel and composite structures

(unprotected or protected). In addition, some specific design rules are given, allowing

simple verification of whether the behaviour of the steel structure of single-storey

industrial buildings in fire situation fulfils the safety objectives on the basis of

performance-based requirement.




7 - 1

1 INTRODUCTION

Due to the particularities of single-storey buildings, the life safety objective in

case of fire can be met easily without onerous fire resistance requirement for the structure. However, other safety objectives have to be taken into account if

the collapse of these buildings or a part of them may be accepted. In

consequence, many European fire safety building regulations are moving

toward acceptance of alternative fire safety engineering designs. Prescriptive

rules can then be replaced with performance based requirements, such as

adequate fire behaviour of the structure, aimed at satisfying fire safety

objectives that include life safety of people (occupants and fire-fighters),

protection of environment, property protection and business continuity.

Benefits and successful application of the performance-based approach to

building fire safety designs have already been well demonstrated for single-

storey buildings, especially where fire resistance was required, allowing insome cases more innovative, cost effective and safer solutions to be adopted.

To help the structural fire design of buildings, a new set of European Standards

has been developed, the Eurocodes. The Parts of the Eurocodes that are

relevant to the fire design of single-storey building consist of EN 1991-1-2[1]

(which includes principal concepts and rules necessary for describing thermal

and mechanical actions on structures exposed to fire) and Parts of material –

specific Eurocodes dealing with the fire design of structures, such as

EN 1993-1-2,[2], related to steel structures and EN 1994-1-2[3] related to

composite steel and concrete structures.

The fire parts of Eurocodes provide at present a wide range of calculation

methods. They allow engineers to follow either a prescriptive approach to meet

the fire safety requirements, as specified in national building regulations, or to

carry out on the basis of performance-based rules, a fire safety engineering

design that involves in general more complex computational analysis and

provides more accurate answers to fire safety objectives.

The present guide provides an overview of the current design methods

available for evaluating the fire performance of single-storey buildings

composed of either steel or composite structure as well as their application

fields. Simple calculations methods, easy to use, and more advancedcalculations models are dealt with separately. Moreover, to allow quick

assessment, simple design rules are given to assess quickly whether the

structural behaviour of steel structures of storage and industrial buildings fulfils

the fire safety objectives required by the fire safety regulations for industrial

buildings.

This guide aims also to help the engineer to understand more clearly the

different calculation methodologies and to carry out the structural fire design of

single-storey building according to the Eurocodes, from a relatively simple

analysis of single members under standard fire conditions to a more complex

analysis under real fire conditions.




7 - 2

2 FIRE RISKS IN SINGLE-STOREY BUILDINGS

2.1 Fire safety objectives

The primary objective of most fire safety regulations is to ensure the protectionof life (building occupants and fire fighters), environment and to some extent

property (building contents and building itself). Through a lot of measures

including a combination of active and passive fire protection systems, the

objectives are:

To reduce and prevent the incidence of fire by controlling fire hazards in

the building.

To provide safe escape routes for evacuation of building occupants.

To prevent fire spread from the fire compartment to others parts of the

building and to neighbouring buildings.

To ensure that the building remains structurally stable for a period of time

sufficient to evacuate the occupants and for the fire-fighters to rescue

occupants, if necessary.

2.2 Fire risk analysisSingle-storey buildings used as factories, warehouses or commercial centres

constitute a very common type of steel construction today. In the specific case

of warehouses, according to the storage arrangement (including free standing

storage, palletised rack storage, post-pallet storage or storage with solid or slatted shelves) and the combustibility of materials being stored, fire may

develop very quickly and then might endanger occupants long enough before

the structural collapse of the building. Indeed, fire growth may be extremely

important, as the upward flame propagation is usually very rapid. Vertical and

horizontal shafts formed between adjacent pallets and racking behave as

chimneys, which increase the spread of flames up to the roof. The smoke

quickly forms a hot layer under the roof and then descends progressively with

fire development. Obviously, the rate at which this occurs varies according to

the combustible contents and the building arrangement. In unventilated

conditions, single-storey buildings can become smoke-logged in few minutes.

Although the smoke is largely made up of ‘entrained’ air, it contains enoughtoxic substances and asphyxiates to incapacitate or kill within minutes people

exposed to them. Moreover, the hot smoke layer will also radiate high heat flux

to people escaping from fire area. A hot gas layer at 500°C leads to a heat flux

of about 20 kW/m² (corresponding to the radiant energy emitted by a

blackbody at the temperature of 500°C) and, under such thermal conditions,

skin burn will occur after only a few seconds4. Generally, it is agreed that the

tenability threshold is 2.5 kW/m2, which is much lower than heat flux needed

to lead to the failure of structural members. Consequently, buildings will

survive longer than occupants and the structural collapse of steel structures of

single-storey buildings generally does not provide additional threat to people

escaping from the fire area.




7 - 3

Regarding fire service operations, it is commonly accepted that fire-fighters

should not enter a single-storey building because of fast fire growth. Usually

they are forced to fight the fire from outside, covering neighbouring walls with

water. Hazard in this case for fire-fighters is then reduced to zero in the event

of structural collapse since it occurs at a level of temperature at which fire-

fighters can not withstand (provided that the progressive collapse, in the caseof compartmented buildings, and the collapse of the structure toward outside

do not occur [5,6]). In the event of, at the beginning of fire, they need to enter

within the building to rescue people, they cannot last within the building after

the heat flux is more than 7 kW/m², which is also very far for the risk of

collapse of the structure.

For these reasons, an increase of the intrinsic fire resistance of single-storey

buildings is unnecessary. However, the overall stability of the structure and the

stability of fire walls need to be accurately considered, to avoid any

progressive collapse. A single-storey building undergoes progressive collapse

when local failure of the heated part of the structure leads for the failure of adjoining cold structures. In addition, to provide a safe situation to fire-fighters

located around the building, the structure of single-storey building (including

façade elements) must collapse towards the inside of the building.

Many National Regulations have taken into account previous remarks for

industrial single-storey buildings as well as for public buildings by not

requiring any fire resistance rating for such works but introducing specific

safety requirements in terms of overall structural behaviour and concentrating

requirements on egress facilities and early fire detection and/or suppression.

With regards to other single-storey buildings with relatively low fire loads, therisk of life in the event of fire is reduced as egress of occupants and fire-ground

operations are straightforward.

2.3 Main requirements of current fire regulations2.3.1 Fire resistance of structural members

Despite the comments above, fire resistance ratings are sometimes required for

the structure of single-storey buildings[7].

The fire resistance is expressed as the time during which a building elementcan withstand exposure to fire without losing its function (load-bearing

elements or separating element). Usually, building elements are classified

using three performance criterion:

The load bearing capacity, R, which is the ability for a load-bearing element

to resist a fire without losing its structural stability

The integrity, E , which is the ability of a separating element, when exposed

to fire on one side, to prevent the passage through it of flames and hot gases

The insulation, I , is the ability of a separating element, when exposed to fire

on one side, to restrict the temperature rise on the unexposed face below

specified levels (in general a average value of 140°C).




7 - 4

In prescriptive fire regulations, required fire resistance for a building element is

expressed in terms of the minimum period of time during which the building

element would function satisfactorily while subject to the standard fire.

When fire stability requirements are given for single-storey buildings, they

usually range from 15 minutes (R15) to 60 minutes (R60), depending on theoccupancy class of the building, the provision of sprinklers, the building height

and the compartment size.

2.3.2 Compartmentation and building separation

Single-storey building must be subdivided into compartments separated by fire

walls when the floor area of the building exceeds the allowed maximum

compartment size. Limits on the compartment size may be removed by fitting

the building with sprinklers.

The effects providing compartmentation on property loss is that direct damage

is confined to the content of the compartment in which the fire starts, reducingthe chances of the fire growing large. As regards the life safety, people in other

parts of the building can use escape routes to get out safely without being

exposed to the smoke or gases from the fire.

When considering fire walls between compartments, fire resistance is generally

in the range of REI 60 to REI 120.

Fire spread to neighbouring buildings also needs to be prevented. This is

achieved traditionally by sufficient separating distances or façade elements

with adequate fire resistance. In the French research project Flumilog, a design

method has been recently developed to assess the thermal radiant effects of fires in single-storey storage buildings. The method allows calculation of the

safe separating distances, taking into account the main characteristics of the

building, such as the building content, the type of façade elements and roof,

etc.

2.3.3 Fire suppression

Sprinklers may be required by national fire regulations. In addition to their

obvious effect in the reduction of the fire growth, their use leads usually to a

reduction of the fire resistance rating required for the structure. They allow also

larger fire compartment sizes.

2.3.4 Smoke control systems

National fire regulations may require that smoke control systems are

implemented in public buildings, storage building and industrial buildings in

order to facilitate escape, by minimising risks of smoke inhalation and injury

and to some extend to enable fire-fighters to better see the fire and therefore to

extinguish it more speedily and effectively. Smoke control systems help in

removing smoke from the fire area, and in limiting the spread of hot gas

beneath the roof, which increases the time for the compartment to become

smoke-logged, giving people more time to escape safely from the building.

This can be achieved by a combination of smoke exhaust systems (mechanicalor natural) and screens (which contain the smoke in specific areas).




7 - 5

2.3.5 Fire detection and fire alarms

Adequate measures are necessary for detecting any outbreak of fire and for

alerting the building occupants and the fire department of the occurrence of

fire. In small single-storey buildings where all exits are visible, it is likely that

any fire will be quickly detected by the occupants and a shout of ‘Fire!’ may be

sufficient. In larger single-storey buildings, a simple sounder such as a battery powered alarm or rotary bell may be adequate. In an industrial building, the

ambient noise has to be considered, to ensure that the alarm will be heard by

the occupants.

2.3.6 Egress facili ties

For safe evacuation, appropriate means of escape are needed, such as a proper

number and width of emergency exits and proper length, width and height of

passages and evacuation accesses. Escape routes in small single-storey

buildings generally lead directly to a safe location outside the building; they do

not normally require any special treatment. In larger buildings, where traveldistances are greater and where the fire is likely to make a single escape routes

unusable, an alternative means of escape may be necessary. Consideration of

disabled people must also be made




7 - 6

3 PRACTICAL FIRE ENGINEERING OPTIONSIN THE EUROCODES

3.1 Current design approachesUsing the fire parts of Eurocodes[8,9], single-storey buildings can be designed

using either the prescriptive approach or the performance-based approach

applying fire safety engineering principles[10].

The prescriptive approach is mostly applied to fulfil standard fire resistance

requirements usually prescribed in national fire regulations. It gives a safety

level that is relatively easy to achieve and implement. However it may be

conservative, in requiring the use of important passive fire protection to fulfil

the required fire resistance rating. This approach is usually carried out for the

design of relatively simple buildings and structures.

As an alternative or when allowed by national regulation, the performance-

based approach can allow to assess adequate measures to satisfy a set-out of

defined fire safety objectives, such as stated in paragraph 2.1, and the

corresponding performance criteria. Using structural fire engineering,

engineers can assess the necessary fire resistance to structure in order to avoid

the spread of fire and/or to prevent a premature structural collapse. As regards

the single-storey buildings, the main structure could be designed to remain

stable under fire exposure conditions long enough for the occupants to escape.

Such an approach takes into account the severity of fire exposure by

appropriate estimations of actual fire loads and fire development parameters,which may be calculated from the building activity.

The performance-based approach provides flexibility when selecting technical

solutions to meet the fire safety objectives, but usually requires the use of

sophisticated design tools. Engineers and designers using advanced

calculations models need to be properly educated in their use and in their

limitations. As fire safety engineering allows for highly efficient designs, with

little unassigned reserve capacity, an experienced user is required to ensure that

appropriate models are used.

Where national fire regulations authorise the performance-based approach,regulatory bodies may require that the fire design is checked by a third party.

The fire performance of a whole structure, or a part of it, is carried out by

following, for a given design fire scenario, three successive steps of structural

fire engineering[1].

Fire Analysis. To calculate the thermal actions/exposure - Fire models.

Thermal analysis. To determinate the heating rate and temperatures on

structural members - Thermal models.

Structural analysis. To calculate the mechanical response of structural

members- Structural models.




7 - 7

Available design methods to evaluate the fire performance of structure are

briefly described below. These methods range from simple hand calculations to

the use of sophisticated computer models. The overall complexity of the fire

safety design will depend on the assumptions and methods adopted to predict

each of the three design steps.

3.2 Fire analysisThe main objective of the fire modelling is the simulation of the fire

development and the prediction of thermal actions (gas temperature, heat flux)

on the structural members (in order to determinate, in a following step, the

temperature in the structural members).

Although common practice is to represent a fire by a standard fire curve,

structural fire design may be based on a design fire that provides more realistic

conditions in fire compartment. In this way, parameters such as the magnitude

of the fire load, the rate of heat release and the ventilation factor, which play an

important role in fire severity, are taken into account. Moreover, the

identification of relevant and realistic design fire scenarios is a crucial aspect of

the fire safety design. The design fire scenarios used for the analysis of a

building fire have to be deduced from all the possible fire scenarios. In most

buildings, the number of possible fire scenarios is infinite and need to be

reduced. Only ‘credible worst case’ fire scenarios will need to be studied.

When the design fire scenarios are chosen, a number of fire models are

available to assess the fire severity and calculate the corresponding thermal

actions

Different levels of fire models are relevant to the various stages of fire

development. When a fire is initiated, it is localised within a compartment and,

according to the characteristics of the compartment and of the fire load, it can

remain localised or becomes generalised to the whole compartment. In the case

of small compartments or compartments with small ventilation openings

relative to the size of the compartment, the fire develops into to a fully

engulfed fire.

Three levels of modelling are available to describe both localised and fully

generalised fires, as shown in Table 3.1.

Table 3.1 Levels of fire models

Levels of the model Localised fire Generalised fire

Simplified model Hasemi modelHeskestad model

Parametrical fires

Zone models 2 zone model 1-zone model

Field model CFD CFD

The simplified models are generally empirical models based on conventional

assumptions. The zone models take into account the main parameters

controlling the fire, but introduce simplified assumptions that limit the domain

of application. They would be used in simple easily defined compartment

geometries. The field models are more accurate but are rather complex for use




7 - 8

as a general design tool; they would be required in compartments with complex

geometries or with high and irregular ceilings.

Conditions of use will be briefly detailed in Chapter 6.

3.3 Heat transfer analysisOnce the thermal actions are calculated, the thermal transfer to the structural

elements has to be calculated. Thermal models, which will be used, should be

based on the acknowledged principles and assumptions of the theory of heat

transfer.

Different modelling can be used according to the assumptions and needs. In the

thermal models, there are the analytical rules allowing obtaining an estimation

of uniform temperature across-section, mainly for steel elements. There are

also advanced calculation methods based on either finite elements or the finite

difference method, allowing determination of the 2D or 3D temperaturedistribution in structural members (through the cross-section and along the

length). Advanced models can be applied for any type of structural member

analysis in fire design.

Thermal models will be briefly detailed in following chapters.

3.4 Structural analysisFrom the temperature fields previously obtained in the structural members and

from the combination of the mechanical actions loads in case of fire thestructural behaviour can be assessed following one of the three possible

approaches:

Member analysis, in which each member of the structure will be assessed

by considering it fully separated from other members. The connection

condition with other members will be replaced by appropriate boundary

conditions.

Analysis of parts of the structure, in which a part of the structure will be

directly taken into account in the assessment by using appropriate boundary

conditions to reflect its links with other parts of the structure

Global structural analysis, in which the whole structure will be used in the

assessment




7 - 9

Memberanalysis

Analysis of partof the structure Global structural

analysis

Figure 3.1 Different design approaches for mechanical response of structures in fire

Member analysis is easy to use particularly with simplified calculation methods

and therefore largely used under standard fire condition. The analysis of thewhole structure or its subassemblies considers at least several structural

members together, so that the interaction effect between them will be directly

dealt with. In this way, load redistribution from heated parts (weakened parts

inside fire compartment) to cold parts (stronger parts outside fire compartment)

can be taken into account in accurate way and global analysis provides

therefore a much better understanding of overall behaviour of structure under

fire condition.

According to the Eurocodes, three types of design methods can be used to

assess the mechanical behaviour of structures under fire situation in the

different design approaches explained above. Fire design can be carried out bymeans of:

A simple calculation method, based on predefined tabulated data, as given

in EN 1994-1-2[3]. This method is only applicable to steel and concrete

composite structures. The tables were evaluated by numerical models and

experiments for basic types of structures, such as slabs, beams and

columns, for certain time of fire resistance, for heating according to the

nominal fire curve and for defined level of loading. The tables are easy to

use and safe but cover only a limited range of section types.

Simple calculation models. This type of design method can be divided into

two different families. The first one is the critical temperature method

widely applied to steel structural member analysis. The second is the use of

simple mechanical models (verification in strength domain) developed for

both steel and composite structural member analysis. Models have been

developed for standard structural elements, e.g. slabs, beams, and columns.

Advanced calculation models. This kind of design method can be applied to

all types of structures and the models are, in general, based on either finite

element method or finite difference method. They should provide a realistic

analysis of structures. The results of the analysis are generally obtained in

terms of deformation of structure during the whole fire period.

Structural models will be briefly detailed in following chapters.




7 - 10

4 GUIDANCE ON APPROPRIATE FIRE ENGINEERING SOLUTIONS

4.1 Field of application of different design methodsThe following table shows the field of application of the available fire design

methods, considering either design according to prescriptive requirements

based on the standard fire or a performance-based fire design[11].

Table 4.1 Field of application of different design methods

Approach ToolsThermal actions

Thermal modelling

Structural modelling

Pre-engineered datafrom standard fire

tests (Data frommanufacturers)

Tabulated data fromEN 1994-1-2

EN 1994-1-2, §4.2

SteelEN 1993-1-2§4.2.5

SteelEN 1993-1-2§4.2.3 §4.2.4

Simplified calculationmodels given inEurocodes

Composite EN 1994-1-2 §4.3

Steel and composite

P r e s c r i p t i v e a p p r o a c h

( S t a n d a r d f i r e d e s i g n )

Advanced calculationmodels

Standard ISOcurve

EN 1991-1-2

FEA* or FDA** FEA*

Simplified calculationmodels

Fully engulfedfire (Parametricfire, standardISO curve***)

Localized fire

Steel

EN 1993-1-2§4.2.5

Steel

EN 1993-1-2§4.2.3 §4.2.4

Specific rulesbased on fully

engulfed fire§5.4

Steel and composite P e r f o r m a n c e b a

s e d a p p r o a c h

( n a t u r a l f i r e d e s i g n )

Advanced calculationmodels

Zone models

Field models FEA* or FDA** FEA*

*FEA : Finite element Analysis **FDA : Finite Difference analysis

*** Collapse of single-storey buildings usually occurs when the building structure (a part of it orthe whole structure) is fully engulfed in fire. In such fire condition, because the gas temperature

rise has no significant effect on the failure mode of the building structure, a performance-basedapproach referring to thermal actions based on standard fire curve is appropriate to investigatethe fire behaviour of single-storey buildings. This approach can be used to demonstrate thenon-progressive collapse and the failure inwards of the building structure.




7 - 11

4.2 Choice of optimum design approachThe choice of the design approach depends on the type of building (storage

building, industrial building, commercial building, etc.), the requirements

specified in the corresponding national fire regulation and the acceptance or

not by the regulatory authorities of applying a performance-based approach as

an alternative to prescriptive rules.

Some suggestions on the choice of fire design approach are given below.

With the diversity of requirement, the most important first step is to answer the

following:

What is the required fire resistance rating, if any?

Is it possible to carry-out a performance-based approach?

When a prescriptive approach is to be used (with reference to standard fire

design):

It may be appropriate to use simplified calculation models where low fire

resistance ratings (R15 or R30) are required for structural members

Advanced calculation models must be used where structural members are

not covered by the simplified calculation models. They can also be

employed with some economic benefits for steel structure where high fire

resistance ratings (higher than R60) are required, reducing the thickness of

fire protection on steel members.

Where the performance-based approach is accepted by the regulatory

authorities and structural stability is needed:

A performance-based approach is most likely to be beneficial where the

structure is unusual and may not be well covered by traditional prescriptive

methods

Localised fire protection may be needed, considering the overall behaviour

of the whole structure in a real fire, to ensure adequate life safety for the

building occupants and firemen.

National fire regulations may require the use of the performance based

approach for single-storey buildings with significant fire risks (high fire loads).

National fire regulations may allow a performance-based fire safety design to

refer to simple rules and design recommendations for single-storey buildings.

Such approaches are given in §5.4 and Appendix A. Other design guidance and

recommendations can be found in reference[12].

Active fire protection measures (installation of sprinklers, fire detectors, fire

alarms, smoke exhaust systems) and passive fire protection measures

(compartmentation, egress facilities, etc.) are usually implemented in buildings

in accordance with the requirements in fire national regulations.




7 - 12

5 DIRECT USE OF SIMPLE ENGINEERINGOPTIONS FOR USE BY NON SPECIALISTS

This chapter gives an overview of current easy-to-use ‘simple’ calculationdesign rules, for assessing the fire resistance of steel and composite steel and

concrete structural members.

Specific simple design rules and design recommendations to satisfy specific

safety requirements in terms of structural behaviour introduced recently in fire

safety regulations of many European countries for single-storey storage and

industrial buildings are given. It is noted that these methods are also applicable

to other type of single-storey buildings.

5.1 Fire models5.1.1 Nominal temperature-time curves

EN 1991-1-2[1] provides three standard fire curves, defining arbitrary hot gas

temperature-time relationships in which no physical parameters of the fire load

or fire compartment are taken into account. The most commonly used

relationships in building design and in regulation prescriptions is the standard

temperature-time curve (standard ISO fire) which represents a fully developed

compartment fire. The second curve, the external fire curve, is intended for

façade elements and the third curve is the hydrocarbon fire curve, representing

a fire with hydrocarbon or liquid type fuel.

The nominal temperature-time curves are defined as follows:

For standard temperature-time curve (standard ISO fire ):

)18(log34520 10 t g (1)

For the external fire curve:

20)313,0687,01(660 8,332,0g t t ee (2)

For the hydrocarbon fire curve:

20)675,0325,01(1080 5,2167,0g t t ee (3)

where:

θ g is the gas temperature in the fire compartment [°C]

t is the time [min]

It is important to note that the previous curves are reference curves. They do

not represent the real thermal effect of a fire. The temperatures given by these

curves always increase with time, without considering the limited fire load.

The standard fire resistance rating required for structural members (expressed

in terms of time) does not therefore indicate the actual time for which they willsurvive in a building fire.




7 - 13

5.1.2 Parametric fires

Parametric fire models provide a rather simple design method to estimate gas

temperature in fire compartment, taking into account in a simplified way the

main parameters that influence the fire development, such as the fire

compartment size, the fire load (corresponding to the mass of combustible

materials in the fire compartment), ventilation conditions (openings) andthermal properties (such as thermal conductivity and specific heat) of the

compartment walls and ceilings.

Like nominal temperature-time curves, parametric temperature-time curves

provide gas temperature-time relationships for design. They are based on the

hypothesis that the temperature is uniform in the compartment, which limits

their field of application to post-flashover fires (fires generalised to the whole

compartment) in compartments of reasonable dimensions. The predicted fire

curve comprises a heating phase represented by an exponential curve up to a

maximum temperature, followed by a linearly decreasing cooling phase to a

residual temperature that is usually the ambient temperature. The maximumtemperature and the corresponding fire duration are the two main parameters

affecting the fire behaviour of structural members. Consequently, they were

adopted as the governing parameters in the design formulae for the parametric

fires.

Such a model is given in Annex A of EN 1991-1-2. It is valid for

compartments up to 500 m² of floor area, without openings in the roof, and a

maximum compartment height of 4 m, for compartment linings with thermal

inertia between 100 and 2200 J/m2s1/2K, for an opening factor in the

range 0,02 to 0.20 and for compartments with mainly cellulosic type fire loads.

Due to these limitations, the model is mainly used for the office part of single-

storey buildings.




7 - 14

Time

g

max

t* max

heating

phase

cooling

phase

g=20+1325(1-0,324e-0,2t*-0,2e-1,7t*-0,427e-19t*)

with t*= t.C where t is the time (hours) and

)²1160/04.0/(]² b/O[R

Main parameters:

- Wall characteristics : thermal inertia c b

- Opening characteristics: opening factor tv

A/hAO

max= g (t*max) = g (tmax . ) (°C)

with tmax = max{ (0.2.10-3 qt,d / O). / O, tlim } (hours)

where tlim is a function of fire growth rate (according to building type):- tlim =25 min if slow fire growth rate

- tlim =20 min if medium fire growth rate,

- tlim =15 min if fast fire growth rate,

- qt,d is the design value of the load density [MJ/m²]

g = g (t*, t*max, x) (°C)

= max – 625.(t* - t*max.x) if t*max 0,5

= max – 250.(3- t*max).(t* - t*max.x) if 0,5 < t*max 2

= max – 250.(t* - t*max) if t*max > 2

with t*= t. t*max = (0.2.10-3 qt,d / O).

and x is a function of tmax as follows:

x = 1 if tmax > tlim x = tlim. / t*max if tmax = tlim

Figure 5.1 Parametric Fire (Annex A of EN 1991-1-2)

The inputs for the parametric fire curves are the design fire load density, the

fire growth rate, the ventilation conditions (described by the size and the

position of the openings) and the thermal properties (heat capacity, the density

and the conductivity) of walls to evaluate the heat losses which occur byconvection and radiation at the compartment boundaries. For the fire load

density, it is common practice in design to refer to the characteristic values

given in EN 1991-1-2.

Even though these parametric fire curves offer a significant improvement

compared to the standard “ISO-fire”, the parametric fires are not yet able to

provide a very accurate evaluation of the fire severity. Consequently, some

European countries recommend their use only for pre-design calculation.

5.1.3 Localised fire

EN 1991-1-2 provides simple approaches for determining thermal actions of

localised fires in Annex C. Two situations are distinguished according to the

height of the fire flame relative to the ceiling of the compartment: where the

flame is not impacting the ceiling (based on Heskestad’s method); and where

the flame is impacting the ceiling (based on Hasemi’s method).




7 - 15

Flame axis

L

z D

f

H

Z 0 = 1,02 D + 0,00524 Q2/5

z 0

Flame axis

L

z D

f

H

Z 0 = 1,02 D + 0,00524 Q2/5

z 0

Flame axis Lh

D

H

r

Flame axis Lh

D

H

r

The flame is not impacting the ceiling The flame is impacting the ceiling

Required data:- Rate of heat realase: Q (W)

- Distance fire Source-ceiling: H (m)- Diameter of the fire: D (m)

Results:

- Flame length Lf (m) :

Lf = -1,02 D + 0,0148 Q2/5

-Temperature (z) in the plume along

the symmetrical vertical axis:

(z) = 20 + 0,25 (0.8Q)2/3 (z-z0)-5/3

(z) 900°C

Results:

- Horizontal flame length Lh

- heat flux received by the fire exposed unit surface

area at the level of the ceiling at the distance r fromthe flame axis:

h =100000 if y 0,30

h =136300-121000y if 0,30 <y <1,0

h =15000 y -3,7

if y 1,0

with

'

'

h z H L

z H r y

where

r: is the distance from the flame axis to the

point where the thermal flux is calculated (m)

z: is the vertical position of the virtual heat

source (m)

D: is the diameter of the fire (m)

Figure 5.2 Localised Fires (Annex C of EN 1991-1-2)

For situations where the fire is not impacting the ceiling, a design formula is

given to calculate the temperature in the plume at heights along the vertical

flame axis. For situations where the fire is impacting the ceiling, some simple

steps are given to calculate the heat flux received by the fire-exposed surfaces

at the level of the ceiling.

These models are most often used to calculate thermal actions (expressed in

terms of heat flux resulting from a radiation part and a convection part) on

horizontal structural members, such as beams. At the present time, no method

is available for vertical steel members affected by a localised fire.

The input data are the rate of heat release (RHR), the distance between the fire

source and the ceiling, and the diameter of fire. The RHR is usually determined

by using EN 1991-1-2 section E.4.

These approaches are limited to cases where the diameter of fire D is less than

10 m and the rate of heat release of fire Q is less than 50 MW.




7 - 16

5.2 Thermal ModelsConsidering the high thermal conductivity of steel and the small thickness of

steel profiles commonly used in the construction, it is sufficiently accurate to

ignore thermal gradients within the cross-section of structural members and

assume a uniform temperature when uniformly heated.

Consequently, simple design equations can be used to predict the temperatures

of steel members that are fully exposed to fire or steel members that support a

concrete slab and are exposed on three sides. Similar rules exist for fire-

protected steel sections, although the thermal properties of the proposed

protection material are needed, which can be difficult to obtain.

For the composite steel-concrete members, strictly speaking, there are no

simplified models to estimate the evolution, as a function of time, of

temperature distribution through members. To simplify the design, information

on temperature distribution at current time of standard fire exposure (i.e. 30,

60, 90 and 120 minutes) is given in EN 1994-1-2.

5.2.1 Unprotected steel member

Heating of the unprotected steel members can be determined by means of the

simple analytical approach given in EN 1993-1-2. In this method, the

temperature rise depends on the thermal actions (expressed in terms of net heat

fluxes), the thermal properties of the steel and the section factor of the element

Am/V defined as the ratio between the surface area exposed to the heat flux Am

[m²/m] and the volume of the element by unit length V [m3/m]. The section

factors for some unprotected steel members are shown in Figure 5.3.

b

h

t t

t

Am/V=Perimeter exposed to fire/Cross-section area Am/V=1 / t Am/V=2 / t

Figure 5.3 Example of section factor for unpro tected steel members

Assuming an equivalent uniform temperature distribution in a cross-section,

the increase of temperature a,t in an unprotected steel member during a time

interval t may be determined from:

t hc

/V Ak

dnet,aa

mshta,

with t 5 s (4)

where:

shk is the correction factor for the shadow effect caused by local

shielding of radiant heat transfer due to shape of steel profile

aC is the specific heat of steel [J/kgK]




7 - 17

a is the unit mass of steel [kg/m3]

h dnet, is the net heat flux per unit area [W/m²]

Solving the incremental equation step by step gives the temperature

development of the steel element during the fire. In order to assure thenumerical convergence of the solution, some upper limit must be taken for the

time increment t. In EN 1993-1-2, it is suggested that the value of t should

not be taken as more than 5 seconds.

The thermal actions are determined by the net heat flux r net,h absorbed by the

steel member during the fire exposure. It is expressed in terms of the hot gas

temperature as the sum of two distinct fluxes: a convective component cnet,h

and a radiant component r hnet, .

Convective heat flux is expressed as:

)( mgccnet, h (5)

where:

c is the coefficient of heat transfer by convection [W/m²K] g is the gas temperature [°C]

m is the surface temperature of the member [°C] Radiant heat flux is given by:

)273)()273(( 4m

4r m0r net, h (6)

where:

is the configuration factor, including position and shape effect (<1)

m is the surface emissivity of the member

r is the radiation temperature of the fire environment [°C] ( r ≈ g)

m

is the surface temperature of the member [°C]

0 is the Stephan Boltzmann constant [= 5,67 10-8 W/m2 K 4]

According to EN 1991-1-2, for many practical cases the configuration factor

may be taken equal to unity. The coefficient of convection ( c ) varies from

25 W/m²K (standard fire conditions) to 50 W/m²K (hydrocarbon fire

conditions). The emissivity of carbon steel and composite steel and concrete

members may be taken as 7,0m .




7 - 18

For cross-section with a convex shape, such as hollow steel sections, fully

embedded in fire, the shadow effect does not play a role and it can be taken

that k sh = 1. Otherwise, the correction factor for the shadow effects k sh is given

by:

casesothersfor

actionsfirenominalunder sections-Ifor

/

]/[

/

]/[9,0

m

bm

m

bm

sh

V A

V A

V A

V A

k (7)

where:

bm ]/[ V A is the box value of the section factor [m -1].

Application of the EN 1993-1-2 calculation method with standard ISO fire

exposures of 15 and 30 minutes leads to the temperature curves illustrated in

Figure 5.4 and given in Table 5.1 as function of the section factor including

shadow effect ( Am/V )sh = k sh Am/V .

Figure 5.4 Temperature of unprotected steel members after 15 and 30 minutes of standard ISO fire exposure

0

100

200

300

400

500

600

700

800

900

0 50 100 150 200 250 300 350 400 450 500 ( Am / V ) sh= k sh ( Am / V ) (m

-1 )

Temperature (°C)

15 minutes

30 minutes

10




7 - 19

Table 5.1 Temperature of unprotected steel members after 15 and 30 minutes of standard ISO fire exposure

Steel temperature (°C) Steel temperature (°C)Sectionfactor

( Am / V )sh 15 min 30 min

Sectionfactor

( Am / V )sh 15 min 30 min

10 113 257 130 621 802

20 194 431 140 634 809

30 265 554 150 646 815

40 328 636 160 655 819

50 383 690 170 664 822

60 432 721 180 671 825

70 473 734 190 677 827

80 509 741 200 682 828

90 539 753 250 699 833

100 565 767 300 708 835

110 586 781 400 716 837

120 605 792 500 720 838

5.2.2 Protected steel member

EN 1993-1-2 also provides a simple design approach for insulated members

with passive fire protection materials. In such cases, the temperature rise

depends on the section factor A p/V for the steel member insulated by fire

protection material ( A p is the appropriate area of fire protection material per

unit length and V is volume of the steel member per unit length) and the

insulation characteristics. The insulating materials can be in form of profiled or

boxed systems, but this simple approach does not cover intumescent coatings.

Assuming uniform temperature distribution, the temperature increase a,t in

an insulated steel member during a time interval t may be determined from:

tg,10/

ta,tg, p

aa

p pta, 1e

3/1

1/

t

V

A

c

d (8)

with

V

Ad

c

c p p

aa

p p

(9)

where:

pd is the thickness of fire protection material [m]

pC is the specific heat of fire protection material [J/kgK]

p is the thermal conductivity of the fire protection material [W/mK]

p is the unit mass of the fire protection material [kg/m3]




7 - 20

g is the gas temperature [°C]

Figure 5.5 gives expressions to calculate the section factor of protected steel

members.

Am/V=(P-b) / As

b

Am/V=(2h+b) / As Am/V=2(2+b) / As Am/V=P / As

h

bb

h

P : perimeter ; As : cross-section area

Figure 5.5 Example of section factor for insu lated steel members

It is important to note that thermal characteristics of fire protection materials

are usually determined from fire tests performed under standard fire conditions.

Consequently, referring to thermal actions based on natural fires, the use of

Equation (8) for the fire design situation of protected steel members should be

handled with some caution. The calculation should be performed only if

appropriate data are available or if it can be shown that fire conditions have no

significant effects on thermal characteristics and integrity of fire protection

materials. Nevertheless, it is commonly assumed that thermal properties of an

insulation material can be used under natural fire conditions when the

temperatures of hot gases remain lower than the maximum temperature reached

during the standard fire test for the insulation material (For example, about1100°C for 4 hours of the standard temperature-time curve).

The material properties given in Table 5.2 may be used as a first approximation

to calculate heating of protected steel members. These average values are

derived from fire tests by material manufacturers.




7 - 21

Table 5.2 Average materials properties of main fire protection materials

Material Density

p [(kg/m3 ]

Conductivity

p [W/mK]

Specific heat

pC [J/kgK] )

Mineral fibre 300 0,12 1200

Vermiculite andcement

350 0,12 1200Sprays

perlite 350 0,12 1200

vermiculite (orperlite) and cement

550 0,12 1100High density

sprays vermiculite (or

perlite) and gypsum 650 0,12 1100

vermiculite (orperlite) and cement

800 0,2 1200

fibre-silicate or fibrecalcium-silicate

600 0,15 1200

fibre-cement 800 0,15 1200

Boards

gypsum board 800 0,2 1700

Compressedfibre boards

fibre-silicate,mineral, stone-wool

150 0,2 1200

5.3 Structural ModelsAccording to the Eurocodes, several simple design methods can be used to

assess the fire resistance of structures under fire conditions. The first one is the

critical temperature method widely applied to steel structural member analysis

and the second one is the simple mechanical models developed for both steel

and composite steel and concrete structural members.

It is important to remember that the design methods available for composite

members are only valid for the standard fire exposure. Moreover, design

methods given for columns should be only applied to members of braced

frames (where the column ends have no horizontal displacement).

5.3.1 Critical temperature method

The critical temperature is calculated by using applied mechanical actions,

design resistance in the normal temperature condition and the strength loss of

steel at elevated temperature. This critical temperature generally varies

between 500°C and 800°C. It can be obtained by calculation according to the

simple rules given in the EN 1993-1-2 or by referring to default values.

According to the critical temperature method, the fire resistance of a steel

member without instability effect is satisfied after a time t if the steel

temperature t,a does not exceed the critical temperature cr of the element:

cr t,a (10)




7 - 22

The critical temperature of the member can be calculated from the degree of

utilization 0 as follows:

48219674.0

1ln19,39

833.30

cr (11)

The degree of utilization 0 is obtained from:

d,0fi,

dfi,0

R

E (12)

where:

dfi, E is the design effect of actions for the fire design situation, according

to EN 1991-1-2

d,0fi,

R is the corresponding design resistance of the steel member, for the

fire design situation, at time t = 0 (at normal temperature) but with

safety factor fi,M in fire situation

The expression for θ cr can be used for all classes of section except the very

slender Class 4 sections, for which a single conservative critical temperature of

350°C should be used.

In principle, Expression (11) applies for members in pure bending, short

columns without buckling and members in tension, heated uniformly or with

slight temperature gradient. However, in situations of instability (slender

columns, unrestrained beams), the method becomes applicable by calculatingthe design resistance for the fire design situation at time t = 0 with a value of

the slenderness that takes into account temperature effects on the slenderness

of structural members. As a simplification, the slenderness in fire situations can

be taken as 3.1 (where is the non dimensional slenderness at

normal temperature).

As an alternative, to relation (11) nationally determined critical temperatures

can be given in the National Annex to EN 1993-1-2.

A simple conservative expression for 0 can also be used for tension members

and restrained beams (where lateral-torsional buckling is not a potential failure

mode):

21

M

fi,M

t,fi0

(13)

where:

t,fi is the load level at time t

fi,M is the relevant partial safety factor for fire situation ( 1fi,M )

0M is the partial safety factor at normal temperature ( 10M )




7 - 23

κ 1, κ 2 are the adaptation factors to take account a non-uniform

temperature distribution on steel member.

The load level at time t is defined as:

d

dfi,tfi,

R

E

(14)

where:


to EN 1991-1-2

d R is the ultimate resistance in room temperature

For a given fire duration t , assuming that cr t,a , the maximum value of

utilization level 0 of unprotected steel members to satisfy the required fire

resistance may be easily calculated from (11), as function of section factor including the shadow effect ( Am/V )sh. In this way, it may be assumed that fire

resistance of unprotected steel members is satisfied after a time t if:

max0 (15)

Maximum degrees of utilisation max calculated for standard fire resistance

R15 and R30 are given in Figure 5.6. It should be noted that for a fireresistance R30, unprotected members with a section factor ( Am/V )sh higher than

50 m-1

can only achieve very low values of the degree of utilisation.

Figure 5.6 Maximum utilization level as a function of section factor ( Am / V )sh

0

0.1

0.2 0.3

0.4

0.5

0.6

0.7

0.8

0.9

1

0 50 100 150 200 250 300 350 400 450 500

( Am V ) sh = k sh ( A m V )(m-1 )

max

10

15 minutes

30 minutes

pract ical f ield of 0




7 - 24

5.3.2 Simple design method for steel members

According to EN 1993-1-2, the load-bearing function of a steel member should be assumed to be maintained at a time t if:

tfi,d,dfi, R E (16)

where:


to EN 1991-1-2

tfi,d, R is the corresponding design resistance of the steel member, for the

fire design situation, at time t

The following simplified calculation methods allow the designer to assess the

design fire resistance (buckling resistance, resistance moment) of steelmembers. They are mainly based on the assumption of constant temperature

within the section.

Steel columns under compression only

The design resistance for the fire design situation at time t of a compression

member with a Class 1, 2 or 3 cross-sections at a uniform temperature θ a should be determined from:

Rdθy,f M,

M0f Rdt,fi, N k N

ii

(17)

where:

θy,k is the reduction factor for the yield strength of steel at the steel

temperature θ reached at time t

fi,M is the partial safety factor for fire situation ( 1fi,M )

0M is the partial safety factor at normal temperature ( 10M )

Rd N is the design resistance of the cross-section N pl,Rd for the normal

temperature design according to EN 1993-1-1

fi is the reduction factor for flexural buckling in the fire design

situation

The reduction factor fi for flexural buckling is obtained from the non-

dimensional slenderness at temperature θ using:

2θ

2θθ

f

1

i but fi 1.0 (18)

with

2

θθθ

12

1




7 - 25

where:

is the imperfection factor for the appropriate buckling curve given

by y/23565.0 f with f y is characteristic yield strength of steel.

The non dimensional slenderness at temperature θ is given by:

θE,θy,θ / k k (19)

where:

θy,k is the reduction factor for the yield strength of steel at the

temperature

θE,k is the reduction factor for the slope of the linear elastic range at the

temperature

The non dimensional slenderness at normal temperature, according

to EN 1993-1-1

The non dimensional slenderness at normal temperature is given by:

E

f

i

ycr

π

1 (20)

where:

cr is the buckling length in the buckling plane considered

i is the radius of gyration about the relevant axis, determined usingthe properties of the gross cross-section

For a practical use, the reduction factor if for flexural buckling can be directly

calculated from values given in Table 5.3, according to the steel grade and the

non dimensional slenderness of steel member at normal temperature . Values

of reduction factor fi in Table 5.3 were calculated assuming a slenderness in

the fire situation equal to 3.1θ . For intermediate value of non-

dimensional relative slenderness, linear interpolation may be used.




7 - 26

Table 5.3 Values of reduction factor fi as function of non dimensional

slenderness at normal temperature and the steel grade

Steel grade Steel grade

S235 S275 S355

S235 S275 S355

0,2 0,8480 0,8577 0,8725 1,7 0,1520 0,1549 0,1594

0,3 0,7767 0,7897 0,8096 1,8 0,1381 0,1406 0,1445

0,4 0,7054 0,7204 0,7439 1,9 0,1260 0,1282 0,1315

0,5 0,6341 0,6500 0,6752 2 0,1153 0,1172 0,1202

0,6 0,5643 0,5800 0,6050 2,1 0,1060 0,1076 0,1102

0,7 0,4983 0,5127 0,5361 2,2 0,0977 0,0991 0,1014

0,8 0,4378 0,4506 0,4713 2,3 0,0903 0,0916 0,0936

0,9 0,3841 0,3951 0,4128 2,4 0,0837 0,0849 0,0866

1 0,3373 0,3466 0,3614 2,5 0,0778 0,0788 0,0804

1,1 0,2970 0,3048 0,3172 2,6 0,0725 0,0734 0,0749

1,2 0,2626 0,2691 0,2794 2,7 0,0677 0,0686 0,0699

1,3 0,2332 0,2387 0,2473 2,8 0,0634 0,0642 0,06531,4 0,2081 0,2127 0,2200 2,9 0,0595 0,0602 0,0612

1,5 0,1865 0,1905 0,1966 3 0,0559 0,0565 0,0575

1,6 0,1680 0,1714 0,1766

Steel beams

The design moment resistance for the fire design situation of a laterallyunrestrained beam with a Class 1, 2 or 3 cross-section, at a uniform

temperature a is given by:

Rdθy,fiM,

M0

f LT,Rdt,fi, M k M i

(21)

where:

θy,k is the reduction factor for the yield strength of steel at the steel

temperature θ reached at time t

Rd M is the moment resistant of the gross cross-section (plastic moment

resistant Rd pl, M or elastic plastic moment resistant Rdel, M for the

normal temperature design calculated using EN 1993-1-1

fiLT, is the reduction factor for lateral-torsional buckling in the fire

design situation. It may be calculated in the same way as the

reduction factor for flexural buckling but using the appropriate non-dimensional slenderness

For laterally restrained beams, the same design method can be used, adopting

1fiLT, .

Often structural members will not have a uniform temperature. An adaptationfactor κ 1 can be introduced to take account a non-uniform temperature

distribution over the height of the steel section. A further adaptation factor κ 2 can be also introduced to account for variations in member temperature along

the length of the structural member when the beam is statically indeterminate.




7 - 27

The value of the adaptation factors κ 1 and κ 2 should be taken according to

EN 1993-1-2.

Members subject to combined bending and axial compression

A simplified design method is also available to verify the fire resistance of

steel members subjected to combined bending and axial compression, such asslender columns under eccentric load and long beams with lateral buckling. For

this situation, the simple calculation model takes into account the combinationeffect of bending and compression by combining above two models for the

simple loading condition. Detailed information is given in EN 1993-1-2.

5.3.3 Determination of fire protection material thickness

In situations where requirements with respect to fire resistance are high

(generally more than R30), the application of prescriptive rules usually leads to

the fire protection of steel structures. When passive fire protection is necessary,the knowledge of the critical temperature, the section factor and the fire

resistance time required, allow for a given fire protection system (spray, board,intumescent coating), determination of the thickness to apply. Only products

which were tested and assessed in standard fire tests according to the Europeanstandard EN 13881 may be used in practice.

The required thickness can usually be determined from manufacturer’s

published data. Such manufacturer’s data can be given in form of table or diagram as illustrated in Figure 5.7. The data generally relates the thickness of

fire protection material to the section factor of the steel member ( A p/V ), the

critical temperature and the fire resistance time required. For typical buildingconstruction using standard I and H steel profiles, the value of Am/V is usually

in the range 30 – 450 m-1.

Fire resistance rating R60

Section factor A p / V (m- )

S t e e l

t e m p e r a t u r e ( ° C )

Figure 5.7 Example of French diagram for boarded fire protection

In practical design, for a given fire protection material, the thickness may be

determined according to following steps:

Choose the data related to the fire resistance time required

Calculate the section factor according to the shape of the steel profile, the presence of any shading of the structural member against heat transfer from




7 - 28

the fire during the fire duration (for example a concrete slab put on the

upper flange of the profile), the type of fire protection (according to theoutline of the steel profile or in box)

Determine the thickness from the manufacturer’s data using the criticaltemperature and the section factor. Linear interpolation is permissible to

determine thickness.

The European Convention for Constructional Steelwork (ECCS) has developed

so-called Euro-nomograms[13]

, which relate for a given time of standard fireexposure, the temperature reached by insulated steel members to the factor

( λ p/d p) ( A p/V ) depending on the fire protection characteristics ( λ p and d p) andthe section factor A p/V . Note that these Euro-nomograms are determined on the

basis of the ENV version of the fire part of Eurocode 3. Also for this reasonthey should be used with some caution. Other nomograms based on

EN 1993-1-2 have been recently developed[14]

.

5.3.4 Design tables for composite members

Design tables for composite members are given in EN 1994-1-2. They are

applicable only to steel and concrete composite members (composite beamswith partially or fully concrete encasement of steel beam, composite columns

with partially or fully concrete encased profiles, composite columns withconcrete filled rectangular or circular steel hollow sections). They use

predefined values, based mainly on standard fire test results, improved withanalytical investigation. The tables allow the designer to quickly obtain the

member size (minimum dimensions of cross-section, the necessary reinforcing

steel area and its minimum concrete cover) as a function of the load level for

common standard fire resistances. The most important advantage of thismethod is the ease of application. However it is limited by a very strict set of geometrical rules and it gives more conservative results compared to other

simple calculation models or advanced calculation models. As a consequence,it should only be applied for the pre-design of a building.

Detailed information is given in EN 1994-1-2.

5.3.5 Simplified calculations models for composite members

The following design methods have been developed to predict the resistance of

individual members when exposed to a standard fire curve. Therefore they are

not applicable to “natural” fires.

Only the design methods for the most commonly used composite members in

single-storey building (composite columns and partially encased concrete beams) are described here.

Composite columns

The simple design methods for columns allow the designer to assess the fireresistance of a composite column by calculating its buckling resistance using

the temperature distribution through the cross-section and the correspondingreduced material strength defined at the required fire resistance time. This

method is based on the buckling curve concept: the plastic resistance to axialcompression N fi,pl,Rd and the effective flexural stiffness ( EI )fi,eff , are used to

derive a reduction factor for buckling. The method is applicable to all types of




7 - 29

composite column provide that an appropriate buckling curve is used.

Checking the column consists of proving that the axial compression (for thecombination of actions considered in fire situation according to EN 1991-1-2)

is less than the buckling resistance of the column.

For a given temperature distribution across the cross-section, the designresistance of a composite column N fi,Rd can be determined from the appropriate buckling column curve relating the load capacity N fi,Rd to the plastic load

N fi,pl,Rd and the elastic critical load N fi,cr as follows:

Rd pl,fi,θRdfi, . N N (22)

is the reduction factor for flexural buckling depending on the slenderness in

fire situation θ .For composite columns, θ may be defined as:

cr fi,R pl,fi,θ / N N (23)

where:

cr fi, N is the Euler buckling load

R pl,fi, N is the value of N fi,pl,Rd according to (24) when the partial security

factors M,fi,a, M,fi,s, and M,fi,c,of the materials are taken as 1.0

The reduction factor is determined as for normal temperature design butusing an appropriate buckling curve defined as function of column type

(partially encased steel section, filled hollow steel section).

The ultimate plastic load, N fi,pl,Rd of the cross-section is determined bysumming the strengths of every part of the cross-section (yield stress for steel

parts, compressive strength for concrete parts) multiplied by the correspondingareas, taking into account the effect of temperature on these elements, withoutconsidering their interaction (due to differential thermal stresses), i.e.:

m

c

k

s

j

f A

f A

f A N )()().(

cfi,M,

θc,

sfi,M,

θs,

afi,M,

θay,aRd pl,fi,

(24)

N fi,cr is the Euler buckling load calculated as a function of the effective flexural

stiffness of the cross-section eff fi,)( EI and the buckling length

of the

column in fire situation, i.e.:

2θ

eff fi,2cr fi,

)(π

EI N (25)

The effective rigidity ( EI )fi,eff is determined from:

mk j

I E I E I E EI )()()()( θc,θsec,c,θc,θs,θs,θs,θa,θa,θa,eff fi, (26)

where:

θ,i E is the characteristic modulus of material i at the temperature . For

steel, it is the modulus of elasticity. For concrete: 2/3 secc,c, E E




7 - 30

where θsec,c, E is the characteristic value for the secant modulus of

concrete in the fire situation, given by the ration between f c,θ and

cu,

I i is the second moment of area of material i related to the central axis

(y or z) of the composite cross-section

a, (for steel profile), s, (for reinforcements) and c, (for concrete) are

reduction coefficients due to the differential effects of thermal stresses.

Detailed information is given in EN 1994-1-2 §4.3.5.

Partially encased steel beams

The simple design method for partially encased steel beams allows the designer

to assess the fire resistance by calculating its bending resistance at the required

fire resistance time. It is based on the simple plastic moment theory. The

method requires the calculation of the neutral axis and corresponding bendingresistance, taking into account temperature distribution through the cross-

section and corresponding reduced material strength. Distinction is made

between sagging moment capacity (usually at mid-span) and the hogging

moment capacity (at the support, if appropriate). If the applied moment is less

than the bending resistance of the beam, the member is deemed to have

adequate fire resistance.

The plastic neutral axis of the beam is determined such that the tensile and

compressive forces acting in the section are in equilibrium:

01 cfi,M,

c,,θc,

1 afi,M,

,y,y,

m

j

j j j

n

i

iii

f k A

f k A

(27)

where:

f y,i is the nominal yield strength for the elemental steel area Ai taken as

positive on the compression side of the plastic neutral axis and

negative on the tension side

f c, j is the nominal compressive strength for the elemental concrete area

A j taken as positive on the compression side of the plastic neutral

axis and negative on the tension side

The design moment resistance Rdt,fi, M may be determined from:

m

1 j c,fi,M

jc,

j,c, j j

n

1i a,fi,M

i,y

i,y,iiRd,t,fi

f k zA

f k zAM

(28)

where:

z i, z j are the distances from the plastic neutral axis to the centroid of the

elemental area Ai and A j




7 - 31

For the calculation of the design value of the moment resistance, the cross-

section of the beam is divided into various components, namely:

the flanges of the steel profile

the web (lower and upper parts) of the steel profile

the reinforcing bars

the encased concrete.

To each of these parts of the cross-section, simple rules are given which define

the effect of temperatures and allow calculation of the reduced characteristic

strength in function of the standard fire resistance R30, R60, R90 or R120.

Detailed information is given in EN 1994-1-2 §4.3.4.

5.4 Specific design rules for single-storey bui ldings National fire regulations of many European countries have been changed

recently to introduce, for single-storey storage and industrial buildings with

significant fire risks (high fire loads), specific safety requirements in terms of

structural behaviour as an alternative to standard prescriptive requirements.

The following criteria relating to the structural behaviour of storage and

industrial buildings (load-bearing structure, façade elements, roofing and fire

walls) must be satisfied to ensure adequate life safety for building occupants

and firemen:

In case of fire occurring in one of the cells of the building, its structure

(including façade elements) must not collapse towards the outside.

In case of fire occurring in one of the cells of the building, the localized

failure of the cell in fire must not lead to the collapse of the neighbouring

cells.

To help the design of storage and industrial buildings with a steel structure,

several simple design methods can be used5,6. These design methods allow the

designer to easily prove that the behaviour of the steel structure of these

buildings in fire situations fulfils the above criteria. The methods are

implemented in the LUCA software[15].

The design methods enable the designer to:

Evaluate forces induced by the collapse of the heated part of the structure.

These forces should be used as additional horizontal load for the stability

check of the part of the frame that remains cold during the fire. That part

can be assessed using normal conditions design tools for structure analysis.

Provide maximum horizontal displacements developed at the ends of the

compartment affected by the fire. These displacements are used to ensure

that movements of the structure in the event of fire do not adversely affect

the stability of fire walls or building façades. Design methods used for this

verification depend on the type of the wall (such as in lightweight concrete,reinforced concrete, hollow block, steel sheeting with insulator,

plasterboard, bricks, etc.) and connection to the steel frame.




7 - 32

The following buildings can de designed by these methods:

Storage and industrial buildings with steel structure. Either steel portal

frames with standard H or I hot rolled profiles or equivalent welded plate

girders, or steel frames based on lattice beams with columns in standard H

or I hot rolled profiles or equivalent welded plate girders

Storage and industrial buildings of portal frame construction divided in

several cells, separated one from each other by fire walls. These walls can

be either perpendicular to the steel portal frames or parallel to the steel

portal frames (see Figure 5.8).

These methods were specifically developed for storage and industrial buildings

but they can also be applied to other type of single-storey buildings.

fire wall perpendicularto the steel frame

fire wall parallel to thesteel frame

Figure 5.8 Location of fire wall compared to steel frames

Calculation methods (see Section 5.5) are only required when fire walls are

perpendicular to steel frames of the building and the building height exceeds

20 m5. When fire walls are parallel to steel frames, the risks of collapse

towards the outside and progressive collapse (between different fire

compartments) can be simply avoided by following the recommendations in

Section 5.5.3.




7 - 33

5.5 Simplif ied design methodsA flowchart showing simplified calculation methods is given in Figure 5.9.

yes

No

No

yes

(*) For all the possible fire scenarios according to the building arrangement

Industrial hall

Checking of failure modes

Choice of fire scenario(*)

(see Figure 5-14)

Calculation of displacements of the steel structure δi

(see expression (30))

Calculation of tensile force F i (see expression (29))

Checking of the compatibility of displacements

Steel structure and partition elements

Steel struc ture and facade elements

Checking of the stability at the

ultimate limit states of the cold par ts of the st eel st ruc tu re

End of

checking

yes

Change in the

steel structure

Change in the design of partitionor facade elements to ensure the

compatibility of displacements

No

Is it a simple isolated po rt al f ram e?

Design recommendations

at the bottom of columns(see end of §5.6.2)

yes

Figure 5.9 Application flowchart of calculation methods

The calculations of tensile force and lateral displacements at compartment ends

must be performed for all possible fire scenarios. Examples of scenarios are

given in Section 5.5.3. Calculation methods are given in Sections 5.5.1 and

5.5.2.

5.5.1 Tensile force at compartment ends

m1 = 1 m2 = 2n = 1

K 2 F F

K 1

Figure 5.10 Horizontal tensile force at the fire compartment ends

When a fire occurs in a compartment of the building, the horizontal tensile

force F at the compartment ends resulting from the collapse of the roof

structure (see Figure 5.10), which is needed to verify the stability of the cold

part of the structure can be obtained from:

qnc F eff p (29)




7 - 34

where:

pc is an empirical coefficient (depending on the slope of the roof and

the type of steel structure)

FramesLatticefor

FramesPortalfor

45,1

slope10%for 10,1

slope5%for 16,1

slope0%for 19,1

pc

neff is a coefficient related to the total number of heated bays n in the

fire compartment (see Table 5.4)

q is the linear load on roof [N/m] (equal to the load density multiplied

by the spacing between frames) applied on the beam and calculated

in fire situation (q = G + 1 S n), where G is the permanent loadincluding self-weight of the steel frame and service overloads, S n is

the snow load and 1 is the load factor according to load

combination coefficients defined in EN 1990 and corresponding

national annexes.

is the span of on heated bay connected to the column [m]

Table 5.4 Values of coeffi cient neff

Portal frame Lattice Frame

Setting of compartment in fire Setting of compartment in fire

Number of bay in fire

end middle end middlen =1 neff =0,5 neff =1,0 neff =0,6 neff =1,0

n 2 neff =1,0 neff =2,0 neff =1,0 neff =1,0

Where columns of the steel frame support a boundary fire wall, columns

should be designed (providing adequate robust base to columns) to resist a

horizontal force calculated according to equation (29) but using neff = 1,0.

5.5.2 Lateral disp lacements at the fire compartment ends

In the event of fire, movements of steel single-storey buildings can be of the

order of several tens of centimetres and therefore could lead to the failure of

façade or the partition element if it is not sufficiently ductile or not accurately

fixed. So it is important to check that façade elements and fire walls in contact

with the steel structure are compatible with the lateral displacements developed

at the ends of fire compartments and that they keep their integrity to avoid the

collapse towards outside and the progressive collapse between different fire

compartments




7 - 35

Maximum lateral displacements δi (i = 1, 2) induced at the top of columns

located at the compartment ends can be obtained using the following

expression (see Figure 5.11):

buildingtheof middlein theisfirewhen the;Max

buildingtheof endat theisfirewhen the

tht

tht

ii

ii

K

F nl c

K

K

nl c

K

K

(30)

where:

n is the number of heated bays

K i is the equivalent lateral stiffness of the considered part i of the

structure [N/m]

K t is the equivalent stiffness (depending on equivalent stiffnesses

1 K and 2 K ) given by:

21

21t

K K

K K K

is the span of one heated bay connected to the column [m]

F is the tensile force [N]

cth is an empirical coefficient (dependent on the slope of the roof and

the type of steel structure)

FramesLatticefor

FramesPortalfor

009,0

slope10%for015,0

slope5%for011,0

slope0%for01,0

thc

Lateral stif fness K for fi re in the middle of a frame

If the fire compartment is in the middle of the frame as illustrated in

Figure 5.11 , K 1 and K 2 should be calculated by an elastic method.

1 2

m1 = 1 m2 = 2n = 1

K 2 K 1

Figure 5.11 Fire located in a cell at the middle of the building

However, for usual steel frames (constant range, even standard steel profiles

from one span to another), the equivalent lateral stiffness i K on either side of

the fire can be calculated approximately according to the number of cold spans

on that side (mi) using the following relationships:




7 - 36

2for

1for

i

ii

mck

mk K (31)

)6,0

1(

21

12

22

1

)(

12

21

with

c

b

3c

h

f

l

f h

I

I

j

im

j

jc

f h

EI

k

(32)

where, for each side in turn (i = 1, 2):

h is the height of the columns

f is the ridgepole

l is the length of the span

I b is the second moment of area of the beam

I c is the second moment of area of the column

E is the modulus of elasticity of steel for normal temperature

f

h

mi=2

I b

I c

Figure 5.12 Definition of parameters of cold parts on side i of the frame

Lateral stif fness K for f ire at the end of a frame

If fire compartment is at the end of the frame, K 2 should be calculated as for

fire in the middle compartment. K 1, which is defined as the lateral stiffness of

the steel frame of the heated fire compartment, should be calculated as follows:

frameslatticefor 2for 3,0

1for 2,0

frames portalfor

2for 13,0

2for 13,0

1for 065,0

2

2

1

n K

n K

nk c

nk

nk

K (33)

where k and c are calculated from equation (32) with m1 = n − 1, where n is thenumber of heated bays, as shown in Figure 5.13.




7 - 37

K 1 1 2

n = 1 m2 = 3

K 2

Figure 5.13 Fire in a compartment at the end of the building

5.5.3 Example of fire scenarios

The above calculations must be performed for all possible fire scenarios. These

scenarios are defined in accordance with the arrangement of the storage

building (structure and partitioning) as illustrated in the example in

Figure 5.14.

Configuration of storage building: 5 spans and 3 cells

Cell 1 Cell 2 Cell 3

Fire wall Fire wall

Scenario 1: fire in cell 1



3 fire scenarios need to be co nsidered

Figure 5.14 Fire scenarios accord ing to the arrangement of the building

5.6 Design recommendationsAdditional design recommendations for fire walls, façade elements and bracing

systems must be put into practice to avoid the collapse toward the outside of

the building and the progressive collapse of the steel structure. Obviously,

recommendations allow also the collapse of the steel structure under fire

condition on either side of fire wall without causing any damage to this wall.




7 - 38

5.6.1 Fire walls

To limit the fire spread to a neighbouring compartment from the fire

compartment, a solution that requires the building to be subdivided into

independent compartments can be achieved by implementing one of the

following construction details:

Two independent fire walls (such as sandwich panels, prefabricated panels,

etc.) each fixed to an independent structural frame (see Figure 5.15 (a)). In

this case, when one structure and its fire wall collapse during a fire, the fire

cannot spread to the neighbouring structure, which remains stable and fire

protected by the second fire wall

A single fire wall inserted between both structures. This fire wall can be a

self-stabilized wall and fully independent. The fire wall can be also fixed at

its top to both structures by means of “fusible” ties (see Figure 5.15 (b))

which, in case of fire near the wall, releases the connection to the ‘hot’

structure (usually when a temperature from 100 to 200°C is reached in

bolts) without causing any damage to the wall (it one remains attached to

the steel structure located on the ‘cold’ side) and the stability of the

neighbouring cold structure.

Self-stabilized walls are commonly used in practice. However during a fire,

this solution can be dangerous for people (occupants and firemen) because they

collapse away from the fire as a consequence of thermal bowing effect. So,

they should be used only if their behaviour has been evaluated by advanced

calculation model taking into account second order effects. Moreover, where

spacing from the self-stable wall to the neighbouring steel structure is not

sufficient, it is important to make sure that the fire wall can bear the force

which may be induced by the movements of the building due to the thermal

elongation of the roof structure (beams and purlins) due to the increase of

temperature in the cell with the fire.

As an alternative to the previous solutions, it is possible to insert the fire wall

into the steel structure of the single-storey building as illustrated in

Figure 5.15(c). Such wall can be either perpendicular to the steel frame or

parallel to the steel frame. Several solutions can be then considered: fire wall

inserted into a line of columns, fire wall attached to columns or fire wall

moved from a line of columns. For these solutions, adequate measures must be

implemented to avoid the collapse of the wall as a result of significant lateraldisplacements of the steel structure. These measures concern:

The attachment of fire walls to the steel structure

The fire protection of the steel structure near fire walls,

The roof system above fire walls

The bracing system.




7 - 39

a) Doubling of the structure aswell as fire walls

b) Doubling of structure withfire wall fixed by “fusible”ties

c) Example of fire wallinserted into the steelstructure

Figure 5.15 Some solu tions of fire walls

At tachment of façade elements and fi re walls to s teel st ructure

Fire walls and façade elements fixed to steel structure of single-storey-

buildings have to remain solidly attached in order to prevent any failure of

these elements due to significant lateral displacements of structure in the event

of fire, and so to avoid risks of progressive collapse and collapse towards the

outside of the building.

3m

3m

3m

3m

Fire wallFacade element

Figure 5.16 Design detail for façade elements and fire walls

One solution consists of fixing these elements to the columns of the load-

bearing structure by means of suitable attachment systems uniformly

distributed over the building height. The maximum spacing of these

attachments will be fixed by the manufacturer of the walls; it is recommended

that the spacing should not exceed 3 m for walls constructed on-site walls

(concrete, masonry, etc.).

In addition, fastenings used to connect fire walls and façade elements on the

columns must be designed to resist the forces produced due to wind and self-

weight of partition elements under the effect of the lateral displacement

induced by the steel frame of the building. If these fastenings are in steel and

unprotected against fire, each of them must be designed at ambient temperature

to resist the following force:

nd pW F i /5 (34)

where:

W is the characteristic wind load used for the design at ambient

temperature and applied to each fastening [N]

p is the self-weight of the wall [N/m²]

d is the spacing between frames [m]




7 - 40

n is the total number of fastenings (uniformly distributed along the

height)

i is the maximum lateral displacement obtained from relation

(26) [m]

Fire protection of steel elements near to fi re walls

The requirement that there should be no fire propagation between different

compartments and no progressive collapse (i.e. the integrity condition of fire

walls must be preserved and the cold parts of the structure must remain stable),

leads to the requirement that that columns used as supports of fire walls must

achieve the same fire resistance as required for fire walls. In common cases,

these fire requirements lead to the application of fire protection to the columns.

On the other hand, columns which do not support fire walls will not require fire

protection.

Additionally, structural members that could damage fire walls (such as beamsand purlins near or crossing the walls) will also have to be fire protected.

5.6.2 Recommendations for steel portal frames

Fire wall perpendicular to steel frame

Figure 5.17 illustrates the situation where the fire wall is perpendicular to the

steel frame. For this situation:

Columns that are built into or near a wall must be fire protected.

Where fire wall is inserted between the flanges of the columns, no

additional fire protection is needed for the roof beams (Figure 5.17 (a)).

Where portal frames do not have haunches and fire wall is fixed to one

flange of columns, fire protection must be applied to any beam crossing the

fire wall (on the side of the wall) over a minimum length of 200 mm

beyond the wall limit. This protection allows a shift of the plastic hinges

away from the walls and thus prevents damage to the wall as a result of the

collapse of the beam (see Figure 5.17 (b)). Where portal frames have

haunches, no fire protection is needed for the beams.

Purlins do not cross the fire wall in this situation and no special

considerations are required.

The thickness of fire protection material applied to columns may be calculated

assuming a critical temperature of 500°C and the same required fire resistance

as the fire walls. Fire protection should be provided over the full height of

columns.

If beams are partially protected, the thickness of fire protection material may

be calculated assuming a steel section exposed on four faces for the section

factor, a standard fire exposure of one hour and a critical temperature of 500°C.




7 - 41

beam purlin

firewall

protected

column

purlin

firewall

protectedcolumn

fire

protection

d 200 mm

beam

a) Wall inserted between the flanges of columns

b) Wall fixed to one flange of columns

Figure 5.17 Design detail near fire walls perpendicular to portal steel frame

Fire wall parallel to steel frameFigure 5.18 illustrates the situation where the fire wall is parallel to the steel

frame.

For this situation:

The fire wall either be located between two frames or in the plane of the

frame, between faces of the columns and beams.

Columns and beams that within the fire wall or near a fire wall must be fire

protected.

Purlins will cross the fire walls. It is therefore necessary to fire protect

continuous purlins (over a distance of 200 mm from the wall) or to design anon-continuous purlin system. For example, where fire wall is in the plane

of a frame, steel elements fixed to the beams should be inserted through the

wall to support the purlins.

The thickness of fire protection material applied to columns and beams may be

calculated assuming a critical temperature of 500°C and the same required fire

resistance as the fire walls. Fire protection should be provided over the full

height of columns.




7 - 42

fire

wall

protectedbeam

purlin

protectedcolumn

purlin

flexible fireprotectionmaterial

continuous purlin

protectedbeam

d 200 mm

protectedcolumnfire wall

fireprotection

purlin

Rigidsupport

fire wall

purlin

protectedbeam

protectedcolumn

a) Fire wall inserted betweenthe flanges of columns

b) Fire wall fixed to one flange of columns

Figure 5.18 Design detail near fire walls parallel to portal steel frame

If purlins are partially protected, the thickness of fire protection material may

be calculated assuming a steel section exposed on four faces for the section

factor, a standard fire exposure of one hour and a critical temperature of 500°C.

Addi tional design recommendations for s imple portal steel frames

In the case of single-storey buildings with simple portal steel frame where the

column height/beam span ratio of the frame (h/l ) is greater than 0,4, the failure

mode towards the outside can be avoided by designing the connections

between columns and foundation, and the foundation itself, to have sufficient

resistance to sustain the vertical loads in the fire situation together with an

additional bending moment equal to 20% of the ultimate plastic moment of thecolumn at normal temperature.

Fire wall

simpleportal steel frame

simple portal steel frame

h

L

Figure 5.19 Single-storey bu ildings with simple portal steel frame

Examples of fire walls

Illustrations of fire walls adopting some of the above recommendations are

shown in Figure 5.20. They show clearly that the fire walls were not damaged,

despite the collapse of the steel structure.




7 - 43

a) Self-stable fire wall inserted between twoindependent steel framework

b) Partially fire protected steel beam crossinga fire wall fixed to steel columns

Figure 5.20: Views of fi re walls after fire disaster in steel single-storey building

5.6.3 Recommendations for steel frames based on lattice beams

Fire wall perpendicular to steel frame

Figure 5.21 illustrates the situation where the fire wall is perpendicular to the

steel frame. For this situation:

Columns that are built into or near a wall must be always fire protected.

Where fire wall is inserted between the flanges, the lattice beams should befire protected on both side of the wall (see Figure 5.21 (a)).

Were the fire wall is fixed to one flange, only the lattice beams on the wall

side have to be protected. Fire protection must be applied to the beams over

a minimum length equal to the distance separating the wall with the first

vertical member of lattice frame (see Figure 5.21 (b)).

Purlins do not cross the fire wall in this situation and no special

considerations are required.

The thickness of fire protection material applied to columns may be simply

calculated assuming a critical temperature of 500°C and the same fireresistance as required for fire walls. Fire protection should be provided over the

full height of the columns.

If lattice beams are partially protected, the thickness of fire protection material

may be calculated assuming for the section factor: a steel section exposed on

four faces for bottom chords, vertical members and diagonals; and on three

faces for top chords. A standard fire exposure of one hour and a critical

temperature of 500°C may be used.




7 - 44

first verticalmember

fire wall

protectedcolumn

fire protectionlattice beam

fireWall

protectedcolumn

fire protection

first verticalmember

lattice beam

a) Fire wall inserted between the flangesof columns


Figure 5.21 Design detail near fire walls perpendicular to steel frame wi thlattice beam

Where fire wall is parallel to steel frame

Figure 5.22 illustrates the situation where the fire wall is parallel to the steel

frame. For this situation:

It is not practical to provide a wall in the plane of a frame, because it is

difficult to make it continuous through the depth of the lattice beam. roof,

Fire walls parallel to a frame are therefore usually either beside and in

contact with the steel frame or between two independent steel structures.

Where the fire wall is attached to a steel frame, the columns and beams

must be fire protected (see Figure 5.22 (b)). Moreover purlins and beamstays near the wall must be fire protected over a minimum length

corresponding to the distance from the wall to the joint purlin/beam stay

when the roof structure is made of purlins.

Where the fire wall is inserted between two independent steel structures, no

fire protection is needed (see Figure 5.22 (a)).

If columns are protected, the thickness of fire protection material may be

calculated assuming a critical temperature of 500°C and the same fire

resistance as required for fire walls. Fire protection should be provided over the

full height of the columns.

If lattice beams are protected, the thickness of fire protection material may be

calculated assuming for the section factor: a steel section exposed on four faces

for bottom chords, vertical members and diagonals; and three faces for top

chords. A standard fire exposure of one hour and a critical temperature of

500°C may be assumed. Fire protection should be provided over the full length

of the lattice beams.

The thickness of fire protection material applied to purlins and beam stays may

be simply calculated assuming a steel section exposed on four faces for the

section factor, a standard fire exposure of one hour and a critical temperature of

500°C.




7 - 45

purlinfire wall

beam stay

columnprotectedcolumn

purlin

firewall

protectedlattice beam beam stay

a) Fire wall inserted between two independentsteel framework


Figure 5.22 Design detail near fire walls parallel to steel frame with lattice

beam

5.6.4 Recommendations for bracing system

The requirement for no collapse towards the outside of the building in the the

longitudinal direction (perpendicular to steel frames) can be satisfied using

appropriate bracing systems. Specifically, each compartment must have its own

bracing system.

Fire wall perpendicular to the steel frame

Figure 5.23 (a) illustrates the situation where the fire wall is perpendicular to

the steel frame. For this situation:

Use additional vertical bracing systems at both ends of fire wall, to ensure

integrity of wall. These bracing systems should be designed to support a

lateral load taken as 20% of that due to normal wind actions (according to

the combination of actions for the fire situation), calculated for a gable area

that is limited to the width between gable posts.

Provide double bracing systems (i.e. have bracing systems on both sides of

fire walls) or protect the bracing system.

The bracing systems must be arranged in a way that they will not cause

problems for normal temperature design, for example by compromising

movement of an expansion joint.

Fire wall parallel to the steel frame

Figure 5.23 (b) illustrates the situation where the fire wall is parallel to the steel

frame. For this situation:

Install bracing systems (vertical bracing and horizontal bracing on roof) in

each compartment. This solution may lead to additional bracing systems for

normal conditions.

Design each bracing system to provide adequate stability in normal

condition and to support in fire condition a horizontal uniform load [N/m]

taken as F = 1,19 (G + ψ 1 S n)l f , where l f is the spacing between steelframes, G is the permanent action, including service overloads, S n is the




7 - 46

snow load and ψ 1 is the frequent combination factor according given in the

relevant National Annex to EN 1990.

Where the fire wall is fixed to one flange of the columns, the elements of

bracing systems must be fixed to rigid steel elements supporting the purlins

on the side of the wall.

Fire wall

Building end

Doubling of additionalbracing system put at theend of fire wall

Bracing system fornormal temperature

Buildingend

Wall perpendicular to steel frame

Fire wall

Bracing system

Wall parallel to steel frame

Figure 5.23 Recommendations for bracing system

5.6.5 Recommendations for roof systems above the separationelements

The roof should be independent from one compartment to the next, adopting

the following recommendations (see Figure 5.24 (a)):

Purlins should be provided either side of the fire wall.

The roof should be stopped on both sides of the fire wall

The roof should be provided with fire protection over a width of 2,50 meither side of the wall.

Alternatively, the wall may be extended above the roof, up to a specific

distance d (see Figure 5.24 (b)).

National regulations may specify other special requirements for roof covering

adjacent to fire walls.

protectedcolumn

Fire wall

beam

purlin

roof with fireproof material 2x2,50m

roof

part of roof between purlins

fireprotection

Protectedcolumn

fire wall

beam

purlin

roof

d

fireprotection

a) Roof with fireproof material b) wall above the roof

Figure 5.24 Roof system above the separation elements




7 - 47

6 GUIDANCE ON THE USE OF MORE ADVANCED SOLUTIONS

This chapter gives an overview of advanced calculation models available for fire modelling, thermal modelling, and structural modelling that can be used in

fire engineering design [9,16].

6.1 Fire modelsTwo kind of numerical models are available to model the development of real

fires: zone models and field models. These models and allow temperatures,

smoke descent, flame spread, time to flashover and many other effects to be

calculated.

6.1.1 Zone models

The simplest model is a one-zone model for fully developed fires (post-

flashover fires), in which the conditions within the compartment are assumed

to be uniform and represented by a single temperature.

Two-zone models may be used for pre-flashover situations, mainly in the

growth phase of a fire. The model is based on the hypothesis of smoke

stratification, separating the fire compartment into two distinct layers: a hot

upper layer (containing most of the fire’s heat and smoke), and a cool lower

layer (which remains relatively uncontaminated by smoke). A fire plume feeds

the hot zone just above the fire. The temperature of each layer is calculatedfrom conservation of energy; the amount of toxic combustion products in each

layer is calculated from conservation of chemical species; and the size of each

zone is calculated from conservation of mass. Simple rules govern plume

entrainment, heat exchange between zones and mass flow through openings to

adjoining compartments. As a result of the simulation the evolution of gas

temperature in each of the two layers, the evolution of wall temperatures,

evolution of flux through the openings and the evolution of the thickness of

each layer are given as a function of time. The thickness of the lower layer,

which remains at rather cold temperature and contains no combustion products,

is very important to assess the tenability of the compartment for the occupants.

Often, the local effect near the fire may be studied using a simple model suchas Hasemi methodology with the two-zone models. The combination of both

models then allows the determination of the gas temperature field near and far

from the fire (see Figure 6.1).

When the thickness of the lower layer is too small compared to the height of

the compartment, the two-zone assumption becomes inapplicable and a one

zone model becomes more appropriate. Moreover if the fire area is big

compared to the floor area, the one-zone model assumption is usually better

than the two-zone one.

Some zone models include the possibility of a switch from a two-zone modelto a one-zone model when some conditions for temperatures, fire area and

smoke layer thickness corresponding to flashover) are encountered.




7 - 48

It is also still possible to choose to follow a two-zone or a one-zone strategy for

the entire duration of a fire. With these strategies, the whole simulation is made

considering two or one zones, from the initial time to the end of the calculation.

No modification of the rate of heat release is made, except via the combustion

models.

Localised fire2 zon emodel

roof

g

Beam

Hasemi method’s

20°C

z

g

2-zone model

at beam level

x

Figure 6.1 Combination of two-zone model with Hasemi method’s

Some of the more complex zone models allow radiation calculations between

the upper layer and room objects. They may also allow multiple fire plumes

and multiple compartment analysis with mass exchange between each

compartment (see Figure 6.2).

The input data are usually the room geometry, room construction (including allwalls, floors and ceilings), number of vents (or holes) and their sizes, room

furnishing characteristics, and fire data (such as RHR curve, pyrolisis rate,

combustion heat of fuel). The output data are usually the prediction of sprinkler

and fire detector activation time, time to flashover, upper and lower layer

temperature, smoke layer height, and species yield.

The fire load can be considered to be uniformly distributed if the combustible

material is present more or less over the whole floor surface of the fire

compartment and when the fire load density (quantity of fuel per floor area) is

more or less uniform. By contrast, the fire load should be “localised” if the

combustible material is concentrated on quite a small surface compared to thefloor area with the rest of the floor area being free of fuel.

An essential parameter in advanced fire models is the rate of heat release. For

design it is common practice to refer to the values given in EN 1991-1-2.

For irregular or complex building geometry, complex ventilation systems, or

where more detail is required on convective or radiant heat exposure levels at

specific targets, the use of a field model should be considered.




7 - 49

A7

12.5m*9m

A13

25m*54m

A3

12.5m*9m

A4

12.5m*9m

A8

12.5m*9m A2

25m*18m

A1

25m*18m

A12

25m*18m

A9

12.5m*9m

A14

25m*54m

A10

12.5m*9m

A5

12.5m*9m

A6

12.5m*9m

A11

25m*18m

Fire source

Figure 6.2 Example of fire modelling using zone models for an indus trial building

6.1.2 Field models

Field models (computational fluid dynamics models) are the most sophisticated

deterministic models for simulating enclosure fires. They incorporate sub-models for turbulence, heat transfer and combustion.

The CFD modelling technique is based on a complete, time-dependent, three-

dimensional solution of the fundamental conservation laws (conservation of

mass, momentum, and energy). The volume under consideration, usually a fire

compartment, is divided into a very large number (sometimes hundreds of

thousands or even millions) of cells. The approximate number of cells

appropriate for the studied compartment will depend on the compartment

geometry, the accuracy required, and from a practical standpoint, the computer

speed and memory.

Three cases of field models, according to the turbulence method implemented

in model, exist:

Direct numerical simulations (DNS): The basic equations are directly

solved but need very short time and spatial steps in order to simulate all

time and spatial scales coming from the turbulent and the chemical

processes. DNS require particularly powerful computers and are used for

academic studies or are confined to simple applications.

Large Eddy Simulation (LES): Large scale motions of the flow are

calculated while the effect of smaller scales is modelled using sub-grid

scale model. The most commonly used sub-grid model is the Smazorinskymodel.




7 - 50

Reynolds-averaged Navier Stokes (RANS): The basic equations are

averaged and all turbulent scales are modelled. The most frequently model

used is k model.

The input data are the same as those required for a zone model but they have to

be supplied with a higher degree of detail. They are the detailed roomgeometry, room construction (including all walls, floors and ceilings), number

of vents (or holes) and their sizes, room furnishing characteristics,

fuel/combustion characteristics, turbulence parameters, and radiation

parameters.

The output data are the smoke and heat movements, prediction of sprinkler and

fire detector activation time, time to flashover, temperatures in the domain,

velocities, smoke layer height, and species yield.

Due to their complexity and the CPU time needed, field models are very little

used for evaluating fire resistance of structures, particularly for fully developedfire. In the fire domain, the use of a field model is often reduced to specific

cases with sophisticated geometry.

6.2 Thermal ModelsAdvanced heat transfer models can be used to calculate temperature

distribution in a structure in a fire. They are mostly based on either finite

difference methods or finite element methods. They are often used to estimate

temperature gradients through structural members primarily made of materials

with a low thermal conductivity and/or high moisture content, such as concrete.

Moreover, they can be applied to structural members under nominal fire

conditions or natural fire conditions.

Such methods have to take into account non-linearity due to temperature

dependence of material properties and boundary conditions. As commonly

assumed in fire design, heat transfer from fire to exposed surfaces is essentially

by convection and radiation. Inside homogeneous materials such as steel, heat

is only transferred by conduction. On the other hand, for porous materials such

as concrete or where internal cavities exist, heat transfers are more complex.

The three processes: conduction, convection and radiation can occur together,

to which may be added mass exchange. However, by way of simplification,

only the dominating process is explicitly introduced in thermal analysis, taking

into account secondary processes through adequate adjustment. In fire design,

it is usually assumed that concrete is a homogeneous material and that heat

transfer occur mainly by conduction. Heat transfer by convection and radiation

occurring in pores are considered as secondary processes and are implicitly

taken into account in thermal properties available for concrete (conductivity,

specific heat). Moreover, mass-exchange is generally neglected and only

moisture evaporation in concrete is taken into account. The effects of moisture

(assumed uniformly distributed in the concrete) is treated in a simplified way,

assuming that when the temperature in a concrete part reaches 120°C, all of the

heat transferred to that part is used to evaporate water. Moisture movementsare rarely modelled. For composite members, contact between steel parts and

concrete parts can be assumed to be perfect (no gap). Radiation in internal




7 - 51

voids (such as hollow steel section) should be considered in the thermal

analysis.

In principle, where the effects of a fire remain localised to a part of the

structure, temperature distributions along structural members can be strongly

non-uniform. So a precise calculation of temperatures should be determined bya full 3D thermal analysis. However, due to the prohibitive computing time of

such analysis, it is often considered an acceptable simplification to perform a

succession of 2D thermal analyses through the cross-sections of the structural

members. Calculations are then performed at relevant location along the length

of each structural member and the temperature gradients are obtained, assuming

linear variation between adjacent temperature profiles. This approach gives

usually a reasonable approximation to the actual temperature profile through

members and allows significant reduction of the modelling and numerical

effort. In 2D thermal analysis, cross-sections of members are commonly

discretised by means of triangular or quadrilateral plane elements with thermal

conduction capability. All sections encountered in civil engineering can thus bemodelled. Each plane element describing the cross-section can have its own

temperature-dependent material such as steel, concrete or insulation materials.

Boundary conditions can be either prescribed temperatures or prescribed

impinging heat flux to simulate heat transfer by convection and radiation from

fire to the exposed faces of structural members. Effects of non-uniform thermal

exposure may be introduced in modelling with appropriate boundary

conditions.

Effects of mechanical deformations (such as buckling of steel element,

cracking and crushing of concrete, etc.) on the temperature rise of structuralmembers is neglected, which is the standard practice. Consequently geometry

of structural members does not vary during the analysis

As for simple models, the use of advanced models require knowledge of the

geometry of structural members, thermal properties of the materials (thermal

conductivity, specific heat, density, moisture...) and heat transfer coefficients at

the member’s boundaries (emissivity, coefficient of heat transfer by

convection).

Usually for fire design, temperature-dependent thermal material properties of

concrete and steel are taken from EN 1992-1-2 and EN 1993-1-2 and heattransfer coefficients are those given in EN 1991-1-2 respectively.

6.3 Structural modelsAdvanced numerical models for the mechanical response should be based on

the acknowledged principles and assumptions of the theory of structural

mechanics. They are usually finite element models. They can simulate a partial

or a whole structure in static or dynamic modes, providing information on

displacements, stress and strain states in structural members and the collapse

time of whole building if collapse occurs within the period of the fire. The

changes of mechanical properties with temperature, as well as non-linear

geometrical and non-linear material properties, can be taken into account in the

structural fire behaviour. The transient heating regime of structures during fire




7 - 52

is modelled by use of step-by-step iterative solution procedures, rather than a

steady state analysis.

This Section outlines some of the primary considerations in modelling the

behaviour of single-storey buildings with steel or composite frames in the fire

situation, notably features related to material models, computation procedure,structural modelling, etc.

Advanced calculation models can be used in association with any heating

curve, provided that the material properties are known for the relevant

temperature range and that material models are representative of real

behaviour. At elevated temperature, the stress-strain curve of steel is based on a

linear-elliptic-plastic model, in contrast to the elasto-plastic model adopted for

normal temperature design. The steel and concrete stress-strain relationships

given in EN 1993-1-2 and EN 1994-1-2 are commonly used.

In the fire situation, the temperature field of structural members varies withtime. As stress-strain relationships of materials are non-linear and temperature

dependant, an appropriate material model has to be adopted in advanced

numerical modelling to allow the shift from one behaviour curve to another, at

each step of time (and thus of temperature). The so-called kinematical material

model is usually used for steel structures, assuming that the shift from one

stress-strain curve to another one due to the change of temperature is made by

staying at a constant plastic strain value (see Figure 6.3). This model can be

used at any stress state of steel (tension or compression). For concrete, it is

much more complicated, since the material has a different behaviour in tension

and in compression. Therefore, different shift rules are needed for when the

material is in tension or in compression. Generally, this kinematic model isused in most advanced calculation models for fire safety engineering

applications.

Behaviour of steel is often modelled with a Von Mises yield contour including

hardening. Behaviour of concrete in compression is modelled with a

Drucker-Prager yield contour, including hardening.

) ,ε( θ d ε

d σ 01

θ (t)θ 1

Δt)θ (t θ 2

a) Behaviour law of structural steel

Parallel to

)02

, ε( θ

d ε

d σ Parallel to

Compression

b) Behaviour law of concrete

θ (t)θ 1

Δt)θ (t θ 2

tensile

Figure 6.3 Kinematic material models for steel and concrete

Another aspect to be noted in the application of advanced calculation models

for steel and composite structures under natural fire conditions is the material behaviour during cooling phase. It is well known that for commonly used steel

grades, the variation of mechanical properties with temperature are considered




7 - 53

as reversible, which means that once they cool down they will recover their

initial mechanical properties. However, this phenomenon is not true with

concrete, whose composition will be totally modified when heated to an

elevated temperature. After cooling down, it cannot recover its initial strength.

Indeed, its strength might even be less after cooling than at maximum

temperature.

The effects of thermal expansion should be taken into account. This is done by

assuming that the total deformation of structural members is described by the

sum of independent terms:

r tr cσtht )( (30)

where th ,σ , r and c are the strains due to thermal expansion, stress, residual

stress and creep, respectively. tr is the strain due to transient and non uniform

heating regime for concrete (usually neglected).

In Eurocodes, the creep strain is considered to be included implicitly in stress-strain

relationships of steel and concrete. The residual stress is usually neglected except for

some special structural analysis. The thermal strain is the thermal expansion ( L/ L)

that occurs when most materials are heated. Thermal strains are not important for fire

design of simply supported steel members, but they must be considered for

composite members, frames and complex structural systems, especially where

members are restrained by other parts of the structure (as for single-storey

building divided into cells separated from one another by fire walls) since

thermally induced strains, both due to temperature rise and temperature

differential, can generate significant additional internal forces.

Distribution of temperaturefor z =cte

Unit strainCross-section(x =cte)

z

y

G

c th

t

r

Figure 6.4 Strain composit ion o f material in advanced numerical modelling

In general, the structural analysis in the fire situation is based on ultimate limitstate analysis, at which there is equilibrium of the structure between its

resistance and its applied loading. However, significant displacement of the

structure will inevitably occur, due to both material softening and thermal

expansion, leading to large material plastification. Therefore, advanced fire

analysis is a non-linear elasto-plastic calculation in which both strength and

stiffness vary non-linearly. From a mathematical point of view, the solution of

such analysis cannot be obtained directly and has to be achieved using an

iterative procedure:

A step-by-step analysis is carried out in order to find the equilibrium state

of the structure at various instants (at different temperature fields).

Within each time step, an iterative solution procedure is carried out to find

the equilibrium state of the structure behaving in elasto-plastic way.




7 - 54

Different types of convergence procedure are usually employed, such as the

pure Newton-Raphson procedure and the modified Newton-Raphson

procedure. The pure Newton-Raphson procedure is recommended for

structures made of beam elements, and the modified Newton-Raphson

procedure is recommended for structures made of shell elements.

Static analysis is normally sufficient for modelling the behaviour of a structure

in fire. However, local failure or instability of a structural member (such as

lateral buckling of purlin) does not lead to overall structural failure.

Consequently, analysis should be performed by a succession of subsequent

static and dynamic analyses to pass instabilities and to obtain the complete

failure mechanism to predict the influence of a local failure on the global

behaviour of the structure and to follow eventually progressive collapse. It has

to be kept in mind that here the aim is not the precise modelling of dynamic

effects. So, default values of the main parameters fixed in models to

determinate acceleration and damping effects can be used.

Existing boundary conditions should be rightly represented. It is common to

design structure by assuming pinned support conditions at the column bases.

However, as fully pinned bases of columns are never achieved in reality, it is

also possible, when data are available, to introduce semi-rigid connections.

Where only a part of the structure is modelled, some restrained conditions from

unmodelled part of the structure should be taken into consideration in

appropriate way. The choices of restrained conditions that have to be applied at

the boundaries between the modelled substructure and the rest of the structure

have to be chosen by the designer. For example, in case of symmetry boundary,

restraints to translation across the symmetry boundary and rotational restraint

about the two major axes on the plane of symmetry are introduced inmodelling.

Usually, beam-to-column joints are assumed to be fully rigid in the fire design

of steel and steel-concrete composite frames. However, in the case of steel

frames based on lattice beams, joints between members of lattice beams and

connections between top and bottom chords of lattice beams and columns can

be assumed pinned or fully rigid according to the type of truss.

Two types of action need to be applied to heated structures. The first type is

static loading. It must correspond to that for fire situation. The second type

consists of the temperature increase (above ambient) of the structural membersobtained, from previous thermal analysis. Boundary conditions at supports as

well as applied gravity loads are assumed to remain unchanged throughout the

fire exposure

It is important to choose an appropriate structural modelling strategy.

Simulation of the mechanical behaviour of single-storey building in fire

conditions can be performed either by a 2D or a 3D analysis.

In a 2D analysis, simulation are performed in the plane of each portal frame,

assuming a three dimensional behaviour of the frame to take into account the

lateral instability of the members (columns, beams). In such modelling,adequate restraint conditions should be introduced to stabilize the frame

laterally. In reality, these out-of-plane restraints are provided by roof structure




7 - 55

(as purlins) as well as façades elements fixed on columns (concrete walls,

sandwich panels, steel sheeting), so that out-of-plane collapse does not occur.

In a 3D analysis, several parallel portal frames, the roof structure (purlins) and

eventually bracing system are explicitly modelled (see Figure 6.5). The main

difference in this 3D analysis is that the interaction effects between memberswill be directly dealt with; load redistribution from heated parts (weakened

parts inside fire compartment) to cold parts (stronger parts outside fire

compartment) can be taken into account in an accurate way and the global

behaviour of structures will be analysed, providing a more realistic situation of

mechanical response of structures in fire. Computation cost with a three-

dimensional analysis is high because of significant number of elements used in

the modelling.

The choice between 2D and 3D analysis will depend on several parameters,

such as the type of structure (steel or composite frame), the dimensions of the

single-storey building, the fire scenario and objectives of structural fire design(to fulfil a prescriptive requirement, or to verify a failure mode).

Firewall

Figure 6.5 Example of 3D mechanical modell ing

The basic finite element set-ups used to represent the structural members of

frame are given below. Solid elements are omitted. as they are numerically too

expensive.




7 - 56

REFERENCES

1 EN 1991-1-2:2002 Eurocode 1: Actions on structures - Part 1-2: General rules -

Actions on structures exposed to fire

2 EN 1993-1-2:2003 Eurocode 3: Design of steel structures - Part 1-2: General rules – Structural fire design

3 EN 1994-1-2:2003 Eurocode 4: Design of composite steel and concrete structures

– Part 1-2: General rules - Actions on structures exposed to fire

4 HOCKEY, S.M., and REW, P.J.

Human response to thermal radiation

HSE Books, UK, 1996.

5 VASSART, O., CAJOT, L-G., ZHAO, B., DE LA QUINTA, J.MARTINEZ DE

ARAGON, J. and GRIFFIN, A.

Fire Safety of industrial halls and low-rise buildings: Realistic fire design, active

safety measures, post-local failure simulation and performance based requirements

ECSC research project 7210-PR-378.

6 RFCS Research: Fire safety of industrial hall, Design Guide, Arcelor Mittal,

CTICM, Labein tecnalia, ULG, Directorate-General for research, Research Fund

for Coal and Steel Unit, RFS2-CR-2007-00032, Luxembourg, 2007.

7 Report to ECCS: Fire building regulations for single-storey buildings in 9

European countries. Document RT915. Version 02 June 2002.

8 LENNON, T., MOORE,D., WANG, B. Y. C. and BAILEY, G.

Designers’ Guide to EN 1991-1-2, EN 1992- 1-2, EN 1993-1-2 and EN 1994-1-2

Actions on Structures Exposed to Fire and Structural Fire Design

Thomas Telford, 2007.

9 DIFISEK - Dissemination of Structural Fire Safety Engineering Knowledge

ECSC research project RFS-C2-03048.

10 PURKISS, J.A.

Fire safety design of structures

Butterworth-Heinemann, Oxford, UK

11 Risk Based Fire Resistance Requirements Competitive (RISK -REI), ECSC

research project 7210-PR-378.

12 SIMMS, W.I., and NEWMAN, G.M.

Single-storey steel framed building in fire boundary conditions (P313)

The Steel Construction Institute, 2002.

13 ECCS TC3: Euro-monograms for fire exposed steelwork.

14 SD005a-EN-EU, Data: Nomogram for protected members, www.steel-access.com

15 RFCS Research: Fire safety of industrial hall, Design Guide, Arcelor Mittal,

CTICM, Labein tecnalia, ULG, Directorate-General for research, Research Fund

for Coal and Steel Unit, RFS2-CR-2007-00032, Luxembourg, 2007.

16 FRANSSEN J. M., KODUR V. and ZAHARIA R.

Designing steel structures for fire safety

Balkema Book, 2009.




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APPENDIX A German fire safety procedure for single-storey industrial and commercial buildings

In Germany, buildings for commercial and industrial use must conform to the“Musterbauordnung” (MBO) and to all federal state building regulations

“Bauliche Anlagen und Räume besonderer Art und Nutzung” (“Structural

facilities and spaces with special requirements and uses”). In such cases, and in

order to meet essential requirements (concerning human safety, public security,

and protection of the natural environment), it is possible to adopt alternative

solutions to the prescriptive federal state building regulations.

This general statement has to be considered in the context of physical and

technical fire protection requirements for a building with reference to of

“Wohngebäude und vergleichbare Nutzungen” (“residential and similar uses”)

according to the federal state building regulations. For commercial andindustrial uses, it is neither necessary nor appropriate to apply the requirements

of the federal state building regulations. When it comes to meeting general

structural fire protection objectives, it is more important to consider each

building on an individual basis.

A standard procedure for assessing requirements, using scientifically based

methods, is recommended.

Since industrial buildings are considered “Sonderbauten” (“special buildings”)

within the definition of §51 Abs.1 MBO and cannot usually be exempt from

the applicable regulations, the goal of MIndBauRl (the technical constructionregulation) is to determine the minimum requirements for structural fire

prevention. The MIndBauRl also uses design procedures according DIN

18230-1: Structural fire protection in industrial buildings –fire resistance

design.

Regarding §3 Abs. 3, Satz 3 MBO, which permits variations from technical

construction standards, the procedure limits this to accepted methods for fire

protection engineering and requires that these are listed in accordance with

Annex 1.

The aim of the procedure is to regulate the minimum requirements for fire protection of industrial buildings, in particular regarding:

the fire resistance of components and the flammability of building materials

the size of fire compartments and fire-fighting areas

the availability, location and length of emergency escape routes.

The procedure will facilitate design for building owners, designers, draftsmen

and specialists; for the authorities it will provide justification for relaxation or

deviation from the alternatively applicable rules of the MBO. It offers building

control and approval bodies a benchmark for equivalent risks. A design method

that requires no detailed engineering analyses and no particular calculation has been established. This responds to legal responsibilities and offers a

straightforward form of approval.




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MIndBauRl applies to all industrial buildings regardless of their size. It does

not apply to:

industrial buildings which are only used for storing technical equipment or

facilities and where only access is temporarily needed for maintenance and

inspection purposes

industrial buildings that are mostly open, such as covered outdoor areas or

open warehouses

buildings which can be assimilated due to their behaviour in fire.

In addition, the procedure does not apply to storage shelves more than 9.0 m

high (to the top of stored material).

This procedure may also be used for allowing and justifying relaxation of the

regulations according to §51 MBO for buildings and structural facilities, which

are not directly covered by the scope of MIndBauRl, although they are

comparable to industrial structures in respect to fire risk.

Justification for relaxation of conditions under §51 Abs. 1 MBO may be

provided with one of the following procedures.

Simplified procedure

In the procedure according to Abs. 6, the maximum fire compartment

surface for a fire section area will depend on the fire-resistance

classification of the supporting and stiffening components as well as the

structure’s fire technical protection infrastructure.

Complete verification procedureIn the procedure according to Abs. 7, the maximum surface area and the

requirements for the components in accordance with the fire safety classes

for a fire compartment will be based on the calculation procedure according

to DIN 18230-1.

Engineering methods

Instead of proceeding according to Abs. 6 and 7, standard fire protection

engineering design methods may also be used.

The initiator of a fire protection concept has the choice which method (Abs. 6

or 7) will be implemented when using the MIndBauRl. However it is not

permissible to combine procedures.

Concerning the fire engineering methods, the MIndBauRl identifies the

principles and conditions for the hypotheses of such designs. It regulates the

verification and checking as well as documentation.

The MIndBauRl, which has been introduced as a standard in the Building

Regulations in all German states, is legally applicable. As part of the

application of IndBauRl, there are several procedural methods. The same

general requirements apply for all verifications; these are identical for all

procedures and must be respected. These include fire-fighting water requirements, smoke evacuation, location and accessibility, emergency exits

and fire spread.




7 - 59

Fire-fighting water requirements must be agreed with the responsible fire

department taking into account the surface areas and fire loads. These

requirements should be assumed to last for a period of two hours.

minimum 96 m³/h for a surface area up to 2500 m²

minimum 192 m³/h for a surface area greater than 4000 m².

Intermediate values can be linearly interpolated.

For industrial buildings with automatic fire extinguishing systems, a water

quantity of at least 96 m³/h over a period of one hour is sufficient to extinguish

the fire.

Any factory or warehouse with an area of more than 200 m² must have wall or

ceiling openings to allow smoke evacuation.

Individual spaces which are bigger than 1600 m² must have a smoke evacuator,so that fire fighting operations are possible. This is because a smoke layer of

2,5 m height has been mathematically proven.

In addition to the location and accessibility of each fire compartment, at least

one side has to be located at one outside wall and be accessible from there for

the fire department. This is not applicable for fire compartments which have an

automatic fire extinguishing system.

Stand-alone and linked industrial structures with foundations of greater than

5,000 m² have to be accessible from all sides by fire fighting vehicles. These

access routes must meet the requirements for fire brigade usage.

The fire service access roads, operating areas and other routes should be kept

continuously free. They have to be permanently and easily recognizable.

Included in the emergency exits in industrial buildings are the main production

corridors and storage areas, the exits from these areas, staircases and exits to

the outside. Each room with an area of more than 200 m² must have at least

two exits.

Regarding the maximum allowable length for emergency escape routes,

equipment and structural fire protection both influence each other.

The maximum length of emergency escape routes is limited as a rule to 35 m

for a clear height up to 5 m. However, if a fire alarm system is installed, then

this increases to 50 m.

The maximum increase in length in relation to free height up to 50 is 70 m.

The distances are measured as distances in space, but not through construction

elements or components. The real length should not be more than 1.5 times the

distance that was measured in space. Attention should be paid to the fact that

from any point in a room, a main gangway must be reachable within a

maximum of 15 minutes.




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In case of fire, roofs often contribute significantly to fire spread; damage will

depend on which structural fire prevention measures were implemented for the

roof.

Regarding fire propagation in case of a fire from below, then the following

failure mechanisms are typical:

The “Durchbrand” burn- through. This is the worst case, with fire spreading

on top of the roof, followed by the spread of fire down into other areas

through existing roof openings.

Failure of the load-bearing roof shell by slipping from the supports, for

example with large spans.

Fire propagation below the roof.

Fire propagation within the roof shell. This is very dangerous because it

will not be seen from below. It becomes very critical when the fire services

are fighting at the fire source and suddenly it begins to burn behind them.

Table A.1 Fire compartment sizes

Maximum fire compartment size (m²)

Safety category Without fire resistancerequirement

“R0”

With fire resistancerequirement

R30

K1Without requirements

1800* 3000

K2Fire detection

2700* 4500

K3Rescue service

3200 - 4500* 5400-7500

K4Fire suppression

(Sprinkler system)10000 10000

* heat extraction area 5% and building width 40m

The simplified method is based on the relationship between the permitted

surface area of the fire compartment and the safety category, the number of

storey and the fire rating classification of the components.

The surface area is given in Table A.1 and is well within extreme safety

measures.

For industrial buildings with an existing sprinkler system (safety category K4),

a maximum fire compartment surface area of 10000 m² can be realized without

requirements for the fire resistance of structural components.

Without any fire protection requirements, surface areas up to 1800 m² can be

left unprotected.

For industrial buildings which cannot be evaluated using the simplified

procedure, the entire verification procedure will be based in accordance with

DIN 18230-1.




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First, the equivalent fire duration is determined using this method. With the

equivalent fire duration, a relationship between the incendiary effect of a

natural fire and the “Einheitstemperaturzeitkurve” (ETK standard temperature

time curve) is generated. The equivalence refers to the maximum temperature

of structural components under a natural fire.

Once the equivalent fire duration has been determined, two different methods

are available.

The first method is to determine the maximum floor surfaces using Table A.2.

No requirements for fire resistance of structural components are needed when

using this table.

The second method requires somewhat more effort. First, the maximum floor

surface is calculated using a formula. In this procedure, the fire resistance

rating of the structural components has to be proven. This is done with the

necessary fire resistance.

Table A.2 Maximum floor area (m2 ) according to safety category and

equivalent fire duration

Equivalent fire durationSafety category

15 30 60 90

K1Without requirements

9000* 5500* 2700* 1800*

K2Fire detection

13500* 800* 4000* 2700*

K3

Rescue service

1600-22500* 10000-13500* 5000-6800* 3200-4500*

K4Fire suppression

(Sprinkler system)

30000 20000 10000 10000

Minimum heat extractionarea

1 1 3 4

Maximum building width 80 60 50 40

In Table A.2, the maximum admissible floor surface can be defined with

reference to its safety category and the equivalent fire duration. In addition, the

corresponding heat extraction surface can be identified, indicated as a % of the

floor surface and the corresponding maximum width of the building.

Using the second method for the entire verification procedure, the maximum

floor area (m²) is calculated using the base value for the surface area of

3000 m² and factors F1 to F5.

A = 3000 F1 F2 F3 F4 F5

where:

F1 the equivalent fire duration

F2 the safety categoryF3 : the height of the lowest floors




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F4 : the number of storey

F5 : the type of floor openings

The sum of the total surface area shall not exceed 60000 m².

According to the Table A.2, when the simplified procedure is used for structural components without requirements, the result is a maximum possible

surface area of 10 000 m².

When using the full verification procedure according to this table, a maximum

surface of 30000 m² is possible. When using the full verification procedure in

addition to the fire resistance calculation, then a 60000 m² surface area is

possible.

Under very special conditions, even larger surfaces, up to 120000 m² can be

achieved.

Example:

The procedure and possibilities associated with MIndBauRl can best be shown

and explained by an example:

Building parameters

Length: 100 m

Width: 50 m

Average height: 6 m

Size: 5000 m²

Number of storey: 1

Openings in the roof: 135 m²

Doors, windows: 132 m²

Fire load: qR = 126 kWh/m²

Automatic fire alarm systems: Safety category K2

No internal fire walls

The first possibility is the simplified method according to Table A.1. The

industrial building must be equipped with an automatic sprinkler in order to

meet the above conditions.

In order to apply fully the full verification method, the equivalent fire duration

must first be determined. In this case, the heat extraction factor w is needed.

The heat extraction factor is determined by taking into account the related

opening surfaces. The related opening surfaces are auxiliary values. This is

simply a question of dividing the roof openings by the ground surface and then

the wall openings by the ground surface.

Determination of the related horizontal opening surface ah:

ah = Ah / A = 135 m² / 5000 m² = 0,027




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Determination of the related vertical opening surface av:

av = Av / A = 132 m² / 5000 m² = 0,026

The values of the related opening surfaces are introduced in Figure A.1 and the

value w0 can be defined. In Figure A.2, the height of the hall is considered.

vertical opening area av

horizontal opening area ah

Figure A.1 Factor w 0 according to opening areas

height of the hall (m)

Figure A.2 Factor w according to height of the hall

The heat extraction value of the buildings is:

w = w0 = 1,70 1,0 = 1,70

The equivalent fire duration (t ä) is based on the following factors: the fire loaddensity, the heat extraction factor and a factor c which takes into account the




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heat extraction surface of the peripheral construction elements. In this example

c is given, for simplicity, the worst value.

t ä = qR c w = 126 0,25 1,70 = 54,0 min

Through interpolation in Table A.2, in the safety category K2 for an equivalentfire duration of 54 minutes, a maximum surface area of 4800 m² can be

defined. At this point, some additional work by the designer could be useful in

reviewing the input data. Is the fire load case too high? What will happen when

the opening surfaces are modified and the ground floor is also modified at the

same time? Alternatively, what about the surfaces? Can the surface be reduced

by 200m²? The onus is on the designer to present and explain the different

opportunities to the client and to list the comparison costs.

The second possibility using the full verification method is more precise. The

maximum floor surface is calculated using the basic value for the surface of

3000 m² times factors F1 to F5. The factor values are taken from tables of DIN18230-1 and do not need to be determined.

According to table 3 of DIN 18230-1 the factor F1 is: 1,9




According to table 7 of DIN 18230-1the factor F5 is: 0,7.

Inserted into the formula:

A = 3000F1F2F3F4F5 = 3000 1,9 1.5 1,0 1,0 0,7

A = 5989 m².

In this method, the fire resistance classification of the structural components

has to be calculated with the following equation:

Required fire resistance duration t f = t ä L

The design of the fire resistance duration includes the following factors:

the equivalent fire duration of 54 minutes

the safety factor of 0,6 according to Table 2 of DIN 18230-1, and

the factor alpha L takes into account the fire related infrastructure of 0,9

according to Table 4 of DIN.

Hence: t f = 54 0,6 0,9 = 29,16 min => R30




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Table A.3 Summary of maximum compartment sizes

Area gi ven by simp li fied method (m2 )

Safety category Without fire resistance

requirement With fi re resistance requirement

K1

K2 2700 4500

K3 5400-7500

K4 10000

R0 R30

A comparison of these methods, the options available and responsibilities of

the designer, can be seen in table A.3. In order to contain the industrial

building in one single fire compartment without requirements for the load-

bearing structure, it is necessary to install an automatic sprinkler system when

using the simplified method. When using the full verification method and

respecting the given conditions, a fire compartment of 4800 m² is possible. To

achieve one fire compartment of 5000 m², at least one plant fire service must

be present.

With a fire resistance requirement of R30 for the load bearing structure, at least

one plant fire service is required for the simplified method (according to the

table). With a fire detector system, however, only one fire compartment area of

4500 m² is possible. With the full verification method, a fire compartment

surface of 5989 m² is possible.

Based on the results of the different methods, the designer’s task is clearly

defined. He should not only develop one fire protection concept, but has to

demonstrate alternative and more economical procedures to the client in

relation to the various production processes.





Part 8: Building Envelope









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FOREWORD

This publication is part eight of the design guide, Single-Storey Steel Buildings.




Part 3: Actions















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ContentsPage No

FOREWORD iii

SUMMARY vii 1 INTRODUCTION 1

1.1 The building envelope 1 1.2 The functions of building envelope 3

2 TYPES OF METAL CLADDING SYSTEMS 4 2.1 Single-skin trapezoidal sheeting 4 2.2 Built-up double skin cladding 5 2.3 Insulated (composite or sandwich) panels 8 2.4 Standing seam systems 9 2.5 Structural liner trays 10 2.6 Structural deck and membrane roof systems 10

3 SPECIFICATION OF THE CLADDING 12 3.1 Weathertightness 13 3.2 Building appearance 14 3.3 Thermal performance 15 3.4 Interstitial condensation 18 3.5 Acoustics 18 3.6 Fire performance 20 3.7 Durability 21 3.8 Structural performance 21

4 COLD ROLLED SECONDARY STEELWORK 24

4.1 Purlin and side rail options 24 4.2 Loading 30 4.3 Deflections 31 4.4 Purlin and side rail selection 31 4.5 Restraint provided to the rafters and columns 32 4.6 Restraint of purlins and cladding rails 33

5 HOT-ROLLED SECONDARY STEELWORK 35

REFERENCES 37




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8 - vii

SUMMARY

This publication provides guidance on selection of the building envelope for single-storey buildings. The building envelope is generally formed of secondary steelwork(often cold-rolled steel members) and some form of cladding. In addition to providing a

weathertight barrier, the envelope may also have to meet thermal, acoustic and fireperformance requirements. In some arrangements, the building envelope may have animportant structural role in restraining the primary steel frames.

The document describes the common forms of cladding for single storey buildings, andoffers advice on how an appropriate system is specified. The document also describesthe systems of secondary steelwork that support the cladding.




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8 - 1

1 INTRODUCTION

Metal cladding systems provide an efficient, attractive and reliable solution to

the building envelope needs of single storey buildings (steel, concrete or woodframed structures). Over the years, these systems have evolved from the singleskin metal cladding often associated with agricultural buildings to highlydeveloped systems used in industrial, retail and leisure applications. However,as with all construction components, the ability of the cladding to satisfy itsfunctional requirements is dependent on its correct specification andinstallation and, equally as important, on its interaction with the other elementsof the building envelope and structure.

This publication provides guidance relating to the secondary structures andbuilding envelope types used in single storey buildings. Description is given of

the common types of profiled metal cladding systems currently used in Europe. These systems include insulated panels, built-up systems, deck and membrane,and liner trays. Guidance is also given on key issues that should be consideredwhen specifying either the building envelope or its supporting structure.

Reference is made to a selection of technical documents published by TheMetal Cladding and Roofing Manufacturers Association (MCRMA). Thesetechnical documents provide comprehensive guidance on various associatedtopics, which are applicable throughout Europe and can be readily downloadedfrom www.mcrma.co.uk. Additional information can also be found on theFrench language website Acier Construction at

http://www.acierconstruction.com

Guidance has been included in this document which considers the restrainingaction of the secondary steelwork to primary steelwork and the restraintprovided by cladding sheeting to secondary steelwork. However, in certaincountries within Europe (e.g. in France), this restraining behaviour cannot beutilised, and a footnote has been added highlighting where this is the case.

1.1 The bui lding envelope The principal components of a modern metal-clad industrial type building are

shown in Figure 1.1.




8 - 2

2

1

3

4

5

1 Profiled steel roof cladding

2 Wall cladding

3 Purlins

4 Side rails

5 Primary steel frame

Figure 1.1 Principal building components

There are essentially three layers to the structure:

1. The primary steel frame, consisting of columns, rafters and bracing. Theexample shown in Figure 1.1 is a portal frame, but the guidance given inthis publication is also applicable to other types of structure.

2. The secondary steelwork, consisting of side rails for the walls and purlinsfor the roof. These members serve three purposes:

- To support the cladding

- To transfer load from the cladding to the primary steel frame

- To restrain the primary steel frame members (see Section 4.5 onlimitations on such use) .

3. The roof and wall cladding, whose functions include some or all of thefollowing:

- Separating the enclosed space from the external environment

- Transferring load to the secondary steelwork

- Restraining the secondary steelwork

- Providing thermal insulation

- Providing acoustic insulation

- Preventing fire spread- Providing an airtight envelope




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- Providing ventilation to a building (ventilated or unventilated roofs andwalls).

The cladding will also normally include ancillary components such aswindows, rooflights, vents and gutters.

As an alternative to the layout shown in Figure 1.1, some types of claddingmay be installed directly to the primary steelwork without the need for purlinsor cladding rails. Examples of this type of construction are deck and membranefor roofs and liner trays for walls. Where such solutions are chosen, thecladding must be designed to:

- Span directly between the rafters, roof beams or trusses. This is usuallyachieved by the use of deep profiled decks or trays, but where these areinsufficient for the required span, intermediate supports in the form of secondary beams or hot-rolled purlins will need to be installed.

- Restrain the primary steel members. Structural decks and liner trays, if

fastened correctly, should be able to provide sufficient lateral restraintto the outer flange of the supporting rafter or column. This should allowthe columns and rafters to be designed as fully restrained under gravityloads or positive wind pressure. However, additional restrainingmembers will need to be included in the structure in order to provideintermediate restraint against wind suction (uplift on the roof).

1.2 The functions of bui lding envelopeAll buildings, whatever their use, must provide a controlled internal

environment that is protected from the variable and uncontrollable externalclimate. The nature of the internal environment will depend on the intended useof the building and this will naturally determine the requirements for thebuilding envelope.

Generating and maintaining a controlled internal environment is a complexprocess, requiring a combination of mechanical and electrical services to heatand/or cool the building and a well-designed building envelope to regulate theheat gain and loss. The design of the building envelope is an important factorin specifying the Mechanical and Electrical (M&E) plant and in determiningthe energy performance of the building. With pressure to reduce energy

consumption now being placed on the construction industry across Europe, thebuilding envelope has never before been under such close scrutiny.

In addition to forming the building envelope, the roof and wall cladding mayalso have an important role to play in the structural performance of thebuilding, by providing restraint to the secondary steelwork againstlateral-torsional instability. Where such restraint is assumed (as is often thecase in the purlin and side-rail manufacturers’ load/span tables), it is essentialthat the cladding is capable of providing this restraint in practice.




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2 TYPES OF METAL CLADDING SYSTEMS

There are a number of proprietary types of cladding that may be used in

industrial buildings. These tend to fall into a few broad categories as describedin this Section.

The steel sheet is generally coated with a zinc or zinc-aluminium alloy in ahot-dip process. The top coating is an organic coating to provide an attractivefinish, typically based on Polyvinyl–chloride (PVC or Plastisol),Polyvinylidene–fluoride (PVDF or PVF2), Polyester or Polyurethaneformulations. Aluminium cladding sheets are also available.

For hot-dip galvanised sheeting, typical design lives are shown in Table 2.1.

Table 2.1 Typical design life for coated steel sheet

Coating Design life (years)

PVC – 200 microns 10 – 30

PVC – 120 microns 10 – 25

PVDF or PVF2 – 25 microns 10 – 15

Polyester – 25 microns 5 – 10

Polyurethane – 50 microns 10 – 15

2.1 Single-skin trapezoidal sheeting

Single-skin sheeting is widely used in agricultural and industrial structureswhere no insulation is required. The sheeting is fixed directly to the purlins andside rails as shown in Figure 2.1. The cladding is generally made from 0,7 mmgauge pre-coated steel with a 32 mm to 35 mm trapezoidal profile depth.

1

1 Slope

Figure 2.1 Single-skin trapezoidal sheeting




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2.2 Buil t-up double skin cladding This common type of cladding consists of a metal liner, a layer of insulationmaterial, a spacer system and an outer metal sheet, as illustrated in Figure 2.2.

The span of such systems is limited by the spanning capability of the claddingsheets, which is typically in the order of 2 m to 2,5 m depending on the appliedloading. Built-up cladding systems must, therefore, be supported by secondarysteelwork (purlins or side rails). As the name suggests, these systems are builtup from their constituent parts on site.

21

6

54

3

1 Weather sheet2 Slope

3 Bar

4 Liner sheet5 Bracket

6 Insulation

Figure 2.2 Built-up roof cladding

2.2.1 Liner sheet

The liner sheet serves several purposes:

It supports the thermal insulation

It provides an airtight layer

It provides restraint to the purlins.

Liner sheets are usually manufactured from cold formed pre-coated steel oraluminium and possess a shallow trapezoidal profile (i.e. height 18 mm to20 mm is illustrated in Figure 2.3). For steel liners, the sheet thickness isusually either 0,4 mm or 0,7 mm, while aluminium liner sheets are slightlythicker at 0,5 mm or 0,9 mm. The choice of liner will depend on the requiredspanning capability, the cladding installation method and the acousticrequirements of the cladding. Where required, the acoustic performance of thecladding, in particular its ability to absorb sound and minimise reverberation,may be enhanced by the use of a perforated liner sheet.




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1

2

1 Thickness (0,4 – 0,7 mm)

2 Profile height (18 – 20 mm)

Figure 2.3 Liner sheet prof ile

The shallow liner sheets are not strong enough to walk on, so it is essential thatthe insulation, spacer system and weather sheet are installed from boards oraccess platforms, as illustrated in Figure 2.4. However, they do provide anon-fragile barrier against falling once they have been fully fastened. Wherewalking access is required, it is common practice to replace the shallow linerprofile with a more substantial sheet (i.e. 32 mm to 35 mm trapezoidal profilein 0,7 mm gauge steel).

Figure 2.4 Liner sheet installation progressing into the span of the purlins.

2.2.2 Insulation

The primary function of the insulation layer is to provide a barrier to the flowof heat between the interior of the building and the external environment. Thethickness of the insulation layer in roof and wall assemblies has increasedsignificantly in recent years from approximately 80 mm in the 1980s to valuesapproaching 200 mm in 2009. Further increases in thickness are expected overthe next few years as the regulations on energy use in buildings become moreonerous.

The most common form of insulation in built-up cladding systems is mineralwool quilt, which is favoured due to its light weight, low thermal conductivity,ease of handling and relatively low cost. Rigid mineral wool slabs areavailable, but are less deformable than mineral wool quilts, giving rise to the




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potential for air gaps between the insulation and the profiled metal sheets.Rigid mineral wool slabs are also much heavier than mineral wool quilts, withconsequences for the loading on the supporting steelwork and manual handlingon site.

2.2.3 Spacer system The primary function of the spacer system is to support the weather sheet at therequired spacing from the liner sheet. The components of the system must,therefore, possess sufficient strength and stiffness to safely transmit therequired loading through to the purlins, without excessive deformation. Acommon form of spacer is a bar and bracket system, as shown in Figure 2.5.

The system consists of cold formed steel bars, which provide continuoussupport to the weather sheet, supported at intervals by steel brackets firmlyattached to the purlins through the liner. Many bar and bracket systems alsoincorporate plastic pads (which act as thermal breaks) in order to minimisethermal bridging. Other types of spacer systems are also available, for example

Z spacers supported on thermally insulating plastic blocks.

1

2

3

4

1 Bar

2 Bracket

3 Sway bracket

4 Purlin

Figure 2.5 Bar and bracket spacer system

2.2.4 Weather sheet

The outer sheet of a double skin built-up cladding system is known as the

weather sheet. As the name suggests, its primary function is to protect thebuilding from the exterior climate by forming a weather-tight envelope.However, the weather sheet should also be regarded as a structural element, asit plays an important role in transferring externally applied loads (e.g. fromwind, snow and foot traffic) through to the other cladding components,secondary steelwork and the primary load-bearing frame.

The weather sheets are usually made from either steel or aluminium and areavailable in a wide variety of finishes and colours. Steel weather sheets aremanufactured from pre-coated steel coil. Aluminium weather sheets areavailable in a mill finish or in a range of painted finishes. Detailed

requirements for the weather sheets for roof and wall cladding applications aregiven in EN 14782[1].




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2.2.5 Fasteners

A wide variety of proprietary fasteners are available, which where required,can be watertight. Most fasteners used for metal cladding applications are bothself-tapping and self-drilling, although screws which are only self-tapping arealso available for use in pre-drilled holes. Fasteners can be used to connect

sheeting to supporting steelwork (or other materials) or to connect adjacentsheets. For most fastener applications, a choice between plated carbon steel andstainless steel (typically grade 304 austenitic stainless steel is used) is made.Visible fasteners have the option of factory coloured plastic heads to suit theweather sheet. Further information describing these and other fasteners (e.g.secret fix fasteners) is available from MCRMA Technical Paper No 12Fasteners for Metal Roof and Wall Cladding: Design, Detailing andInstallation Guide[2].

2.3 Insulated (composite or sandwich) panelsInsulated roof and wall cladding panels consist of a rigid layer of insulationsandwiched between two metal skins, as shown in Figure 2.6. The result is astrong, stiff, lightweight panel with good spanning capabilities due tocomposite action in bending. These panels are commonly used on industrialbuildings and retail ‘sheds’ in place of the built-up cladding described inSection 2.2. In this case, the panels span between cold formed purlins or siderails, which in turn span between the primary frame members. However, forcommercial buildings, where the secondary steelwork is not needed forrestraint purposes, it is quite common for composite wall cladding panels tospan directly between the columns.

Standing seam and through-fixed systems are available, with either atrapezoidal weather sheet and shallow profiled liner, as shown in Figure 2.6, ortwo flat / micro-ribbed sheets. Profiled composite panels are used for roofs toallow rainwater to run off without penetrating the fastener holes, while flatpanels are favoured for walls due to their better appearance.

1

2 1 Insulation

2 Metal sheets

Figure 2.6 Insulated panel

Unlike built-up systems, there is no need for a spacer system, as the rigidinsulation is strong and stiff enough to maintain the correct spacing of thesheets. Any loads applied in the plane of the cladding (e.g. down-slope loads




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on a pitched roof) are transferred from the external sheet through the twoadhesive bonds and the layer of insulation to the internal sheet and thesupporting structure.

Polyisocyanurate (PIR) is the most common insulation material used in

foam-insulated panels. PIR expands rapidly when sprayed onto the metalprofile and bonds to it without the need for an adhesive. This property makes itideally suited to the type of continuous manufacturing process employed by thelarger manufacturers of foam-filled panels. Alternatively, rigid slabs of mineralwool or other insulating materials may be bonded to the metal sheets using anadhesive. This method is commonly used for flat-faced wall panels.

2.4 Standing seam systems‘Standing seam’ or ‘secret fix’ systems use a specially designed profile for theweather sheet, which incorporates a clipped joint between adjacent sheets. Thiseliminates the need for exposed fasteners and improves the weather tightnessof the cladding system. Consequently, standing seam systems may be used onvery low roof slopes (down to 1º compared to 4º for systems with exposedfasteners). Insulated panel systems are also available with a standing seam jointin the weather sheet. Standing seam sheeting can be manufactured from steel oraluminium.

A typical standing seam system is shown in Figure 2.7.

5

3

4

1 2

1 Outer sheeting

2 Slope

3 Standing seam clip

4 Inner sheeting

5 Insulation

Figure 2.7 Standing seam roof cladding

The disadvantage of this system is that significantly less restraint is provided tothe purlins than with a conventionally fixed system. Nevertheless, a correctlyfixed liner will provide adequate restraint.

Further information on standing seam cladding systems may be obtained fromMCRMA Technical Paper 3 Secret fix roofing design guide[3] and also fromECCS-TC7 Publication 41Good practice in steel cladding and roofing[6].




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2.5 Structural liner traysStructural liner trays are a popular alternative to composite wall panels. Theycomprise a deep structural profile into which a slab of insulation is inserted onsite. The assembly is completed with the addition of an external profiled metalsheet, as shown in Figure 2.8. Unlike built-up systems, liner trays span directlybetween the main structural columns, thereby removing the requirement forsecondary cladding rails. This is possible because of the depth of the liner trayprofile and its resulting bending stiffness. The lack of secondary steelworktherefore can have clear advantages in terms of the speed and cost of theconstruction process and installation tolerances.

However, consideration, should be given to thermal bridging that can existwith liner trays. This issue may be partially overcome by placing an additionallayer of rigid insulation on the outside of the tray.

Where plastic design of portal frames is a common design approach, theabsence of side rails can create issues when attempting to provide restraint tothe inside flange of the columns (e.g. in the hogging region of a portal frame),since traditional knee bracing cannot easily be attached to the liner tray profile.

Structural liner trays can also be specified with perforations where improvedacoustic performance is required.

1

2

3

1 External profile sheeting

2 Insulation

3 Liner tray

Figure 2.8 Structural liner tray cladding systems

2.6 Structural deck and membrane roof systemsStructural deck and membrane systems provide a long spanning alternative tobuilt-up cladding on cold formed purlins and are especially popular on ‘flat’ orvery low pitch roofs on which a waterproof membrane is required. The roof construction comprises a trapezoidal profiled metal deck of sufficient depthand gauge to span directly between the rafters, roof beams or trusses. Acommon metal deck typically has a profile height of 100 mm and a steel




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thickness between 0,75 mm and 1,0 mm. The deck supports a layer of rigidinsulation on top of which the waterproof membrane is placed, as shown inFigure 2.9. The use of a high density rigid membrane permits the loads fromfoot traffic and snow to be carried through the insulation layer to the structuraldeck without the need for an external metal sheet or spacer system. The deck is

capable of restraining the top of the beam or truss, making it ideal for buildingdesigns that have simply supported roof structures. However, structural decksare not suitable for plastically designed portal frames due to the need to restrainthe inner flange of the rafter in the hogging region.

1

2

3

4

5

6

1 Structural deck

2 External membrane

3 Rigid gypsum roof boards

4 Insulation

5 Vapour retarder

6 Supporting steelwork

Figure 2.9 Structural deck and membrane cladding system




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3 SPECIFICATION OF THE CLADDING

The specification of roof and wall cladding has implications well beyond the

aesthetics and weathertightness of the building. The choice of cladding canaffect many aspects of the building’s performance, from its construction rightthrough to its eventual demolition and disposal. Indeed, the fitness for purposeof the whole building could be compromised if sufficient care is not taken whenspecifying the cladding. Listed below are the factors that should be taken intoconsideration when specifying profiled metal cladding systems. Further detailson the principal technical considerations are given in Sections 3.1 to 3.8.

Weathertightness

Strength and rigidity

Thermal insulation Control of condensation

Control of thermal movement

Sound insulation

Fire resistance

Appearance

Durability

Cost

Daylighting

External attachments

Lightning protection

Design detailing

Maintenance, remedial work and renewal.

Control of air leakage.

Minimum performance requirements for a number of these factors are laiddown by legislation in Europe. Other factors, such as appearance and daylighting, may not seem to be as critical from an engineering viewpoint, butmight be crucial to the success of the building in terms of the well-being of theoccupants and the acceptance of the building by the local community. It shouldnot be forgotten that the cost of the insulated cladding in a typical commercialor industrial building is usually a significant proportion of the overallconstruction cost, so decisions related to the cladding could influence theeconomic success or failure of the project. The cladding also has a significantimpact on the operational energy requirements and, therefore, the operatingcosts of the building in service, specifically heating, cooling and lighting.




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3.1 Weathertightness The primary function of the cladding system is to provide a weathertightbuilding envelope, suitable for the intended use of the building. With this inmind, the cladding specifier must give careful consideration to the selection of the cladding components and the detailed design of the system. The location of the building, its orientation and the external climate should all be consideredwhen specifying the cladding. The satisfactory performance of the system alsodepends on the correct assembly of the components in the factory and/or onsite.

In general, roofs are at greater risk of leakage than walls, and this risk increasesas the roof pitch decreases. This is an important factor in the design of modernnon-domestic buildings, since many have low pitch or flat roofs in order tominimise the volume of empty roof space. Not all types of roof cladding aresuitable for use on low pitch roofs. Specifiers must, therefore, pay careful

attention to the minimum pitch recommended by the manufacturers, togetherwith the published guidance on detailing and installation.

Trapezoidal metal roof sheets with through fix fasteners are generally suitablefor slopes of 4º (7%) or steeper. This 4º limit is critical to the performance of the cladding and should take into account deflections in the supportingsteelwork and localised cladding deformations that may lead to ponding.Where the primary steelwork is precambered to off-set the deflections due topermanent actions, great care must be taken to ensure that excessive precamberdoes not result in local high points, as these could also cause ponding. Forshallower pitches, down to 1,5º (1,5%), a secret fix system with no exposed

through fasteners, special side laps and preferably no end laps should be used.Secret fix systems may also be used on steeper roofs where increased reliabilityis desired.

For low pitch roofs, ponding is a potential problem that must be considered atthe design stage in order to avoid the deleterious effects of prolonged soakingand the increased loading due to the weight of the water. Where pondingoccurs on rooflights, there is also the additional problem of the water leavingdirt deposits as it evaporates.

Side and end laps in profiled sheeting are weak points in the building envelope,

where the wind and rain could potentially penetrate the cladding. The designand construction of the laps is therefore critical to the weathertightness of thecladding system. End laps typically consist of two continuous butyl sealantstrips, which are compressed to form a weathertight seal by the clampingaction of the fasteners. The pitch of fasteners required to achieve a proper sealwill depend on the profile geometry, but one fastener per trough is common. Atypical side lap between trapezoidal sheets is formed by overlapping theprofiles with a strip of butyl sealant positioned on the weather side of thefastener to provide a weather-resistant seal. The side laps should be stitched at500 mm centres or closer using steel stitcher fasteners. Further information onside and end lap details is given in MCRMA Technical Paper No. 6 Profiled

metal roofing design guide[4]

and Technical Paper No. 16 Guidance for theeffective sealing of end lap details in metal roofing constructions[5]. Referencecan also be made to ECCS-TC7 Publication 41Good practice in steel claddingand roofing[6].




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3.2 Building appearance The choice of wall and roof cladding can have a significant impact on theappearance of a building. The following factors are particularly important:

Profile shape

Colour

Fasteners.

The profile shape can have a significant impact on the appearance of a buildingdue to its effect on the perceived colour and texture of the cladding (caused bythe reflection of light). The orientation of the cladding (ribs horizontal or ribsvertical) will also influence the appearance of the building, due to the effects of shadow and reflection. A potential disadvantage of horizontal ribs is that theytend to suffer from an accumulation of dirt over time, unless the cladding iscleaned regularly. Where the location and function of the building demand a

smooth flat exterior, insulated wall panels with flat facing sheets may be used,however, it should be noted that any defect on the surface will be readilynoticeable.

The steel from which profiled cladding sheets are made is available pre-coatedin a wide range of colours and textures, allowing architects to choose a finishthat best suits the location and function of the building. In choosing the finish,the architect should bear in mind the influence of the profile shape on theoverall appearance by making an allowance for the effects of reflection andshadow on the perceived shade of colour.




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Figure 3.1 Typical wall cladding with a mix of flat panels and profiled sheeting

The overall appearance of the building can also be affected by the choice of fasteners, especially on wall cladding or on steeply pitched roofs. Claddingspecifiers should, therefore, give careful consideration to the size, shape, colourand locations of the fasteners and washers. Fasteners with factory colouredplastic heads are available to match the colour of the weather sheet. Where

exposed fasteners are considered detrimental to the appearance of the building,the architect may consider the use of secret fix insulated panels or standingseam systems in which all fasteners are hidden from view. Further informationon fasteners is available from MCRMA Technical Paper No 12 Fasteners forMetal Roof and Wall Cladding: Design, Detailing and Installation Guide[2].

3.3 Thermal performance3.3.1 Energy consumpt ion

The increase in public awareness of global climate change and the association

with human activity has placed energy consumption and carbon dioxideemissions high on the political agenda. Under the terms of the Kyoto Protocol,European countries are now legally bound to reduce their carbon dioxide




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emissions and meeting this obligation will require significant changes in manysectors of industry, especially construction.

A significant proportion of carbon dioxide emissions in Europe is related to theoperational energy requirements of buildings (heating, lighting, ventilation

etc.). This issue is addressed by European Directive 2002/91/EC: Energyperformance of buildings[7]. Although many factors influence a building’senergy efficiency, the thermal performance of the building envelope issignificant. Consequently, it has been sought to reduce energy consumption byimproving the thermal performance of the cladding and associated components.

The main sources of heat loss through the building envelope are shown inFigure 3.2.

1 2 3 1 Thermal bridge (metal spacer)

2 Thermal transmittance through insulation

3 Air leakage through joints

Figure 3.2 Main sources of heat loss through the building envelope

3.3.2 Thermal transmittance

Thermal transmittance through the building envelope can be a significantsource of energy loss within a building, especially if there is insufficientinsulation. One measure of thermal transmittance is the “U-value”, which isdefined as the rate of heat transfer through an element of the building envelope(e.g. a wall, window, section of roof or rooflight) per square metre. The SI unitfor the U-value is W/m2K. For an individual component such as a claddingpanel, the elemental U-value depends on the conductivity and thickness of theinsulation, the profile shape and the presence of thermal bridges. Cladding andinsulation manufacturers usually quote U-value for their products for a range of insulation thicknesses. Alternatively, the U-value of a given built-up of envelope may be calculated using software.

National regulations generally specify maximum U-values. These are often theweighted average (or similar “overall” figure) for the whole of the roof or wall,with maximum values for individual elements such as doors. The individualelements tend to have much higher U-values than the cladding.

Typical limiting U-values are shown in Table 3.1.




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Table 3.1 Limi ting U-values

Element Area weigh ted average

( Wm-2 K

-1 )

Wall 0,35

Roof 0,25

Window 2,2

Pedestrian door 2,2

Roof ventilator 6

Over recent years, the drive to improve the energy performance of buildingshas resulted in a significant reduction in the U-values for building envelopeelements, resulting in a considerable increase in insulation thickness. This hashad important implications for the structural performance of the claddingsystem and its relationship with other structural elements. Of particular concern

to the structural engineer are the increased depth and weight of the claddingand its ability to adequately restrain the purlins or side rails. Inevitably thetrend will continue towards improved thermal efficiency. However, thediminishing returns obtained from further reductions in U-values means that infuture more emphasis is likely to be placed on airtightness and the performanceof mechanical services, rather than ever increasing insulation thicknesses.

While some countries have adopted the U-value as the preferred means of quantifying the performance of the envelope, elsewhere the chosen parameteris the R-value or thermal resistance. The R-value is simply the reciprocal of theU-value and the points noted in the preceding paragraphs are equally

applicable in these countries.

Typical U-values for different cladding systems are shown in Table 3.2.

Table 3.2 Typical U-values for cladding

Element U-value( Wm

-2 K

-1 )

Built-up system, 180 mm insulation 0,25

Built-up system, 210 mm insulation 0,2

Composite panel, mineral fibre, 120 mm 0,34

Composite panel, mineral fibre, 150 mm 0,27

Composite panel, PIR, 60 mm 0,33

Composite panel, PIR, 100 mm 0,20

3.3.3 Thermal bridges

Thermal bridges are areas or components within the roof or wall claddingassembly whose thermal insulation properties are lower (often much lower)than those of the surrounding material, thereby permitting local high heat flowsthrough the building envelope. A common example of a thermal bridge wouldbe an all-metal spacer in a built-up cladding system. In general, all metalcomponents will act as thermal bridges, because of their high thermalconductivity, unless specific measures are taken to interrupt the heat flow byintroducing a layer of thermal insulation. Thermal bridging increases the heat




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loss from a building, thereby increasing the operational energy requirement. Itcan also lead to a reduction in the internal surface temperature of the cladding,causing condensation to form under certain conditions.

3.3.4 Airtightness

The airtightness of a building is central to the requirements of the buildingregulations and is likely to become even more important as architects strive toimprove the thermal performance of the building envelope without significantincreases in insulation thickness. The airtightness of a building is quantified interms of its air permeability, which is defined as the volume flow rate of air persquare metre of building envelope and floor area at a given pressure. Themaximum permissible air permeability for a given building will depend on anumber of factors including the requirements of the building regulations, thespecified CO2 rating for the building and the means by which this rating is tobe achieved (e.g. the architect may specify a very low level of air permeabilityas an alternative to increasing the thickness of insulation). In many countries,

achievement of the specified air permeability must be demonstrated bypost-construction testing.

3.4 Intersti tial condensationInterstitial condensation occurs within the layers of the cladding constructionand is due to warm moist air from within the building penetrating the liner andcondensing on the cold outer sheet and other components. The severity of theproblem will depend on the relative humidity of the air within the building, theexternal air temperature and humidity, and on how well the liner is sealed.

Buildings in cold climates and those containing swimming pools, laundries orother similar applications are most at risk, as are cladding systems thatincorporate a perforated liner and separate vapour control barrier. In extremecases, the condensation could result in corrosion of steel components within theroof assembly or in wetting of the insulation.

Recommendations for avoiding interstitial condensation are usually given inNational Standards.

3.5 Acoustics

Depending on the application, acoustic performance can be an importantconsideration when specifying roof and wall cladding. There are threecategories of acoustic performance to consider, as illustrated in Figure 3.3.

3.5.1 Airborne sound transmission

Where there is a need to limit the passage of sound through the buildingenvelope, the cladding specifier needs to consider the Sound Reduction Index(SRI) of the cladding. The SRI is a measure of the reduction in sound energy(in decibels) as sound passes through a construction at a given frequency. Theacoustic performance of a particular cladding system will depend on the

insulation material, the weather sheet and liner sheet profiles and the method of assembly. Of these, the insulation is the dominant factor, with soft mineralwool insulation giving better sound insulation than rigid board (dependentupon density).




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3

1

2

1 Impact noise from rain

2 Reverberation3 Airborne sound transmission

Figure 3.3 Categories of acoust ic performance

3.5.2 Reverberation

In certain applications, such as offices or residential accommodation, internalacoustic performance might be critical to the functionality of the building. Of particular interest is the reverberation caused by sound waves reflecting off hard internal surfaces, including elements of the building envelope. Typically,the internal finishes of the building will be used to limit reverberation, but

architects may also take advantage of the sound absorbing properties of thecladding insulation layer by replacing the standard liner sheet with a perforatedliner. Where the envelope consists of insulated sandwich panels, it is notuncommon to install a perforated liner and a layer of mineral wool insulationon the inside of the envelope in order to reduce reverberation.

3.5.3 Impact noise

The noise created by the impact of rain or hail on metal roof sheeting cansometimes create a nuisance for the building occupants. Where impact noise isconsidered to be important, it can sometimes be reduced by placing a flexibleinsulation layer directly below the outer sheet to act as a damper.

3.5.4 Noise associated with building services equipment

Consideration should also be given to attenuating noise emanating fromservices equipment. These include providing sound enclosures for noise pronemachinery and/or including equipment supports with dampers. Reduction of noise from services is particularly appropriate in industrial buildings.

National regulations may specify acoustic performance standards in terms of reducing noise coming into a building – but these are often for residentialbuildings. 65 dB is generally considered a suitable indoor noise level inindustrial buildings, whereas 50 to 55 dB is considered a suitable indoorambient noise level for commercial, retail and leisure buildings. For industrialbuildings, noise break-out is usually a greater concern. Local regulations may




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specify acoustic requirements to reduce noise break-out from within a building(for example if the building is sited adjacent to a residential area).

Cladding system manufacturers will be able to provide acoustic performancedata for different constructions, and be able to recommend a system to meet the

specification.

A built-up system comprising an inner and outer sheet of pre-finished steelwith mineral wool insulation generally achieves over 40 dB of sound reduction.Rock mineral wool has a greater density than glass mineral wool, and generallyimproves the sound insulation. Sound insulation can be improved by includinga layer of dense acoustic mineral wool slab, in addition to the insulation quilt.

In general, factory insulated foam filled composite systems are not as effectiveas built up systems, because of the low mass of the foam core and the directcoupling of the inner and outer skins.

The sound reduction index Rw for various systems is shown in Table 3.3. Ahigher index indicates higher sound reduction.

Table 3.3 Sound reduction index for typical cladding systems

Cladding type Sound reduction index R w

Built-up system – with rock wool and acousticinsulation

47

built-up system with rock wool 45

built-up system with glass mineral wool 41

composite panel with mineral wool 31

composite panel with foam 25

single skin 24

3.5.5 Further information

Further guidance is available in MCRMA Technical paper No. 8 Acousticdesign guide for metal roof and wall cladding[8] and also from ECCS-TC7Publication 41Good practice in steel cladding and roofing[6].

3.6 Fire performanceIn general, any concerns about the reaction of cladding to fire are faroutweighed by concerns about the smoke and gas generated by the contents of the building, not the envelope.

Single sheet cladding is considered to contribute significantly to any fire.Single sheet cladding is generally assumed not to make any contribution to fireresistance, although in practice some integrity and resistance will be provided.Single skin sheeting is generally not used on boundaries, when prevention of fire spread to neighbouring structures is important.

Built-up systems that use mineral wool or glass wool insulation are notconsidered to contribute significantly to any fire. Built-up systems may also be




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specified to meet the requirements for external envelope applications.Composite panels that use mineral wool fall in the same category.

Factory insulated composite panels may use polyurethane (PUR) orpolyisocyanurate (PIR). It is generally considered that PIR panels have

improved performance in fire compared to PUR panels. The core of either typeof panel is difficult to ignite. Panels with appropriate joint designs with eitherPUR or PIR filling do not present an undue fire risk, and PUR panels are thestandard core in many European countries.

Polystyrene filled panels present a fire risk, and their use is diminishing.

3.7 Durability All cladding systems suffer a certain degree of degradation over time due tomoisture, atmospheric pollution and UV radiation. However, the cladding

specifier can have a considerable influence on the long term performance of thecladding through careful selection of materials and good detailing. Once inservice, regular maintenance will prolong the life of the building envelope.

The metal from which the weather sheet is made is available with several typesof coating with a wide variety of colours and finishes. Guidance on theexpected design lives of these coatings is available from MCRMA Technicalpaper No. 6 Profiled metal roofing design guide[4] and also from ECCS-TC7Publication 41Good practice in steel cladding and roofing[6]. It is worth notingthat the colour of the coating has a very significant impact on its design life.Light colours reflect thermal radiation more efficiently than dark colours,

resulting in lower surface temperatures and a reduction in the degradationexperienced by the coating.

When detailing the building envelope, particular attention should be given tothe avoidance of water and dirt traps by specifying suitable slopes and end laps.Careful detailing is needed at the external interfaces to avoid the ingress of water and at the internal interfaces to prevent water vapour from within thebuilding entering the cladding assembly (resulting in interstitial condensation).

In order to ensure that the building envelope remains fully functionalthroughout its design life, it is important that it receives regular maintenance,

including inspection, removal of debris, cleaning and repair of damage. Sincemaintenance usually involves access by workmen, often carrying equipment, itis essential that this is allowed for in the design of the building envelope andthe supporting structure. The need for maintenance may be greatly reduced byspecifying a coating for the weathersheet with a ‘maintenance free’ guaranteefor the expected design life of the cladding (typically 20 to 30 years). Suchcoatings can provide significant benefits to the client in terms of whole lifecosts and improved safety.

3.8 Structural performanceMetal cladding systems are required to carry externally applied loads, such assnow and wind loading without deflecting excessively or compromising theother performance requirements. The individual characteristic loads (actions)




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should be obtained from the appropriate part of EN 1991[9], taking into accountthe building geometry and location as applicable. These individual actionsshould then be combined using the appropriate safety factors from EN 1990[10] to obtain the load cases used in design.

3.8.1 ActionsPermanent actions

For most industrial and commercial applications of metal cladding technology,the only permanent action for which the roof cladding needs to be designed isits own self-weight, including the weight of the insulation. Typical weights of insulated panels and built-up cladding systems are given in Table 3.4. Forinformation on specific cladding products, designers should consult thetechnical literature available from manufacturers or suppliers. For wallcladding, it is not normally necessary to consider permanent actions, since theself-weight acts in the plane of the cladding. However, where a rainscreensystem is attached to the outer face of the cladding panel or assembly, it will benecessary to consider the impact of the rainscreen system weight whenspecifying the fasteners.

Table 3.4 Typical cladding system weights

Sheet t hicknessSystem Insulation Depth*

Inner Outer

Weight kN/m

2

Built-up Mineral wool 180 mm 0,4 mm 0,7 mm 0,16

Built-up Mineral wool 180 mm 0,7 mm 0,7 mm 0,20

InsulatedPanels

PIR 80 mm 0,4 mm 0,5 mm 0,12

* The depths chosen in Table 3.1 correspond to a U-value of 0,25 W/m2K for typical cladding

systems using the insulation shown.

Variable actions

In addition to its self-weight, the roof cladding must also be designed for thefollowing variable actions as specified in the appropriate parts of EN 1991:

Access for cleaning and maintenance

A uniformly distributed load due to snow over the complete roof area. Thevalue of this load will depend on the building’s location

Asymmetric snow load and loading due to snow drifts

Wind pressure and suction.

Care should be taken when ‘green’ roofs are specified, as they tend to beconsiderably heavier than traditional metal roofs and, in the case of roof gardens, must be designed for the presence of garden furniture and people.

Wall cladding should be designed for wind loading according toEN 1991-1-4[9]. Positive wind pressure and wind suction will need to beconsidered, with special attention paid to the areas of high wind suction close

to the corners of the building. The wind suction design case is often governedby the resistance of the fasteners connecting the cladding panels or sheets tothe supporting steelwork.




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3.8.2 Deflections

The cladding must be capable of carrying the specified design loads withoutdeflecting excessively, if the other performance requirements such asweathertightness, airtightness and durability are to be achieved. The predicteddeflections are normally calculated for the unfactored variable actions only.

Loading at the construction stage is not normally included in the serviceabilityload cases and is not normally considered when specifying cladding systems.However, care must be taken on site to avoid excessive local deflections,especially those caused by concentrated loads such as foot traffic or stackedmaterials on roof liner sheets, as these could result in permanent damage to thecladding. Typical deflection limits imposed on the cladding are dependent onthe loading regime considered (imposed load only or permanent plus imposedloading), the location (wall or roof) of the structural component and whether abrittle material is present. Deflection limits may be specified by Nationalregulations. Common deflection limits are:

Span/150 for wall cladding, spanning between secondary steelwork

Span/200 for roof cladding, spanning between purlins

Span/180 for purlins or side rails.

3.8.3 Use of safe load tables

The manufacturers of profiled metal sheeting and insulated panels provide safeload tables for their products, which may be used either to select a suitableprofile or, where the profile has already been chosen, to determine themaximum permissible purlin spacing. It is important to note that the load tables

often assume that the loading is uniformly distributed and that safe workingloads are usually specified. If in doubt, specifiers should seek guidance fromthe cladding manufacturers.




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4 COLD ROLLED SECONDARY STEELWORK

For steel portal framed industrial type buildings with low pitch roofs (5 to 10

degrees), the cladding panels or sheets are normally supported by a system of light steel purlins and side rails spanning between the rafters and columnsrespectively. See Figure 4.1 showing secondary steelwork in the roof where thepurlins span between the rafters of the main frame. The primary function of these secondary members is to transfer load from the cladding to the primarysteel frame, including cladding self-weight, wind loads and, for roofs, imposedloads due to snow and maintenance access. The purlins and side rails may alsobe used to provide restraint to the rafters and columns and to transfer horizontalloads into the bracing system.

Figure 4.1 Purlins spanning between rafters in the roof

This Section presents guidance on some of the key issues relating to the use of cold formed purlins and cladding rails.

4.1 Purlin and side rail opt ionsPurlins and side rails are generally cold formed light gauge galvanized steelmembers, supplied as part of a proprietary cladding support system, togetherwith fittings, fasteners and other associated components.

4.1.1 Section options

Purlins and side rails are available in a variety of shapes and a wide range of sizes. The depth of the section typically lies between 120 mm and 340 mm,with the profile thickness varying between 1,2 mm and 3,2 mm. Some of the

more common section shapes are shown in Figure 4.2. Purlins and side rails,because of their high length/thickness values, are typically classed as Class 4




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sections as defined in EN 1993-1-3[11], hence section properties will be need tobe based on effective values (reduced gross properties).

Further information on these sections may be obtained from the manufacturers’technical literature.

1 2 3 4

1 Zed

2 Ultrazed

3 Zeta

4 Sigma

Figure 4.2 Common types of purlin

4.1.2 Purlin and side rail layout opt ions

Most manufacturers produce guidance on typical purlin layouts that areefficient for various situations. These layouts are governed by such aspects asmaximum purlin length (generally not more than 16 m for transport and siteaccess reasons) and the ability to provide semi continuity by the use of sleevesor overlaps for maximum efficiency. The most commonly used layouts areshown in Figure 4.3 to Figure 4.7. Specifiers seeking further information onwhen and how to use a particular layout should consult the purlinmanufacturers for detailed information relating to their specific systems. In anyevent, the purlin manufacturer should be consulted before the layout isfinalised.

Single-span lengths - sleeved system

In sleeved systems, each purlin is the length of a single span but sleeves areprovided at alternate supports so that each purlin is effectively continuousacross two spans (Figure 4.3). At the penultimate support, sleeves are providedat each purlin, to provide semi continuity and additional strength in the endbay. This system is considered to be the most efficient for buildings with baycentres between 5 m and 7 m. Heavier sections can be provided in the end bayif necessary.

1 23

4

1 Sleeved purlin

2 Penultimate support

3 Raf ter4 Sleeve

Figure 4.3 Single-span lengths – sleeved system




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Single-span lengths - butted system

Single-span butted systems have a lower capacity than the other systems, butare simpler to fix either over the rafters or between rafter webs (Figure 4.4).

This layout may be used for small buildings with close frame centres, such asagricultural applications.

1 Single-spanpurlin

2 Rafter

Figure 4.4 Single-span lengths - butted system

Single-span lengths - overlapping system

An overlapping system provides greater continuity and can be used for heavyloads and long spans (Figure 4.5). It is best suited to buildings with a largenumber of bays.

1 Purlin

2 Rafter

Figure 4.5 Single-span lengths - overlapping system

Double-span lengths – non sleeved system

In this system, the double-span lengths are staggered (Figure 4.6). Sleeves areprovided at the penultimate supports to ensure semi continuity. The capacitywill generally be less than for the equivalent double span sleeved system, butdouble-span purlins use fewer components and lead to faster erection. This

system is limited to bay sizes less than 8 m, for reasons of transport anderection of the purlins.




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1 Double-span

purlin2 Penultimate

support

3 Rafter

4 Sleeve

Figure 4.6 Double-span lengths – non sleeved system

Double-span lengths - sleeved system

In double-span sleeved systems, the double-span lengths are staggered andsleeves are provided at alternate supports (Figure 4.7). Sleeves are provided toevery purlin at the penultimate support to ensure semi continuity. A doublespan sleeved system has a slightly higher capacity than the double-spannon-sleeved system and has the advantages of semi continuity at all sleevepositions. This system is limited to bay sizes less than 8 m, for reasons of transport and erection. Heavier purlins can be provided in the end bays, if necessary.

1 Sleeveddouble-spanpurlin

2 Sleeve

Figure 4.7 Double span lengths - sleeved system

4.1.3 The use of anti-sag rods for purl ins

Anti-sag rods are small rods or angles that are bolted or clipped between thepurlins. A typical arrangement is shown in Figure 4.8; other systems are alsoavailable. When used, they are commonly placed either at mid-span or at thirdpoints along the purlin and serve the following functions:

They provide restraint to the purlins against lateral-torsional buckling underwind uplift conditions




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They provide restraint to the purlins in the construction condition (beforethe installation of the cladding)

They provide additional support to the down-slope component of theapplied loads

They help to maintain the alignment of the purlins.

The anti-sag rods are assisted in these functions by eaves beam struts and apexties, both of which are also illustrated in Figure 4.8.

6

10

8

117

9

1

2

3

4

5

1 Purlin

2 Eaves beam

3 Column

4 Eaves beam

5 Column

6 Eaves beam strut

7 Purlin

8 Anti-sag ties (at 1/2 or 1/3 span)

9 Rafter

10 Apex tie

11 Rafter

Figure 4.8 Typical anti-sag ties and eaves beam strut layout

The need for anti-sag rods is dependent on a number of factors, including thechosen purlin section, the spacing between the purlins, the span of the purlinsand the magnitude of the applied loads. Advice on this issue may be obtainedfrom the purlin manufacturers’ technical literature. In some instances, thespecifier may have a choice between the use of anti-sag rods or the selection of a heavier purlin that does not require intermediate restraint or support. There isclearly a trade-off between the cost of a heavier purlin section and the time(and corresponding cost) associated with the installation of additionalcomponents.

Anti-sag rods only provide restraint at discrete locations along the span of thepurlin. The purlins should only be considered to be ‘fully’ restrained under




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gravity loading in the finished condition, when continuous restraint is providedto the compression flange of the purlin by the cladding.

4.1.4 The use of side rail supports for wall cladding

Support for wall cladding is provided by a framework of horizontal cladding

side rails that span between the columns of the building’s primary steelwork.Vertical restraints are connected to the side rails at discrete locations (similar tothe anti-sag rods in roofs). These restraints prevent the occurrence of lateral-torsional buckling (due to bending of the side rails under wind suctionloading) and also prevent the side rails from sagging under the weight of thecladding and its supporting steelwork. These vertical restraints are typicallylight gauge steel sections (tubes, angles or channels) or steel bars/rods.

In order to channel the forces generated in the side rail supports efficiently tothe primary structure (columns) and to prevent the side rails from sagging priorto the installation of the cladding, it is customary to provide a vertical braced

bay arrangement between the lowest two side rails, as shown in Figure 4.10. These bracing members operate in tension, so it is common to use steel wiresrather than cold formed light gauge steel sections. To restrict the forces in thetie wires, it is common practice to restrict the angle of the tie wire to thecladding rail to a minimum of 25° or 30° (refer to the manufacturers’recommendations). With this restriction imposed on the diagonal tie wires, thenumber of side rail supports is predetermined, based on the spacing of the siderails and the spacing of the columns.

For column spacings up to 6 m with a typical side rail spacing of 1,8 m, asingle central vertical restraint will normally be sufficient (see Figure 4.10).

However, for greater column spacings, two or even three vertical restraintsmay be required. In many cases, the uppermost side rail is connected to theeaves beam. This arrangement will reduce the forces in the tie wires, but theadditional force in the eaves beam will need to be considered when thismember is sized. It is also worth noting that, once installed, the cladding willstiffen up the wall substructure and transfer a significant proportion of thevertical load to the columns by diaphragm action. The cladding will also fullyrestrain the side rails against lateral-torsional buckling in the sagging case andwill provide partial restraint in the hogging case.

4.1.5 Cleats

Purlins are attached to rafters using cleats that are usually welded to the rafterin the shop before delivery to site. However, the use of bolted cleats (seeFigure 4.9) is becoming popular due to savings in transportation (as the raftersstack more compactly) and the opportunity they present to adjust the alignmentof the purlins on site (with beneficial consequences for the installation of thecladding). The cleats are often provided by the purlin manufacturer, in whichcase it is likely that they will have been designed specifically for that design of purlin. However, generic bolted cleats made from an angle section or simpleflat plates welded to the rafter may also be used in many cases, eitherunstiffened or stiffened.




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1

2

3

4

5

1 Eaves beam2 Main column

3 Tension wire

4 Anti sag bar(section or tube)

5 Side rail

Figure 4.9 Cleat supporting a purlin using a bolted connection

1

2

1 Purlin

2 Cleat

Figure 4.10 Side rail support for wall cladding

4.2 Loading The purlins and cladding rails need to be designed to carry all of the loadsapplied to them from the cladding and to transfer these loads into the structuralframe. These loads will include the permanent actions due to the weight of thecladding and secondary steelwork together with the variable actions describedin Section 3.7.1. It will usually be acceptable to consider these actions as actinguniformly over the purlins, but account must be taken of high local forces suchas the wind suction forces close to the edges of the building. In addition to thecladding loads, the purlins may also be required to support the weight of




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services or suspended ceilings. The structural engineer responsible forspecifying the purlins will frequently play little or no part in the specificationof the services or ceilings. Nevertheless, it is important that an accurateestimate of these loads is obtained together with the nature of the loading(whether concentrated or distributed), since they could form a significant

proportion of the overall gravity loading on the purlins. Particular care shouldbe taken where the purlins are required to support concentrated loads. Guttersand their supporting structure require special attention, as the loads associatedwith them are often very high. Designers need to consider the weight of thegutters plus that of their contents (water or snow). Specific information on thespecified gutter system should be sought from the gutter manufacturers.

During the construction stage, the purlins may still be required to carrysignificant gravity loads, but without the benefit of any restraint provided bythe cladding. The magnitude of the construction load will depend largely on thecladding installation procedure and the materials, plant and labour used. The

cladding installation sequence, in particular, can have a significant effect on thebuckling resistance of a purlin, due to its influence on the unrestrained lengthof the purlin and the location of the load within the span. It is thereforeessential that the designer takes account of the proposed method of workingwhen specifying the purlins. Preferably, this should be achieved by dialoguebetween the roofing contractor and the designer at the time of the purlinspecification.

4.3 Deflections

The deflection limits for the purlins and side rails are generally governed bythe choice of roof and wall cladding, since the governing factor is the ability of the cladding to deflect without compromising weathertightness, airtightness,non-fragility or any other performance requirement. In general, the greater theflexibility of the cladding, the larger the allowable purlin or side-rail deflection.In this respect, profiled metal cladding systems are far more tolerant of deflections than brittle materials such as masonry. By contrast, windows areoften critical and further guidance should be sought from the glazingmanufacturers.

Excessive deflection under purlin or rail self-weight, or under the action of

construction loads prior to the fixing of the cladding, can lead to difficulties forthe cladding installation. This should be addressed by careful consideration of the likely construction loading and by specifying a method of claddinginstallation that avoids overloading the unrestrained purlins. Gutters areespecially sensitive to deflections, due to the need to avoid backfalls.

4.4 Purlin and side rail selection The major purlin and cladding rail suppliers have invested heavily over manyyears in the development and testing of their systems and all publish designguidance and load/span tables for their products. In many cases, design

software is also available. Thanks to these design tools, the structural engineeris spared the complexities of the design of light steel members and can simplyselect the most suitable section from the available range. However, specifiers




8 - 32

should note that in using the load/span tables they are automatically acceptingthe assumptions made by the purlin and cladding rail manufacturers, includingassumptions regarding the level of restraint provided by the cladding to thesupporting steelwork. If in doubt, the secondary steelwork specifiers shouldcontact the manufacturers for advice on the suitability of the chosen section for

the application in question, taking into account the proposed cladding type andany other circumstances likely to invalidate the manufacturer’s assumptions,e.g. heavy point loads.

4.5 Restraint provided to the rafters and columns The structural efficiency of any steel framed building depends not only on theselection of light and efficient sections, but also on the interaction between theframe members, the secondary steelwork and the cladding system. For thisreason, it is common practice to use the secondary steelwork (the purlins andrails) to restrain the primary steelwork.

It is generally accepted that purlins and rails need not be checked for forcesarising from the lateral restraint of rafters in either roof trusses or portal framesprovided that the following conditions are met:

The purlins are adequately restrained by sheeting

There is bracing of adequate stiffness in the plane of the rafters oralternatively the roof sheeting is capable of acting as a stressed-skindiaphragm

The rafters carry predominantly roof loads.

In certain European countries, the assumption that the secondary members canrestrain the primary frame is acceptable as long as the secondary memberproviding the restraint is connected to a node point of the bracing system. Inother countries, it is presumed that the roof system supplies a sufficiently stiff diaphragm to relax the requirement. In this case, roof bracing is still required,but need not intersect with every secondary member providing restraint. If apurlin or side rail cannot be used with stays (as shown in Figure 4.11) as atorsional restraint, a hot rolled member may be provided to meet thisrequirement.

Ideally, the compression flange of the rafter or column should be laterallyrestrained by direct attachment of the purlins or cladding rails. However, underthe action of wind uplift, or close to the haunches of a portal frame undergravity loading, the inner flange of the member (i.e. the one to which thecladding is not attached) will be in compression and cannot be restraineddirectly by the purlins or cladding rails. In this situation, the frame designer caneither introduce an additional hot-rolled steel member (often a structuralhollow section) to laterally restrain the compression flange or, alternatively, thecompression flange can be effectively held in position by a combination of lateral restraint to the tension flange (provided by the purlins or rails) andtorsional restraint provided by rafter or column stays. Recommendations for

the provision and design of restraints are given in EN 1993-1-1[12], § 6.3.5.2and Annex BB.3.




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Rafter or column stays, as shown in Figure 4.11, may be used to providetorsional restraint to the rafter or column provided that they are connected to asuitably stiff purlin or cladding rail. Thin cold formed steel straps (working asties) are often used, although angles may be used if the stay must work incompression (for example, if a stay can only be provided on one side of a

member).

2

1

3

4

1 Built up orcomposite cladding

2 Cold-rolled eavesbeam

3 Rafter stay4 Column stay

Figure 4.11 Details of column and rafter stay and connection

In order to provide the required level of torsional restraint to the rafters or

columns, the purlins or cladding rails must possess sufficient flexural stiffness.Otherwise, there is a risk that the restraining member will bend and allow therestrained members to rotate, as shown in Figure 4.12. As a rule of thumb, it isnormally adequate to provide a purlin or cladding rail of at least 25% of thedepth of the member being restrained. In practice, this generally means that thepurlins and side rails will be sufficiently stiff for portal frames with spans up to40 m and frame spacings of 6 to 8 m. However, as the span increases relative tothe frame spacing (and the rafter size increases relative to that of the purlins),the purlin stiffness may become insufficient to provide adequate torsionalrestraint and should, therefore, be checked.

Figure 4.12 The importance of adequate purl in stif fness

4.6 Restraint of purlins and cladding rails

Cold formed steel purlins and cladding rails are extremely efficient at carryingloads by bending action, but they are susceptible to failure throughlateral-torsional buckling unless they are adequately restrained. The economic




8 - 34

and safe design of the cladding and its supporting steelwork relies on theinteraction between the individual components that make up the whole system.

Purlins and cladding rails are normally selected from manufacturer’s load/spantables, which are derived from analytical models supported by test data. In

producing their design data, all purlin manufacturers have to make a judgementregarding the degree of restraint that is available from the cladding systemunder gravity and wind uplift conditions. These assumptions are central to thedesign model and can have a significant effect on the design resistance of thepurlin or rail. It is therefore essential that an equal or greater level of restraint isachieved in practice. This will depend on the choice of sheeting and thespacing of the fasteners.

In the gravity load case (or positive wind pressure in the case of a wall),restraint is provided directly to the top flange of the purlin (or side rail) by theliner sheet or insulated panel, as shown in Figure 4.13(a). Built-up cladding

and insulated panels are generally capable of providing sufficient lateralrestraint for the gravity loading case. In general, perforated liners are notconsidered to be restraining and the supporting purlins should, therefore, bedesigned as unrestrained members.

C

T

(a)

(b)

T

C

1

2

1 Lateral restraint provided to

compression flange by cladding2 Cladding provides lateral restraint to

tension flange and partial torsionalrestraint

Figure 4.13 Purlin restraint

For wind uplift (or negative pressure on a wall), the cladding cannot providelateral restraint directly to the compression flange. In this case, the purlin (orcladding rail) is restrained by a combination of lateral restraint to the tensionflange and torsional restraint, as shown in Figure 4.13(b). The ability of thecladding to provide restraint is dependent not only on its in-plane shearstiffness (including the fasteners), but also its flexural stiffness. EN 1993-1-3includes a method in Section 10 for assessing the degree of restraint providedby the cladding in this case. Unlike the gravity load case, the cladding onlyprovides partial restraint to the purlin or rail. Consequently, the purlinmanufacturers’ technical literature should always give a lower capacity forpurlins subjected to wind uplift loading (or suction on cladding rails).

EN 1993-1-3[11] covers the design of purlins, liner trays and sheeting inSection 10.




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5 HOT-ROLLED SECONDARY STEELWORK

As an alternative to cold formed steel, purlins and cladding rails may also be

made from hot-rolled steel sections. At one time, this type of purlin wascommon in industrial buildings, often used in conjunction with steel roof trusses. The development of cold formed purlins (which are considerablylighter and cheaper) and the trend towards plastically designed portal frameswith their onerous restraint requirements meant that the use of hot-rolledpurlins became unusual in the UK and Ireland. However, hot-rolled purlinscontinue to be used in Continental Europe, often with long spanning claddingsolutions such as deck and membrane or composite panels. They areparticularly useful for providing an intermediate support to structural decking,where the decking by itself is incapable of spanning rafter to rafter.

Hot-rolled purlins have a higher load-carrying capacity than all but the largestcold formed purlins. This means that they are generally used at much greaterspacings than their cold formed counterparts, typically 3 m or more. This widespacing makes them unsuitable for plastically designed portal frames, whichcommonly require restraint to the rafters at approximately 1,8 m intervals.However, they are suitable for elastic frames and also for spans beyond therange of standard cold formed purlins (above 8 m). Hot-rolled purlins could of course be used at closer centres, but this would be uneconomic in mostcircumstances.

A considerable advantage of hot-rolled purlins over their cold formed rivals is

their resistance to lateral-torsional bucking, especially where rectangularhollow sections are used. This property is essential if the cladding is unable toprovide adequate restraint against lateral-torsional buckling. By contrast, coldformed purlins are only able to span as far as they do (typically 6 m to 8 m)because of the continuous restraint provided by the cladding. Similarly, wherethe local building regulations forbid using the cladding to restrain the structure,hot-rolled purlins are the only viable alternative to long spanning decksrunning rafter to rafter. Of course, apart from square hollow sections, hot-rolledpurlins are not immune to lateral-torsional buckling and must, therefore, bedesigned with this mode of failure in mind.

Unlike cold formed purlins, it is not common for the manufacturers to producesafe load tables for hot-rolled beams. Their capacities must, therefore, becalculated by a structural engineer according to the recommendations of EN 1993-1-1[12], taking account of the cross section resistance, lateral-torsionalbuckling and deflections. This process must be repeated for gravity and upliftload cases. If lateral-torsional buckling is the critical design criterion, theresistance of the member could be enhanced by the introduction of tubularrestraints either at the mid-span or third points of the purlin. However, this willadd cost to the structure in terms of additional steelwork and erection time.

Hot-rolled purlins can be designed as single or double-span beams. The latter

option will significantly increase the bending stiffness of the purlin and shouldbe used where deflection is the governing criterion. However, the high reaction

at the intermediate support (1,25 load in one span) can cause web crushing atthis location. Sleeves are not generally used with hot-rolled purlins.




8 - 36

Hot rolled purlins have the added advantage of better fire resistance than lightgauge cold formed purlins. This is shown by the noticeably higher inherentMassivity factor (cross section area/perimeter) which is used as a measure todefine the fire resistance of a structural section.




8 - 37

REFERENCES

1 EN 14782:2006 Self-supporting metal sheet for roofing, external claddingand internal lining. Product specification and requirements

2 MCRMA Technical Paper No 12: Fasteners for metal roof and wallcladding: Design, detailing and installation guide

The Metal Cladding and Roofing Manufacturers Association, 2000

3 MCRMA Technical Paper No. 3: Secret fix roofing design guide. The Metal Cladding and Roofing Manufacturers Association, 1999

4 MCRMA Technical Paper No. 6: Profiled metal roofing design guide The Metal Cladding and Roofing Manufacturers Association, 2004

5 MCRMA Technical paper No. 16: Guidance for the effective sealing of endlap details in metal roofing constructions

The Metal Cladding and Roofing Manufacturers Association, 2004

6 ECCS Publication 41 European recommendations for steel construction:Good practice in steel cladding and roofingEuropean Convention for Constructional Steelwork – Recommendations forsteel construction Technical Committee TC7, 1983.

7 European Directive 2002/91/EC: Energy Performance of Buildings The European Commission, 2002

8 MCRMA Technical paper No. 8: Acoustic design guide for metal roof andwall cladding.

The Metal Cladding and Roofing Manufacturers Association, 19949 EN 1991:2002: Eurocode 1 Actions on structures

10 EN 1990: 2002: Eurocode Basis of structural design

11 EN 1993-1-3:2006: Eurocode 3 Design of steel structures. General rules.Supplementary rules for cold-formed members and sheeting

12 EN 1993-1-1:2005: Eurocode 3 Design of steel structures. General rulesand rules for buildings





Part 9: Introduction to Computer

Software






Part 9: Introduction to Computer

Software



9 - ii



Part 9: Introduction to Computer Software

9 - iii

FOREWORD

This publication is part nine of the design guide, Single-Storey Steel Buildings.




Part 3: Actions











The design guides have been prepared under the direction of Arcelor Mittal,Peiner Träger and Corus. The technical content has been prepared by CTICM and SCI,collaborating as the Steel Alliance.




9 - iv




9 - v

ContentsPage No

FOREWORD iii

SUMMARY vi

1 INTRODUCTION 1 1.1 Software listing 1 1.2 Use of software 2

2 AVAILABLE FREE SOFTWARE 3 2.1 Member design, such as beams and columns 3 2.2 Composite construction 4 2.3 Cellular beam design 6 2.4 Portal frames 6 2.5 Simple connections 7 2.6 Moment resisting connections 8

2.7 Fire 8 2.8 Seismic 10




9 - vi

SUMMARY

This document contains details of freely available software to assist in design of single-storey steelbuildings according to the Eurocodes.




9 - 1

1 INTRODUCTION

Design in accordance with the Eurocodes may be facilitated by the use of

software. In many cases, the verifications required by the Standard can bereadily programmed into simple spreadsheets or into more complexprogrammes, which minimise the manual effort and reduce the risk of numerical errors.

In many countries, software has been written for the purpose of facilitatingdesign to the Eurocodes and has been made freely available. This publicationpresents a summary of software that is available, at March 2010. All thesoftware listed in this document is freely available.

No endorsement of any of the software programmes listed in this documentshould be presumed. Equally, the omission of existing software from the listingdoes not imply that it is inappropriate, inaccurate or non-endorsed. Moresoftware will undoubtedly become available as design to the Eurocodesbecomes more widespread.

Apart from the list of freely available software presented here, there arenumerous software houses that provide comprehensive analysis and designpackages, covering all aspects of steel building design, as described in thisguide.

1.1 Software lis ting In Section 2, software is listed under the following headings:

Member design, such as beams and columns

Composite construction

Cellular beam design

Analysis of frames

Portal frames

Simple connections

Moment resisting connections Fire

Seismic

For each item of software, the following details are listed:

Scope. A general description of the software

Design Standard. The design standard may be the published Eurocode, butmay be early versions of the Standard. Users must ensure that the version of the Eurocode is appropriate.

National Annex. Which National Annex is covered in the software, if any Source. Where the software can be obtained (web site)

Language. The language used in the software




9 - 2

1.2 Use of softwareNo systematic review of the software listed in this document has beenundertaken, so the user must verify that the software is appropriate for thedesign situation.




9 - 3

2 AVAILABLE FREE SOFTWARE

2.1 Member design, such as beams and

columnsSoftware Verifica di profi li sotti li piegati a freddo Scope Design and analysis of cold formed sections

Design Standard EN 1993-1-3, EN10162

National Annex Italian NTC2008

Source http://www.promozioneacciaio.it/costruttori_schede.php

Language Italian

SoftwareCorus sections interactive "blue book"

Scope The Corus sections interactive "blue book" comprises design data for theAdvance®, Celsius®and Hybox®ranges of sections. All design data isgenerated from the root software functions used to populate SCI P363:Steel Building Design: Design Data, in accordance with Eurocodes andthe UK National Annexes and SCI P202: Steelwork Design Guide toBS 5950-1: 2000. Volume 1 - Section Properties - Member Capacities.

Design Standard BS 5950 and BS EN 1993-1-1

National Annex UK only

Source http://www.corusconstruction.com/en/design_guidance/the_blue_book/

Language English

Software A3C (Arcelo rMi ttal CTICM Col umns Calculator)

Scope A3C is a new software that allows a structural designer to check theresistance of a member under bending moment and axial force accordingto EN 1993-1-1.

The field of application covers rolled profiles.

The ULS verifications include classification of the cross-sections, sectionresistance, flexural buckling, lateral torsional buckling, shear buckling andall interactions (M+N, M+V, M+N+V). Various design options areavailable (for example: Annex A or Annex B for interaction factors inEN 1993-1-1).

A detailed calculation sheet can be edited and printed.Design Standard EN 1993-1-1

National Annex French National Annex as option

Source http://www.arcelormittal.com/sectionshttp://www.cticm.com

Language English, French




9 - 4

Software LTBeam

Scope LTBeam software has been designed to calculate the critical moment forLateral Torsional Buckling (LTB), in simple or complex situations.

Even for simple cases, the critical moment is often a complex step in theprocess of verification of the LTB resistance. Moreover usual formulae donot allow the designer to take into account the specific restraintconditions of real cases. So they lead the designer to chooseconservative assumptions. That is why LTBeam can be used todetermine a more realistic value of the critical moment.

LTBeam software is based on a modelling by beam elements thatpermits to take into account specific aspects like warping stiffness,position of the transverse loads from the shear centre, position of thelateral restraints, etc.

LTBeam aims at facilitating the application of Eurocode 3, but it can beused with other codes, for a LTB verification based on the concept of critical moment.

Even though the calculations are complex, LTBeam is very simple to useand it does not require special training provided that the phenomenon is

well known by the user.Design Standard n/a

National Annex n/a

Source http://www.cticm.com/spip.php?rubrique6

Language French, English

2.2 Composite construction

Software ABC V2.11

Scope ABC Software allows a structural designer to check the resistance of beams according to the European standards EN 1993-1-1 andEN 1994-1-1.

The field of application covers simply supported beams, composite ornon composite, made from a I-rolled profile.

For composite beams, the connection can be ensured by either weldedstuds or HILTI connectors. Partial connection is allowed. At theconstruction stage, the composite beam can be fully propped or apropping can be defined. Appropriate verifications at the constructionstage are carried out when necessary.

The ULS calculations include the verification of the section resistanceunder bending moment and shear force, the resistance to lateral torsionalbuckling, the shear buckling resistance where necessary. The resistance

to lateral torsional buckling is based on the critical moment calculated bya modal analysis performed by the LTBeam engine.

A detailed calculation sheet can be edited and printed.

Design Standard EN 1993-1-1 and EN1994-1-1

National Annex n/a

Source http://www.arcelormittal.com/sections/index.php?id=119

Language French, English




9 - 5

Software ACP V1.02

Scope Construction phase for composite solution. To check the LTB behaviourof composite and/or partially encased beams during erection

Design Standard EN 1993-1-1 and EN1994-1-1

National Annex n/a


Language English, French, German, Spanish, Portuguese

Software ACD V3.06 Scope ArcelorMittal composite column design according to Eurocode 4.

Replaces CDD

Design Standard ENV 1994-1-1

National Annex n/a


Language English, French, German, Spanish

Software Software compendium for steel and composit e structures Scope This new software (currently a Beta version) for the analysis, calculation

and design of steel and composite structures, has been developed byConsulting Engineers FHECOR with funding from the Association for theAdvancement of Steel Technology (APTA) and ArcelorMittal. It is meantas a tool for use in design offices to facilitate the pre-design of structuresor verification of existing projects and designs. It is not intended tocompete with commercial software and can be used as a teaching toolfor steel structures (levels of deformation, stresses, effective widths,

section grade, etc.). as well as the development of checking examples.Design Standard It complies with Spanish CTE code and Eurocode 3, according to user’s

selection.

National Annex n/a

Source http://piem.fhecorconocimiento.es/

Language Spanish




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2.3 Cellular beam design

Software ACB+ V2.01

Scope Cellular beams design

ACB+is a piece of software dedicated to the design of cellular beams

made up from rolled profiles. It covers composite and non compositecellular beams, including curved beams.

ACB+includes practical tools for selecting the diameter and the spacingof the openings in accordance with fabrication requirements.

ULS verifications are performed according to the principles of theEurocodes (EN 1993-1-1 and EN 1994-1-1), with specific verifications forcellular beams (Vierendeel effect, web post buckling, etc).

For SLS verifications, the deflections are calculated by taking intoaccount the local bending due to the Vierendeel effect.

ACB+allows the designer to assess the fire resistance according to theprinciples of EN 1993-1-2 and EN 1994-1-2.

Design Standard EN 1993-1-1, EN 1994-1-1, EN 1993-1-2, EN 1994-1-2

National Annex n/a


Language English, German, French, Italian

Software AngelinaTM

Scope Angelina software has been especially designed for the calculation of aspecial type of beams with sinusoidal web openings, called Angelinabeams, fabricated from hot rolled I-profiles. This new software coversboth composite and non composite beams.

ULS verifications are carried out according to the principles of theEurocodes. They take into account the specific aspects of such beams,like local bending by Vierendeel effect. The deflections are alsocalculated by appropriate methods, in view to SLS verifications.

Design Standard EN 1993-1-1, EN 1994-1-1

National Annex



2.4 Portal frames

Software PORTAL Version 1.1 Scope PORTAL is a pre-design software for portal frames with single span,

made of rolled sections. It includes an automatic calculation of the snowload and the wind action, elastic global analysis of the frame, verificationsof the members, calculations of the deflections. The calculations arecarried out according to Eurocodes (ENV 1993-1-1).

The automatic pre-design is based on the weight criterion for a givensteel grade, but sections can be defined by the user for performingverifications.


National Annex Not suitable for National Annex application. Only partial safety factorsmay be user defined.






9 - 7

Software Pre-design of one span of a portal frame

Scope Pre-design of one span of a portal frame

Design Standard EN 1993-1-1

National Annex EN 1993-1-1 ANB 2008

Source Online calculation on www.infosteel.beLanguage Dutch and French

Software Pre-deSsign of a roof structure for residential buildings

Scope Pre-design of a roof structure for residential buildings


National Annex EN 1993-1-1 ANB 2008

Source Online calculation on www.infosteel.be

Language Dutch and French

2.5 Simple connections

Software ACOP V1.02

Scope Connection programme to design joints in steel building structures.


National Annex n/a


Language English, French, German

Software Unioni bullonate

Scope Bolted joints. Scheda di calcolo (ZIP – 4 Mb)



Source http://www.promozioneacciaio.it/costruttori_schede.php

Language Italian

Software Unioni saldate

Scope Welded joints.



Source http://www.promozioneacciaio.it/costruttori_schede.phpScheda di calcolo (ZIP 500 kb).

Language Italian




9 - 8

Software Verifica collegamenti a squadretta

Scope J oint design.

Design Standard EN 1993-1-1 and EN 1993-1-8


Source http://www.promozioneacciaio.it/costruttori_schede.phpScheda di calcolo (ZIP – 600 Kb).

Language Italian

Software Dimensionamiento unioni travature reticolari

Scope J oint verification of the trusses, bolted and welded



Source http://www.promozioneacciaio.it/costruttori_schede.phpScheda di calcolo (ZIP – 650 Kb)

Language Italian

2.6 Moment resisting connections

Software PlatineX

Scope PlatineX is an on-line software that covers the design of momentconnections made of rolled profiles (European I and H sections),according to EN 1993-1-8. Various geometries are possible for beam-to-beam connections (apex connections) and beam-to-column connections.

This piece of software checks the validity of the dimensions defined bythe user (edge distances, distance between bolts, etc). If the geometry is

valid, it calculates the moment resistance, the shear resistance, the axialresistance and the flexural stiffness.A detailed calculation sheet can be edited and saved as PDF file.


National Annex French NA

Source http://www.steelbizfrance.com/prog/platinex/

Language French

2.7 Fire

Software Arcelo rMittal Ozone 2.2.6

Scope Gas temperature in the event of fire according to EN 1991-1-2,corresponding steel temperature according to EN 1993-1-2 and simplifiedresistance check.


National Annex n/a


Language English




9 - 9

Software Software LUCA

Scope LUCA is software accompanying a design guide for industrial halls in fireconditions. This tool calculates displacements and additional horizontalforces that appear in industrial halls during fire enabling the engineers toconsider their effect in the design in order to avoid collapse or risk of human life. Software was developed within RFCS project RFS2-CR-

2007-00032.

Design Standard EN 1991-1, EN 1993-1-2

National Annex n/a


Language English, French, Spanish

Software AFCB V3.08

Scope Composite beam design in case of fire


National Annex n/a



Software AFCC V3.06

Scope Composite column design in case of fire


National Annex n/a



Software Fracof Scope Composite floor slabs

This software designs composite floor slabs at elevated temperatures bytaking into account the enhancing effects of the membrane action in slab.FRACOF also checks perimeter beams and provides a criticaltemperature for each of them.

Design Standard EN 1994-1-1, EN 1990, EN1991-1

National Annex n/a


Language English and French




9 - 10

2.8 Seismic

Software INERD 1.0.0 Scope Innovation for earthquake design.

INERD concept is a composite constructive system to improve therobustness and the safety of reinforced concrete frame structure

Design Standard

National Annex


Language English





Part 10: Model Construction

Specification






Part 10: Model Construction

Specification



10 - ii



Part 10: Model Construction Specification

10 - iii

FOREWORD

This publication is the tenth part of the design guide, Single-Storey Steel Buildings.




Part 3: Actions















10 - iv




10 - v

ContentsPage No

FOREWORD iii

SUMMARY vii 1 INTRODUCTION 1

1.1 Scope 2

2 NORMATIVE REFERENCES 4

3 BASIS OF STRUCTURAL DESIGN 9 3.1 General assumptions according to EN 1990 9

4 ACTIONS ON STRUCTURES 10 4.1 Self-weight and imposed loads for buildings 10 4.2 Snow loads 10

4.3 Wind loads 11 4.4 Thermal actions 11 4.5 Actions during execution 11 4.6 Accidental actions 13 4.7 Actions induced by cranes 14 4.8 Seismic actions 15

5 DESIGN OF STEEL STRUCTURES 17 5.1 Rules for single-storey buildings – EN 1993-1-1 17 5.2 Supplementary rules for sheeting – EN 1993-1-3 18 5.3 Design of plated structural elements – EN 1993-1-5 18 5.4 Design of joints – EN 1993-1-8 18

5.5 Fatigue – EN 1993-1-9 19 5.6 Material toughness and through-thickness properties – EN 1993-1-10 19 5.7 Crane supporting structures – EN 1993-6 20

6 EXECUTION SPECIFICATION 21 6.1 General 21 6.2 Execution classes 21 6.3 Preparation grades 21 6.4 Geometrical tolerances 21

7 CONSTITUENT PRODUCTS 23 7.1 Identification, inspection documents and traceability 23

7.2 Structural steel products 23 7.3 Welding consumables 23 7.4 Mechanical fasteners 23 7.5 Grouting materials 24

8 PREPARATION AND ASSEMBLY 25 8.1 Identification 25 8.2 Handling and storage 25 8.3 Cutting 25 8.4 Shaping 25 8.5 Holing 25 8.6 Assembly 26

9 WELDING 27 9.1 General 27 9.2 Qualification of welding procedures 27




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9.3 Welders and welding operators 27 9.4 Welding coordination 27 9.5 Preparation and execution of welding 27 9.6 Acceptance criteria 29

10 MECHANICAL FASTENING 30

11 ERECTION 31

12 CONSTRUCTOR’S DOCUMENTATION 34

13 INTERFACES OF THE STEEL STRUCTURE 35 13.1 Interface to concrete surfaces 35 13.2 Interface to neighbouring constructions 36

Appendix A MODEL PROJ ECT SPECIFICATION 37




10 - vii

SUMMARY

This guide is a Model Construction Specification to be used in contract documents for atypical construction project of a single-storey building. Its main objectives are toachieve greater uniformity in steelwork contract specifications in Europe and to provide

a guide to specification of appropriate standards for the design, fabrication and erectionof steelwork structures for buildings.

It deals with structural steelwork designed in accordance with applicable parts of theEurocode Standards, to be executed in accordance with applicable parts of EN 1090. Allthe relevant Sections of the model specification are included in an appendix that can bedirectly copied and used in contracts, with any additional project-specific informationthat may be required.




10 - viii




10 - 1

1 INTRODUCTION

This guide is a Model Construction Specification to be used in contract

documents for a typical construction project of a single-storey building. Itsmain objectives are:

To achieve greater uniformity in steelwork contract specifications inEurope.

To provide a guide to specification of appropriate standards for the design,fabrication and erection of steelwork structures for buildings.

It is essential that the designer and the steelwork contractor receive, on time, allinformation necessary for them to carry out the contract. This ModelConstruction Specification gives guidance on the items and information that

should be included in the Project Specification.

The Member States of the EU and EFTA recognise that Eurocodes serve asreference documents for the following purposes:

As a means to prove compliance of building and civil engineering workswith the essential requirements of Construction Products Directive89/106/EEC of 21 December 1988 (amended by Directive 93/68/EEC of 22

July 1993), particularly Essential Requirement No. 1 – Mechanicalresistance and stability – and Essential Requirement No. 2 – Safety in caseof fire.

As a basis for specifying contracts for construction works and relatedengineering services.

As a framework for drawing up harmonised technical specifications forconstruction products (ENs and ETAs).

The Eurocodes, as far as they concern the construction works themselves, havea direct relationship with the Interpretative Documents referred to in Article 12of the Construction Products Directive, although they are of a different naturefrom harmonised product standards. There is a need for consistency betweenthe harmonised technical specifications for construction products and thetechnical rules for works.

The steel construction industry in Europe will have to use CE marked products. The performances of these products can be declared by reference torequirements given in:

The harmonised European Standards such as the standards EN 10025 andEN 1090. Parts 1 of these Standards (i.e. EN 10025-1 and EN 1090-1respectively) include a special Annex ZA relating to CE marking.

A European Technical Approval (ETA).

CE Marking of steel products to EN 10025 has been mandatory since 2006.

The use of CE marked products according to EN 1090 will be mandatory fromthe first semester 2011 for most of the European countries. Once it appears inthe European Official Journal, the standard will be in the application phase.




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In EN 1090-1, for some special types of construction products (modularconstruction for example), reference is made to the Eurocodes. In this case, itshall be mentioned which Nationally Determined Parameters have been takeninto account.

Much of the information noted in this Model Construction Specification isbased upon that given in these Standards, but it must not be inferred that thefull details of the standards are not relevant.

References to applicable parts of European Standards have been madethroughout this Model Construction Specification.

1.1 Scope This Model Construction Specification deals with structural steelwork designedin accordance with applicable parts of the Eurocode Standards and executed in

accordance with applicable parts of EN 1090.

It can be used for all types of single- storey building construction designed forstatic loading, including cases where the dynamic effects are assessed usingequivalent quasi-static loads and dynamic amplification factors, including windactions and actions induced by hoists and cranes and cranes on runway beams.

It is not intended to be used for steelwork in dynamically loaded structures.

This Model Construction Specification covers structural steelwork producedfrom hot rolled structural steel products only. It does not cover structural

steelwork produced from cold formed structural steel (only cold formedprofiled steel sheeting and cold formed stressed-skin sheeting used as astructural diaphragm are herein covered), structural hollow sections, channelsand tubes, and stainless steel products.

This Model Construction Specification should be introduced into a steelworkcontract by a Project Specification, the contents of which are detailed inAppendix A of this document and completed with project-specific information.

The Project Specification should also include any additions or modificationsthat may be required by the National Structural Steelwork Specification by theClient for a particular contract if the form of behaviour or other aspects of the

structure are unorthodox.

Contract documents (which include architectural and/or structural designdrawings, specifications and addenda) vary considerably in intricacy andcompleteness. Nonetheless, the designer, the fabricator and the erector must beable to rely upon the accuracy of the contract documents, in order to allowthem to provide the Client with bids that are adequate and complete. It alsoenables the preparation of the general arrangement drawings and the shop anderection drawings, the ordering of materials and the timely fabrication anderection of construction components.

Critical requirements that are necessary to protect the Client’s interest, thataffect the integrity of the structure or that are necessary for the designer, the




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fabricator and the erector to proceed with their work, must be included in thecontract documents. Non-exhaustive examples of critical information include:

Standard specifications and codes that govern structural steel design andconstruction, including bolting and welding

Material specifications Welded-joint configuration and weld-procedure qualification

Surface preparation and shop painting requirements

Shop and field inspection requirements

If any, non-destructive testing (NDT) requirements, including acceptancecriteria

Special requirements on delivery and special erection limitations.




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2 NORMATIVE REFERENCES

The European Standards incorporate, by dated or undated reference, provisions

from other publications. These normative references are cited at the appropriateplaces in the text and the publications are listed in Tables 2.1 to 2.3.

Table 2.1 Design and structural engineering

Title

EN 1990:2002 Basis of structural design

EN 1991-1-1:2003 Actions on structures – Part 1-1: General actions – Densities,self- weight, imposed loads for buildings

EN 1991-1-2:2002 Actions on structures – Part 1-2: General actions – Actions onstructures exposed to fire

EN 1991-1-3:2003 Actions on structures – Part 1-3: General actions – Snowloads

EN 1991-1-4:2005 Actions on structures – Part 1-4: General actions – Wind loads

EN 1991-1-5:2003 Actions on structures – Part 1-5: General actions – Thermalactions

EN 1991-1-6:2005 Actions on structures – Part 1-6: General actions – Actionsduring execution

EN 1991-1-7:2006 Actions on structures – Part 1-7: General actions – Accidentalactions

EN 1991-3:2006 Actions on structures – Part 3 : Actions induced by cranes andmachinery

EN 1993-1-1:2005 Design of steel structures – Part 1-1: General rules and rulesfor buildings

EN 1993-1-2:2005 Design of steel structures – Part 1-2: General rules –Structural fire design

EN 1993-1-3:2006 Design of steel structures – Part 1-3: General rules –Supplementary rules for cold-formed members and sheeting

EN 1993-1-4:2006 Design of steel structures – Part 1-4: General rules –Supplementary rules for stainless steels

EN 1993-1-5:2005 Design of steel structures – Part 1-5: Plated structuralelements

EN 1993-1-8:2005 Design of steel structures – Part 1-8: Design of joints

EN 1993-1-9:2005 Design of steel structures – Part 1-9: Fatigue

EN 1993-1-10:2005 Design of steel structures – Part 1-10: Material toughness andthrough-thickness properties

EN 1993-6:2007 Design of steel structures – Part 6: Crane supportingstructures

EN 1998-1:2004Design of structures for earthquake resistance – Part 1:General rules, seismic actions and rules for buildings




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For each European country, each part of the Eurocode applies with its NationalAnnex when the latter is available.

Table 2.2 Execution, fabrication and erection

Title

EN 1090-1:2009 Execution of steel structures and aluminium structures.Part 1: Requirements for conformity assessment of structuralcomponents

EN 1090-2:2008 Execution of steel structures and aluminium structures.Part 2: Technical requirements for steel structures

EN ISO 12944 Paints and varnishes – Corrosion protection of steel structuresby protective paint systems

EN 1461 Hot dip galvanized coatings on fabricated iron and steelarticles – specifications and test methods

EN ISO 17659:2004 Welding - Multilingual terms for welded joints with illustrations

EN ISO 14555:1998 Welding - Arc stud welding of metallic materials

EN ISO 13918:1998 Welding - Studs for arc stud welding

EN ISO15609-1:2004

Specification and qualification of welding procedures formetallic materials - Part 1: Welding procedure specification forarc welding of steels

EN ISO15614-1:2004

Specification and qualification of welding procedures formetallic materials – Welding procedure test - Part 1: Arc andgas welding of steels and arc welding of nickel and nickelalloys

EN 1011-1:1998 Welding – Recommendations for welding of metallic materialsPart 1: General guidance for arc welding

EN 1011-2:2001 Welding – Recommendations for welding of metallic materialsPart 2: Arc welding of ferritic steels

EN ISO 25817:2003 Arc-welded joints in steel - Guidance for quality levels forimperfections

ISO 286-2:1988 ISO system of limits and fits - Part 2: Tables of standardtolerance grades and limit deviations for hole and shafts




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Table 2.3 Products

Title

EN 10025-1:2004 Hot-rolled products of structural steels - Part 1: Generaldelivery conditions.

EN 10025-2:2004 Hot-rolled products of structural steels - Part 2: Technical

delivery conditions for non-alloy structural steels.

EN 10025-3:2004 Hot-rolled products of structural steels - Part 3: Technicaldelivery conditions for normalized rolled weldable fine grainstructural steels.

EN 10025-4:2004 Hot-rolled products of structural steels - Part 4: Technicaldelivery conditions for thermo-mechanical rolled weldable finegrain structural steels.

EN 10025-5:2004 Hot-rolled products of structural steels - Part 5: Technicaldelivery conditions for structural steels with improvedatmospheric corrosion resistance.

EN 10025-6:2004 Hot-rolled products of structural steels - Part 6: Technical

delivery conditions for flat products of high yield strengthstructural steels in the quenched and tempered condition.

EN 10164:2004 Steel products with improved deformation propertiesperpendicular to the surface of the product - Technical deliveryconditions.

EN 10210-1:2006 Hot finished structural hollow sections of non-alloy and finegrain structural steels – Part 1: Technical deliveryrequirements.

EN 10219-1:2006 Cold formed hollow sections of structural steelPart 1: Technical delivery requirements.

EN 10029:1991 Hot rolled steel plates 3 mm thick or above - Tolerances on

dimensions, shape and mass

EN 10034:1993 Structural steel I- and H-sections - Tolerances on shape anddimensions

EN 10051:1991 Continuously hot-rolled uncoated plate, sheet and strip of non-alloy and alloy steels - Tolerances on dimensions and shape

EN 10055:1995 Hot rolled steel equal flange tees with radiused root and toes -Dimensions and tolerances on shape and dimensions

EN 10056-1:1995 Structural steel equal and unequal leg anglesPart 1: Dimensions

EN 10056-2:1993 Structural steel equal and unequal leg angles

Part 2: Tolerances on shape and dimensionsEN 13001-1:2004 Cranes – General design – Part 1 : General principles and

requirements

EN 13001-2:2004 Crane safety – General design – Part 2 : Load effects

EN 14399-1:2002 High strength structural bolting for preloadingPart 1 : General Requirements

EN 14399-2:2002 High strength structural bolting for preloadingPart 2 : Suitability Test for preloading

EN 14399-3:2002 High strength structural bolting for preloadingPart 3 : System HR - Hexagon bolt and nut assemblies




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Table 2.3 Continued…

Title

EN 14399-4:2002 High strength structural bolting for preloadingPart 4 : System HV - Hexagon bolt and nut assemblies

EN 14399-5:2002 High strength structural bolting for preloading

Part 5 : Plain washers for system HR

EN 14399-6:2002 High strength structural bolting for preloadingPart 6 : Plain chamfered washers for systems HR and HV

EN ISO 898-1:1999 Mechanical properties of fasteners made of carbon steel andalloy steel - Part 1: Bolts, screws and studs (ISO 898-1:1999)

EN 20898-2:1993 Mechanical properties of fastenersPart 2: Nuts with special proof load values - Coarse thread(ISO 898-2:1992)

EN ISO 2320:1997 Prevailing torque type steel hexagon nuts - Mechanical andperformance requirements (ISO 2320:1997)

EN ISO 4014:2000 Hexagon head bolts - Product grades A and B (ISO4014:1999)

EN ISO 4016:2000 Hexagon head bolts - Product grade C (ISO 4016:1999)

EN ISO 4017:2000 Hexagon head screws - Product grades A and B (ISO4017:1999)

EN ISO 4018:2000 Hexagon head screws - Product grade C (ISO 4018:1999)

EN ISO 4032:2000 Hexagon nuts, style 1 - Product grades A and B (ISO4032:1999)

EN ISO 4033:2000 Hexagon nuts, style 2 - Product grades A and B (ISO4033:1999)

EN ISO 4034:2000 Hexagon nuts - Product grade C (ISO 4034:1999)EN ISO 7040:1997 Prevailing torque hexagon nuts (with non-metallic insert), style

1 - Property classes 5, 8 and 10

EN ISO 7042:1997 Prevailing torque all-metal hexagon nuts, style 2 - Propertyclasses 5, 8, 10 and 12

EN ISO 7719:1997 Prevailing torque type all-metal hexagon nuts, style 1 -Property classes 5, 8 and 10

ISO 1891:1979 Bolts, screws, nuts and accessories - Terminology andnomenclature – Trilingual edition

EN ISO 7089:2000 Plain washers- Nominal series- Product grade A

EN ISO 7090:2000 Plain washers, chamfered - Normal series - Product grade AEN ISO 7091:2000 Plain washers - Normal series - Product grade C

EN ISO 10511:1997 Prevailing torque type hexagon thin nuts (with non-metallicinsert)

EN ISO 10512:1997 Prevailing torque type hexagon nuts thin nuts, style 1, withmetric fine pitch thread - Property classes 6, 8 and 10

EN ISO 10513:1997 Prevailing torque type all-metal hexagon nuts, style 2, withmetric fine pitch thread - Property classes 8, 10 and 12

When manufactured construction products, with Harmonised Standards (i.e.

EN 10025, EN 1090), are to be used, CE marking shall be placed on theproducts according to the relevant European Harmonised Standards.Harmonised Standards are European Standards adopted by the European




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Committee for Standardisation (CEN), following a mandate issued by theEuropean Commission (mandate M/120 for structural metallic products). Notall European Standards (ENs) are harmonised - only those which have beenlisted in the Official Journal.

When manufactured construction products, without Harmonized Standards, areto be used (i.e. metal anchors, fire protective products, metal frame buildingkits, fire stopping and fire sealing products, prefabricated building units, etc.),European Technical Approval Guidelines (ETAG) require manufacturers toplace CE marking on their products in accordance with the relevant European

Technical Approval (ETA).

The relevant ETAs shall be specified in the contract documents.

An full list of valid ETAs is available on the official website of the EuropeanOrganisation for Technical Approvals (EOTA): www.eota.be.

The latest edition of the publication referred to applies.

National Standards Bodies publish up-to-date versions on their officialwebsites.

Table 2.4 National Standards Bodies

Country Standards body Web site

Belgium NBN www.nbn.be

France AFNOR www.afnor.org

Germany DIN www.din.de

Italy UNI www.uni.comNetherlands NEN www.nen.nl

Poland PKN www.pkn.pl

Spain AENOR www.aenor.es

Switzerland SNV www.snv.ch

Luxembourg ILNAS www.ilnas.lu

Austria ASI www.as-institute.at




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3 BASIS OF STRUCTURAL DESIGN

EN 1990 establishes the Principles and Requirements for safety, serviceability

and durability of structures, describes the basis for their design and verificationand gives guidelines for related aspects of structural reliability.

For the design of new structures, EN 1990 is intended to be used, for directapplication, together with Eurocodes EN 1991 to 1999.

EN 1990 is applicable for the structural appraisal of existing construction, indeveloping the design of repairs and alterations or in assessing changes of use.

Design of steel structures shall conform to the basic requirements of § 2.1 of EN 1990.

Reliability, durability and quality management shall conform to § 2.2, § 2.4and § 2.5 of EN 1990.

National choice is allowed through clauses listed in the Foreword to EN 1990.

3.1 General assumptions according to EN 1990 The choice of structural system and the design of the structure is made by

appropriately qualified and experienced personnel

Execution is carried out by personnel having the appropriate skill andexperience

Adequate supervision and quality control is provided during the executionof the work, i.e. in design offices, factories, plants and on site

The construction materials and products are used as specified in EN 1990 orin the relevant execution standards or reference material or productspecifications

The structure will be adequately maintained

The structure will be used in accordance with the design assumptions.

Addi tional contract document requirements

According to § 2.1(4)P of EN 1990, relevant additional specific events (impact,

explosion, etc.), defined by the Client and the relevant authority, must be takeninto account in the design and the execution of a structure.

According to § 2.3 of EN 1990, the contract documents should specify thedesign working life of the structure.

According to § 3.3(2) of EN 1990, the contract documents should state anyrelevant additional specific circumstances where the limit states that concernthe protection of the contents are to classified as ultimate limit states.

According to § 3.4(1) of EN 1990, the contract documents shall specify theserviceability requirements of the project.




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4 ACTIONS ON STRUCTURES

4.1 Self-weight and imposed loads for bui ldings

EN 1991-1-1 gives design guidance and actions for the structural design of buildings, including the following aspects:

Densities of construction materials and stored materials

Self-weight of construction elements

Imposed loads for buildings.

National choice is allowed through clauses listed in the Foreword toEN 1991-1-1.


According to § 3.3.2(4) of EN 1991-1-1, the contract documents shall specifythe imposed loads to be considered for serviceability limit state verifications, inaccordance with the service conditions and the requirements concerning theperformance of the structure.

According to § 4.1(1) and 4.1(2) of EN 1991-1-1, characteristic values of densities of construction and stored materials shall be specified in the contractdocuments, especially for materials which are not covered by the Tables inAppendix A.

According to § 6.1(4) of EN 1991-1-1, loads for heavy equipment (e.g. in

communal kitchens, radiology rooms, boiler rooms, etc.) shall be agreedbetween the Client and the relevant authority and specified in the contractdocuments.

4.2 Snow loadsEN 1991-1-3 gives guidance to determine the values of loads due to snow, tobe used for the structural design of buildings.

National choice is allowed through clauses listed in the Foreword to

EN 1991-1-3.


According to § 1.5 of EN 1991-1-3, in some circumstances tests and provenand/or properly validated numerical methods may be used to obtain snow loadson the construction works. These circumstances are those agreed with theClient and the relevant authority, and specified in the contract documents.

According to § 4.1(1) of EN 1991-1-3, to cover unusual local conditions, theNational Annex may additionally allow the Client and the relevant authority toagree upon different characteristic values of snow load which have to be

specified in the contract documents.




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4.3 Wind loadsEN 1991-1-4 gives guidance on the determination of natural wind actions forthe structural design of buildings (with heights up to 200 m) for each of theloaded areas under consideration.



According to § 7.2.2 of EN 1991-1-4, the rules for the velocity pressuredistribution for leeward wall and sidewalls may be given in the National Annexor be defined for the individual project and specified in the contract documents.

4.4 Thermal actions

EN 1991-1-5 gives design guidance, principles and rules for calculatingthermal actions arising from climatic and operational conditions for thestructural design of buildings. Principles needed for cladding and otherappendages of buildings are also provided.

EN 1991-1-5 describes the changes in the temperature of structural elements.Characteristic values of thermal actions are presented for use in the design of structures which are exposed to daily and seasonal climatic changes. Forstructures not exposed to climatic conditions, thermal actions may not need tobe considered.

National choice is allowed through clauses listed in the foreword toEN 1991-1-5.


According to § 5.2(2)P of EN 1991-1-5, operational effects (due to heating,technological or industrial processes) shall be considered in accordance withthe particular project, and thus specified in the contract documents.

According to § 5.2(3)P of EN 1991-1-5, values of TM and Tp may beprovided for the particular project, and thus specified in the contractdocuments.

4.5 Actions during executionEN 1991-1-6 gives principles and general rules for the determination of actionsto be taken into account during the execution of buildings. EN 1991-1-6 can beused as guidance for the determination of actions to be taken into accountduring structural alterations, reconstruction, partial or full demolition, and forthe determination of actions to be used for the design of auxiliary constructionworks (false-work, scaffolding, propping system, etc.) needed for the executionphases. Rules and additional information are given in Annexes A1 and B, and

can also be defined in the National Annex or in the contract documents for theindividual project.




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National choice is allowed through clauses listed in the foreword toEN 1991-1-6.


The rules concerning the safety of persons, on and around the construction site,

shall be specified in the contract documents for the individual project, and areoutside the scope of EN 1991-1-6.

EN 1991-1-6 also provides rules for determining the actions that can be usedfor the calculation of auxiliary construction works needed for the executionphases.

The contract documents shall classify construction loads in accordance with Tables 2.2 and 4.1 of EN 1991-1-6.

Loads due to construction equipments, cranes and/or auxiliary structures can be

classified as fixed or free loads, depending on their possible spatial variation;contract documents shall specify the loads and their classification.

If construction loads are classified as fixed, then the contract documents shalldefine tolerances for the possible deviations to the theoretical position.

If construction loads are classified as free, then the contract documents shalldefine the limits of the potential area of spatial variation.

In the absence of any specific requirement in the National Annex, the contractdocuments shall specify:

Return periods for the assessment of the characteristic values of variable(climatic, seismic, etc.) actions during execution phases (see § 3.1(5) of EN 1991-1-6)

A minimum wind velocity during execution phases (see § 3.1(5) of EN 1991-1-6)

Rules of combination of snow loads and wind action with the constructionloads (see § 3.1(7) of EN 1991-1-6)

Geometric imperfections of the structure and the structural elements, for theselected design situations during execution (see § 3.1(8) of EN 1991-1-6)

Criteria associated with serviceability limit states during execution (see§ 3.3(2) of EN 1991-1-6)

When appropriate, frequent values of particular loads to be taken intoaccount (see § 3.3(5) of EN 1991-1-6)

Requirements of suitability for service of auxiliary structures in order toavoid excessive deformation and/or deflection that affect the durability,fitness for use or aesthetic appearance in the final stage (see § 3.3(6) of EN 1991-1-6).

Concerning the wind actions, the contract documents shall specify whether ornot a procedure is needed for calculating dynamic response of the structureduring the various stages of execution, taking into account the degree of




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completion and stability of the structure and its components (see § 4.7(1) of EN 1991-1-6).

The contract documents shall specify the maximum allowable wind velocityduring crane operations or other short term execution stages (see § 4.7(1) of

EN 1991-1-6).

The contract documents shall specify, when relevant, accidental designsituations due to cranes or exceptional conditions applicable to the structure orits exposure, such as impact, local failure and subsequent progressive collapse,fall of structural or non-structural parts, and abnormal concentrations of building equipment and/or building materials, water accumulation on steelroofs, fire, etc. (see § 4.12(1) and (3) of EN 1991-1-6).

The contract documents shall specify, when relevant, the design values of the

ground acceleration as well as the importance factor I to be taken into account

for the assessment of seismic actions, given the reference period of theconsidered transient situation (see § 4.13 of EN 1991-1-6).

The contract documents shall specify the characteristic values of horizontalactions due to imperfections or deformations related to horizontaldisplacements to be taken into account during execution phases (see § A1.3(1)of EN 1991-1-6).

4.6 Accidental actionsEN 1991-1-7 describes Principles and Application rules for the assessment of

accidental actions on buildings and bridges. The following actions areincluded:

Impact forces from vehicles, rail traffic, ships and helicopters

Actions due to internal explosions

Actions due to local failure from an unspecified cause.

EN 1991-1-7 does not specifically deal with accidental actions caused byexternal explosions, warfare and terrorist activities, or the residual stability of buildings damaged by seismic action or fire.



According to § 2(2)P of EN 1991-1-7, the contract documents may specify thetreatment of accidental actions which are not classified as free actions.

According to § 3.1(2) of EN 1991-1-7, the contract documents shall specify thestrategies and rules to be considered for accidental design situations.

According to § 3.1(2) of EN 1991-1-7, notional values for identified accidentalactions may be specified in the contract documents.




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According to § 3.4(1) of EN 1991-1-7, the strategies for accidental designsituations may be based on the Consequence Classes as set out in EN 1990.

Thus, these Consequence Classes shall be specified in the contract documents.

According to § 4.3.1(2) of EN 1991-1-7, the contract documents shall specify

whether or not the equivalent static design forces due to vehicular impact onmembers supporting structures over or adjacent to roadways, Fdx and Fdy, actsimultaneously.

According to § 4.5.1.2 of EN 1991-1-7, if the building may be subject toimpact from derailed railway traffic, the contract documents shall definewhether it is a Class A or Class B structure.

According to § 4.5.2(1) of EN 1991-1-7, frontal and lateral dynamic designforces due to impact from river and canal traffic, as well as the height of application of the impact force and the impact area shall be specified in the

contract documents.

4.7 Actions induced by cranesEN 1991-3 gives design guidance and specifies imposed loads (models andrepresentative values) induced by hoists and cranes on runway beams, whichinclude dynamic effects and braking, acceleration and accidental forces.

National choice is allowed through clauses listed in the Foreword toEN 1991-3.

Addi tional contract document requirementsUnless more accurate data (concerning the crane characteristics) is specified inthe contract documents (the crane supplier shall therefore be known at the timeof writing the contract documents), provisions of Section 2 of EN 1991-3apply.

According to § 2.3(6) of EN 1991-3, the contract documents shall specifywhether or not tests are performed with cranes on the supporting structures forthe serviceability limit state verification.

According to § 2.5.2.2(2) of EN 1991-3, the contract documents shall specify

whether one or several forces of the five horizontal types (a) to (e) listed in2.5.2.2(1) shall be included in the same group of simultaneous crane loadcomponents.

According to § 2.5.2.2(4) of EN 1991-3, the contract documents shall specifythe way the longitudinal horizontal forces HL,i and the transverse horizontalwheel forces H T,i, caused by acceleration and deceleration of masses of thecrane or the crab, shall be applied. Otherwise, provisions given in Figure 2.3 of EN 1991-3 shall apply.

According to § 2.5.3(2) of EN 1991-3, the contract documents shall define the

maximum number of cranes to be taken into account as acting simultaneously.




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The Hoisting Class (HC1 to HC4) of the crane shall be specified in the contractdocuments, unless it is specified in the crane supplier specification. Referencecan be made to Annex B (informative) of EN 1991-3.

According to § 2.9.1(1) of EN 1991-3, the contract documents shall specify the

vertical load to be applied to access walkways, stairs and platform. Otherwise,provisions given in § 2.9.1(2), 2.9.1(3) or 2.9.1(4) shall apply.

According to § 2.9.2(1) of EN 1991-3, the contract documents shall specify thehorizontal load to be applied to the guard rail. Otherwise, provisions given in§ 2.9.2(1) or 2.9.2(2) shall apply.

To make allowance of relevant accidental actions, the contract documents shallspecify:

Whether buffers are used or not

Whether or not a crane with horizontally restrained loads can tilt when itsload or lifting attachment collides with an obstacle.

To make allowance for fatigue effects, the contract documents shall providesufficient information on the operational conditions; the fatigue loads can thenbe determined according to EN 13001 and Annex A of EN 1993-1-9.Otherwise, provisions of § 2.12 of EN 1991-3 apply.

Where a simplified approach for determining the fatigue loads is favoured inthe contract documents, the latter shall specify:

the class of load spectrum (Q0 to Q5) for all tasks of the crane

the class of total number of working cycles (U0 to U9) during the design lifeof the crane

the crane classification (S0 to S9). If the crane classification is not includedin the crane supplier specification, reference can be made to Annex B(informative) of EN 1991-3.

According to § A.3.2(1) of the normative Annex A of EN 1991-3, the contractdocuments shall specify the partial factor for actions on crane supportingstructures to be used in serviceability limit states. Otherwise, this partial factorshall be taken as 1,0.

4.8 Seismic actionsEN 1998-1 applies to the design and construction of buildings and civilengineering works in seismic regions. Its purpose is to ensure that in the eventof earthquakes:

Human lives are protected

Damage is limited

Structures important for civil protection remain operational (special

structures such as nuclear power plants, offshore structures and large dams,are beyond the scope of EN 1998-1).




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One fundamental issue in EN 1998-1 is the definition of the seismic action.Given the wide difference of seismic hazard and seismo-genetic characteristicsin the various member countries, the seismic action is herein defined in generalterms. The definition allows various Nationally Determined Parameters whichshall be confirmed or modified in the National Annexes.



According to § 2.1(2) and (3) of EN 1998-1, target reliabilities for the no-collapse requirement and for the damage limitation requirement are establishedby the National Authorities for different types of buildings on the basis of theconsequences of failure. Contract documents shall specify the ImportanceClass of the individual project (see 4.2.5 of EN 1998-1).

Depending on the Importance Class of the structure and the particularconditions of the project, contract documents shall specify whether or notground investigations and/or geological studies shall be performed to identifythe ground type (A, B, C, D, E, S1 or S2), according to Table 3.1 of EN 1998-1.

Contract documents shall specify the seismic zone of the individual project(according to the zonation map, decided by the National Authority, and foundin the National Annex to EN 1998-1).

Contract documents shall specify according to which concept earthquake

resistant steel buildings shall be designed to (DCL, DCM or DCH).

According to 6.2(8) of EN 1998-1, the required toughness of steel and weldsand the lowest service temperature adopted in combination with the seismicaction shall be defined in the contract documents.




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5 DESIGN OF STEEL STRUCTURES

Eurocode 3 is intended to be used in conjunction with:

EN 1990 Basis of structural design

EN 1991 Actions on structures

ENs, ETAGs and ETAs for construction products relevant for steelstructures

EN 1090 Execution of Steel Structures – Technical requirements

EN 1992 to EN 1999 when steel structures or steel components are referredto.

Eurocode 3 is concerned only with requirements for resistance, serviceability,durability and fire resistance of steel structures. Other requirements, e.g.concerning thermal or sound insulation, are not covered.

5.1 Rules for single-storey buildings – EN 1993-1-1EN 1993-1-1 gives basic design rules for steel structures with materialthicknesses t >3 mm. It also gives supplementary provisions for the structuraldesign of single-storey steel buildings.

Material properties for steels and other construction products and the

geometrical data to be used for design shall be those specified in the relevantENs, ETAGs or ETAs unless otherwise indicated.



The design working life shall be taken as the period for which a buildingstructure is expected to be used for its intended purpose. For the specificationof the intended design working life of a permanent building see Table 2.1 of EN 1990.

The effects of deterioration of material, corrosion or fatigue where relevantshall be taken into account by appropriate choice of material, see EN 1993-1-4and EN 1993-1-10, and details, see EN 1993-1-9, or by structural redundancyand by the choice of an appropriate corrosion protection system.

The dimensional and mass tolerances of rolled steel sections and plates shallcomply with the relevant product standard, ETAG or ETA unless more severetolerances are specified.

Any semi-finished or finished structural product used in the structural design of

buildings shall comply with the relevant EN Product Standard or ETAG orETA.




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With reference to Annex A1.4 of EN 1990, limits for vertical deflectionsaccording to Figure A1.1, for horizontal deflections according to Figure A1.2and for vibrations of structures on which the public can walk, shall be specifiedin the contract documents and agreed with the Client.

5.2 Supplementary rules for sheeting – EN 1993-1-3EN 1993-1-3 gives, among other, design requirements for profiled steelsheeting. Methods are also given, in this part of Eurocode 3, for stressed-skindesign using steel sheeting as a structural diaphragm.



According to § 2(6) of EN 1993-1-3, contract documents shall define theStructural Class (I to III) of the construction, associated with failureconsequences according to Annex B of EN 1990:

Structural Class I: construction where sheeting is designed to contribute tothe overall strength and stability of a structure

Structural Class II: construction where sheeting is designed to contribute tothe strength and stability of individual structural elements

Structural Class III: construction where sheeting is used as an element thatonly transfers loads to the structure.

5.3 Design of plated structural elements –EN 1993-1-5 EN 1993-1-5 gives design requirements of stiffened and unstiffened plateswhich are subject to in-plane forces.


5.4 Design of joints – EN 1993-1-8 EN 1993-1-8 gives design methods for the design of joints subject topredominantly static loading using steel grades S235, S275, S355 and S460.



According to § 3.4.1 of EN 1993-1-8, the category of bolted connections(Category A, B or C for joints loaded in shear, and Category D or E for joints

loaded in tension) shall be specified in the contract documents.




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According to § 3.9 of EN 1993-1-8, the contract documents shall specify theclass of friction surfaces for slip-resistant connections using pre-loaded 8.8 or10.9 bolts.

According to § 4.1 of EN 1993-1-8, the contract documents shall specify the

quality level of welds according to EN ISO 25817. The frequency of inspectionof welds shall be specified in the contract documents and shall conform to therequirements of EN 1090-2.

5.5 Fatigue – EN 1993-1-9EN 1993-1-9 gives methods for the assessment of fatigue resistance of members, connections and joints subjected to fatigue loading.

According to § 2(1) of EN 1993-1-9, structures designed using fatigue actionsfrom EN 1991 (i.e., EN 1991-3) and fatigue resistance according to

EN 1993-1-9 are deemed to satisfy an acceptable level of probability that theirperformance will be satisfactory throughout their design life.



According to § 3(1) of EN 1993-1-9, contract documents shall specify whetherfatigue assessment shall be undertaken using either ‘damage tolerant method’or ‘safe life method’. If the ‘damage tolerant method’ is specified, a prescribedinspection and maintenance regime for detecting and correcting fatigue damageshall be implemented throughout the design life of the structure. The ‘safe lifemethod’ shall be specified in cases where local formation of cracks in onecomponent could rapidly lead to failure of the structural element or structure.

According to § 3(7) of EN 1993-1-9, contract documents shall specify theFailure Consequence classification (Low Consequence or High Consequence)in order to determine the partial factor for fatigue strength, in conjunction withthe specified fatigue assessment method.

5.6 Material toughness and through-thickness properties – EN 1993-1-10 EN 1993-1-10 contains design guidance for the selection of steel for fracturetoughness and for through-thickness properties of welded elements where thereis a significant risk of lamellar tearing during fabrication, for constructionsexecuted in accordance with EN 1090.

The guidance given in Section 2 of EN 1993-1-10 shall be used for theselection of material for new construction. The rules shall be used to select asuitable steel grade from the European Standards for steel products listed inEN 1993-1-1.

The choice of Quality Class shall be selected from Table 3.1 EN 1993-1-10depending on the consequences of lamellar tearing.




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Depending on the Quality Class selected from Table 3.1, either:

through thickness properties for the steel material shall be specified fromEN 10164, or

post-fabrication inspection shall be used to identify whether lamellar

tearing has occurred.

Guidance on the avoidance of lamellar tearing during welding is given inEN 1011-2.


5.7 Crane supporting structures – EN 1993-6 EN 1993-6 provides design rules for the structural design of runway beams and

other crane supporting structures. It covers overhead crane runways insidebuildings and outdoor crane runways for:

Overhead travelling cranes, either:

- supported on top of the runway beams or

- underslung below the runway beams

Monorail hoist blocks.



According to § 2.1.3.2(2) of EN 1993-6, the design working life of temporarycrane supporting structures shall be agreed with the Client and the PublicAuthority, taking account of possible re-use.

According to § 4(3) of EN 1993-6, where crane rails are assumed to contributeto the strength or stiffness of a runway beam, contract documents shall specifythe appropriate allowances for wear to be made in determining the properties of the combined cross-section.

According to § 4(4) of EN 1993-6, where actions from soil subsidence orseismic actions are expected, tolerances for vertical and horizontal imposeddeformations shall be specified in the contract documents, agreed with thecrane supplier, and included in the inspection and maintenance plans.

According to § 7.3(1) of EN 1993-6, the specific limits for deformations anddisplacements, together with the serviceability load combinations under whichthey apply, shall be specified in the contract documents for each project.




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6 EXECUTION SPECIFICATION

6.1 General

The necessary information and technical requirements for execution of eachpart of the works shall be agreed and complete before commencement of execution of that part of the works. Execution of works shall comply with therequirements of EN 1090-2.

6.2 Execution classesExecution Classes (EXC1 to EXC4) may apply to the whole structure or to apart of the structure or to specific details. A structure can include severalExecution Classes. A detail or group of details will normally be ascribed one

Execution Class. However, the choice of an Execution Class does notnecessarily have to be the same for all requirements.

If no Execution Class is specified EXC2 shall apply.

The list of requirements related to Execution Classes is given in Annex A.3 of EN 1090-2.

Guidance for the choice of Execution Classes is given in Annex B of EN 1090-2.

The choice of Execution Classes is related to Production Categories andService Categories, with links to Consequence Classes as defined in Annex Bof EN 1990.

6.3 Preparation gradesPreparation grades (P1 to P3 according to ISO 8501-3) are related to theexpected life of the corrosion protection and corrosivity category as defined in§ 10 of EN 1090-2.

Preparation grades may apply to the whole structure or to a part of the structure

or to specific details. A structure can include several preparation grades.A detail or group of details will normally be ascribed one preparation grade.

6.4 Geometrical tolerances Two types of geometrical tolerances are defined in § 11 of EN 1090-2:

a) Essential tolerances shall be in accordance with Annex D.1 of EN 1090-2. The values specified are permitted deviations.

- Manufacturing tolerances are described in § 11.2.2 of EN 1090-2;

- Erection tolerances are described in § 11.2.3 of EN 1090-2.




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b) Functional tolerances in terms of accepted geometrical deviations shall be inaccordance with one of the following two options:

- The tabulated values described in § 11.3.2 and Annex D.2 of EN 1090-2;

- The alternative criteria defined in § 11.3.3 of EN 1090-2.

If no option is specified the tabulated values shall apply.

Tolerances on products are defined in the standards:

- EN 10034 for structural steel I and H sections,

- EN 10056-2 for angles,

- EN 10210-2 for hot-finished structural hollow sections,

- EN 10219-2 for cold formed hollow sections.




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7 CONSTITUENT PRODUCTS

7.1 Identif ication, inspection documents and

traceability If constituent products that are not covered by the European Standards listed in

Table 2 of EN 1090-2, are to be used, their properties shall be specified in thecontract documents.

The properties of supplied constituent products shall be documented in a waythat enables them to be compared to the specified properties. Their conformitywith the relevant product standard shall be checked in accordance with § 12.2of EN 1090-2.

For metallic products, the inspection documents according to EN 10204 shallbe as listed in Table 1 of EN 1090-2.

For Execution Classes EXC3 and EXC4, constituent products shall betraceable at all stages from receipt to hand over after incorporation in theworks.

For Execution Classes EXC2, EXC3 and EXC4, if differing grades and/orqualities of constituent products are in circulation together, each item shall bedesignated with a mark that identifies its grade.

Methods of marking shall be in accordance with that for components given in

§ 6.2 of EN 1090-2.

7.2 Structural steel productsStructural steel products shall conform to the requirements of the relevantEuropean product standards as listed in Table 2 of EN 1090-2, unless otherwisespecified. Grades, qualities and, if appropriate, coating weights and finishes,shall be specified together with any required options permitted by the productstandard, including those related to suitability for hot dip zinc-coating, if relevant.

7.3 Welding consumablesAll welding consumables shall conform to the requirements of EN 13479 andthe appropriate product standard, as listed in Table 5 of EN 1090-2. The type of welding consumables shall be appropriate to the welding process (defined in§ 7.3 of EN 1090-2), the material to be welded and the welding procedure.

7.4 Mechanical fasteners

All mechanical fasteners (connectors, bolts, fasteners) shall conform to therequirements of § 5.6 of EN 1090-2.




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7.5 Grouting materials The grouting materials to be used shall conform to the requirements of § 5.7 of EN 1090-2.




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8 PREPARATION AND ASSEMBLY

This Section specifies the requirements for cutting, shaping, holing and

assembly of constituent steel components.

Structural steelwork shall be fabricated considering the surface treatmentrequirements in § 10 of EN 1090-2, and within the geometrical tolerancesspecified in § 11 of EN 1090-2.

8.1 IdentificationAt all stages of manufacturing, each piece or package of similar pieces of steelcomponents shall be identifiable by a suitable system, according to therequirements of § 6.2 of EN 1090-2.

8.2 Handling and storageConstituent products shall be handled and stored in conditions that are inaccordance with product manufacturer's recommendations. Structural steelcomponents shall be packed, handled and transported in a safe manner, so thatpermanent deformation does not occur and surface damage is minimized.

Handling and storage preventive measures specified in Table 8 of EN 1090-2shall be applied as appropriate.

8.3 Cutting Known and recognized cutting methods are sawing, shearing, disc cutting,water jet techniques and thermal cutting. Hand thermal cutting shall be usedonly if it is not practical to use machine thermal cutting. Cutting shall becarried out in such a way that the requirements for geometrical tolerances,maximum hardness and smoothness of free edges as specified in § 6.4 of EN 1090-2 are met.

8.4 Shaping Steel may be bent, pressed or forged to the required shape either by the hot orby the cold forming processes, provided the properties are not reduced belowthose specified for the worked material.

Requirements of § 6.5 of EN 1090-2 shall be applied as appropriate.

8.5 Holing Dimensions of holes, tolerances on hole-diameters and execution of holing

shall comply with the requirements of § 6.6 of EN 1090-2.




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8.6 Assembly Assembly of components shall be carried out so as to fulfil the specifiedtolerances. Precautions shall be taken so as to prevent galvanic corrosionproduced by contact between different metallic materials.

Requirements of § 6.9 and § 6.10 of EN 1090-2 shall be applied as appropriate.




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9 WELDING

9.1 General

Welding shall be undertaken in accordance with the requirements of therelevant part of EN ISO 3834 or EN ISO 14554 as applicable.

A welding plan shall be provided as part of the production planning requiredby the relevant part of EN ISO 3834. The content of a welding plan isdescribed in § 7.2.2 of EN 1090-2.

Welding may be performed by the welding processes defined in EN ISO 4063,and listed in § 7.3 of EN 1090-2.

9.2 Qualification of welding proceduresWelding shall be carried out with qualified procedures using a WeldingProcedure Specification (WPS) in accordance with the relevant part of EN ISO 15609 or EN ISO 14555 or EN ISO 15620. If specified, specialdeposition conditions for tack welds shall be included in the WPS.

Qualifications of welding procedures, depending on welding processes, aredescribed in § 7.4.1.2 and § 7.4.1.3 of EN 1090-2.

9.3 Welders and welding operatorsWelders shall be qualified in accordance with EN 287-1 and welding operatorsin accordance with EN 1418. Records of all welder and welding operatorqualification tests shall be kept available.

9.4 Welding coordinationFor Execution Class EXC2, EXC3 and EXC4, welding coordination shall bemaintained during the execution of welding by welding coordination personnelsuitably qualified for, and experienced in the welding operations they supervise

as specified in EN ISO 14731.

With respect to the welding operations being supervised, and for structuralcarbon steels, welding coordination personnel shall have a technical knowledgeaccording to Table 14 of EN 1090-2.

9.5 Preparation and execution of welding Precautions shall be taken to avoid stray arcing, and if stray arcing does occurthe surface of the steel shall be lightly ground and checked. Visual checkingshall be supplemented by penetrant or magnetic particle testing.

Precautions shall be taken to avoid weld spatter. For Execution Class EXC3and EXC4, it shall be removed.




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Visible imperfections such as cracks, cavities and other not permittedimperfections shall be removed from each run before deposition of furtherruns.

All slag shall be removed from the surface of each run before each subsequent

run is added and from the surface of the finished weld.

Particular attention shall be paid to the junctions between the weld and theparent metal.

Any requirements for grinding and dressing of the surface of completed weldsshall be specified.

Joint preparation shall be appropriate for the welding process. If qualificationof welding procedures is performed in accordance with EN ISO 15614-1,EN ISO 15612 or EN ISO 15613, joint preparation shall comply with the type

of preparation used in the welding procedure test. Tolerances for jointspreparations and fit-up shall be given in the WPS.

Joint preparation shall be free from visible cracks. Visible cracks shall beremoved by grinding and the joint geometry corrected as necessary.

If large notches or other errors in joint geometry are corrected by welding, aqualified procedure shall be used, and the area shall be subsequently groundsmooth and feathered into the adjacent surface.

All surfaces to be welded shall be dry and free from material that wouldadversely affect the quality of the welds or impede the process of welding (rust,organic material or galvanizing).

Prefabrication primers (shop primers) may be left on the fusion faces only if they do not adversely affect the welding process. For Execution Class EXC3and EXC4, prefabrication primers shall not be left on the fusion faces, unlesswelding procedure tests in accordance with EN ISO 15614-1 or EN ISO 15613have been completed using such prefabrication primers.

Other special requirements are described in EN 1090-2, as indicated in Table 9.1:




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Table 9.1 Special requirements

Clause

Storage and handling of welding consumables 7.5.2

Weather protection 7.5.3

Assembly for welding 7.5.4

Preheating 7.5.5

Temporary attachments 7.5.6

Tack welds 7.5.7

Fillet welds 7.5.8

Butt welds 7.5.9

Stud welding 7.5.12

Slot and plug welds 7.5.13

9.6 Acceptance criteriaWelded components shall comply with the requirements specified in § 10 and§ 11 of EN 1090-2.

The acceptance criteria for weld imperfections shall conform to therequirements of § 7.6 of EN 1090-2.




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10 MECHANICAL FASTENING

Section 8 of EN 1090-2 covers requirements for shop and site fastening,

including the fixing of profiled sheeting; it refers to bolting assembliesconsisting of matching bolts, nuts and washers (as necessary).

Contract documents shall specify if, in addition to tightening, other measuresor means are to be used to secure the nuts.

Minimum nominal fastener diameter, bolt length, length of protrusion, lengthof the unthreaded bolt shaft and clamp length shall comply with therequirements of § 8.2.2 of EN 1090-2.

Requirements given in § 8.2.3 of EN 1090-2 for washers shall apply.

Tightening of non-preloaded bolts shall comply with the requirements of § 8.3of EN 1090-2.

Precautions and preparation of contact surfaces in slip resistant connectionsshall comply with the requirements of § 8.4 and Table 18 of EN 1090-2. Slipfactor shall be determined by test as specified in Annex G of EN 1090-2.

Tightening methods of preloaded bolts shall comply with the requirements of § 8.5 of EN 1090-2, and shall be specified in the contract documents.




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11 ERECTION

Section 9 of EN 1090-2 gives requirements for erection and other work

undertaken on site including grouting of bases as well as those relevant to thesuitability of the site for safe erection and for accurately prepared supports.

Erection shall not commence until the site for the construction works complieswith the technical requirements with respect to the safety of the works. Safetyitems related to site conditions are listed in § 9.2 of EN 1090-2.

If the structural stability in the part-erected condition is not evident, a safemethod of erection, on which the design was based, shall be provided. Itemsrelated to the design basis method of erection are listed in § 9.3.1 of EN 1090-2.

A method statement describing the steelwork contractor's erection method shallbe prepared and checked in accordance with design rules. The erection methodstatement shall describe procedures to be used to safely erect the steelwork andshall take into account the technical requirements regarding the safety of theworks. The erection method statement shall address all relevant items in § 9.3.1of EN 1090-2; additional items are listed in § 9.3.2 of EN 1090-2.

Erection drawings or equivalent instructions, in accordance with therequirements of § 9.6.1 of EN 1090-2, shall be provided and form part of theerection method statement.

Site measurements for the works shall be in accordance with the surveyrequirements of § 9.4 of EN 1090-2.

The condition and location of the supports shall be checked visually and byappropriate measurement before the commencement of erection. If supports areunsuited to erection, they shall be corrected prior to the commencement of erection. Nonconformities shall be documented.

All foundations, foundation bolts and other supports for the steelwork shall besuitably prepared to receive the steel structure. Installation of structuralbearings shall comply with the requirements of EN 1337-11. Erection shall notcommence until the location and levels of the supports, anchors or bearingscomply with the acceptance criteria in § 11.2 of EN 1090-2, or an appropriateamendment to the specified requirements.

If foundation bolts are to be pre-stressed, arrangements shall be made that theupper 100 mm of the bolt, as a minimum, has no adhesion to the concrete.Foundation bolts intended to move in sleeves shall be provided with sleevesthree times the diameter of the bolt with a minimum diameter of 75 mm.

Whilst erection is proceeding, the supports for the steelwork shall bemaintained in an equivalent condition to their condition at the commencement

of erection.




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Areas of supports that require protection against rust staining shall be identifiedand appropriate protection provided.

Compensation for settlement of supports is acceptable, unless otherwisespecified in the contract documents. This shall be done by grouting or packing

between steelwork and support. The compensation will generally be placedbeneath the bearing.

Shims and other supporting devices used as temporary supports under baseplates shall be placed in accordance with the requirements of § 8.3, 8.5.1,§ 9.5.4 and § 9.6.5.3 of EN 1090-2.

Grouting, sealing and anchoring shall be set in accordance with theirspecification and the requirements of § 5.8, 9.5.5 and § 9.5.6 of EN 1090-2.

Components that are individually assembled or erected at the site shall be

allocated an erection mark, in accordance with the requirements of § 6.2 and§ 9.6.2 of EN 1090-2.

Handling and storage on site shall comply with the requirements of § 6.3 and§ 9.6.3 of EN 1090-2.

Any site trial erection shall be performed in accordance with the requirementsof § 6.10 and § 9.6.10 of EN 1090-2.

The erection of the steelwork shall be carried out in conformity with theerection method statement and in such a way as to ensure stability at all times.

Foundation bolts shall not be used to secure unguyed columns againstoverturning unless they have been checked for this design situation.

Throughout the erection of the structure, the steelwork shall be made safeagainst temporary erection loads, including those due to erection equipment orits operation and against the effects of wind loads on the unfinished structure.

At least one third of the permanent bolts in each connection should be installedbefore that connection can be considered to contribute to stability of the partcompleted structure.

All temporary bracing and temporary restraints shall be left in position untilerection is sufficiently advanced to allow its safe removal.

All connections for temporary components provided for erection purposes shallbe made in accordance with the requirements of EN 1090-2 and in such a waythat they do not weaken the permanent structure or impair its serviceability.

If backing bars and draw cleats are used to support the structure duringwelding, it shall be ensured that they are sufficiently strong and that theirretaining welds are appropriate for the erection load conditions.

If the erection procedure involves rolling or otherwise moving the structure, orpart of the structure, into its final position after assembly, provision shall be




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made for controlled braking of the moving mass. Provision for reversing thedirection of movement may need to be considered.

All temporary anchoring devices shall be made secure against unintentionalrelease.

Only jacks that can be locked in any position under load shall be used unlessother safety provisions are made.

Care shall be taken that no part of the structure is permanently distorted orover-stressed by stacking of steelwork components or by erection loads duringthe erection process.

Each part of the structure shall be aligned as soon as practicable after it hasbeen erected and final assembly completed as soon as possible thereafter.

Permanent connections shall not be made between components until sufficientof the structure has been aligned, levelled, plumbed and temporarily connectedto ensure that components will not be displaced during subsequent erection oralignment of the remainder of the structure.

Alignment of the structure and lack-of-fit in connections may be adjusted bythe use of shims (see above). If lack-of-fit between erected components cannotbe corrected by the use of shims, components of the structure shall be locallymodified in accordance with the methods specified in EN 1090-2. Themodifications shall not compromise the performance of the structure in thetemporary or permanent state. This work may be executed on site. Care shall

be taken with structures built of welded latticed components and spacestructures to ensure that they are not subjected to excessive forces in an attemptto force a fit against their inherent rigidity.

Unless otherwise prohibited in the contract documents, drifts may be used toalign connections. Elongation of holes for bolts used for transmission of loadsshall not be more than the values given in § 6.9 of EN 1090-2.

In case of misalignment of holes for bolts, the method of correction shall bechecked for consistency with the requirements of § 12 of EN 1090-2.

Realigned holes may be proven to comply with the oversize or slotted holerequirements specified in 8.1 of EN 1090-2, provided the load path has beenchecked.

Correction of misalignment by reaming or using a hollow milling cutter ispreferred, but if the use of other cutting methods is unavoidable, the internalfinish of all holes formed by these other methods shall be specifically checkedfor consistency with the requirements of § 6 of EN 1090-2.

Completed site connections shall be checked in accordance with 12.5 of EN 1090-2.

Erection tolerances are detailed in § 11.2.3 and Tables D.1.11 to D.1.15 and Tables D.2.19 to D.2.28 of Annex D of EN 1090-2.




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12 CONSTRUCTOR’S DOCUMENTATION

Quality documentation, mandatory for Execution Classes EXC2 to EXC4, is

defined in § 4.2.1 of EN 1090-2.

If required, a quality plan (defined in EN ISO 9000) for the execution of theworks is described in § 4.2.2 of EN 1090-2. Annex C of EN 1090-2 gives acheck-list for the content of a quality plan recommended for the execution of structural steelwork with reference to the general guidelines in ISO 10005.

Method statements giving detailed work instructions shall comply with thetechnical requirements relating to the safety of the erection works as given in§ 9.2 and § 9.3 of EN 1090-2.

Sufficient documentation shall be prepared during execution and as a record of the as-built structure to demonstrate that the works have been carried outaccording to the execution specification.

Design and structural engineering documentation shall be prepared beforeexecution of the works, and approved by any approval body designated by theOwner. The documentation should contain:

Design assumptions

Software used (if any)

Member and joint design verification

General Arrangement drawings and joint details.




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13 INTERFACES OF THE STEEL STRUCTURE

13.1 Interface to concrete surfaces

Information showing holding-down bolts and the interface of steelworkcomponents to foundations shall include a Foundation Plan showing the baselocation, position and orientation of columns, the marks of all columns, anyother components in direct contact with the foundations, their base location andlevel, and the datum level.

Similar information shall also be provided for components connecting to wallsand other concrete surfaces.

Complete details of fixing steel and bolts to the foundations or walls, methodof adjustment and packing space shall be provided.

Before erection of steelwork starts, the steelwork contractor shall inspect theprepared foundations and holding-down bolts for position and level; if he findsany discrepancies which are outside the deviations specified in § D.2.20 of EN 1090-2, he shall request that remedial work be carried out before erectioncommences.

Shims and other supporting devices used as temporary supports under baseplates shall present a flat surface to the steel and be of adequate size, strengthand rigidity to avoid local crushing of the substructure concrete or masonry.

If packings are subsequently to be grouted, they shall be placed so that thegrout totally encloses them with a minimum cover of 25 mm unless otherwisespecified.

If packings are left in position after grouting, they shall be made from materialswith the same durability as the structure.

If adjustment to the position of the base is achieved using levelling nuts on thefoundation bolts under the base plate, these may be left in position unlessotherwise specified. The nuts shall be selected to ensure that they are suitableto maintain the stability of the part-erected structure but not to jeopardize the

performance of the foundation bolt in service.

If spaces under base plates are to be grouted, fresh material shall be used inaccordance with § 5.8 of EN 1090-2.

Grouting shall not be carried out under column base plates until a sufficientportion of the structure has been aligned, levelled, plumbed and adequatelybraced.

Grouting material shall be used as follows:

The material shall be mixed and used in accordance with product

manufacturer's recommendations notably regarding its consistency whenused. Material shall not be mixed or used below 0°C unless themanufacturer's recommendations permit it.




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The material shall be poured under a suitable head so that the space iscompletely filled.

Tamping and ramming against properly fixed supports shall be used if specified and/or recommended by the grout manufacturer.

Vent holes shall be provided as necessary.

Immediately before grouting, the space under the steel base plate shall be freefrom liquids, ice, debris and contaminants.

If treatment of steelwork, bearings and concrete surfaces is required beforegrouting, it shall be specified in the contract documents.

Care shall be taken that the external profile of grouting allows water to bedrained away from structural steel components. If there is a danger of water orcorrosive liquid becoming entrapped during service, the grout around base

plates shall not be surcharged such that it rises above the lowest surface of thebase plate and the geometry of the concrete grout shall form an angle from thebase plate.

If no grouting is needed, and the edges of the base plate are to be sealed, themethod shall be specified.

Anchoring devices in concrete parts of the structure or adjacent structures shallbe set in accordance with their specification. Suitable measures shall be takento avoid damage to concrete in order to achieve the necessary anchoringresistance.

Foundations shall be adequately designed by a qualified foundation engineer tosupport the building reactions and other loads which may be imposed by thebuilding use. The design shall be based on the specific soil conditions of thebuilding site.

13.2 Interface to neighbouring constructions The mutual influence of neighbouring constructions for wind or snow actionsmust be carefully considered. Design wind and snow loads may varyconsiderably regarding the site and the construction environment, hence,

precise indications shall be given, in the contract documents, concerning thesurrounding constructions.




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APPENDIX A MODEL PROJECT SPECIFICATION

The execution of steelwork for single-storey buildings in Europe will generally

be specified to be in accordance with EN 1090-2, and the design to be inaccordance with applicable parts of the Eurocode Standards. These Standards,which cover technical requirements for a wide range of steel structures, includeclauses where the execution/design specification for the works is required togive additional information or where it has the option to specify otherrequirements.

Appendix A offers a set of clauses that may be used for single-storey steelbuilding projects to supplement and quantify the rules of the EuropeanStandards.

The clauses are arranged in a two-column format. The left column contains theproposed clauses. The right column gives a commentary to several clauses, forthe information of the person drawing up project documents; thosecommentaries are not intended to be included within the executionspecification. The model specification must be made specific to theconstruction project by completing the relevant clauses with appropriateinformation.

The model project specification proposed in this Appendix covers structuralsteelwork produced from hot rolled structural steel products only. It does notcover structural steelwork produced from cold formed structural steel (only

cold formed profiled steel sheeting and cold formed stressed-skin sheeting usedas a structural diaphragm are herein covered), structural hollow sections,channels and tubes and stainless steel products. This model projectspecification relates principally to conventional construction using constituentproducts to the standards referenced in EN 1090-2. If more complex forms of construction are involved or other products are used, designers need to considerany modifications that might be needed to the execution specification to ensurethat the desired quality and/or functionality are achieved.

For consistency, in Appendix A, those clause headings that are numbered andin bold, correspond to the Section headings of this document.




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Proposed Clauses Commentary

3 BASIS OF STRUCTURAL DESIGN

3.1 Design of steel structures shall conformto the basic requirements of § 2.1 of EN 1990.

3.2 Reliability, durability and qualitymanagement shall conform to § 2.2, 2.4and 2.5 of EN 1990.

3.3 The following additional specific eventsshall be taken into account for thedesign and the execution of thestructure: (insert list)

§ 2.1(4) of EN 1990.

3.4 The design working life of the structureshall be equal to ... years.

§ 2.3 of EN 1990.For the specification of the intended designworking life of a permanent building, seeTable 2.1 of EN 1990.

A working life of 50 years will provideadequate durability for common single-storey

buildings.

3.5 For the following additional specificcircumstances, the limit states thatconcern the protection of the contentsshall be classified as ultimate limitstates: (insert list)

§ 3.3(2) of EN 1990.

3.6 The serviceability requirements of theproject shall be as follows: (insert requirements)

§ 3.4(1) of EN 1990.

4. ACTIONS ON STRUCTURES

4.1 Self-weight and imposed loads

4.1.1 The following imposed loads shall beconsidered for serviceability limit stateverifications: (insert list)

§ 3.3.2(4) of EN 1991-1-1.In accordance with the service conditionsand the requirements concerning the

performance of the structure.

4.1.2 The characteristic values of densities of construction and stored materials shallbe taken as follows: (insert list)

§ 4.1(1) and 4.1(2) of EN 1991-1-1.Especially for materials which are not covered by the Tables in Annex A of EN 1991-1-1.

4.1.3 Loads of heavy equipments shall be asspecified on the relevant drawings.

§ 6.1(4) of EN 1991-1-1.e.g. in communal kitchens, radiology rooms,boiler rooms, etc.

4.2 Snow loads

4.2.1 In the following circumstances, testsand proven and/or properly validatednumerical methods may be used toobtain snow loads on the constructionworks: (insert particular circumstances,

if any)

§ 1.5 of EN 1991-1-3.These circumstances should be agreed uponwith the Client and the relevant authority.

4.2.2 Particular snow loads shall comply withthe following requirements: (insert special requirements, if any)

§ 4.1(1) of EN 1991-1-3.To cover unusual local conditions, theNational Annex may additionally allow theClient and the relevant authority to agreeupon different characteristic values of snow load.




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4.3 Wind loads

4.3.1 (Optional) The following rules for thevelocity pressure distribution for leewardwall and sidewalls shall apply: (insert

rules)

§ 7.2.2 of EN 1991-1-4.Certain rules may also be given in theNational Annex.

4.4 Thermal actions

4.4.1 The following specific operationalthermal effects shall apply: (insert list of specific thermal actions)

§ 5.2(2)P of EN 1991-1-5.due to heating, technological or industrial

processes.

4.4.2 The following specific values of T M and

T P shall apply: (insert values)

§ 5.2(3)P of EN 1991-1-5.

T M : linear temperature differencecomponent;

T P : temperature difference betweendifferent parts of a structure given by thedifference of average temperatures of these

parts.

4.5 Actions during execution4.5.1 The following rules concerning the

safety of persons, on and around theconstruction site, shall apply: (insert rules)

These rules are outside the scope of EN 1991-1-6.

4.5.2 Construction loads shall be as specifiedon the relevant drawings

See Tables 2.2 and 4.1 of EN 1991-1-6.

4.5.3 Tolerances for the possible deviations tothe theoretical position of constructionloads shall be as specified on therelevant drawings

If construction loads are classified as fixed loads.

4.5.4 The limits of the potential area of spatialvariation of construction loads shall beas specified on the relevant drawings

If construction loads are classified as freeloads.

4.5.5 The following minimum wind velocityduring execution phases shall apply: ...

§ 3.1(5) of EN 1991-1-6.In the absence of any choice in the National

Annex.

4.5.6 The following rules of combination of snow loads and wind action with theconstruction loads shall apply: (insert rules)


Annex.

4.5.7 The geometric imperfections of thestructure and the structural elementsduring execution shall be as follows:(insert values)


Annex.

4.5.8 Criteria associated with serviceabilitylimit states during execution shall be asfollows: (insert criteria)


Annex.

4.5.9 The maximum allowable wind velocityduring crane operations shall be ...

§ 4.7(1) of EN 1991-1-6.

4.6 Accidental actions

4.6.1 The following notional accidental loadsshall apply: (insert accidental actions)

Equivalent static design forces due tovehicular impact;Frontal and lateral dynamic design forcesdue to impact from river and canal traffic, aswell as the height of application of the impact force and the impact area;

Classification of structures subject to impact from derailed railway traffic ( § 4.5.1.2 of EN 1991-1-7);




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4.7 Actions induced by cranes

4.7.1 For the serviceability limit stateverification, tests shall (or may not) beperformed with the cranes on the

supporting structures (specify thealternative to be recommended)

§ 2.3(6) of EN 1991-3

4.7.2 The following forces shall be included inthe same group of simultaneous craneload components: (insert list of forces)

§ 2.5.2.2(2) of EN 1991-3

Insert one or several forces among the fivehorizontal types (a) to (e) listed in§ 2.5.2.2(1) of EN 1991-3.

4.7.3 The longitudinal horizontal forces H L,i and the transverse horizontal wheelforces H T,i, caused by acceleration anddeceleration of masses of the crane orthe crab, shall be applied according tothe following provisions: (insert

provisions)

§ 2.5.2.2(4) of EN 1991-3

Otherwise, provisions given in Figure 2.3 of EN 1991-3 should apply.

4.7.4 The maximum number of cranes to betaken into account as actingsimultaneously shall be: (insert number)

§ 2.5.3(2) of EN 1991-3

4.7.5 The Hoisting Class of the crane shallbe: (specify class from HC1 to HC4)

Hoisting class to be specified unless it isspecified in the crane supplier specification.

Reference can be made to Annex B(informative) of EN 1991-3

4.7.6 The vertical load to be applied to accesswalkways, stairs and platform shall beequal to: (insert provisions)

§ 2.9.1(1) of EN 1991-3

Otherwise, provisions given in § 2.9.1(2),2.9.1(3) or 2.9.1(4) should apply.

4.7.7 The horizontal load to be applied to theguard rail shall be equal to: (insert

provisions)

2.9.2(1) of EN 1991-3

Otherwise, provisions given in § 2.9.2(1) or 2.9.2(2) should apply.

4.7.8 To make allowance of relevantaccidental actions:- Buffers are (or are not) used;- A crane with horizontally restrainedloads can (or cannot) tilt when its loador lifting attachment collides with anobstacle.(specify construction conditions)

4.7.9 To make allowance for fatigue effects,

the following operational conditions shallapply: (insert information)

If sufficient information is provided, the

fatigue loads can then be determined according to EN 13001 and Annex A of EN 1993-1-9.

Otherwise, provisions of § 2.12 of EN 1991-3should apply.




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(Optional clause in case a simplified approachfor determining the fatigue loads isfavoured)

4.7.10- The class of load spectrum for all tasks

of the crane shall be: (specify class fromQ0 to Q5);- The class of total number of workingcycles (U0 to U9) during the design life of the crane shall be: (specify class fromU0 to U9);- The crane classification shall be:(specify class from S0 to S9)

If the crane classification is not included inthe crane supplier specification, referencecan be made to Annex B (informative) of EN 1991-3.

4.7.11 The partial factor for actions on cranesupporting structures to be taken inserviceability limit states shall be equalto: (specify factor value)

Clause A.3.2(1) of normative Annex A of EN 1991-3

Otherwise, this partial factor should be taken

as 1,0.

4.8 Seismic actions

4.8.1 The Importance Class of the projectshall be ...

Table 4.3 of EN 1998-1.Ordinary buildings (other than schools, firestations, power plants, hospitals, etc.)correspond to Importance Class II;

4.8.2 The Ground Type shall be as specifiedon the relevant documents.

Table 3.1 of EN 1998-1.Depending on the particular conditions of the

project, contract documents should specify whether ground investigations and/or geological studies should be performed toidentify the ground type;

4.8.3 The seismic zone of the project shallbe....

According to the zonation map, decided by the National Authority, and found in theNational Annex of EN 1998-1

4.8.4 Earthquake resistant steel building shallbe designed according to ... concept

DCL, DCM or DCH.

5. DESIGN OF STEEL STRUCTURES

5.1 General rules

5.1.1 To ensure durability, the building and itscomponents shall either be designed forenvironmental actions (and fatigue if relevant) or else protected from them.

§ 2.1.3.3(1)B of EN 1993-1-1.

5.1.2 The effects of deterioration of materialand corrosion (and fatigue whererelevant) shall be taken into account byappropriate choice of material (seeEN 1993-1-4 and EN 1993-1-10), anddetails (see EN 1993-1-9), or bystructural redundancy and by the choiceof an appropriate protection system.

§ 2.1.3.3(2)B of EN 1993-1-1.

5.1.3 For the following components, thepossibility of their safe replacementshall be verified as a transient designsituation (insert list of the components of the building that need to be replaceable)

§ 2.1.3.3(3)B of EN 1993-1-1.




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5.1.4 With reference to Annex A1.4 of EN 1990, vertical deflections (accordingto Figure A1.1), horizontal deflections(according to Figure A1.2) andvibrations of structures on which the

public can walk shall comply with thefollowing limits: (insert serviceability limits states)

§ 7 of EN 1993-1-1.

5.2 Rules for sheeting

5.2.1 The Structural Class of the construction(Class I to III), associated with failureconsequences according to Annex B of EN 1990, shall be as specified on therelevant documents.

§ 2(6) of EN 1993-1-3Structural Class I: construction wheresheeting is designed to contribute to theoverall strength and stability of a structure;Structural Class II: construction wheresheeting is designed to contribute to thestrength and stability of individual structural elements;Structural Class III: construction where

sheeting is used as an element that only transfers loads to the structure.

5.4 Design of joints

5.4.1 Bolted connections Category shall be asspecified on the relevant documents.

§ 3.4.1 of EN 1993-1-8.

5.4.2 Friction surfaces for slip-resistantconnections using pre-loaded 8.8 or10.9 bolts shall be as specified on therelevant documents.

§ 3.9 of EN 1993-1-8.

5.4.3 According to EN ISO 25817, the qualitylevel of welds shall be as specified onthe relevant documents.

§ 4.1 of EN 1993-1-8.

5.4.4 The frequency of inspection of weldsshall conform to the requirements of EN 1090-2 and shall be as specified onthe relevant documents.

§ 4.1 of EN 1993-1-8.

5.5 Fatigue

5.5.1 Fatigue assessment shall beundertaken using ‘damage tolerantmethod’ or ‘safe life method’ (specify assessment method to be used).

§ 3(1) of EN 1993-1-9

If the ‘damage tolerant method’ is specified, a prescribed inspection and maintenanceregime for detecting and correcting fatiguedamage should be implemented throughout the design life of the structure.The ‘safe life method’ should be specified in

cases where local formation of cracks in onecomponent could rapidly lead to failure of thestructural element or structure.

5.5.2 In order to determine the partial factorfor fatigue strength, in conjunction withthe specified fatigue assessmentmethod, the failure Consequenceclassification shall be taken as ‘LowConsequence’ or ‘High Consequence’(specify the consequence class).

§ 3(7) of EN 1993-1-9

5.6 Material toughness and through-thickness properties

5.6.1 The guidance given in section 2 of

EN 1993-1-10 shall be used for theselection of materials for fracturetoughness.




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5.6.2 The guidance given in section 3 of EN 1993-1-10 shall be used for theselection of materials for through-thickness properties.

5.7 Crane supporti ng structures5.7.1 Where crane rails are assumed to

contribute to the strength or stiffness of a runway beam, the properties of thecombined cross-section shall bedetermined as follows:(Specify the appropriate allowances for wear to be made).

§ 4(3) of EN 1993-6

5.7.2 Where actions from soil subsidence orseismic actions are expected,tolerances for vertical and horizontalimposed deformations shall be taken asfollows:(Specify the appropriate allowances).

§ 4(4) of EN 1993-6

These allowances should be agreed with thecrane supplier, and included in the inspectionand maintenance plans.

5.7.3 The limits for deformations anddisplacements shall be taken as follows:(specify the specific limits together withthe serviceability load combinationsunder which they apply).

§ 7.3(1) of EN 1993-6

6. EXECUTION SPECIFICATION

6.1 General

6.1.1 The requirements for the execution of structural steelwork for the project aregiven in the following documents: (Insert list)

Insert a list of the relevant drawings and other documents, including reference toEN 1090-2.

6.2 Execution Class

6.2.1 For building structures, EXC2 shallgenerally apply, except where specifiedotherwise on the drawings.

The use of EXC2 as the default class will provide adequate reliability for most elementsof ordinary buildings. For some structures, agreater scope of inspection and testing and/or higher quality level acceptance criteriamay be required, either generally or for

particular details. Particular details where thisis required, such as where special inspectionand testing is required, should be indicated on the drawings.Table A.3 of EN 1090-2 gives a list of requirements related to execution classes;

Annex B of EN 1090-2 gives guidance for thechoice of execution classes;The choice of execution classes is related to

production categories and service categories,with links to consequence classes as defined in Annex B of EN 1990.

6.3 Preparation grades

6.3.1 The preparation grade of all surfaces towhich paints and related products are tobe applied shall be ...Otherwise,

The expected life of the corrosionprotection shall be ... years or corrosivity

category shall be ...

Preparation grades (P1 to P3 according toISO 8501-3) are related to the expected lifeof the corrosion protection and corrosivity category as defined in § 10 of EN 1090-2.




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6.4 Geometrical tolerances

6.4.1 For essential tolerances, the tabulatedvalues in Annex D.1 of EN 1090-2 shallapply.

If the steelwork is not within tolerance, itshall be reported to the designer of thepermanent works and shall be adjusted,if necessary, to maintain the structuraladequacy in accordance with the designrules.

Manufacturing tolerances are described in§ 11.2.2 of EN 1090-2;Erection tolerances are described in § 11.2.3

of EN 1090-2;

6.4.2 For functional tolerances (in terms of accepted geometrical deviations), either the tabulated values in § 11.3.2 andAnnex D.2 of EN 1090-2 shall apply, or ,the alternative criteria defined in§ 11.3.3 of EN 1090-2 shall apply.

7. CONSTITUENT STEEL PRODUCTS

7.1 Identification, inspection documentsand traceability

7.1.1 Properties for (...) shall comply with therequirements given in (...).

§ 5.1 of EN 1090-2 Insert details for any constituent product not covered by the European Standards listed inTable 2 of EN 1090-2.

7.1.2 The inspection documents (according toEN 10204) shall be as listed in Table 1of EN 1090-2.

§ 5.2 of EN 1090-2.

(Optional clause) 7.1.3 For Execution Classes EXC3 and

EXC4, constituent products shall be

traceable at all stages from receipt tohand over after incorporation in theworks.

§ 5.2 of EN 1090-2.

7.1.4 For Execution Classes EXC2, EXC3and EXC4, if different grades and/orqualities of constituent products are incirculation together, each item shall bedesignated with a mark that identifies itsgrade.

§ 5.2 of EN 1090-2.Methods of marking should be in accordancewith that for components given in § 6.2 of EN 1090-2.If marking is required, unmarked constituent

products should be treated as nonconforming product.

7.2 Structural steel products

7.2.1 The grade and quality of structural steelshall be as specified on the drawings.

7.2.2 For structural steel plates, thicknesstolerances class A, in accordance withEN 10029, shall be used.

§ 5.3.2 of EN 1090-2.Class A is usually sufficient, even whereEXC4 is specified, but if class C is required by the technical authority or for other reasons, that class should be specified instead.




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7.2.3 Structural carbon steels shall conform tothe requirements of the relevantEuropean product standards as listed in

Table 2 of EN 1090-2, unless otherwisespecified on the drawings. Grades,

qualities and, if appropriate, coatingweights and finishes, together with anyrequired options permitted by theproduct standard, including thoserelated to suitability for hot dip zinc-coating, if relevant, shall be as specifiedon the drawings.

§ 5.3.1 of EN 1090-2.

7.2.4 For carbon steels, surface conditionshall be as follows:Class A2, for plates in accordance withthe requirements of EN 10163-2;Class C1, for sections in accordancewith the requirements of EN 10163-3.If relevant, surface imperfections (such

as cracks, shell or seams) or repair of surface defects by grinding inaccordance with EN 10163, shallcomply with the following restrictions :(insert list of special restrictions)

§ 5.3.3 of EN 1090-2.

(Optional clause) 7.2.5 For EXC3 and EXC4, the locations (and

width) where internal discontinuityquality class S1 of EN 10160 isrequired, are specified on the relevantdrawings.

§ 5.3.4 of EN 1090-2.Especially for welded cruciform jointstransmitting primary tensile stresses throughthe plate thickness, and for areas close tobearing diaphragms or stiffeners.

7.2.6 Areas where material shall comply withrequirements for improved deformation

properties perpendicular to the surface(according to EN 10164) are specifiedon the drawings.

§ 5.3.4 of EN 1090-2.Consideration should be given to specifying

such material for cruciform, T and corner joints. Should only be invoked wherenecessary; specify only those parts of thestructure which need these properties.

7.3 Welding consumables

7.3.1 All welding consumables shall conformto the requirements of EN 13479 andthe appropriate product standard, aslisted in Table 5 of EN 1090-2. The typeof welding consumables shall beappropriate to the welding process(defined in § 7.3 of EN 1090-2), thematerial to be welded and the welding

procedure.

§ 5.5 of EN 1090-2.

7.4 Mechanical fasteners

7.4.1 All mechanical fasteners (connectors,bolts, fasteners) shall conform to therequirements of § 5.6 of EN 1090-2.Studs for arc stud welding includingshear connectors for steel/concretecomposite construction shall complywith the requirements of EN ISO 13918.

7.4.2 The property classes of non-preloadedbolts and nuts, and surface finishes,shall be as specified on the drawings.

7.4.3 The property classes of preloaded boltsand nuts, and surface finishes, shall beas specified on the drawings.

HV bolts are sensitive to over-tightening, sothey require a greater level of site control.It is not advisable to use both HR and HV assemblies on the same project.




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7.4.4 The chemical composition of weatherresistant assemblies shall comply withthe requirements for Type 3 Grade Afasteners to ASTM standard A325, orequivalent.

7.4.5 Reinforcing steels may be used forfoundation bolts. In this case, they shallconform to EN 10080 and the steelgrade shall be as specified on thedrawings.

(Optional clause) 7.4.6 Where locking devices are specified on

the drawings, they shall comply with therelevant standards listed in § 5.6.8 of EN 1090-2, and additionally ... (Insert any particular requirements for locking devices).

7.5 Grouting materials

7.5.1 Grouting materials to be used shall beas specified on the relevant drawings.

8. PREPARATION AND ASSEMBLY

8.1 Identification

8.1.1 Soft or low stress stamps may be used,except in any areas specified on thedrawings.

Soft or low stress stamp marks can easily beobliterated by the protective system. Thefabricator will usually mask the stamped areaafter application of primer and complete thecoating locally after erection.

8.1.2 Areas where identification marks are not

permitted or shall not be visible aftercompletion are specified on thedrawings.

8.2 Handling and storage

8.2.1 Structural steel components shall bepacked, handled and transported in asafe manner, so that permanentdeformation does not occur and surfacedamage is minimized.Handling and storage preventivemeasures specified in Table 8 of EN 1090-2 shall be applied asappropriate.

8.3 Cutting

8.3.1 Hand thermal cutting shall be used onlyif it is not practical to use machinethermal cutting.Cutting shall be carried out in such away that the requirements forgeometrical tolerances, maximumhardness and smoothness of freeedges, as specified in § 6.4 of EN 1090-2, are met.

8.4 Shaping

8.4.1 Requirements of § 6.5 of EN 1090-2

shall be applied as appropriate.




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8.5 Holing

8.5.1 Dimensions of holes, tolerances onhole-diameters and execution of holingshall comply with the requirements of

§ 6.6 of EN 1090-2.8.5.2 Where specified on the drawings, holes

with special dimensions shall beprovided for connections of movement

joints.

8.5.3 Special tolerances on hole diametersshall be as specified on the drawings.

Special tolerances would only be needed inexceptional conditions.If pins are used, tolerances should bespecified for both holes and pins.

8.5.4 Holes for fasteners shall be formed bydrilling or by punching followed byreaming.

8.5.5 Long slotted holes shall be executed asspecified on the drawings.

This option is only needed for special cases,such as slotted holes for pins in movement

joints. Details must then be given on thedrawings.

8.6 Assembly

8.6.1 Requirements of § 6.9 and 6.10 of EN 1090-2 shall be applied asappropriate.

8.6.2 Holes for which elongation is notpermitted are shown on the relevantdrawings.

This option is needed for fit bolts for instance.

8.6.3 The acceptability of the addition of anywelded temporary attachments and the

making of any butt welds additional tothose specified on the drawings shall beverified according to the design rules.A record of the details of suchattachments and butt welds shall beprovided as part of the constructor’sexecution documentation.Areas where temporary attachmentshave been made shall be made good.If weld repairs are necessary these shallbe carried out in accordance with therequirements of the appropriateStandard.

If there are any restrictions on positioning of temporary attachments, they should be

specified, either in this clause or on thedrawings.In general, temporary welded attachmentsare not acceptable within 25 mm of theedges of flange plates.

9. WELDING

9.1 General

9.1.1 Welding shall be undertaken inaccordance with the requirements of therelevant part of EN ISO 3834 orEN ISO 14554 as applicable.

9.1.2 A welding plan shall be provided as partof the production planning required bythe relevant part of EN ISO 3834.

The content of a welding plan is described in§ 7.2.2 of EN 1090-2.

9.1.3 Welding may be performed by thewelding processes defined inEN ISO 4063.

Welding processes are listed in § 7.3 of EN 1090-2.




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9.2 Qualification of welding procedures

9.2.1 Welding shall be carried out withqualified procedures using a WeldingProcedure Specification (WPS) in

accordance with the relevant part of EN ISO 15609 or EN ISO 14555 orEN ISO 15620.

Qualifications of welding procedures,depending on welding processes, aredescribed in § 7.4.1.2 and 7.4.1.3 of

EN 1090-2.

9.3 Welders and welding operators

9.3.1 Welders shall be qualified in accordancewith EN 287-1 and welding operators inaccordance with EN 1418.Records of all welder and weldingoperator qualification tests shall be keptavailable.

9.4 Welding coordination

9.4.1 Welding coordination shall bemaintained during the execution of welding by welding coordinationpersonnel suitably qualified for, andexperienced in the welding operationsthey supervise as specified inEN ISO 14731.

This option is needed for Execution ClassEXC2, EXC3 and EXC4.With respect to the welding operations being supervised, and for structural carbon steels,welding coordination personnel should havea technical knowledge according to Table 14of EN 1090-2.

9.5 Preparation and execution of welding

9.5.1 Precautions shall be taken to avoidstray arcing, and if stray arcing doesoccur the surface of the steel shall belightly ground and checked. Visualchecking shall be supplemented bypenetrating or magnetic particle testing.

9.5.2 Precautions shall be taken to avoid weldspatter. For Execution Class EXC3 and EXC4, weld spatter should be removed.

9.5.3 Visible imperfections such as cracks,cavities and other not permittedimperfections shall be removed fromeach run before deposition of furtherruns.

9.5.4 All slag shall be removed from thesurface of each run before eachsubsequent run is added and from thesurface of the finished weld.

9.5.5 Particular attention shall be paid to the junctions between the weld and the

parent metal.

9.5.6 Special requirements for grinding anddressing of the surface of completedwelds are shown on the relevantdrawings.

9.5.7 J oint preparation shall be free fromvisible cracks. Visible cracks shall beremoved by grinding and the jointgeometry corrected as necessary.

9.5.8 If large notches or other errors in jointgeometry are corrected by welding, aqualified procedure shall be used, andthe area shall be subsequently ground

smooth and feathered into the adjacentsurface.




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9.5.9 All surfaces to be welded shall be dryand free from material that wouldadversely affect the quality of the weldsor impede the process of welding.

Such as rust, organic material or galvanizing.

9.5.10 Requirements of § 7.5.1 to 7.5.16 of EN 1090-2 shall be applied asappropriate.

9.6 Acceptance crit eria

9.6.1 Welded components shall comply withthe requirements specified in § 10 and11 of EN 1090-2.

9.6.2 The acceptance criteria for weldimperfections shall conform to therequirements of § 7.6 of EN 1090-2.

10. MECHANICAL FASTENING

10.1 General

10.1.1 Minimum nominal fastener diameter,bolt length, length of protrusion, lengthof the unthreaded bolt shaft and clamplength shall comply with therequirements of § 8.2.2 of EN 1090-2.

10.1.2 Requirements given in § 8.2.3 of EN 1090-2 for washers shall apply.

10.1.3 Tightening of non-preloaded bolts shallcomply with the requirements of § 8.3 of EN 1090-2.

The bolt shall protrude from the face of the nut, after tightening, not less than

one full thread pitch.

10.1.4 P recautions and preparation of contactsurfaces in slip resistant connectionsshall comply with the requirements of § 8.4 and Table 18 of EN 1090-2. Slipfactor shall be determined by test asspecified in Annex G of EN 1090-2.

10.1.5 Tightening methods of preloaded boltsshall comply with the requirements of § 8.5 of EN 1090-2; specialrequirements are specified on therelevant documents.

10.2 Bolts10.2.1 Bolt sizes for structural bolting shall be

as specified on the drawings.

10.2.2 Where the structure has been designedto utilise the shear resistance of theunthreaded shank of bolts, this isspecified on the drawings and thedimensions of the bolts are given.

The locations and dimensions must be givenon the drawings. Reliance on the resistanceof the unthreaded shank, rather than thethreaded part, is inadvisable because it requires a higher level of control on bolt supply and installation to ensure that only unthreaded parts exist in the part of theconnection where the resistance to shear isrequired.

10.3 Nuts

10.3.1 Nuts shall be assembled so that theirdesignation markings are visible forinspection after assembly.




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10.3.2 Nuts shall run freely on their partneringbolt, which is easily checked duringhand assembly.

Any nut and bolt assembly where the nut does not run freely should be discarded.

10.4 Washers

10.4.1 Washers shall be provided under thenut or the bolt head of non-preloadedbolts, whichever is to be rotated.

10.4.2 For preloaded bolts :- for 8.8 bolts, a washer shall be usedunder the bolt head or the nut,whichever is to be rotated;- for 10.9 bolts, washers shall be usedunder both the bolt head and the nut.

10.5 Preparation of contact surfaces in slip-resistant connections.

10.5.1 The area of contact surfaces inpreloaded connections shall be asspecified on the drawings.For contact surfaces in slip-resistantconnections shown on the relevantdrawings, the following particulartreatment shall apply: ... (Insert requirements).

The treated surfaces shall beadequately protected until they arebrought together.

10.5.2 Preparation of contact surfaces in slip-resistant connections shall comply withthe requirements of § 8.4 of EN 1090-2;special requirements are specified on

the relevant documents.10.6 Tightening of preloaded bolts

10.6.1 The nominal minimum preloading forceF p,C shall be taken as indicated on therelevant drawings.

Usually, F p,C = 0,7.f ub.As.

10.6.2 The following tightening method(s) shallbe used: ... (insert specific tightening

methods)

The different tightening methods aredescribed in Table 20 of EN 1090-2.

10.6.3 As an alternative to Table 20 of EN 1090-2, calibration to Annex H of EN 1090-2 may be used:- for all tightening methods;- for all tightening methods, except forthe torque method.(choose one of the above options)

10.6.4 When bolts are tightened by rotation of the bolt head, the following specialprecautions shall be taken: ... (insert special precautions depending on thetightening method adopted).

10.6.5 For thick surface coatings shown on therelevant drawings, the followingmeasures shall be taken to offsetpossible subsequent loss of preloadingforce: ... (insert specific measures,depending on the tightening method

adopted).

If torque method is used, this may be by retightening after a delay of some days.




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10.6.6 For the combined method, when usingthe value M r,1 for the first tighteningstep, the simplified expression of M r,1 (in§ 8.5.4 of EN 1090-2) may (or may not)be used. (choose one of the above

options)

10.6.7 For the combined method, values otherthan those given in Table 21 of EN 1090-2 shall not be used unlesscalibrated in accordance with Annex Hof EN 1090-2.

10.6.8 For the HRC method, the first tighteningstep shall be repeated as necessary if the pre-tightening is relaxed by thesubsequent tightening of the remainderof the bolts in the connection.

This first step should be completed for all bolts in one connection prior tocommencement of the second step.Guidance of the equipment manufacturer may give additional information on how toidentify if pre-tightening has occurred, e.g.sound of shear wrench changing, or if other

methods of pre-tightening are suitable.10.7 Fit bolts

10.7.1 Where permitted on the drawings, thelength of the threaded portion of theshank of a fit bolt may exceed 1/3 of thethickness of the plate, subject to thefollowing requirements: ... (Insert details)

Insert this clause if such permission is to begiven and specify on the drawings for whichbolts the longer thread length is permitted.

11. ERECTION

11.1 The design is based on the constructionmethod and/or sequences given in the

following documents: (Insert list).11.2 Requirements for temporary bracing

compatible with the construction methodand/or sequences are specified on thefollowing drawings: (Insert list)

Insert list of relevant drawings and other documents. Information should include,

amongst other things, allowances for permanent deformations (pre-camber),settlement of supports, assumptions for temporary stability and assumptions about

propped/un-propped conditions in staged construction.The designer has the duty to ensure that the

permanent works can be built safely. Thedrawings will show a construction method and/or sequences and will show either indetail or indicatively the nature and positionsof temporary bracings compatible with thosesequences. These temporary bracings will normally be those required to provide stability in the ‘bare steel’ and ‘wet concrete’

conditions. The elements of the temporary bracing would normally be designed by the

permanent works designer; if that is not thecase, it should be stated in the contract documents (preferably on the drawings) that their design is the constructor’s responsibility.




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11.3 The allowances for permanentdeformation and other associateddimensions specified on the relevantdrawings allow for the quasi-permanenteffects of the following actions, using

the design basis method of erection:i) after steelwork erection:

- Self weight of structural steelwork;ii) after completion of structure:

- Self weight of structural steelwork;- Self weight of structural concrete;- Self weight of non-structural parts;- The effects of shrinkage modifiedby creep.

It is the designer’s responsibility to determinethe allowances (i.e. the addition to thenominal profile) required to offset the effectsof permanent actions, including shrinkageeffects. These allowances have often been

termed, somewhat loosely, ‘pre-camber’.

11.4 If the constructor proposes to adopt analternative construction method and/orsequences to that referred to 11.1, theconstructor shall verify, in accordancewith the design rules, that the alternative

method and/or sequences can be usedwithout detriment to the permanentworks.

The constructor shall allow a period of at least ... (insert number) weeks for theverification of the erection method inaccordance with the design rules, to thesatisfaction of the permanent worksdesigner.

For major single-storey structures, the designbasis method of erection will normally be

produced through a close working betweenthe designer and the constructor because themethod of erection will often dictate aspects

of the design.Even for lesser or minor structures, thefundamental issue is that the constructor'serection method must be compatible with thedesign basis method of erection or, if it isdifferent, for whatever reason, the design of the permanent works must be re-verified, for that erection method.

11.5 The steelwork dimensions on thedrawings are specified for a referencetemperature of ... °C (Insert referencetemperature)

The steelwork contractor will makeadjustments to suit the calibrationtemperature of his measuring equipment.

11.6 Compensation for settlement of supports shall be made by theconstructor if such settlement differsfrom the design assumptions.

The designer should state the range of settlement of the supports (including temporary supports) that was considered inthe design.

11.7 The finished cover to steel packings(comprising a total thickness of groutand any concrete) shall comply with thecover requirements of EN 1992.

It is normal practice to remove steel packings. Softer packings may be left in place.

11.8 Packings and levelling nuts may be leftin position, provided that it can beverified, in accordance with the designrules, that there is no detriment to thepermanent works.

The implications of introducing a hard spot into the bearing area should be checked withrespect to both steel and concrete elements.

11.9 The treatment of steelwork, bearingsand concrete surfaces before groutingshall be as specified on the drawings.

11.10 Areas where the edges of the baseplate are to be sealed, without grouting,are specified on the drawings.

If grouting is not specified in bearing areas,the perimeter of base plates should besealed. The locations for sealing must beshown on the drawings.

11.11 Surfaces that are to be in contact withconcrete, including the undersides of baseplates, shall be coated withprotective treatment applied to thesteelwork, excluding any cosmeticfinishing coat, for the first ...mm(insert

length, minimum 50 mm) of theembedded length, and the remainingsurfaces need not be coated (or shall becoated, choose one option).

Additional requirements are given in § 10.7 of EN 1090-2.





Part 11: Moment Connections









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11 - iii

FOREWORD

This publication is part eleven of the design guide, Single-Storey Steel Buildings.




Part 3: Actions



















11 - iv




11 - v

ContentsPage No

FOREWORD iii

SUMMARY vi

1 INTRODUCTION 1 1.1 Design approach 1 1.2 Tension zone 1 1.3 Plastic distribution 4 1.4 Resistance of the compression zone 5 1.5 Resistance of the column web panel 6 1.6 Calculation of moment resistance 6 1.7 Weld design 7 1.8 Vertical shear 8 1.9 Stiffeners 9

2 J OINT STIFFNESS 10 2.1 Classification by calculation 10 2.2 Classification boundaries 11

3 BEST PRACTICE GUIDELINES FOR MOMENT CONNECTIONS 12 3.1 Eaves haunch 12 3.2 End plate 12 3.3 Stiffeners 13 3.4 Bolts 13 3.5 Apex connections 14 3.6 Welds 14

4 J OINT DESIGN TABLES 16 4.1 General 16 4.2 Main design assumptions 17 4.3 Notes to the tables 18 4.4 Apex connections 21 4.5 Eaves connections 37

REFERENCES 53




11 - vi

SUMMARY

This publication provides an introduction to the design process for moment-resisting

bolted connections in single storey steel framed buildings. It explains that the design

process is complex, involving many steps to determine the resistance of individual bolt

rows in the tension zone, checking whether the resistance of the bolt group has to bereduced on account of the performance of the connected elements, and evaluating the

bending resistance from the tensile resistances of the rows. To simplify design, a series

of design tables for standard connections are given, for eaves and apex connections in

portal frames, with haunched and un-haunched rafters.




11 - 1

1 INTRODUCTION

Manual design of moment-resisting bolted connections is laborious,

particularly when there are several bolt rows acting in tension. Any iteration of connection geometry or connection component (such as changing the bolt

setting out or bolt size) necessitates a full re-design. For these reasons, the

design of moment-resisting bolted connections is generally carried out by using

appropriate software.

This Section aims to provide an introduction to the verification process

described in EN 1993-1-8[1].

1.1 Design approach

The verification of a bolted moment resisting connection involves three distinctsteps:

1. Determine the potential resistance of the bolt rows in the tension zone, in

isolation.

2. Check whether the total tension resistance can be realised, as it may be

limited by the web panel shear resistance of the column, or the resistance of

the connection in the compression zone.

3. Calculate the moment resistance as the sum of the tension forces multiplied

by their respective lever arms.

The key features of the approach are firstly that a plastic distribution of bolt

row forces is allowed, as long as either the end plate or the column flange is

sufficiently thin. The second key feature is that the complex yield line patterns

in the tension zone are replaced by an equivalent, simple T-stub model which is

more amenable to calculation.

1.2 Tension zoneAccording to EN 1993-1-8 § 6.2.7.2(6), the effective design tension resistance

F tr,Rd at each bolt row in the tension zone is the least of the following

resistances:

Column flange bending and bolt strength ( F t,fc,Rd)

Column web in transverse tension ( F t,wc,Rd)

End plate bending and bolt strength ( F t,ep,Rd)

Rafter beam web in tension ( F t,wb,Rd).

For each bolt row, the effective design tension resistance may thus be

expressed as:

F tr ,Rd = min( F t,fc,Rd; F t,wc,Rd; F t,ep,Rd; F t,wb,Rd)

The relevant clauses of EN 1993-1-8 for the above components are given in

Table 1.1.




11 - 2

Table 1.1 Components of the join t to determine the potential design resistance of a bolt row

Component EN 1993-1-8 clause number

Column flange in bending Rd fc,t, F 6.2.6.4 and Table 6.2

Column web in transversetensionRd wc,t, F 6.2.6.3

End-plate in bending Rd t,ep, F 6.2.6.5 and Table 6.6

Rafter web in tension Rd wb,t, F 6.2.6.8

The resistance for each row is calculated in isolation. The connection resistance

may be limited by:

The design resistance of a group of bolts

The stiffness of the column flange or end plate, which may preclude a

plastic distribution of tension forces

The shear resistance of the column web panel

The resistance in the compression zone.

Because the tension resistance of a row may be limited by the effects of forces

in other rows in the bolt group, the effective design tension resistances are

considered to be potential resistances – their full realisation may be limited by

other aspects of the design.

The potential design tension resistance F tr ,Rd for each bolt row should be

determined in sequence, starting from the furthest bolt row from the centre of compression (with the maximum lever arm). In accordance with § 6.2.7.2(4),

the resistance of any bolt rows closer to the centre of compression are ignored

when calculating the resistance of a specific bolt row, or group of rows.

Subsequent rows are verified both in isolation and also as part of a group in

combination with rows above. The resistance of row 2 is therefore taken as the

lesser of:

the resistance of row 2 acting alone, and

the resistance of rows 1 and 2 acting as a group minus the resistance

already allocated to row 1.

Row 1 is furthest from the centre of compression, and rows are numbered

sequentially.

A stiffener in the column, or in the rafter, disrupts any common yield line

pattern, which means that groups containing a stiffener need not be verified on

that side. In a detail with an extended end plate, such as in Figure 1.1, the

flange of the rafter means that there cannot be a common yield line pattern

around the top two bolt rows in the end plate. On the column side, however, a

common yield line pattern around the top two rows is possible, and must be

verified.




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r =1

r =2

r =3

r =4

Figure 1.1 Extended end plate in a haunched eaves connection

1.2.1 End plate and column flange in bending

When determining the potential tension resistance of the end plate in bending,

F t,ep,Rd and the column flange in bending, F t,fc,Rd, EN 1993-1-8 converts the real

yield line patterns into an equivalent T-stub. Generally, a number of yield line

patterns are possible – each with a length of equivalent T-stub. The shortest

equivalent T-stub is taken. When bolts are located adjacent to a stiffener, or

adjacent to the rafter flange, the increased resistance of the flange or end plate

is reflected in a longer length of equivalent T-stub. Bolts adjacent to an

unstiffened free edge will result in a shorter length of equivalent T-stub. .

Effective lengths of equivalent T-stubs eff are given in Table 6.4 of

EN 1993-1-8 for unstiffened flanges, in Table 6.6 for unstiffened end plates

and in Table 6.5 for stiffened flanges (or end plates)..

In all cases, effective lengths of equivalent T-stubs are given for individual bolt

rows and for bolt rows as part of a group – the length of the equivalent T-stub

for a group of bolts is assembled from the contributions of the rows within the

group.

The beneficial effect of stiffeners depends on the geometry of the stiffener, the

location of the bolt and the proximity to the web. This is addressed inFigure 6.11 of EN 1993-1-8, which provides an factor used in determining

the effective length of equivalent T-stub. When the bolt is sufficiently far from

both web and stiffener, the stiffener has no effect – the effective length is the

same as that in an unstiffened zone.

Once the effective length of T-stub has been determined, the resistance of the

T-stub is calculated. Three modes, as illustrated in Figure 1.2, are examined:

Mode 1, in which the flange of the T-stub is the critical feature, and yields

in double curvature bending

Mode 2, in which the flange and the bolts yield at the same load

Mode 3, in which the bolts are the critical component and the resistance is

the tension resistance of the bolts.




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Mode 1 Mode 2 Mode 3

Figure 1.2 Behaviour modes of an equivalent T-stub

The expressions to calculate the resistance in the different modes are given in

Table 6.2 of EN 1993-1-8.

1.2.2 Column web in transverse tension

The design resistance of an unstiffened column web in transverse tension is

given by expression 6.15 in EN 1993-1-8, and is simply the resistance of a

length of web, with a reduction factor for the interaction with shear in thecolumn web panel. For bolted connections, § 6.2.6.3(3) states that the length of

web to be assumed at each row, or for each group of rows, is equal to the

length of the equivalent T-stub determined for that row (or group of rows).

1.2.3 Beam web in tension

The design resistance for a beam web in tension is given by § 6.2.6.8 and is the

same as that for the column web in transverse tension, (see Section 1.2.2), but

without an allowance for shear. The length of the beam web in tension is taken

to be equal to the length of the equivalent T-stub determined for that pair (or

group) of bolts.

1.3 Plastic distributionA plastic distribution of forces in bolt rows is permitted, but this is only

possible if the deformation of the column flange or end plate can take place.

This is ensured by placing a limit on the distribution of bolt row forces if the

critical mode is mode 3, because this failure mode is not ductile.

According to § 6.2.7.2(9) of EN 1993-1-8, this limit is applied if the resistance

of one of the previous bolt rows is greater than 1,9 F t,Rd, where:

F t,Rd is the tensile resistance of a single boltThe limit is applied by reducing the resistance of the row under consideration,

to a value F tr ,Rd , such that:

xRd tx,Rd ,t / hh F r r , where:

F tx,Rd is the design tension of the furthest row from the centre of

compression that has a design tension resistance greater than 1,9 F t,Rd

hx is the lever arm from the centre of compression to the row with

resistance F tx,Rd

hr

is the lever arm from the centre of compression to the row under

consideration.




11 - 5

The effect of this limitation is to apply a triangular distribution of bolt row

forces.

1.4 Resistance of the compression zone

1.4.1 General

The design resistance in the compression zone may be limited by:

The resistance of the column web ( F c,wc,Rd), or

The resistance of the beam (rafter) flange and web in compression ( F c,fb,Rd).

The relevant clauses of EN 1993-1-8 are given in Table 1.2.

Table 1.2 Joint Components in compression

Component EN 1993-1-8 clause number

Resistance of column web F c,wc,Rd 6.2.6.2

Resistance of the beam(rafter) flange and web

F c,fb,Rd 6.2.6.7

1.4.2 Column web without a compression sti ffener

Ideally, stiffeners in the column should be avoided, as they are expensive and

can be disruptive when making connections in the minor axis. However,

stiffeners in the compression zone of a column are usually required, especially

in a portal frame eaves connection. In a portal frame, the bending moment is

large, producing a large compression force, and the column is usually an I-

section with a relatively thin web.

The design resistance of an unstiffened column web subject to transverse

compression is given by EN 1993-1-8, § 6.2.6.2. The design resistance is based

on an effective width of web in compression, with the web verified as a strut,

and with a reduction factor ω for shear and a reduction factor ρ for longitudinal

compressive stress in the column.

1.4.3 Column web with a compression sti ffener

The design resistance of a stiffened column subject to transverse compression

may be calculated in accordance with § 9.4 of EN 1993-1-5.

1.4.4 Beam (rafter) flange and web in compression

The compression resistance of the beam flange and adjacent web in

compression is given in § 6.2.6.7 of EN 1993-1-8 by:

fb

Rdc,Rd,fb,c,

t h

M F

where:

h is the depth of the connected beam




11 - 6

M c,Rd is the design moment resistance of the beam cross-section, reduced if

necessary to allow for shear, see EN 1993-1-1 § 6.2.5. For a haunched

beam, such as a rafter, M c,Rd may be calculated neglecting the

intermediate flange

t fb is the flange thickness of the connected beam.

For haunched beams, such as those commonly used for rafters in portal frames,

the depth h, should be taken as the depth of the fabricated section, and the

thickness t fb should be that of the haunch flange.

If the height of the beam (rafter + haunch) exceeds 600 mm the contribution of

the rafter web to the design compression resistance should be limited to 20%.

This means that if the resistance of the flange is fby,fbfb f bt then:

8,0

fby,fbfbRdfb,c,

f bt F

1.5 Resistance of the column web panel The resistance of the column web panel is given by § 6.2.6.1 of EN 1993-1-8,

which is valid for 69wt d .

The resistance of an unstiffened column web panel in shear, V wp,Rd is given by:

M0

vcwcy,Rd wp,

3

9,0

A f

where:

Avc is the shear area of the column, see EN 1993-1-1 § 6.2.6(3).

1.6 Calculation of moment resistanceHaving calculated potential resistances in the tension zone (Section 1.2), the

design resistance in the compression zone (Section 1.4) and the resistance of

the column web panel in shear (Section 1.5), the effective design resistances in

the tension zone may be determined.

According to EN 1993-1-8 § 6.2.7.2(7), the total design resistance in the

tension zone must not exceed the design resistance in the compression zone.

Similarly, the total design resistance in the tension zone must not exceed the

design resistance of the column web panel, modified by a transformation

parameter, . This is expressed as:

Rd wp,Rd t, V F

The transformation parameter, is taken from § 5.3(7), and may be taken from

Table 5.4 as 1.0 for one-sided connections.

If either the resistance in the column web panel or in the compression zone is

less than the total design resistance in the tension zone, the resistances in the

tension zone must be reduced.




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The resistance of the bolt row nearest the centre of compression is reduced as a

first step, and then the next row, until the total design resistance in the tension

zone is no more than the compression resistance, or the web panel shear

resistance. Reducing the bolt row resistance in this way is satisfactory, as the

design approach presumes a plastic distribution of bolt forces.

As an alternative to reducing the resistance in the tension zone, stiffeners can

be provided to increase the design resistance of the web panel in shear, and the

web in compression.

Once the effective design tension resistances have been calculated, by reducing

the potential resistances if necessary, the design moment resistance of the

connection can be calculated, as the summation of each bolt row tension

resistance multiplied by its lever arm from the centre of compression, i.e.:

r

r r F h M Rd,tRd j, (as given in § 6.2.7.2 of EN 1993-1-8)

The centre of compression is assumed to be in line with the centre of the

compression flange.

1.7 Weld designEN 1993-1-8 § 6.2.3(4) requires that the design moment resistance of the joint

is always limited by the design resistance of its other basic components, and

not by the design resistance of the welds. A convenient conservative solution is

therefore to provide full-strength welds to components in tension. When

components are in compression, such as the bottom flange of a haunch, it is

normally assumed that the components are in direct bearing, and therefore onlya nominal weld is required. If the joint experiences a reversed bending

moment, the weld will be required to carry some tension force, and this should

be considered.

1.7.1 Tension flange welds

The welds between the tension flange and the end plate may be full strength.

Alternatively, common practice is to design the welds to the tension flange for

a force which is the lesser of:

(a) The tension resistance of the flange, which is equal to bf t f f y (b) The total tension force in the top three bolt rows for an extended end plate

or the total tension force in the top two bolt rows for a flush end plate.

The approach given above may appear conservative, but at the ultimate limit

state, there can be a tendency for the end plate to span vertically between the

beam flanges. As a consequence, more load is attracted to the tension flange

than from the adjacent bolts alone.

A full strength weld to the tension flange can be achieved by:

a pair of symmetrically disposed fillet welds, with the sum of the throat

thickness equal to the flange thickness, or

a pair of symmetrically disposed partial penetration butt welds with

superimposed fillet welds, or




11 - 8

a full penetration butt weld.

For most small and medium sized beams, the tension flange welds will be

symmetrical, full strength fillet welds. Once the leg length of the required fillet

weld exceeds 12 mm, a full strength detail with partial penetration butt welds

and superimposed fillets may be a more economical solution.

1.7.2 Compression flange welds

Where the compression flange has a sawn end, a bearing fit can be assumed

between the flange and end plate and nominal fillet welds will suffice. If a

bearing fit cannot be assumed, then the weld must be designed to carry the full

compression force.

1.7.3 Web welds

It is recommended that web welds in the tension zone should be full strength. For beam webs up to 11,3 mm thick, a full strength weld can be achieved with

8 mm leg length (5.6 mm throat) fillet welds. It is therefore sensible to consider

using full strength welds for the full web depth, in which case no calculations

are needed for tension or shear.

For thicker webs, the welds to the web may be treated in two distinct parts,

with a tension zone around the bolts that have been dedicated to take tension,

and with the rest of the web acting as a shear zone.

Tension zone

Full strength welds are recommended. The full strength welds to the web

tension zone should extend below the bottom bolt row resisting tension by a

distance of 1,73 g /2, where g is the gauge (cross-centres) of the bolts. Thisallows an effective distribution at 60° from the bolt row to the end plate.

Shear zone

The resistance of the beam web welds for vertical shear forces should be taken

as:

P sw = 2 a f vw,d Lws

where:

a is the fillet weld throat thickness

f vw,d is the design strength of fillet welds (from EN 1993-1-8, § 4.5.3.3(2)) Lws is the vertical length of the shear zone welds (the remainder of the

web not identified as the tension zone).

1.8 Vertical shear Design for vertical shear is straightforward. Generally, the bolts at the bottom

of the connection are not assumed to be carrying any significant tension, and

are allocated to carry the vertical shear. The bolts must be verified in shear and

bearing in accordance with EN 1993-1-8 Table 3.4.




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1.9 StiffenersComponents of the joint may be strengthened by providing additional material,

although this means additional expense. Table 1.3 summarises the

opportunities to strengthen moment resisting joints. Types of stiffeners are

illustrated in Figure 1.3.

Table 1.3 Stif feners

Stiffener type Effect Comments

Compression stiffener Increases the resistance tocompression

Generally required in portal frameconnections.

Flange stiffener in thetension zone

Increases the bendingresistance of the column flange

Diagonal shearstiffener

Improves the column webpanel resistance and alsostrengthens the tension flange

A common solution – connections onthe minor axis may be morecomplicated.

Supplementary webplate

Increases the column webresistance to shear andcompression

Minor axis connections are simplified.Detail involves much welding. See§6.2.6.1 of EN 1993-1-8.

End plate stiffener Increases the bendingresistance of the end plate

Should not be used – a thicker endplate should be chosen.

Cap plate Increases the bendingresistance of the flange, andthe compression resistance(in reversed momentsituations)

Usually provided in the column,aligned with the top flange of therafter. Generally provided for thereversal load combination, buteffective as a tension stiffener to thecolumn flange.

Flange backing plate Increases the bendingresistance of the flange

Only effective to increase mode 1behaviour. See EN 1993-1-8, §6.2.4.3

1 1

2

3

4

5

6

1 Compression stiffener2 Column flange stiffener

3 Cap plate

4 Shear stiffener5 Supplementary web plate

6 End plate stiffener

Figure 1.3 Types of sti ffeners




11 - 10

2 JOINT STIFFNESS

EN 1993-1-8 § 5.2 requires that all joints are classified, by strength or by

stiffness. Classification by strength is appropriate for plastic global analysis.

According to § 5.2.2.1(1), a joint may be classified according to its rotational

stiffness, which should be calculated in accordance with the method described

in Section 6.3 of EN 1993-1-8. It is recommended that software is used to

calculate the initial joint stiffness. An introduction to the approach is given in

Section 2.1.

In § 5.2.2.1(2) it is noted that joints may be classified on the basis of

experimental evidence, experience of previous satisfactory performance in

similar cases or by calculations based on test evidence. Some countries will

accept classification on the basis of satisfactory performance – this may even

be confirmed in the National Annex, which may point to nationally accepted

design methods or joint details, and allow these to be classified without

calculation.

2.1 Classif ication by calculationIn § 6.3.1(4) the initial stiffness, S j is given as:

ii

2

j 1

k

Ez S

where:

E is the modulus of elasticity

is a stiffness ratio that depends on the ratio of the applied moment to

the moment resistance of the joint

z is the lever arm, given by § 6.2.7

k i is the stiffness of the basic joint component

2.1.1 Stiffness of basic joint componentsTable 6.10 of EN 1993-1-8 identifies the basic joint components to be

considered. For a one-sided bolted end plate connection, such as in a portal

eaves frame, the basic joint components to be considered are given in

Table 2.1.




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Table 2.1 Basic joint components in a portal frame eaves connection

Stiffness coefficient Joint component

k 1 column web panel in shear

k 2 column web in compression

k 3

column web in tension

k 4 column flange in bending

k 5 end plate in bending

k 10 bolts in tension

For a joint with two or more rows of bolts, the basic components for each row

should be represented by a single equivalent stiffness, k eq. For a beam-to-

column joint with an end plate connection, this equivalent stiffness is

determined using k 3, k 4, k 5 and k 10 for each individual bolt row, and an

equivalent lever arm. (see EN 1993-1-8, § 6.3.3.1(4)).

Table 6.11 of EN 1993-1-1 indicates how the individual stiffness coefficients

should be determined.

2.2 Classif ication boundariesClassification boundaries are given in EN 1993-1-8 § 5.2.2.5. They depend on

the initial stiffness, S j,ini, the second moment of area of the beam, I b, the length

of the beam, l b and a factor, k b that depends on the stiffness of the frame.

Joints are classified as rigid when b b bini j, l EI k S

Thus, for a given initial stiffness S j,ini, a minimum beam length, l b, may be

calculated such that the joint is classified as rigid. This is the basis for the

minimum lengths given in Section 4.




11 - 12

3 BEST PRACTICE GUIDELINES FOR MOMENT CONNECTIONS

Any moment-resisting connection will involve additional expense compared tosimple (shear only) details. Connections should be detailed to carry the applied

forces and moments in the most economical way. This may involve providing

larger member sizes, or changing the geometry of the connection, to reduce the

fabrication effort involved in fitting stiffeners.

The following Sections offer guidance on appropriate detailing.

3.1 Eaves haunchThe ‘haunch’ in a portal frame is usually taken to mean an additional triangular

cutting that is welded below the rafter beam at the connection to the column.

The length of the cutting will generally be around 10% of the span, or up to

15% of the span in the most efficient elastic designs. The haunch is generally

cut from the same section as the rafter, or a deeper and heavier section.

Pairs of haunch cuttings are fabricated from one length of member, as shown in

Figure 3.1. If the haunch is cut from the rafter section, the maximum depth of

the haunched section is therefore just less than twice the depth of the rafter

section. Deeper haunches require larger sections, or fabrication from plate.

Figure 3.1 Fabrication of haunch cuttings

3.2 End plateEnd plates are generally fabricated from S275 or S235 steel. For class 8.8 bolts

and S275 steel, the end plate thickness should be approximately equal to the

bolt diameter. Common thicknesses are:

20 mm thick when using M20 class 8.8 bolts

25 mm thick when using M24 class 8.8 bolts

The end plate should be wider than the rafter section, to allow a weld all

around the flanges. The end plate should extend above and below the haunched

section, to allow for the fillet welds. In the compression zone, the end plate

should extend below the fillet weld for a distance at least equal to the thickness

of the plate, as shown in Figure 3.2, to maximise the stiff bearing length when

verifying the column in compression.




11 - 13

t

> t p

t

> t p

Figure 3.2 End plate – compression zone

3.3 StiffenersThe various types of stiffener used in an eaves connection are shown in

Figure 1.3. A compression stiffener is usually provided. Other stiffeners should

be avoided if possible. Stiffeners to the end plate are never needed – a thicker

end plate can be chosen to increase the resistance. Column flange stiffeners are

used to increase the resistance of the connection. In preference to providing

stiffeners, increased resistance can also be achieved by:

Providing more bolt rows

Extending the end plate above the top of the rafter, as shown in Figure 3.3

Increasing the depth of the haunch

Increasing the weight of the column section.

21

1 Extended column – may require skew cut

2 End plate stiffener – not preferred

Figure 3.3 Extended end plate connect ion

3.4 BoltsBolts in moment connections are generally M20 or M24, class 8.8 or 10.9. In

some countries, class 8.8 is standard. Bolts should be fully threaded, which

means that the same bolts may be used throughout a building.

Bolts are generally set out at cross-centres (gauge) of 90 or 100 mm. The

vertical pitch is generally 70 to 90 mm. In some countries, common practice is

to have bolts regularly spaced over the complete depth of the connection. In

other countries there may be a significant distance between the ‘tension’ boltsand the ‘shear’ bolts. EN 1991-1-8 does not preclude either detail. Maximum

bolt spacings are given in the Standard to ensure components do not buckle




11 - 14

between connectors, but this behaviour does not occur in end plate

connections.

Preloaded bolts are not required in portal frame connections.

3.5 Apex connectionsA typical apex connection is shown in Figure 3.4. Under gravity loads the

bottom of the haunch is in tension. The haunch may be fabricated from the

same section as the rafter, or may be fabricated from plate.

Figure 3.4 Typical apex connection

For modest structures and small bending moments, the apex detail may simply

have a stiffening plate, as shown in Figure 3.5, rather than a flanged haunch.

Figure 3.5 Alternative apex detail

3.6 WeldsAs described in Section 1.7, full strength welds are generally required to the

tension flange and adjacent to the tension bolts, as shown in Figure 3.6 for the

eaves connection. The remainder of the weld to the web is designed to carry

shear. Although the ‘shear’ web welds may be smaller than those in the tension

zone, it is common practice to continue the same size weld for the full depth of

the web.

In the compression zone, assuming that the ends of the member have been

sawn, the components are in direct bearing and only a nominal weld is

required. For the reversed moment design situation (with uplift due to wind),

the welds at the bottom of the eaves haunch and at the top of the apex

connection are in tension, and the welds should be verified for adequacy under

this combination of actions.




11 - 15

1 nominal weld (but verified for tension when moment is reversed)

2 continuous fillet weld3 full strength weld

Figure 3.6 Haunch welds

The weld between the haunch cutting and the underside of the rafter is

generally a continuous fillet weld. Although an intermittent weld would be

perfectly adequate structurally, it is usually more convenient to provide a

continuous weld.

1

2

3




11 - 16

4 JOINT DESIGN TABLES

4.1 General

This Section gives design tables for several typical configurations of momentconnections in portal frames. It covers both eaves and apex connections.

Three basic profiles are covered: IPE 300, IPE 400 and IPE 500, in steel grades

S235, S275 and S355. The profile sizes are generally those appropriate to span

lengths of 20, 25 and 30 m respectively.

For each profile, three configurations of apex connections are tabulated for a

typical bolt size and end plate thickness, and three configurations of eaves

connections are tabulated for the same typical bolt size and end plate thickness.

For each profile there are two additional tables, one for a different bolt class

and the other for a different end plate thickness. These two additional tables areonly given for apex connections without external bolts and for eaves

connections with half haunch. Tables 4.1 and 4.2 give the table numbers of all

the configurations.

Table 4.1 Apex connections

ProfileEnd plate

t p (mm)Bolt size

Bolt class

Without external bol ts

With external bolts

With external bolts and stiffener

IPE 300 15 M16 8.8 Table 4.10 Table 4.13 Table 4.14

15 10.9 Table 4.11

20 8.8 Table 4.12


20 10.9 Table 4.16

25 8.8 Table 4.17


25 10.9 Table 4.21

20 8.8 Table 4.22

Table 4.2 Eaves connections

Profile End platet p (mm) Bolt size Bolt class Haunch(a) ½ haunch(b) No haunch


15 10.9 Table 4.26

20 8.8 Table 4.27


20 10.9 Table 4.31

25 8.8 Table 4.32


25 10.9 Table 4.36

20 8.8 Table 4.37

(a) The depth of the haunched beam is twice the depth of the basic profile

(b) The depth of the haunch beam is 1,5 times the depth of the basic profile




11 - 17

In Tables 4.10 to 4.39, the following information is given:

A detailed sketch of the connection

The basic parameters (profile, bolt size, bolt class, end plate thickness)

The main design resistances (moment resistance, axial resistance, shear

resistance).

The tables provide the following results:

The design moment resistance M j,Rd+ for positive moment

The minimum span length L b,min for the connection to be considered as

‘rigid’, for positive moment

The design moment resistance M j,Rd – for negative moment

The minimum span length L b,min for the connection to be considered as

‘rigid’, for negative moment

The design axial resistance N t,j,Rd for tension

The design axial resistance N c,j,Rd for compression

The maximum shear resistance V j,Rd for which no interaction with bending

moment needs to be considered.

When a connection is subjected to a bending moment M Ed and an axial force

N Ed, a linear interaction criterion should be applied from the above mentioned

resistances:

N Ed

/ N j,Rd

+ M Ed

/ M j,Rd

≤ 1,0

The interaction should use the appropriate design resistances, in the same

direction as the internal forces:

N t,j,Rd or N c,j,Rd for the axial force (tension or compression)

M j,Rd+ or M j,Rd

– for the bending moment (positive or negative)

4.2 Main design assumptionsThe tables have been prepared using the PlatineX software available on the

web site www.steelbizfrance.com. This software can be freely used online and

allows the designer to deal with any configuration of connections – apex or

eaves connection.

The tables are based on the following design assumptions:

Calculations according to EN 1993-1-8

S235 end plate and stiffeners with S235 members, S275 otherwise

Bolt classes 8.8 and 10.9

Partial factors M as recommended (not to any particular National Annex).




11 - 18

Sign convention:

The bending moment is positive when it generates compression stresses in the

lower flange and tension stresses in the upper flanges (Figure 4.1).

IPE 300 M >0 IPE 300IPE 300 M >0

Figure 4.1: Sign convention fo r bending moment

4.3 Notes to the tables

4.3.1 Apex connections

Tables 4.4 to 4.6 summarize the design moment resistances for the apex

connections subject to positive moments. They can be compared with the

plastic moment resistance of the cross-section (Table 4.3).

Table 4.3 Plastic moment resistance of the cross section (kNm)

Prof ile S235 S275 S355

IPE 300 148 173 223

IPE 400 307 359 464

IPE 500 516 603 779

Bolts outside the profile have a major influence on the moment resistance when

they are in tension. The stiffener welded to the tension flange always increases

the moment resistance, but not to the same degree.

The moment resistance is lower than the plastic moment of the cross-section.

However this is not a problem since the member resistance is usually reduced

by the buckling effects, including lateral-torsional buckling.

The minimum span length to consider the apex connection as fully rigid is

relatively low. In practice, these connections will always be used for portal

frames with a span length greater than this minimum value, and so can be

considered rigid.

At the apex, the shear force is small and this verification will never be critical

in common practice.




11 - 19

Table 4.4 Apex connections with S235 beams – Moment resistance (kNm)

ProfileEnd plate

t p (mm)Bolt size

Bolt class


With external bolts


IPE 300 15 M16 8.8 75,4 118 123

15 10.9 86,3

20 8.8 78,4

IPE 400 20 M20 8.8 189 258 269

20 10.9 210

25 8.8 197

IPE 500 25 M24 8.8 358 449 472

25 10.9 363

20 8.8 340


ProfileEnd plate

t p (mm)Bolt size

Bolt class


With external bolts


IPE 300 15 M16 8.8 78,4 123,5 132,8

15 10.9 91,7

20 8.8 78,4

IPE 400 20 M20 8.8 199,7 284,3 301,2

20 10.9 231,0

25 8.8 199,7

IPE 500 25 M24 8.8 407,3 504,8 533,6

25 10.9 421,5

20 8.8 360,0


ProfileEnd plate

t p (mm)Bolt size

Bolt class


With external bolts


IPE 300 15 M16 8.8 78,4 123,5 132,8

15 10.9 91,7

20 8.8 78,4

IPE 400 20 M20 8.8 199,7 293,9 318,4

20 10.9 231,3

25 8.8 199,7

IPE 500 25 M24 8.8 426,3 577,1 620,4

25 10.9 479,4

20 8.8 360,0




11 - 20

4.3.2 Eaves connections

The minimum span length to consider the eaves connection as fully rigid is

relatively low when a haunch is provided, and in practice these connections

will always be used for portal frames with a span length greater than this

minimum value. The connections may therefore be considered as rigid.

Without a haunch, the bending resistance is lower and the connection might be

classified as semi-rigid. Therefore it is good practice to design the eaves

connections with a haunch, so that the overall depth is at least 1,5 times the

depth of the rafter.

The shear resistance of the column web is often the critical criterion.

For the eaves connections, the shear force is significant but the verification is

generally not critical for the design.

Table 4.7 Eaves connections (S235 members) – Moment resistances (kNm)

ProfileEnd plate

t p (mm)Bolt size

Bolt class

Haunch ½ haunch No haunch

IPE 300 15 M16 8.8 177,2 134,7 87,4

15 10.9 136,4

20 8.8 134,7

IPE 400 20 M20 8.8 388,0 291,2 186,6

20 10.9 293,9

25 8.8 291,2

IPE 500 25 M24 8.8 683,3 511,0 327,8

25 10.9 514,9

20 8.8 500,2


ProfileEnd plate

t p (mm)Bolt size

Bolt class


IPE 300 15 M16 8.8 204,1 154,3 98,9

15 10.9 158,2

20 8.8 154,3

IPE 400 20 M20 8.8 451,8 338,3 214,820 10.9 341,6

25 8.8 338,3

IPE 500 25 M24 8.8 795,8 593,9 379,0

25 10.9 599,2

20 8.8 580,9




11 - 21


ProfileEnd plate

t p (mm)Bolt size

Bolt class


IPE 300 15 M16 8.8 251,9 187,4 113,6

15 10.9 197,2

20 8.8 189,1

IPE 400 20 M20 8.8 564,0 417,5 258,2

20 10.9 435,2

25 8.8 420,8

IPE 500 25 M24 8.8 1000 739,7 462,3

25 10.9 763,7

20 8.8 716,4

4.4 Apex connections

IPE 300 M >0

Figure 4.2 Sign convention for bending moment in apex connections




11 - 22

Table 4.10 Apex connection – IPE 300

60

M16

60

75

150

300 IPE 3008.8

4

3303x70

15

15

6

8.5

15

Bolts M16 8.8

Hole diameter 18 mm

End plate t p =15 mm

Beam IPE 300 S235 S275 S355

Positive moment

Design moment resistance M j,Rd (kNm) 75,4 78,4 78,4

Minimum span length for ‘rigid’ Lb,min (m) 6,37

Negative moment



Design axial resistance

TensionN t,j,Rd (kN) 567 595 595

Compression N c,j,Rd (kN) 1264 1480 1710

Design shear resistance V j,Rd (kN) 135




11 - 23


60

M16

60

75

150

300 IPE 300

4

3303x70

15

15

6

8.5

10.9

15

Bolts M16 10.9

Hole diameter 18 mm


Beam IPE 300 S235 S275 S355

Positive moment



Negative moment










11 - 24


4

6

8.5

20

60

60

3x70 300 IPE 3008.8M16

75

150

15

15

330

Bolts M16 8.8

Hole diameter 18 mm


Beam IPE 300 S235 S275 S355

Positive moment



Negative moment










11 - 25


4

15

6

8.5

60

35

80

15

300 IPE 300

150

75

8.8

3x70

M16

70

385

Bolts M16 8.8

Hole diameter 18 mm


Beam IPE 300 S235 S275 S355

Positive moment



Negative moment










11 - 26


4

15

6

8.5

60

35

80

300 IPE 300

150

75

8.8

3x70

M16

70

385

158

Min =140

70

7.15

Bolts M16 8.8

Hole diameter 18 mm


Stiffeners t p =8 mm

Beam IPE 300 S235 S275 S355

Positive moment



Negative moment










11 - 27


5

15

7

9.9

8.8

400

75

75

4x70 430

1520

180

90

M20 IPE 400

Bolts M20 8.8

Hole diameter 22 mm


Beam IPE 400 S235 S275 S355

Positive moment



Negative moment










11 - 28


5

15

7

9.9

400

75

75

4x70 430

1520

180

90

M20 IPE 40010.9

Bolts M20 10.9

Hole diameter 22 mm


Beam IPE 400 S235 S275 S355

Positive moment



Negative moment










11 - 29


5

15

7

9.9

8.8

400

75

75

4x70 430

15180

90

M20 IPE 400

25

Bolts M20 8.8

Hole diameter 22 mm


Beam IPE 400 S235 S275 S355

Positive moment



Negative moment










11 - 30


5

15

7

9.9

75

4x70

20

180

90

M20 IPE 400

105

8.8

45

505

90

400

Bolts M20 8.8

Hole diameter 22 mm


Beam IPE 400 S235 S275 S355

Positive moment



Negative moment










11 - 31


5

15

7

75

4x70

20

180

90

M20 IPE 400

105

8.8

45

505

90 90

400

Min =180

9.9

10

68.5

Bolts M20 8.8

Hole diameter 22 mm



Beam IPE 400 S235 S275 S355

Positive moment



Negative moment










11 - 32


6

15

8.8

15

500

90

90

5x70M24

100

200

530

25

IPE 500

4

10.3

Bolts M24 8.8

Hole diameter 26 mm


Beam IPE 500 S235 S275 S355

Positive moment



Negative moment










11 - 33


6

15

15

500

90

90

5x70M24

100

200

530

25

IPE 500

4

10.3

10.9

Bolts M24 10.9

Hole diameter 26 mm


Beam IPE 500 S235 S275 S355

Positive moment



Negative moment










11 - 34


6

15

8.8

15

500

90

90

5x70M24

100

200

530 IPE 500

4

10.3

20

Bolts M24 8.8

Hole diameter 26 mm


Beam IPE 500 S235 S275 S355

Positive moment



Negative moment










11 - 35


6

15

8.8500

90

5x70M24

100

20025

IPE 500

4

10.3

625

110

130

55

Bolts M24 8.8

Hole diameter 26 mm


Beam IPE 500 S235 S275 S355

Positive moment



Negative moment










11 - 36


6

15

8.8500

90

5x70M24

100

200

25

IPE 500

4

10.3

625

110

130

55

12

110

6

8.5

Min =220

Bolts M24 8.8

Hole diameter 26 mm



Beam IPE 500 S235 S275 S355

Positive moment



Negative moment










11 - 37

4.5 Eaves connections

IPE 300IPE 300 M >0

Figure 4.3 Sign convention for bending moment in eaves connections




11 - 38

Table 4.25 Eaves connection – IPE 300

4

56

8.5

3

4.2

300

70

60

IPE 300

IPE 300

M16

7.1

80

35

80

10

10450

150

70

75

150

3x70

535

15

15

8.8

Bolts M16 8.8

Hole diameter 18 mm

Column stiffeners t p =10 mm


Column IPE 300 Beam IPE 300 S235 S275 S355

Positive moment



Negative moment










11 - 39


4

56

8.5

3

4.2

300

70

60

IPE 300

IPE 300

M16

7.1

80

35

80

10

10450

150

70

75

150

3x70

10.9 535

15

15

Bolts M16 10.9

Hole diameter 18 mm




Positive moment



Negative moment










11 - 40


4

56

8.5

3

4.2

300

70

60

IPE 300

IPE 300

M16

7.1

80

35

80

10

10450

150

70

75

150

3x70

535

15

8.8

20

Bolts M16 8.8

Hole diameter 18 mm




Positive moment



Negative moment










11 - 41


4

56

8.5

300 IPE 300IPE 300

7.1

80

35

10

10

70

75

150

3x70

15

8.8

60

M16

15

385

Bolts M16 8.8

Hole diameter 18 mm




Positive moment Design moment resistance M j,Rd (kNm) 87,4 98,9 113,6


Negative moment










11 - 42


4

56

8.5

3

4.2

300

IPE 300

IPE 300

7.1

80

35

80

10

10

855

70

75

150

3x70

3x70

15

15

55

285

670M168.8

Bolts M16 8.8

Hole diameter 18 mm




Positive moment



Negative moment










11 - 43


5

67

9.9

3

4.2

8.5

12

12

600

15

8.8

4590

4x70

105

400

70 200

705

180

90

105

100

M16

IPE 400

IPE 400

20

Bolts M20 8.8

Hole diameter 22 mm




Positive moment



Negative moment










11 - 44


5

67

9.9

3

4.2

8.5

12

12

600

15

4590

4x70

105

400

70200

705

180

90

105

100

M20

IPE 400

IPE 400

20

10.9

Bolts M20 10.9

Hole diameter 22 mm




Positive moment



Negative moment










11 - 45


5

67

9.9

3

4.2

8.5

12

12

600

15

8.8

4590

4x70

105

400

70 200

705

180

90

105

100

M20

IPE 400

IPE 400

25

Bolts M20 8.8

Hole diameter 22 mm




Positive moment



Negative moment










11 - 46


5

67

9.9

8.5

12

12

15

8.8

4590

4x70

105

400

180

90

IPE 400IPE 400

75

505M20

20

Bolts M20 8.8

Hole diameter 22 mm




Positive moment



Negative moment










11 - 47


5

7

9.9

3

4.2

12

121155

8.8

20

IPE 400

IPE 400

90

180

4590

105

4x70

4x70

105

15

890

385

400

75

M20

6

8.5

Bolts M20 8.8Hole diameter 22 mm




Positive moment



Negative moment










11 - 48


6

7

3

4.2

8.8

15

875M24

5x70

2x70

70

500

110

250

130

100

200

14

14

IPE 500

IPE 500

4

10.3

9.9

750

25

55

130

Bolts M24 8.8

Hole diameter 26 mm




Positive moment



Negative moment










11 - 49


6

7

3

4.2

15

875M24

5x70

2x70

70

500

110

250

130

100

200

14

14

IPE 500

IPE 500

4

10.3

9.9

750

25

10.9

55

130

Bolts M24 10.9

Hole diameter 26 mm




Positive moment



Negative moment










11 - 50


6

7

3

4.2

15

875M24

5x70

2x70

70

500

110

250

130

100

200

14

14

IPE 500

IPE 500

4

10.3

9.9

750

55

130

8.8

20

Bolts M24 8.8

Hole diameter 26 mm

Column stiffeners t p =14 mmEnd plate t p =20 mm


Positive moment



Negative moment











6

7

15

5x70 500

110

100

200

14

14

IPE 500 IPE 500

4

10.3

9.9

25

55

130

8.8

90

625M24

Bolts M24 8.8

Hole diameter 26 mm




Positive moment



Negative moment

